Ground Improvement Techniques: Select Proceedings of 7th ICRAGEE 2020 (Lecture Notes in Civil Engineering, 118) 9811599874, 9789811599873

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Lecture Notes in Civil Engineering

T. G. Sitharam C. R. Parthasarathy Sreevalsa Kolathayar   Editors

Ground Improvement Techniques Select Proceedings of 7th ICRAGEE 2020

Lecture Notes in Civil Engineering Volume 118

Series Editors Marco di Prisco, Politecnico di Milano, Milano, Italy Sheng-Hong Chen, School of Water Resources and Hydropower Engineering, Wuhan University, Wuhan, China Ioannis Vayas, Institute of Steel Structures, National Technical University of Athens, Athens, Greece Sanjay Kumar Shukla, School of Engineering, Edith Cowan University, Joondalup, WA, Australia Anuj Sharma, Iowa State University, Ames, IA, USA Nagesh Kumar, Department of Civil Engineering, Indian Institute of Science Bangalore, Bengaluru, Karnataka, India Chien Ming Wang, School of Civil Engineering, The University of Queensland, Brisbane, QLD, Australia

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T. G. Sitharam · C. R. Parthasarathy · Sreevalsa Kolathayar Editors

Ground Improvement Techniques Select Proceedings of 7th ICRAGEE 2020

Editors T. G. Sitharam Indian Institute of Technology Guwahati Guwahati, Assam, India

C. R. Parthasarathy Sarathy Geotechnical Engineering Bengaluru, Karnataka, India

Sreevalsa Kolathayar Department of Civil Engineering National Institute of Technology Karnataka Mangalore, Karnataka, India

ISSN 2366-2557 ISSN 2366-2565 (electronic) Lecture Notes in Civil Engineering ISBN 978-981-15-9987-3 ISBN 978-981-15-9988-0 (eBook) https://doi.org/10.1007/978-981-15-9988-0 © The Editor(s) (if applicable) and The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 This work is subject to copyright. All rights are solely and exclusively licensed by the Publisher, whether the whole or part of the material is concerned, specifically the rights of translation, reprinting, reuse of illustrations, recitation, broadcasting, reproduction on microfilms or in any other physical way, and transmission or information storage and retrieval, electronic adaptation, computer software, or by similar or dissimilar methodology now known or hereafter developed. The use of general descriptive names, registered names, trademarks, service marks, etc. in this publication does not imply, even in the absence of a specific statement, that such names are exempt from the relevant protective laws and regulations and therefore free for general use. The publisher, the authors and the editors are safe to assume that the advice and information in this book are believed to be true and accurate at the date of publication. Neither the publisher nor the authors or the editors give a warranty, expressed or implied, with respect to the material contained herein or for any errors or omissions that may have been made. The publisher remains neutral with regard to jurisdictional claims in published maps and institutional affiliations. This Springer imprint is published by the registered company Springer Nature Singapore Pte Ltd. The registered company address is: 152 Beach Road, #21-01/04 Gateway East, Singapore 189721, Singapore

Preface

Ground improvement techniques are becoming more popular and relevant in recent times because of the scarcity of land which necessitates construction on weak soil. This book volume contains the latest research papers on ground improvement techniques for reduction of seismic hazard, slope stability, retaining walls, and laboratory and field testing, selected from the proceedings of 7th International Conference on Recent Advances in Geotechnical Earthquake Engineering and Soil Dynamics, 2021. We thank all the staff of Springer for their full support and cooperation at all the stages of the publication of this book. We do hope that this book will be beneficial to students, researchers, and professionals working in the field of earthquake hazards. The comments and suggestions from the readers and users of this book are most welcome. Guwahati, India Bengaluru, India Mangalore, India

T. G. Sitharam C. R. Parthasarathy Sreevalsa Kolathayar

v

Acknowledgements

We (editors) want to thank all the authors who have contributed to the book. We could bring this book out due to all the authors’ timely contribution and cooperation. We thank and acknowledge the service of the following reviewers for their valuable time and efforts. Ajay Chourasia, CSIR-CBRI Amarnath Hegde, Indian Institute of Technology Patna Amit Verma, IIT (BHU) Anil Cherian, Strainstall Anitha Kumari S. D., Ramaiah University of Applied Sciences Arvind Kumar Jha, Indian Institute of Technology (IIT) Patna Asha Nair, CMR Institute of Technology, Bengaluru Babloo Chaudhary, NITK Surathkal Bal Rastogi, Indian Society of Earthquake Science Chittaranjan Birabar Nayak, Vidya Pratishthan’s Kamalnayan Bajaj Institute of Engineering & Technology Dauji Saha, Bhabha Atomic Research Centre; Homi Bhabha National Institute Deepankar Choudhury, Indian Institute of Technology Bombay Dhanaji Chavan, IISc Dinesh, S. V., Siddaganga Institute of Technology Thumkur Gopal Santana Phani Madabhushi, University of Cambridge Jagdish Sahoo, IIT Kanpur Kalyan Kumar G., NIT Warangal Karthik Reddy Konala S. K., IIT Hyderabad Ketan Bajaj, Risk Management Solutions Manas Kumar Bhoi, PDPU Md Mizanur Rahman, University of South Australia Padmanabhan G, Indira Gandhi Center for Atomic Research Pradeep Kumar Singh Chauhan, CSIR-Central Building Research Institute, Roorkee Premalatha Krishnamurthy, Anna University Prishati Raychowdhury, IIT Kanpur Purnanand Savoikar, Goa Engineering College vii

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Acknowledgements

Rajib Saha, NIT Agartala Rajib Sarkar, IIT(ISM) Dhanbad Ramkrishnan R., Amrita Vishwa Vidyapeetham Rangaswamy K., NIT Calicut Ravi K., IIT Guwahati Renjitha Varghese, National Institute of Technology, Calicut, Kerala, India Sanjay Verma, Indian Geotechnical Society, Jabalpur Chapter Sarat Kumar Das, Indian Institute of Technology (ISM) Dhanbad Shreyasvi C., National Institute of Technology Karnataka Snehal Kaushik, Girijananda Chowdhury Institute of Management and Technology, Guwahati Supriya Mohanty, Indian Institute of Technology (BHU), Varanasi Surya Muthukumar, Amrita School of Engineering, Amrita Vishwa Vidhyapeetham Vinay Srivastava, Retd. IIT ISM Dhanbad Vipin K. S., Swiss Re Visuvasam Joseph Antony, Vellore Institute of Technology

Contents

Application of Steel Fiber in Soil Stabilization . . . . . . . . . . . . . . . . . . . . . . . . Bharat Singh Chauhan, Bablu Kirar, and Bhagyachand Prajapati

1

Investigations on Mechanical Properties of Bio-Stabilized Soil . . . . . . . . . V. Divya and M. N. Asha

13

Strength Behavior of Sand Reinforced with Treated Sisal Fibers . . . . . . . C. Jairaj and M. T. Prathap Kumar

23

Unconfined Compressive Strength of MICP-Treated Black Cotton Soil . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . R. B. Wath and S. S. Pusadkar Potential of Reuse Options of Rice Husk Ash in Various Applications . . . Nazeema Basheer, Jaskiran Sobti, and Nazra Khanam

33 41

Effect of Sulphate Contamination on Lime-Stabilized Black Cotton Soil . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Shivanshi, Arvind Kumar Jha, Ankush Kumar Jain, and M. Parwez Akhtar

51

Influence of Geofoam Infill Trenches in Attenuation of Ground Vibrations Induced During Dynamic Compaction . . . . . . . . . . . . . . . . . . . . Saptarshi Kundu and B. V. S. Viswanadham

63

A Short Review on Improvement in Soil and Cement Properties by Addition of Bagasse Ash . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Furkhanda Khalid Khan, Sadia Kantroo, and Jaskiran Sobti

73

Enhanced Geotechnical Properties of High Plastic Clay Stabilized with Industrial Waste–Step for Sustainable Development . . . . . . . . . . . . . . Khushbu Gandhi and Shruti Shukla

85

Strength Behaviour of Marine Clay Stabilized with Marble Dust and Quarry Dust . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . B. Manjuladevi and H. S. Chore

99

ix

x

Contents

Effect of Influencing Parameters on the Failure Mechanism of Sand Compaction Column Treated Cohesionless Deposit . . . . . . . . . . . . 111 N. Aarthi Accelerating the Consolidation of Soft Clay Soil Using Pervious Concrete Pile . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 123 U. Umanath, A. R. Nandhagopal, and K. Muthukkumaran Bearing Capacity Analysis of Bagasse Ash Reinforced with Polypropylene Fiber . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 133 Rohin Kaushik, Rajiv Kumar, Ranjodh Singh, and Tarun Sharma Settlement Analysis of Locally Available Soft Soil Reinforced with Prefabricated Vertical Drain . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 145 Ajoy Saha, Rai Bahadur Reang, and Sujit Kumar Pal Preliminary Design of Prefabricated Vertical Drains-Embedded Soft Soils in the Field: An Example . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 155 Rai Bahadur Reang, Sujit Kumar Pal, and Sanjay Paul Ground Improvement of Soft Soil Using Bamboo as Flexible Pile with Stabilized Lime-Soil Pile-Cap Encapsulated by Geogrid and Geotextile . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 165 Malay Kumar Deb, Partha Ghosh, Debjit Bhowmik, Lipika Halder, and Sankar Chakraborti Effect of Reinforcement Form on Bearing Capacity of Sand . . . . . . . . . . . 179 B. Venkatesh and T. Thyagaraj Deformation of Stone Column Subjected to Earthquake Loading by Numerical Analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 187 Maniam Rajan Priyadharshini and Premalatha Krishnamurthy Ground Improvement Technique By Psuedo-Binghamian Grout for Fly Ash and Black Cotton Soil . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 201 Nirali B. Hasilkar, Lalit S. Thakur, and Atul Panchal Ground Improvement for Liquefaction Mitigation of Sand Deposits in Southern Dubai . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 215 B. Soundara and S. Bhuvaneshwari Shake Table Studies to Assess the Effect of Reinforced Backfill Parameters on Dynamic Response of MSE Walls . . . . . . . . . . . . . . . . . . . . . 227 Tirtha Sathi Bandyopadhyay, Pradipta Chakrabortty, and Amarnath Hegde Efficacy of Consolidation Grouting in Improving Dynamic Characteristics of a Nuclear Facility Foundation Strata . . . . . . . . . . . . . . . 239 G. Padmanabhan, R. Mano, Sudipta Chattopadhyay, L. Davy Herbert, V. Manoharn, and BPC Rao

Contents

xi

Liquefaction Mitigation of Silty Sands Using Xanthan Biopolymer . . . . . 247 S. Smitha, K. Rangaswamy, and P. Balaswamy Naik Ground Improvement with Vibro Compaction Method to Mitigate the Liquefaction Potential, Case Study . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 257 Houman Soleimani Fard, Tammo Köhler, and Adhila Haris A Study on Stability of Pond Ash Embankments Improved by Stabilizers . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 275 Gaurav Sharma, Koushik Pandit, and Pradeep Kumar Effect of Chemical Admixtures on Swelling Behavior of Expansive Soil . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 287 Prashant Sudani and Nishi A. Patel Laboratory Liquid Limit Determination Comparison of Expansive Clays by Casagrande and Cone Penetration Method . . . . . . . . . . . . . . . . . . 297 Rohini Chhatrapati Kale and K. Ravi Laboratory Swell Pressure Determination of Expansive Clays . . . . . . . . . 309 Kapil Bhanwariwal and K. Ravi Lateral Load Carrying Mechanism of Fibre Reinforced Concrete Pile in Sandy Soil . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 317 Koushik Sukla Das, Plaban Deb, and Sujit Kumar Pal Centrifuge Modelling of Marginal Soil Slopes Under Rainfall with Hybrid Geosynthetic Inclusions . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 329 Dipankana Bhattacherjee and B. V. S. Viswanadham Study of Geosynthetic Reinforced Retaining Wall under Various Loading . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 339 Ratnesh Ojha, Ananya Srivastava, and Vinay Bhushan Chauhan Numerical Analysis on the Effect of Compaction Induced Stresses on the Performance of MSE Walls . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 353 Akhil B. Alex and Renjitha Mary Varghese Evaluation of Moisture Susceptibility of Pyro-Oil Modified Bitumen by Surface Free Energy Approach . . . . . . . . . . . . . . . . . . . . . . . . . . 363 Shubham D. Suryawanshi, Hemantkumar P. Hadole, and M. S. Ranadive Investigating the Potential of Xanthan Gum for Aeolian Erosion Mitigation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 379 Anant Aishwarya Dubey, Navdeep Kaur Dhami, Abhijit Mukherjee, K. Ravi, and Rituraj Devrani Feasibility of Geopolymer Grout for Granular Soil . . . . . . . . . . . . . . . . . . . 387 Hiral Modha and Hasan Rangwala

About the Editors

Prof. T. G. Sitharam is a KSIIDC Chair Professor in the area of Energy and Mechanical Sciences and Senior Professor at the Department of Civil Engineering, Indian Institute of Science, Bengaluru (IISc). He was the founder Chairman of the Center for Infrastructure, Sustainable Transport and Urban Planning (CiSTUP) at IISc, and is presently the Chairman of the AICTE South Western Zonal Committee, Regional office at Bengaluru and Vice President of the Indian Society for Earthquake Technology (ISET). Prof Sitharam is the founder President of the International Association for Coastal Reservoir Research (IACRR). He has been a Visiting Professor at Yamaguchi University, Japan; University of Waterloo, Canada; University of Dolhousie, Halifax, Canada; and ISM Dhanbad, Jharkhand, and was a Research Scientist at the Center for Earth Sciences and Engineering, University of Texas at Austin, Texas, USA until 1994. Dr. C. R. Parthasarathy holds B.E. and M.E. degrees in Civil/Geotechnical Engineering from Bangalore University and Ph.D. (Geotechnical) from Indian Institute of Science, Bangalore, India. He has been involved in numerous large scale site investigation/engineering studies at various levels for multi-storeyed complexes, industrial buildings, embankments, bridges, water-retaining structures, subways, pavements, offshore platforms, pipelines, mobile drilling units, etc. in India, South East Asia and Middle East, Africa and USA. Dr. Parthasarathy is the founder-director of Sarathy Geotech & Engineering Services Pvt Ltd., a company which provide both offshore and on-land integrated geotechnical engineering services in India and abroad. As a technical partner of Pile Dynamics USA, he is instrumental in promoting quality testing of deep foundations in India and authorised trainer for high strain dynamic pile testing. He is the life fellow of Indian Geotechnical Society and member of several other professional bodies like DFI, ISRM, Indian concrete Institute, Institution of Engineers, etc. He is currently Elected National Executive Member of Indian Geotechnical Society (IGS) and Chairman, Indian Geotechnical Society-Bengaluru Chapter.

xiii

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About the Editors

Dr. Sreevalsa Kolathayar pursued M.Tech. from IIT Kanpur, Ph.D. from Indian Institute of Science (IISc) and served as International Research Staff at UPC BarcelonaTech Spain. He is presently Assistant Professor in the Department of Civil Engineering, National Institute of Technology, Karnataka. Dr. Kolathayar has authored three books and over 65 research papers. His broad research areas are geotechnical earthquake engineering, geosynthetics & geonaturals, and water geotechnics. He is currently the Secretary of the Indian chapter of International Association for Coastal Reservoir Research (IACRR), and Executive Committee Member of Indian Society of Earthquake Technology. In 2017, The New Indian Express honored Dr. Kolathayar with 40 under 40 - South India’s Most Inspiring Young Teachers Award. He is the recipient of ISET DK Paul Research Award from Indian Society of Earthquake Technology, IIT Roorkee. He received “IEI Young Engineers Award” by The Institution of Engineers (India), in recognition of his contributions in the field of Civil Engineering.

Application of Steel Fiber in Soil Stabilization Bharat Singh Chauhan, Bablu Kirar, and Bhagyachand Prajapati

Abstract Soil stabilization is a well-known technique used for improving the soil properties, it helps in increasing the shear strength of the soil and controlling soil swell and shrink properties, thus reducing the soil settlement. The use of steel products in our life is increasing day by day hence the production of steel products is also increasing. The industries producing steel products produce a lot of waste material; this waste material is no longer usable because if waste material is to be reused then industries have to melt it, which is not cost-effective. This study proposes a partial solution to this problem by deploying steel fiber (waste) as a stabilizer for soil. This study focuses on obtaining the optimum length of steel fiber for obtaining the maximum strength of the soil. Literature review suggests that soil stabilization using steel fiber has not been explored to the fullest and needs more attention from the research community. The experimentation is undertaken for steel fibers having three different ratios of length to diameter (L/D). Locally available black cotton soil in the region was used for experimentation, the results obtained displayed variation in soil characteristics for different L/D ratios of steel fiber in the form of X–Y plots. The tests were undertaken in accordance with the Indian Standard Codes of Practice. It was found that the effect of steel fiber is significant on the shear strength of the soil. Thus the effect of steel fibers in the improvement of properties of black cotton soil is investigated and the feasibility of steel fibers for the ground is explored. Keywords Steel fiber · Black cotton soil · Shear strength · Soil stabilizationss

B. S. Chauhan · B. Kirar (B) · B. Prajapati Department of Civil Engineering, SATI, Vidisha, India e-mail: [email protected] B. S. Chauhan e-mail: [email protected] B. Prajapati e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_1

1

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B. S. Chauhan et al.

1 Introduction Several times soil encountered at the site by civil engineers is not appropriate for building structures and hence stabilization of the soil is necessary for enhancing the soil properties. Soil stabilization increases the shear strength of soil and controls its shrink and swell properties, thus improving the strength and reduction in settlement. Soil stabilization methods include Mechanical stabilization, Lime stabilization, Cement stabilization, Bituminous stabilization, Thermal stabilization, Chemical stabilization, Electrical stabilization, stabilization by Geotextile and Fabrics, Stabilization by grouting. But all these methods mentioned incorporate a significant cost of stabilization which can affect the project cost substantially. However, in the recent years stabilization of soil with different types of fibers have been attempted and if these fibers are a part of waste generated by industries, then cost of soil stabilization can be reduced. Literature reveals that several studies for stabilization of soil have been undertaken. Yesilbas [1] studied the effect of using rock powder and aggregate waste with lime in reducing the swelling potential. The expansive soil used in this study is prepared in the laboratory by mixing kaolinite and bentonite. Yuan [2] carried out soil modification and stabilization with combined ground granulated blast furnace slag (GGBS), carbide lime (CL), and ordinary Portland cement (OPC) in this study in order to create a sturdy and stable platform for pavement. Sina and Bujang [3] attempted Soil stabilization by admixture developed in Japan during the 1970s and 1980s. The treated soil was found to have greater strength, reduced compressibility, and lower hydraulic conductivity than the original soil. In 2013, Vysakh and Bindu [4] analyzed the suitability of coconut shell, husk, and leaf ash to stabilize the lateritic soil. In 2015, Mohanty [5] treated black cotton soil obtained from Nagpur with an additive like fly ash obtained from Sesa Sterlite, Jharsuguda, Odisha, with various proportions of this additive, i.e., 10%, 20%, 30%, 40%, and 50%. In 2015, Joe and Rajesh [6] carried out a study dealing with stabilization of soil using industrial waste sand which can no longer be reused in the industry. Other ingredients used include are copper slag, cement, and lime. Ashraf [7] investigated the improvement in physical and mechanical properties of the peat soil by the addition of two creative mixtures. The first mixture consisted of clayey diatomite which is widespread in Egypt, calcium carbonate, lime, and water, while the second mixture has the same ingredients as the first one except clayey diatomite is replaced with cement. Second mixture showed significantly higher improvement than the first one. In 2017, Hatmoko and Suryadharma [8] undertook research to improve the physical and mechanical behavior of expansive soils stabilized with waste materials calcium carbide residue (CCR) and bagasse ash (BA), the study indicated that at 8% of CCR and 9% of BA and on 28 day curing time the maximum dry density (MDD) and Unconfined compressive strength (UCS) were significantly increased. Saikrishnamacharyulu et al. [9] carried out a study to investigate the improvement in strength of soil by reinforcing it with randomly distributed waste coir fiber materials and tire waste with varying percentages of reinforcement. Mirzababaei et al. [10] in 2018 carried out a study

Application of Steel Fiber in Soil Stabilization

3

to understand the effect of the combined addition of fibers and a nontraditional polymer on the mechanical behavior of clay. Poly vinyl alcohol (PVA) used as a solution with concentrations of 0.1%, 0.3%, 0.5%, 1.0%, and 1.5% and Butane Tetra Carboxylic Acid (BTCA) was added as a crosslinking agent at concentration rates of 0.1%, 0.3%, and 0.5%, respectively. Short polypropylene fibers were added to the clay as 0.25% and 0.50% of the dry weight of the soil. The results give significant UCS improvements with combined fiber reinforcement and PVA-BTCA stabilization when samples were cured for 14 days. In 2018, Wei et al. [11] undertook a study, where the soil was strengthened with both lime and fibers to investigate the mechanical properties of soil. For this they used four type of fiber wheat straw, rice straw, jute and polypropylene fiber, these fibers were added to the soil and lime soil, respectively. The results showed optimal fiber content as 0.2% or 0.25%, and the optimal fiber length was 30% or 40% of the sample diameter. Reinforcement significantly increased the cohesion and lightly improved the internal frictional angle. All four kinds of fiber improved the strength of soil and lime soil, in which polypropylene fiber was found to be best for reinforcement. Rahgozar et al. [12] in 2018 undertook a study in which soil was stabilized with agricultural waste materials, such as rice husk ash (RHA). The aim of this study is to experimentally investigate the effects of adding RHA and ordinary Portland cement on the geotechnical properties of the clayey sampled from the Sejzi area, which lies east of the city of Isfahan, Iran. In 2018, Tran et al. [13] carried out a study focused on exploring the effects of corn silk fibers on the mechanical properties of cemented soil by conducting standard compaction, triaxial compression tests. The influences of fiber content (0%, 0.25%, 0.5%, and 1% by weight of dry soil), cement content (4%, 8%, and 12% by weight of dry soil), and curing time (7, 14, and 28 days) were investigated in the study. The experimental results revealed that the addition of corn silk fibers in cemented soil improved the compressive and tensile strength. Literature survey revealed that no study is available which focuses on improving the soil mechanical properties with the help of steel fibers. Hence, an attempt is undertaken here by authors to understand the effect of waste steel fiber as reinforcement for locally available black cotton soil. Several tests were conducted like water content test, consistency limit test, sieve analysis, specific gravity test, standard proctor test, CBR (California bearing ratio) test as per Indian standard specifications and optimum L/D ratio was determined.

2 Material and Methodology Used 2.1 Experimental Materials Soil. Commonly available black cotton soil in the Vidisha district of Madhya Pradesh state (India) is used for experimentation. At a depth of 0.5 m from the natural ground level, the soil sample is extracted from the field. Table 1 shows the obtained characteristic of the soil sample.

4 Table 1 Obtained characteristics of the soil sample

Table 2 Specifications of steel fiber used for experimentation

B. S. Chauhan et al. Properties Natural water content

Numerical value 8.58%

Liquid limit

42.05%

Plastic limit

23.42%

Plasticity index

18.63%

Specific gravity

2.40

Properties

Numerical value

Diameter (mm)

0.32

Shape

Round

Melting temperature

14,000 C

Surface

Smooth

Heat treatment

Annealed (soft)

Steel Fibers. Table 2 shows the specification of steel fiber used for experimentation as mentioned by the supplier. Figure 1 shows the photograph of steel fiber used for experimentation. Corrosion Resistance of Steel fiber. All steel are iron-based alloys that contain around 10.5% chromium. Chromium in the alloy forms a self-healing protective clear oxide layer. This oxide layer gives steel their corrosion resistance. The self-healing nature of the oxide layer means the corrosion resistance remains intact regardless of fabrication methods. Even if the material surface is cut or damaged, it will self-heal and its corrosion resistance would be maintained. Fig. 1 Steel fiber

Application of Steel Fiber in Soil Stabilization

5

2.2 Methods of Testing Preparation of sample. According to IS 2720 Part 1 [14], the collected natural soil is brought to the laboratory then its natural moisture content is determined by the oven-dry method [15]. Further, soil is spread on the ground and allows natural soil sample to dry at room temperature and break lumps present in soil sample to get the uniform size of soil, after drying the soil at room temperature for 24 h, now natural soil is ready for testing. Methodology. Several laboratory tests are performed firstly on natural (sample) soil which includes Natural moisture content [15], sieve analysis [16], Atterberg limits [17], specific gravity [18], modified proctor test [19], and California bearing ratio (CBR) test [20]. The California bearing ratio test is conducted at optimum moisture content. Later on, soil specimens are prepared by adding three different lengths of steel fibers, i.e., 2 cm (L/D = 62.50), 3 cm (L/D = 93.75) and 4 cm (L/D = 125) at a fixed percentage of 10% by weight of the soil, to find the optimum moisture content and maximum dry density at different lengths of fiber by modified proctor test. At a fixed percentage of 10%, California bearing ratio (CBR) test is then undertaken for the obtained optimum length of steel fiber. Few photographs of Laboratory work and tests undertaken are shown in Figs. 2, 3, 4, and 5. Figure 6 shows the test results for CBR undertaken for natural soil.

Fig. 2 Liquid limit test

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Fig. 3 Trimming the excess compacted soil Fig. 4 California bearing ratio test

B. S. Chauhan et al.

Application of Steel Fiber in Soil Stabilization

Fig. 5 Preparation of CBR specimen

Fig. 6 CBR graph for natural soil

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B. S. Chauhan et al.

3 Results and Discussion 3.1 Tests on Natural Soil Natural moisture content. Table 3 shows the moisture content of natural soil. Liquid limit. Table 4 shows the liquid limit determination of natural soil. Plastic limit. Table 5 shows the plastic limit of natural soil. Table 3 Natural moisture content Sample No

1

2

3

4

5

Mass of container W1 (gm)

20.19

24.53

20.66

22.77

23.92

Mass of container + wet soil, W2 (gm)

54.16

45.88

46.73

51.39

43.35

Mass of container + dry soil, W3 (gm)

50.54

44.51

44.42

49.46

42.04

Mass of water, WW = W2 –W3 (gm)

3.62

1.37

2.31

1.87

1.31

Mass of dry soil, WS = W3 –W1 (gm)

30.35

19.98

23.76

26.69

18.12

Moisture content w = (W2 –W3 )/(W3 –W1 ) *100

11.92

6.85

9.72

7.23

7.22

Average moisture content w %:- 8.58%

Table 4 Liquid limit determination Sample No

1

2

3

4

5

(1) No of blows

53

43

35

24

15

(2) Mass of container + wet soil (gm)

33.94

34.36

30.90

39.19

37.10

(3) Mass of container + dry soil (gm)

31.02

30.10

27.10

34.50

31.88

(4) Mass of water(2–3) (gm)

2.92

3.76

3.80

4.63

5.22

(5) Mass of container (gm)

23.06

20.59

17.55

23.55

20.53

(6) Mass of dry soil (3–5) (gm)

7.96

10.01

9.55

10.95

11.65

(7) Moisture content (4)/(6) * 100%

36.68

37.56

39.79

42.28

44.80

1

2

3

4

5

Liquid limit at 25 blows = 42.05%

Table 5 Plastic limit determination Sample No (1) No of blows

53

43

35

24

15

(2) Mass of container + wet soil (gm)

33.94

34.36

30.90

39.19

37.10

(3) Mass of container + dry soil (gm)

31.02

30.10

27.10

34.50

31.88

(4) Mass of water(2–3) (gm)

2.92

3.76

3.80

4.63

5.22

(5) Mass of container (gm)

23.06

20.59

17.55

23.55

20.53

(6) Mass of dry soil (3–5) (gm)

7.96

10.01

9.55

10.95

11.65

Average plastic limit = 23.42%

Application of Steel Fiber in Soil Stabilization

9

Table 6 Specific gravity of natural soil Sample No

1

2

3

(1) mass of pycnometer w1 (gm)

505.00

505.00

505.00

(2) mass of pycnometer + dry soil w2 (gm)

705.00

705.00

705.00

(3) mass of pycnometer + soil + water w3 (gm)

1410

1405

1405

(4) mass of pycnometer + water w4 (gm)

1290

1290

1290

(5) specific gravity

2.5

2.35

2.35

Average Specific gravity of soil at 270 C = 2.40, Hence, Soil Type is organic

Table 7 Optimum moisture content and maximum dry density Sample No

1

2

3

4

5

6

Mass of mold + compacted soil (gm)

5005

5020

5050

5010

4980

4920 1940

Mass of compacted soil wt (gm)

2025

2040

2070

2030

2000

Bulk density (Yt = wt /v) (g/cc)

2.02

2.04

2.07

2.03

2.0

1.94

Water content %

12.31

13.03

14.34

15.47

18.89

22.38

Dry density Yd = Yt / (1 + W) g/cc

1.78

1.80

1.81

1.76

1.68

1.58

Plasticity index (P.I.). P.I. = liquid limit—plastic limit, Hence, P.I. = 18.63%. Since P.I is greater than 17 Hence, it can be classified as soil type = clay. Degree of plasticity = high plasticity and degree of cohesiveness = cohesive. Specific gravity test. Table 6 shows the Specific Gravity of Natural Soil. Modified Proctor Test. Table 7 shows results obtained from the modified proctor test. California Bearing Ratio (CBR) Test on Natural Soil. Table 8 shows the CBR values of Natural Soil.

3.2 Test on Soil Sample with Steel Fiber Modified Proctor Test. This test is undertaken for variation in the length of steel fiber at a fixed percentage of 10% steel fiber in the soil by weight. It can be seen from Tables 9, 10, and 11 the optimum length of fiber is 3 cm. CBR Test on Soil + Steel Fiber. Table 12 displays the results for CBR test of soil stabilized with 10 percent steel fiber of 3 cm length. Table 13 shows a comparison of CBR test results for natural soil with soil stabilized with 10 percent steel fiber of 3 cm length.

10

B. S. Chauhan et al.

Table 8 CBR values for natural soil S. no

Dial gauge penetration (mm)

Proving ring reading

1

0

2

0.5

20

3

1.0

31

4

1.5

40

5

2.0

55

6

2.5

70

7

3.0

79

8

3.5

85

9

4.0

89

10

4.5

95

11

5.0

101

12

5.5

105

13

6.0

109

14

6.5

113

15

7.0

117

Stress (kg/cm2 )

CBR %

4.34

6.20

6.26

5.96

0

Table 9 OMC and MDD for 2 cm length of steel fiber and 10% steel Sample No

1

2

3

4

5

(1) Mass of mold + compacted soil (gm)

4910

5030

5175

5150

5120

(2) Mass of compacted soil w1 (gm)

1930

2050

2195

2170

2140

(3) Wet density (Yt = wt/v) (g/cc)

1.93

2.05

2.19

2.17

2.14

(4) Water content %

5.98

8.67

11.38

13.35

15.25

(5) Dry density Yd = Yt / (1 + W) g/cc

1.82

1.89

1.97

1.91

1.86

Table 10 OMC and MDD for 3 cm length of steel fiber and 10% steel Sample No

1

2

3

4

5

(1) Mass of mold + compacted soil (gm)

5030

5130

5270

5190

5150

(2) Mass of compacted soil w1 (gm)

2050

2150

2290

2210

2170

(3) Wet density (Yt = wt/v) (g/cc)

2.05

2.15

2.29

2.21

2.17

(4) Water content %

6.39

8.34

11.86

12.60

14.40

(5) Dry density Yd = Yt / (1 + W) g/cc

1.92

1.98

2.05

1.96

1.89

Application of Steel Fiber in Soil Stabilization

11

Table 11 OMC and MDD for 4 cm length of steel fiber and 10% steel Sample No

1

2

3

4

5

(1) Mass of mold + compacted soil (gm)

4990

5110

5200

5160

5105

(2) Mass of compacted soil w1 (gm)

2010

2130

2220

2180

2125

(3) Wet density (Yt = wt/v) (g/cc)

2.01

2.13

2.22

2.18

2.12

(4) Water content %

6.28

8.06

11.50

12.34

14.05

(5) Dry density Yd = Yt / (1 + W) g/cc

1.89

1.97

1.99

1.94

1.86

Table 12 CBR value for soil + 10% steel fiber of 3 cm length S. No

Dial gauge penetration(mm)

Proving ring reading

1

0

2

0.5

50

3

1.0

70

4

1.5

84

5

2.0

97

6

2.5

109

7

3.0

116

8

3.5

119

9

4.0

131

10

4.5

137

11

5.0

143

12

5.5

146

13

6.0

151

14

6.5

152

15

7.0

155

Stress (kg/cm2 )

CBR %

6.758

9.65

8.866

8.44

0

Table 13 CBR value comparison with natural soil and soil stabilized with steel fiber Penetration (mm)

CBR value natural soil

CBR value 10% steel fiber

Percentage change in CBR value

2.5

6.20%

9.65%

55.64%

5.0

5.96%

8.44%

41.61%

4 Conclusions • Three different lengths of steel fibers, viz., 2 cm, 3 cm, 4 cm having 0.32 mm diameter were added at affix percentage by weight of soil and the optimum length was found out to be 3 cm in attaining highest maximum dry density (MDD). • Largest change in maximum dry density (MDD) for 3 cm length fiber was found out to be 13.26% higher than that for natural black cotton soil.

12

B. S. Chauhan et al.

• In CBR test for the optimum length of 3 cm steel fiber percentage increase was found to be 55.64% and 41.61% for penetration of 2.5 mm and 5.0 mm in comparison to natural soil.

References 1. Yesilbas G (2004) Stabilization of expansive soils using aggregate waste, rock powder and lime. MSc Thesis, the Graduate School of Natural and Applied Sciences of the Middle East Technical University 2. Yuan X (2010) Silt subgrade modification and stabilization with ground granulated blast furnace slag and carbide lime in areas with a recurring high groundwater. International Conference on Mechanic Automation and Control Engineering, Wuhan, China 3. Sina K, Bujang H (2011) Assessment of stabilization methods for soft soils by admixtures. International Conference on Science and Social Research. pp 118–121 4. Vysakh P, Bindu J (2012) Stabilisation of lateritic soil using coconut shell, leaf and husk ash. International Conference on Green Technologies, Trivandrum, India 5. Kumar MM (2015) Stabilization of expansive soils using fly ash. M.Tech. thesis, Department of Civil Engineering National Institute of Technology, Rourkela, India 6. Joe MA, Rajesh AM (2015) Stabilization of soil using industrial waste and lime. Int J Sci Res Eng Technol 4(7) 7. Abdel-Salam AE (2018) Stabilization of peat soil using locally admixture. HBRC J 14(3):294– 299. https://doi.org/10.1016/j.hbrcj.2016.11.004 8. Hatmoko JT, Suryadharma H (2017) Shear behavior of calcium carbide residue-bagasse ash stabilized expansive soil” Procedia Eng 171:476–483 9. Saikrishnamacharyulu I, Kumar CV, Rao KV, KumarI GH (2017) Experimental study on soil stabilization using waste fibre materials. Int J Technol Res Eng 4(10) 10. Mirzababaei M, Arulrajah A, Horpibulsuk S, Soltani A, Navid K (2018) Stabilization of soft clay using short fiber and poly vinyl alcohol. Geotext Geomembr 46(5):646–655 11. Wei L, Chai SX, Zhang HY, Shi Q (2018) Mechanical properties of soil reinforced with both lime and four kind of fiber. Constr Build Mater 172:300–308 12. Rahgozar MA, Saberian M, Li J (2018) Soil stabilization with non–conventional eco-friendly agricultural waste material: an experimental study. Transp Geotech 14:52–60 13. Tran KQ, Satomi T, Takahashi H (2018) Improvement of mechanical behavior of cement soil reinforced with waste cornsilk fiber. Constr Build Mater 178:204–210 14. IS: 2720-part-1 (1983) Preparation of dry sample for various tests. Bureau of Indian Standards, New Delhi, India 15. IS: 2720-part-2 (1973) Determination of water content. Bureau of Indian Standards, New Delhi, India 16. IS: 2720-part-4 (1975) Grain size analysis. Bureau of Indian Standards, New Delhi, India 17. IS: 2720-part-5 (1970) Determination of liquid and plastic limits. Bureau of Indian Standards, New Delhi, India 18. IS: 2720-part-3 (1980) Determination of specific gravity. Bureau of Indian Standards, New Delhi, India 19. IS: 2720-part-8 (1983) Determination of water content–dry density relation using heavy compaction. Bureau of Indian Standards, New Delhi, India 20. IS: 2720-part-16 (1987) Determination of CBR value. Bureau of Indian Standards, New Delhi, India

Investigations on Mechanical Properties of Bio-Stabilized Soil V. Divya

and M. N. Asha

Abstract Bio-stabilization aims at enhancing the engineering properties of soils through enzymatic activity. The current work focuses to understand the modifications in mechanical properties of laterite soil due to enzymatic stabilization. Laboratory tests were conducted on oven-dried laterite soil to determine its basic properties. The different dosages of enzyme reflected in the present analysis are 100 ml/m3 and 200 ml/m3 , respectively. The mechanical properties investigated in the present work are shear strength and permeability characteristics since they represent significant parameters in the settlement and failure analysis of soil. The effect of microbial activity was also studied for curing periods of 0, 7 and 14 days on the treated soil. It was observed from the detailed experimental investigations that bio-stabilization helps in improving the Bearing ratio of enzyme-treated laterite soil. The experimental studies also indicated a decrease in permeability with an increase in dosage as a result of bio-stabilization. The biologically improved soil can find applications in many civil engineering projects such as canal linings, embankment construction, etc. Keywords Permeability · CBR · Terrazyme · Laterite

1 Introduction Soils often exhibit negative performance characteristics such as low compressive strength, excessive settlement and negligible bearing capacity, which results in significant damage to constructions in the vicinity and potential loss of human life. These problems may also pose a serious concern for the development and maintenance of different infrastructure. Soil stabilization can be defined as the modification of various characteristics of the soil, thereby improving its engineering behaviour V. Divya (B) Department of Civil Engineering (VTU RRC), CMR Institute of Technology, Bangalore, India e-mail: [email protected] M. N. Asha Department of Civil Engineering, CMR Institute of Technology, Bangalore, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_2

13

14

V. Divya and M. N. Asha

through mechanical, chemical or biological methods. The choice of a particular method depends mainly on the type of soil to be improved; its characteristics and the type and degree of improvement desired in a particular application. Many researchers have reported the improvement of ground using conventional materials such as lime, fly ash, coir-fibre, rice husk ash, quarry dust, etc. In recent years, stabilization of soil using waste plastics [1], construction demolition waste [2], geotextiles [3], etc. have also been explored by various scientists. The use of conventional methods was observed to pose several issues to the ecosystem. The decreasing availability and increasing cost of construction materials together with the drawbacks opposed by traditional methods have forced the researchers to work on sustainable stabilizing agents. As a result, recently, biological stabilization has been gaining importance due to its eco-friendly and sustainable nature. In this method, enzymes or micro-organisms are used as catalysts during the modification of soil properties. Bioenzymes are chemical, organic and liquid concentrated substances that are formulated from vegetable extracts and found to increase the stability of soil mass [4]. The bio-soil method is an interdisciplinary field consisting of collaboration with the studies of microbiology, geochemistry and civil engineering to find natural treatments for ground improvement. Different microorganisms that can be used are facultative anaerobic bacteria, microaerophilic bacteria, anaerobic fermenting bacteria, anaerobic respiring bacteria and obligate aerobic bacteria. The comprehensive literature review carried out indicates that the quantity of enzymes required for the stabilization is dependent on the soil type and structure. [5] observed that enzyme treatment changes the adsorbed water layer surrounding the clay particles. The metal ions are forced to move towards the free water from the adsorbed water layer and thereby break down of the electrostatic barrier occurs and hence makes the soil structure more friable. According to [6], bio-clogging and bio-cementation of soils could be used to improve the mechanical properties of soil in situ and thereby replace the existing mechanical compaction methods or the expensive and environmentally unfriendly chemical grouting methods. It was summarized that biological stabilization is feasible in soil with a minimum clay content of 15% by [7]. Experimental investigations were carried out on bioenzyme-stabilized lateritic soil blended with sand by [4]. The properties studied by them were Unconfined Compressive strength (UCC), California Bearing Ratio (CBR), compaction and permeability characteristics. The physical and chemical characteristics of enzymatic stabilized soil highlighted the low environmental and economic impact offered by biological treatment. The swelling behaviour of expansive soils treated with bio-enzyme was studied by [8] and confirmed the effectiveness of bio-enzyme in controlling the swelling of expansive soil on the dry side of optimum moisture content. In the current work, a change in mechanical characteristics of bio-enzyme-treated laterite soil is explored. Laterite soil, which is used in many construction activities, is observed to lose its shear strength on saturation with water. It is proposed to understand the modifications in bearing ratio values and permeability characteristics of Terrazyme-treated laterite soil. The significance of varying enzyme dosages,

Investigations on Mechanical Properties of Bio-Stabilized Soil

15

namely, 100 ml/m3 and 200 ml/m3 are also studied and compared in the present project.

2 Materials and Methodology Laterite Soil (LS) The laterite soil used in the current work is obtained from Hoskote at a depth of 1.5 m from the top surface. As the soil is rich in iron and aluminous oxides, it is red in colour. The identification of the soil was done by placing the soil lumps in water and observing the disintegration of the soil particles. Laterite soil is known to lose shear strength on saturation with water and hence requires modification of its properties. The soil from the field was oven dried and broken into fine particles before subjecting to experimental studies. Table 1 gives a summary of the properties of laterite soil. Terrazyme Terrazyme used in the current study, obtained from a commercial vendor, is brown in colour as shown in Fig. 1a. It is obtained from vegetable extracts and is completely soluble in water. They do not cause any irritation to the eyes or skin during handling and possess the smell of molasses. The enzyme has a specific gravity ranging from 1 to 1.09 and pH value varies between 4.30 to 4.60 (as per the vendor). Terrazyme is diluted in water at optimum moisture content before treating the laterite soil. Two dosages, namely, 100 ml/m3 and 200 ml/m3 , obtained by volume proportioning are used in the present study. Methodology Laboratory tests are conducted for the determination of basic properties of the laterite soil. The oven-dried soil sample is treated with Terrazyme at controlled conditions of maximum dry density and optimum moisture content. The enzyme-blended sample is kept for a curing period of 7 and 14 days and then tested for modification in mechanical properties. In order to ensure the controlled conditions, Table 1 Properties of untreated laterite soil

Properties

Values

Specific gravity

2.68

Liquid limit, %

37.6

Plastic limit, %

27.58

Plasticity index, %

9.03

Soil classification

MI

Maximum dry density, g/cc

1.71

Optimum moisture content, %

18.33

C.B.R, %

9.06

Unconfined compressive strength, KPa

115.14

Coefficient of permeability, cm/s

4.83 × 10–4

16

V. Divya and M. N. Asha

the treated samples are kept in airtight covers for curing. The mechanical properties investigated in the current work are California Bearing Ratio and permeability characteristics, as they play a significant role in the design of flexible pavements. Comparative studies are performed to evaluate the impact of varying dosages and curing periods on Terrazyme-treated laterite soil.

3 Experimental Studies Shear Strength Characteristics The shear strength of laterite soil was determined through bearing ratio test in the current work. The California bearing ratio (CBR) test is a penetration test meant for the evaluation of subgrade strength of roads and pavements. This test indicates the resistance offered by soil against deformation under load. In the current study, dynamic compaction was employed on unsoaked specimen for the determination of bearing ratio. Compressibility Characteristics Compressibility characteristics such as consolidation and permeability play an important role in the settlement and failure analysis of soil. A consolidation test was performed in the present study using the conventional method. The permeability of clayey soils is generally determined with the help of falling head permeability test. The conventional method of measuring permeability does not permit its evaluation under different pressure. In the present work, the permeability of treated and untreated soil was studied under sustained pressure which can be directly related to real time applications. The experimental set up for the current study is shown in Fig. 1b.

4 Results and Discussions 4.1 Variation in Bearing Ratio CBR test was conducted in the present study as per IS 2720-16 on treated and untreated soil in unsoaked condition. The samples were prepared at a maximum dry density and optimum moisture content and kept in airtight bags for curing period of 7 days. The variation of CBR values of lateritic soil treated with 100 ml/m3 and 200 ml/m3 for 7 days curing is represented in Fig. 2. It is observed from the experimental results that when compared with the untreated soil, there is an increase of 40% in the bearing ratio value of laterite treated with 100 ml/m3 and 95% increase when treated with 200 ml/m3 . Hence it is inferred that enzyme stabilization can be successfully applied for increasing the strength of pavement subgrade. The variation in values of bearing ratio due to Terrazyme treatment is given in Table 2.

Investigations on Mechanical Properties of Bio-Stabilized Soil

17

Fig. 1 a Terrazyme. b Experimental setup 7 6

Load, KN

5 4 3 Untreated Soil

2

LS+100ml/m3 1 LS+200ml/m3 0

0

5

10

15

20

Penetration, mm

Fig. 2 Variation in CBR of treated and untreated laterite soil Table 2 Bearing ratio values of Terrazyme treated laterite soil

Specification Untreated LS

CBR value, % 9.06

LS + 100 ml/m3

12.64

LS + 200 ml/m3

17.63

18

V. Divya and M. N. Asha

4.2 Variation in Void Ratio Under Sustained Loading Consolidation test was conducted as per IS 2720-15 on plain and blended laterite soil samples. During the consolidation process, the load which is initially carried by the pore water is shifted to the soil structure and hence there occurs a change in the total volume of soil equal to the volume of water drained. The soil samples were blended with varying dosages of Terrazyme at controlled conditions of maximum dry density and optimum moisture content. These treated samples were then placed in airtight containers for a period of 7 and 14 days curing. The test was performed under sustained pressures of 0.05, 0.1, 0.2, 0.4, 0.8 and 1.6 kg/cm2 . Figures 3, 4 and 5 represent the variation of void ratio under sustained loading for different curing periods. Fig. 3 Variation of void ratio with various enzyme dosages for 0 days curing period

0.58 0.56

Void ratio, e

0.54 0.52 0.5 Untreated Soil 0.48 LS+100ml/m3 0.46 0.44 0.01

LS+200ml/m3 0.1

1

10

Pressure, Kg/cm 2

Fig. 4 Variation of void ratio with various enzyme dosages for 7 days curing period

0.58 0.56

Void ratio, e

0.54 0.52 0.5 Untreated Soil 0.48 0.46 0.44 0.01

LS+100ml/m3 LS+200ml/m3 0.1

1

Pressure, Kg/cm2

10

Investigations on Mechanical Properties of Bio-Stabilized Soil Fig. 5 Variation of void ratio with various enzyme dosages for 14 days curing period

19

0.58 0.56

Void ratio, e

0.54 0.52 0.5 Untreated Soil 0.48 0.46

LS+100ml/m3 LS+200ml/m3

0.44 0.01

0.1

1

10

Pressure, Kg/cm2

Table 3 Variation in void ratio with pressure at 14 days curing period

Pressure, Kg/cm2

Void ratio, e Untreated LS

LS + 100 ml/m3

LS + 200 ml/m3

0

0.5667

0.5584

0.5378

0.05

0.5667

0.5523

0.5369

0.10

0.5667

0.5454

0.5352

0.20

0.5667

0.5375

0.5213

0.40

0.5597

0.5280

0.5120

0.80

0.5519

0.5128

0.4667

1.60

0.5232

0.5012

0.4550

Table 3 compares the void ratio at different pressures for two different proportions of enzyme. From the experimental results, it is observed that the void ratio of enzyme blended soil is less in comparison to conventional soil. This demonstrates the close packing of soil grains in bio-stabilized soil. As the sustained pressure increases, correspondingly void ratio is also found to decrease with increasing enzyme dosages. Probably, the addition of enzyme lubricates the soil and weakens the bond between the soil minerals and water molecules. Hence, water could be easily expelled from the soil with an increase in pressures and thus reduce the compressibility of the soil. Furthermore, the curing period is also found to play a prominent role in decreasing the void ratio under the same sustained pressure.

20

V. Divya and M. N. Asha

4.3 Variation in Permeability Characteristics Permeability studies of enzyme blended soil facilitate the evaluation of resistance offered by Terrazyme against pore water seepage. Falling head permeability test was conducted as per IS 2720-17 on both treated and untreated soil at two different curing periods of 7 and 14 days. The test was performed by preparing the sample at the maximum dry density and saturated at optimum moisture content under sustained loading so as to generate actual field conditions. Figures 6, 7 and 8 represent the variation of permeability under static pressure for various dosages of enzyme. It is evident from experimental studies that permeability is a function of soil density, Terrazyme dosage and curing period. The coefficient of permeability is found to decrease with increasing dosage and curing period. These results are in agreement with compressibility results because it demonstrates the decrease in void 0.01

Coefficient of permeability, cm/s

Fig. 6 Variation of coefficient of permeability with various enzyme dosages for 0 days curing period

0.009

Untreated Soil

0.008

LS+100 ml/m3

0.007

LS+200 ml/m3

0.006 0.005 0.004 0.003 0.002 0.001 0 0

0.5

1

1.5

2

Fig. 7 Variation of coefficient of permeability with various enzyme dosages for 7 days curing period

Coefficient of permeability, cm/s

Pressure, kg/cm2

0.01 0.009

Untreated Soil

0.008

LS+100ml/m3

0.007

LS+200ml/m3

0.006 0.005 0.004 0.003 0.002 0.001 0

0

0.5

1

Pressure, kg/cm2

1.5

2

Fig. 8 Variation of coefficient of permeability with various enzyme dosages for 14 days curing period

Coefficient of permeability, cm/s

Investigations on Mechanical Properties of Bio-Stabilized Soil

21

0.01 0.009

Untreated Soil

0.008

LS+100ml/m3

0.007

LS+200ml/m3

0.006 0.005 0.004 0.003 0.002 0.001 0

0

0.5

1

1.5

2

Pressure, kg/cm2

ratio with the addition of enzymes. This shows that the enzyme has a significant role in preventing the infiltration of water into the soil. This characteristic of enzyme blended soil can be utilized while providing liner material in landfills.

5 Conclusions The following conclusions have been comprehended based on the experimental investigations: • Increasing enzyme dosages and curing periods contribute to significant modifications in the bearing ratio values and compressibility characteristics of treated laterite soil. • Bearing ratio of laterite soil was found to increase by 40% and 95% when treated with 100 ml/m3 and 200 ml/m3 , respectively. This confirms the efficiency of Terrazyme in improving the subgrade strength of pavements. • Terrazyme has the potential to decrease the compressibility of soil especially at higher pressures. This could be due to the weakening of the bond between soil and water particles by enzymatic activity. • Coefficient of permeability decreases with increasing dosage and curing period. Hence Terrazyme can be efficiently used to prevent water permeation in various constructions such as landfill liners, pavements, etc.

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V. Divya and M. N. Asha

References 1. Muntohar AS (2009) Influence of plastic waste fibers on the strength of lime-rice husk ash stabilized clay soil. Civ Eng Dimension 11(1):32–40 2. Arulrajah A, Yaghoubi E, Wong YC, Horpibulsuk S (2017) Recycled plastic granules and demolition wastes as construction materials: resilient moduli and strength characteristics. Constr Build Mater 147:639–647 3. Kumar PS, Rajkumar R (2012) Effect of geotextile on CBR strength of unpaved road with soft subgrade. Electron J Geotech Eng 17(1):1355–1363 4. Shankar AU, Rai HK, Mithanthaya R (2009) Bio-enzyme stabilized lateritic soil as a highwaymaterial. In: Indian Roads Congress Journal 70(2) 5. Marasteanu MO, Hozalski RM, Clyne TR, Velasquez R (2005) Preliminary laboratory investigation of enzyme solutions as a soil stabilizer 6. Ivanov V, Chu J (2008) Applications of microorganisms to geotechnical engineering for bioclogging and biocementation of soil in situ. Rev Environ Sci Bio/Technology 7(2):139–153 doi: https://doi.org/10.1007/s11157-007-9126-3 7. Shukla M, Bose S, Sikdar PK (2003) Bio-enzyme for stabilization of soil in road construction a cost-effective approach 8. Naagesh S, Gangadhara S (2010) Swelling properties of bio-enzyme treated expansive soil. Int J Eng Stud 2(2):155–159

Strength Behavior of Sand Reinforced with Treated Sisal Fibers C. Jairaj and M. T. Prathap Kumar

Abstract Improvement of the ground is an important challenge before geotechnical Engineers world over and the concept of reinforcing the soil using plant fibers has been used to improve the strength of the ground since ancient times. Plant fibers are lignocellulose in nature. Among these lignocellulose fibers, sisal fibers have the potential to replace synthetic fibers in various applications. In the present study model footings of different sizes, viz., 10, 20, and 30 mm were tested with sand admixed with sisal fibers. The Sisal fiber was cut in to have an average length of 10 to 20 mm and admixed randomly in sand with different percentages of 0, 0.5, 1, 2, and 3%. Model footing tests were performed by static loading of the footings in the model tank of size 260 mm length 260 mm breadth 350 mm height which was filled with sand and sisal fiber-reinforced sand. The sand was filled using the sand raining technique to achieve 75% relative density. From the test results, it was found that sand mixed with 2% sisal fiber had a larger bearing capacity compared with all other combinations of sand admixed with other percentages of sisal fibers. Keywords Sisal fiber · Sand · SEM

1 Introduction Henry Vidal introduced the modern concept of reinforced soil in the 1960s indicating that the concept of reinforced earth is cost-effective and can become alternatives for most applications where reinforced concrete or gravity type wall have traditionally been used [1–5]. Soil reinforcement generally uses two types of natural and synthetic material. Of recent origin, in view of sustainable applications, the use of natural reinforcement has been increasingly suggested as a method to improve the ground. C. Jairaj (B) Nitte Meenakshi Institute of Technology, Bengaluru, India e-mail: [email protected] M. T. Prathap Kumar RNS Institute of Technology, Bengaluru, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_3

23

24 Table 1 Properties of river sand

C. Jairaj and M. T. Prathap Kumar Description

Values

Specific gravity G

2.66

Coefficient of uniformity (Cu)

4.16

Coefficient of curvature (Cc)

1.18

Maximum dry density (kN/m3 )

18.67

Minimum dry density (kN/m3 )

15.87

Some of the natural fibers such as coir, jute sisal, and other natural plant products reinforcing materials are increasingly used [6–9]. It shall be noted that the strength of natural fibers is somewhat lower than the synthetic fiber and it is less stiff. However, these raw materials are much less expensive and sustainable in terms of carbon footprint [10–13]. Its bulk strength and weight properties are very favorable when compared to metals, and it can be easily formed using molding processes [14, 15]. Nowadays, natural fiber such as sisal and jute fiber composite materials are replacing glass and carbon fibers owing to their easy availability and cost. The use of natural fibers has improved remarkably in construction industries. Several research and studies on natural fibers are concerned with the layered concept of reinforcement rather than fiber reinforcement [16, 17]. The addition of natural fiber is comparatively cheaper and easy to use. In the present study, the mechanical properties of sisal fiber-reinforced sand are studied. The properties such as tensile, compression, flexural, and impact are studied and presented in detail. The results indicated that the addition of randomly distributed sisal fibers of average length 10 to 20 mm admixed with sand improves the strength properties of sand and enhances the bearing capacity of the footing and the improved properties are in conjunction with available literature [18, 19].

2 Materials and Methods 2.1 Sand Natural river sand conforming IS 383–1970 is used and the properties of sand are shown in Table 1. Based on the test results on the index properties of sand, the sand can be classified as poorly graded sand with group symbol SP.

2.2 Sisal Fibers Natural Sisal fibers brought from the local market is used in the study. Sisal fiber was chopped and cut to have average length of 10 to 25 mm was used, in order for

Strength Behavior of Sand Reinforced with Treated Sisal Fibers Table 2 Properties of sisal fibers

Properties Density

(kN/m3 )

Fiber diameter (mm)

25 Values 13.3 0.1–0.5

Tensile strength (N/mm2 )

31–221

Specific gravity (G)

1.39

the ease of mixing without aggregation/segregation in the sand. Table 2 shows the properties of sisal fiber used in the present study.

2.3 Model Footings and Model Tank Model footings of sizes 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm were prepared and provided with three protruding arms to record the settlement in three different locations and average settlements were recorded. The model tank of size 340 × 260 × 550 mm was used and special care was taken to overcome the tilting effect of the foundation. The size of the test tank in any direction is about 4–5 times the size of the model footing to minimize the boundary confinement.

2.4 Methodology Used To reduce the biodegradability of sisal fiber, alkali treatment of fiber was adopted [10]. Morphological changes due to alkali treatment were studied using SEM analysis on both treated and untreated sisal fibers. Further water absorption tests were carried out on both samples to assess the reduction in water absorption capacity of alkali-treated sisal fiber. Different percentages of treated sisal fibers at percentages varying from 0.5, 1, 2 and 3% were used by mixing fibers with sand and tested with different sizes of model footings of 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm. In order to maintain the constant relative density of compacted sand in the test tank, sand raining technique was used. Static loading tests were conducted using model footings of different sizes and fiber percentages and were compared with each other.

26

C. Jairaj and M. T. Prathap Kumar

(a) Untreated Sisal Fiber

(b) Treated Sisal Fiber

Fig. 1 SEM image of untreated and treated sisal fiber

3 Results and Discussion 3.1 SEM Studies on Treated Sisal Fibers Morphological changes associated with untreated and treated sisal fibers were analyzed with the help of XRD and SEM analysis. Figure 1a, b clearly indicates that treated sisal fiber has a rougher texture than the untreated. Thus, it is to clearly state that this rough surface texture of treated sisal fiber leads to increased friction between sand and fibers that causes increased strength compared to untreated sisal fibers.

3.2 Preparation of Sand Bed Sand was compacted by means of sand raining technique using by trial and error method. A funnel was used to pour sand and the height of poring was varied to achieve the desired relative density is achieved. The height of fall thus determined was maintained during the sand bed preparation in the test tank. The model footing was placed at the center of the compacted surface of the sand bed and the static load was applied vertically in line with the piston and load cell by the hydraulic compressor. Different tests were performed for different footing sizes of 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm with different percentages of sisal fibers of 0, 0.5, 1, 2 and 3%. Series of such a test has been carried out for unreinforced and sisal fiber-reinforced sand bed. The schematic setup is as shown in Fig. 2.

Strength Behavior of Sand Reinforced with Treated Sisal Fibers

27

1.loading devise, 2. Model footing, 3. model tank, 4. Lever, 5. Strain rate for load application, 6. digital meter to record load and settlement Fig. 2 Schematic test set up

4 Effect of Sisal Fiber-Reinforced Sand Figure 3 shows the load versus settlement graph for the unreinforced sand with different sizes of the footings of 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm square footing. The test results clearly indicate that increase in the surface area of Load in kN 0

0

5

10

15

20

25

30

35

40

Settlement in mm

10 20 30 40 50 60 3x3 cm Footing Size

2x2cm footing size

1x1 cm Footing Size

Fig. 3 Unreinforced sand model footing graph for 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm footing sizes

28

C. Jairaj and M. T. Prathap Kumar Load in kN 0

0

10

20

30

40

50

60

70

80

Settlemetn in mm

1 2 3 4 5 6 1x1cm Footing size with 0.5% SF

2x2 cm Footing size with 0.5% SF

3x3 cm Footing size with 0.5% SF

Fig. 4 Model footing graph of 0.5% SF reinforced with sand for 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm footing sizes

the footing increases load-carrying capacity of the footing. Figure 4 shows the model footing with 0.5% sisal fiber (SF) reinforced sand indicates greater load-carrying capacity of 10, 50, and 60 kN for 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm footing size respectively and in comparison to unreinforced sand. The load-carrying capacity of SF-reinforced sand is significantly larger- as the load-carrying capacity of unreinforced sand obtained was around 4, 16, and 20 kN 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm, respectively. Even the maximum settlement recorded at failure was less for the reinforced sand when compared to unreinforced sand. The test results clearly indicated that the reinforced sand has a higher load-carrying capacity when compared to unreinforced sand. Comparative analysis of load settlement curve obtained for SF-reinforced sand is shown in Figs. 3 and 4 shows flattening effect in comparison to unreinforced sand indicating increased ductility due to introduction of fibers, and hence the fiber-reinforced sand has increased earthquake resistance. Figure 5 shows model footing with 1% SF-reinforced sand for 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm footing sizes. It indicated that the load-carrying capacity of 10 × 10 mm footing was 14 kN and 20 × 20 mm was 57 kN and for 30 × 30 mm was 68 kN indicating that even 1% SF admixed sand gives higher load-carrying capacity compared to 0 and 0.5% SF admixed sand for the same corresponding settlement. A similar trend was observed as shown in Fig. 6 for model footing resting on sand admixed with 2% SF corresponding to a settlement of 4 mm. Figure 7 shows model footing resting on sand admixed with 3% SF with corresponding load-carrying capacities being 13, 49, and 60 kN, respectively, at the same 4 mm settlement. The test results summarized and shown in Table 2 and comparative curves in Fig. 8 clearly indicates that higher the surface area of the footing in contact with the underlying soil, higher will be the load-carrying capacity and

Strength Behavior of Sand Reinforced with Treated Sisal Fibers

29

Load in kN 0

0

10

20

30

40

50

60

70

80

Settlement in mm

1 2 3 4 5 6 1x1cm Footing size with 1% SF

2x2cm Footing size with 1% SF

3x3cm Footing size with 1% SF

Fig. 5 Model footing graph of 1% SF reinforced with sand for 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm footing sizes Load in kN 0

0

10

20

30

40

50

60

70

80

Settlemetn in mm

1 2 3 4 5 6 1x1cm Footing Size with 2% SF

2x2cm Footing size with 2% SF

3x3cm Footing size with 2% SF

Fig. 6 Model footing graph 2% SF reinforced with sand for 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm footing sizes

maximum load-carrying capacity occurs with 1% sisal fiber in comparison to all other percentages of sisal fibers (Table 3).

30

C. Jairaj and M. T. Prathap Kumar Load in kN 0

0

10

20

30

40

50

60

70

80

Settlement in mm

1 2 3 4 5 6 1x1cm Footing size with 3% SF

2x2cm Footing size with 3% SF

3x3cm Footing size with 3% SF

Fig. 7 Model footing graph 3% SF reinforced with sand for 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm footing sizes 80 70

Load in kN

60 50 1x1cm

40

2x2cm 30

3x3cm

20 10 0

Unreinforced 0.5%SF + Sand

1%SF + Sand 2%SF + Sand 3%SF + Sand

Fig. 8 Comparative variation of load carried by footings of size 10 × 10 mm, 20 × 20 mm, and 30 × 30 mm Table 3 Properties of sisal fibers 0.5% SF + Sand

1% SF + Sand

2% SF + Sand

3% SF + Sand

Settlemt in mm

Load in kN

Settlemt in mm

Load in kN

Settlemt in mm

Load in kN

Settlemt in mm

Load in kN

4

4

10

4

14

4

14

4

13

15

16

4

55

4

57

4

50

4

49

15

20

4

60

4

68

4

62

4

60

Footing sizes

Unreinforced

10 × 10 mm

15

20 × 20 mm 30 × 30 mm

Settlemt in mm

Load in kN

Strength Behavior of Sand Reinforced with Treated Sisal Fibers

31

5 Conclusions Based on results obtained from the experimental studies, the following broad conclusions are arrived at based on the behavior of square footing resting on sand reinforced with random distribution of sisal fibers • The effect of the percentage of reinforcement affects the load-bearing capacity (LBC) of soil. It has been found that it becomes maximum at 1% sisal fiber and can be considered as optimum. • Epoxy water resilient coated Sisal fiber exhibits water resistance and hence can be considered non-biodegradable. • Layout and configuration of sisal fiber play a vital role in increasing the loadcarrying capacity. Increased fiber content and higher distribution provides much better benefit up to 1% SF. Further increase in the fiber content from 2 and 3%, it was found that the load-carrying capacity of reinforced sand decreases when compared to 1% SF. However, load-carrying capacity of reinforced sand is found to be much higher in comparison to unreinforced sand. • All reinforced samples exhibited improved ductile stress–strain response compared to unreinforced sand at all combinations- indicating that fiberreinforced sand is a viable earthquake-resistant material that has many civil engineering applications.

References 1. Binquet J, Lee KL (1975) Bearing capacity analysis of reinforced earth slabs, J Geotechnical Eng Div 101(ASCE 12 Proc): 1257–1276 2. Consoli NC, Casagrande MDT, Preitto PDM, Thome A (2003) Plate load test on fiber reinforced soil, J Geotech Geoenviron Eng 129(10):0951–0955 3. Callum AS, Hill HPS, Khalil A, Hale MD (1997) A study of the potential of acetylation to improve the properties of plant fibers. Ind Crops Prod 8:53–56 4. Einstein HH, Choudhury D, Viswanadham BVS (2010) Model studies on mechanics of engineering, Jl Geotech Geoenv Eng ASCE 140(3):229–248 5. Freitag DR (1986) Soil randomly reinforced with fibers. J Geotech Eng 112(8):823–826 6. Gray DH, Ohashi H (1983) Mechanics of fiber reinforcement in sand. J Geotech Eng 109(3):335–353 7. Gray DH, Al-Refeai T (1986) Behaviour of fabric-versus fiber-reinforced sand. J Geotech Eng 112(8):804–820 8. Raymond GP (1992) Reinforced sand behavior overlying compressible sub grades. J Geotech Eng 118(11):1663–1680 9. Jairaj, Prathap Kumar MT, Gouda S, (2017) Study of CBR on BC soil admixed with untreated and treated coir fibers. J Geotech. Studies 2(3):1–7 10. Jairaj C, Prathap Kumar MT, Raghunandan ME (2018) Compaction characteristics and strength of BC soil reinforced with untreated and treated coir fibers. Innovative Infrastruct Solutions 3(21). https://doi.org/10.1007/s41062-017-0123-2 11. Jairaj, Prathap Kumar MT (2018) Long-term performance studies on strength characteristics of black cotton soil reinforced with untreated and treated coir fibre. Int J Geosynthetics Ground Eng https://doi.org/10.1007/s40891-018-0143-9

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12. Kolekar YA, Dasaka SM, Venkatachalam G (1995) Review of reliability engineering in geotechnics, 4th Intl. Conf. on Reliability Engineering, Feb. 3–5, 1995, Madrid, J.H.Lu, M.V.Chin, A.Juneja (eds.), Balkema (Pubs.), Vol. 1, 201–210 13. McGown A et al (1978) Effect of inclusion properties on the behaviour of sand. Geotechnique 28(3):327–346 14. Maher MH, Ho YC (1994) Mechanical properties of kaolinite/fiber soil composite. J Geotech Eng 120(8):1381–1392 15. Mohanty AK, Mishra M, Drzal LT (2001) Surface modifications of natural fibers and performance of the resulting bio-composites: an overview. Compos Interaces 8(5):313–343 16. Nataraj MS, McManish KL (1997) Strength and deformation properties of soils reinforced with fibrillated fibers. J Geosynth Int 4(1):65–79 17. Nagraja PS (2006) Behaviour of reinforced sand beds subjected to monotonic, cyclic and dynamic loads Ph.D thesis report UVCE, Bangalore University 18. Ozkul ZH, Baykal G (2007) Shear behaviour of compacted rubber fiber-clay composite in drained and undrained loading. J Geotech Geoenviron Eng 133(7):0767–0781 19. Singh D (1991) Static response of fibre reinforced soil, Doctoral Thesis, Indian Institute of Technology Bombay, Mumbai, India

Unconfined Compressive Strength of MICP-Treated Black Cotton Soil R. B. Wath and S. S. Pusadkar

Abstract Expansive soil which is found in several parts of India possesses high swelling and shrinkage properties. The volume change and uplift pressure generated in these soil deposits cause severe damage to the lightweight structures and pavements. In order to mitigate the problems associated with expansive soils, it is necessary to stabilize this soil. Among all the stabilization techniques, microbial induced calcite precipitation (MICP) is a sustainable biological ground improvement technique that is capable of altering and improving the engineering properties of black cotton soil. The application of microbiological processes for improvement of engineering properties of soil offers sustainable, cost-effective, non-disruptive ground improvement for a variety of geotechnical problems. This paper presents the results of a study in which microbiological processes were used to alter the properties of expansive soil. Microbial induced calcite precipitation (MICP) was achieved using the microorganism Bacillus Pasteurii. The microbes in liquid medium and cementation media were introduced to the soil specimen. The soil specimens were kept for different reaction periods (0, 3, 7 and 14 days) with different percentages of microbes. From the study, Bacillus Pasteurii had shown major influence for all percentages of microbe and reaction times on unconfined compressive strength of expansive soil. Keywords MICP · Expansive soil · Soil stabilization · Engineering properties · Unconfined compressive strength

R. B. Wath (B) · S. S. Pusadkar Government College of Engineering, Jalgaon, Maharashtra, India e-mail: [email protected] S. S. Pusadkar e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_4

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R. B. Wath and S. S. Pusadkar

1 Introduction Expansive soils are also called as highly problematic soil because of their inherent potential to undergo unfavorable volume changes corresponding to changes in the moisture regime. They swell when they absorb water and shrink when water evaporates from them. Because of the alternate swelling and shrinkage, civil engineering structures founded in these soils such as residential buildings, pavements, airports, retaining walls, canal beds and linings are severely damaged. Shrinkage and swelling create many problems such as deformations and stresses in the structures resting on or in the soil. To counteract the problem of heave and shrinkage of expansive soils, which include stabilization with various additives such as fly ash, lime, cement and chemicals. All of these methods are not appropriate due to environmentally detrimental, high energy consumption, emission of CO2 , economic nonviable [1]. Therefore, there is a need for developing a new stabilization technique for expansive soil that is economical and environmentally sustainable. Microbial induced calcite precipitation (MICP) is a biological process, which includes the mainly following two mechanisms, bio-clogging and bio-cementation [2, 7]. Where bio-cementation improves the shear strength of soil through the production of calcite precipitation but bio-clogging is a process where the soil void is filled by the product of calcite precipitated [4, 6]. These mechanisms can be achieved by common metabolic processes such as photosynthesis, denitrification, urea hydrolysis and sulphate reduction. MICP by urea hydrolysis had been used by many researchers by using genera Bacillus as shown in Eqs. 1 and 2. The reaction in Eq. 1 showed that 1 mol of urea decomposes into 2 mol of ammonium. CO(NH2 )2 + 2H2 O → 2NH4 + CO3 2−

(1)

The release of ammonium creates a perfect environment for precipitation of calcite with the availability of calcium ion from supplied calcium chloride at higher pH. Ca2+ + CO3 2− → CaCo3

(2)

2 Materials and Method 2.1 Soil Black cotton soil collected from Amravati (Maharashtra, India) was used for the present study. The physical and engineering properties and soil classification was given in Table 1.

Unconfined Compressive Strength of MICP-Treated Black Cotton Soil Table 1 Properties of soil

35

Description

Value

Free swell index

109%

Specific gravity

2.42

Liquid limit

88%

Plastic limit

43.33%

Plasticity index

44.67%

Shrinkage limit

7.52

Maximum dry unit weight

1.44 gm/cc

Optimum moisture content

29%

Swelling pressure

175 kN/m2

Unconfined compression test

263 kN/m2

Soil classification as per IS

CH

2.2 Microorganism The microorganism used in the study was Sporosarcina pasteurii (formerly Bacillus Pasteurii). It was procured from the National Collection of Industrial Microorganisms (NCIL), Pune. Strain number NCIM 2477 is an equivalent strain of ATCC 14,859. Nutrient medium for the microbial culture was followed according to ATCC specifications of 20 g/L yeast extract, 10 g/L ammonium sulphate in 1L of 0.13 M Tris buffer (pH 9.0). Individual ingredients were autoclaved separately followed by inoculation of strain under sterile conditions as the chances of contamination are more for bacteria. Later, the inoculated media was kept in an incubator for 48 h at 37 °C to initiate the bacterial growth.

2.3 Cementation Solution Cementation media was used to provide chemical compositions for ureolysis including urea, CaCl2 · 2H2 O, NH4 Cl, NaHCO3, and nutrient broth [3, 5, 8]. Table 2 shows the chemical compositions of cementation solution for bacteria experiments. The molar ratio of urea and calcium chloride was fixed as 1:1. All the ingredients were autoclaved separately except urea and NaHCO3 .The NaHCO3 was autoclaved as dry powdered (instead of as a liquid media) and then dissolved into autoclaved DI water. This process was done to avoid off-gassing CO2 [3]. Urea was not autoclaved because it will be decomposed due to heat. The pH of the medium was adjusted to 6.0 with a few drops of HCl prior to autoclaving.

36 Table 2 Details of cementation solution concentrations

R. B. Wath and S. S. Pusadkar Chemical

Chemical concentration g/L

NH4 Cl

10.00

0.25 M NaHCO3

2.12

Nutrient broth

3.0

Urea

15.0

CaCl2 · 2H2 O

27.8

pH

6.0

2.4 Methodology The soil sample was oven dried at 105 °C for 24 h before starting the treatment without any sterilization process. Initially, an adequate amount of water and the incubated bacteria were spread over the soil in four sectors equally and mixed proportionally in individual sector which was followed by mixing of the cementation solution in equal proportion. The uniformity mixing of soil and bacterial solution was obtained by thorough hand mixing in sector-wise and then together. The amount of bacteria was selected as 5%, 10%, 15%, and 20% by the weight of soil sample, cementation solution was fixed 100 ml of 0.25 M. The soil was compacted and unconfined compression (UC) tests were performed on MICP-treated soil samples according to IS: 2720 (Part X). To investigate the effect of treatment duration or reaction period, MICP-treated samples were cured for 0, 3, 7 and 14 days before unconfined compression tests began. The samples were wrapped into plastic bags then stored in airtight container at room temperature (20–30 °C).

3 Results and Discussion The unconfined compression test was used for measuring the shear strength of treated soil. The UCS value for the untreated soil sample was 263 kN/m2 . Figure 1 shows the stress–strain curves obtained from unconfined compression tests on the MICPtreated samples catalyzed by Sp. Pasteurii under different bacteria concentrations with cementation concentration of 0.25 M. The stress–strain curve shows that as curing period increases, the failure strain of soil reduces. Table 3 shows unconfined compressive strength of all treated soils for different criteria. It was observed that as the percentage of bacteria increases, the curing period increases to achieve maximum strength. The unconfined compressive strength of soil treated with 5% bacterial content with curing period 7 days shows maximum strength as 623 kN/m2 . For bacterial concentration 10%, the UC strength achieved was 589 kN/m2 at 14 days curing period. The maximum UCS of 15% bacterial concentration treated soil samples was 553 kN/m2 for 14 days curing. When the

Unconfined Compressive Strength of MICP-Treated Black Cotton Soil

37

(a)5% Bacterial Concentration

(b) 10% Bacterial Concentration

Fig. 1 Comparison of variation of UCS values treated with different % of MICP

bacterial concentration was increased by 20%, the UCS values were increased than untreated soil samples but the rate of precipitation is lower than all the concentrations. It was evident that the MICP treatment with the introduction of bacteria and cementation solution was capable to improve the UC values of black cotton soil for all bacterial concentrations. The maximum unconfined compressive strength of 623 kN/m2 was observed when the soil sample was cured for 7 days with Sp. Pasteurii concentration of 5%. Hence, 5% bacterial concentration seems to be optimum with respect to strength and 7 days curing period was observed to be minimum curing period for the same bacterial content. As the curing period increases, the stiffness of soil mass was observed to be increased. However, the failure strain was decreasing with curing period.

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R. B. Wath and S. S. Pusadkar

(c) 15% Bacterial Concentration

(d) 20% Bacterial Concentration

Fig. 1 (continued)

Table 3 UC test results (kN/m2 ) Reaction period

Bacterial concentration 5%

10%

15%

20%

0

437

386

433

302

3

559

472

508

320

7

623

511

500

433

14

590

589

553

451

21

482

–

–

–

It was observed from the test results that by increasing treatment duration UCS values increased for all selected reaction periods. The highest increment was observed for bacterial concentration of 5% at the 7th day reaction period. This can be achieved due to the presence of finer particles in CH soil, which provided a dense arrangement

Unconfined Compressive Strength of MICP-Treated Black Cotton Soil

39

Fig. 2 CaCO3 precipitation on soil sample

of soil particles and offered more particle to particle contact for the bond formation. It also offered a more specific surface area for the precipitate formation. This bond formation in the soil leads to an increase in cohesion of soil which is an important parameter for shear strength of soil and hence an increase in shear strength. UCS was decreased by 5% bacterial concentration at 14 days reaction period. This may be due to the fact that insufficient bacteria may be available at 5% concentration at 14 days of curing period than other concentrations. The insufficient presence of bacteria results in less precipitation of calcite and results in a decrease of unconfined compressive strength. The precipitation of calcite on the soil sample is shown in Fig. 2.

4 Conclusions The results implied that urease forming bacteria can be utilized in MICP for expansive soil, with sufficient and appropriate nutrient provided. The following conclusions were drawn from the present study • The unconfined compression strength increases for all bacterial concentrations than untreated soil • The unconfined compressive strength increases with the reaction period for a higher concentration of bacteria • For black cotton soil, 5% bacterial concentration was found to be optimum concentration and 7 days as an optimum reaction period to achieve maximum strength • The use of MICP as a stabilization agent for black cotton soil would be green, sustainable, eco-friendly technique.

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References 1. Cheng L (2012) Innovative Ground Enhancement by Improved Microbially Induced CaCO3 Precipitation Technology. PhD Thesis, Murdoch University, Perth, WA, 1–252 2. Ivanov V, Chu J (2008) Applications of microorganisms to geotechnical engineering for bioclogging and biocementation of soil in situ. Rev Environ Sci Biotechnology 7(2):139–153 3. Mortensen BM, DeJong JT (2011) Strength and stiffness of MICP treated sand subjected to various stress paths. Geofrontiers 4012–4020 4. Ng WS, Lee ML, Hii SL (2012) An overview of the factors affecting microbial induced calcite precipitation and its potential application in soil improvement. World Acad Sci Eng Technol 6(2):683–689 5. Shannon SF, Galinat JK, Bang SS (1999) Microbiological precipitation of CaCO3 . Soil Biol Biochem 31, 1563–1571 6. Wath RB, Pusadkar SS (2019) Soil Improvement Using Microbial: A Review. A Chapter in Ground Improvement Techniques and Geosynthetics Book, Springer Nature Singapore Pvt. Ltd., 329–335 7. Wath RB, Pusadkar SS (2018) Influence of Bacteria’s on Physical Properties of Black Cotton Soil. Proceeding of Indian Geotechnical Conference IISc Bangaluru, 1–5 8. Zhao Q, Li L, Li C, Li M, Amini F, Zhang H (2014) Factors Affecting Improvement of Engineering Properties of MICP-Treated Soil Catalyzed by Bacteria and Urease. Journal of Materials Civil Engineering 26 (12)

Potential of Reuse Options of Rice Husk Ash in Various Applications Nazeema Basheer, Jaskiran Sobti, and Nazra Khanam

Abstract India is one of the largest producers of rice in the world and subsequently has the potential of producing rice husk in large amounts. The disposal of this waste material generated from the agricultural sector is a growing concern at present. Rice husk ash, a biomass waste, is acquiring adequate fascination owing to its wide value and probability in ecological protection. It can be put to beneficial use in varied applications in the engineering sector considering the noteworthy properties of rice husk ash such as its affordability, being environment-friendly, pozzolanic properties, and ease of access. A peep in the literature has prompted the, while the use of rice husk for waste management such as it is used as an additive in concrete, as a binder material, in soil stabilization and in natural preservation, e.g., reducing pulping material used in cardboard industries, etc. Apart from these basic interest’s rice husk ash has helped in contamination control, e.g., it has been used as a biobased adsorbent in treating contaminated water by removing organic and inorganic matter, as a catalyst in various chemical reactions, etc. This article presents an overall picture of the effective usage and beneficial properties of rice husk ash in a myriad of applications in different fields. Keywords Rice husk ash · Waste management · Bio-based adsorbent

N. Basheer · N. Khanam Department of Civil Engineering, Lovely Professional University, Phagwara, India e-mail: [email protected] N. Khanam e-mail: [email protected] J. Sobti (B) Department of Civil Engineering, Guru Nanak Dev University, Amritsar, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_5

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1 Introduction Preservation of our environment from the generation of immense waste is a mounting concern. The changing lifestyle of people has made the environment vulnerable to many threatening facets which include global warming, degradation of water, air, soil, etc. About 1.3 billion tons per year of waste is generated worldwide out of which India produces about 62 million tons of solid [14]. It is anticipated that by 2031 the generation of municipal solid waste will be approximately 165 million tons and by 2050 it might be stretched out to 436 million tons per annum About 43 million tons of waste is gathered out of which 11.9 million is treated whereas 31 million tons are discarded into landfills. This directs that 75–80% of waste is accumulated and only 22–28% of this is treated [14]. Wastes have been categorized into various groups depending on their source of the generation which includes domestic waste, sanitary waste, agricultural and dairy waste, biomedical waste, radioactive waste, and industrial waste. Agricultural waste alone contributes about 600 million tons of waste in India [29]. These wastes include sugarcane bagasse ash, wheat straw, rice husk, groundnut shell, various fibers such as coir fiber and jute fiber, cotton husk, sawdust, etc. Whereas industrial waste is approximated to 4.5 million tons which are generated from various processes such as metal extraction, steel casting, refineries, petrochemical industries, pharmaceutical, and pesticide industries. Agriculture is the backbone of the Indian economy which contributes to around 15% of gross domestic product (GDP). About 70% of the population in India is dependent on agriculture. The availability of agricultural resources in India has led to the development of agro-industries. Besides the upliftment of the economy, these industries have been a source of enormous waste generation. One of the major reasons for global warming is the burning of these wastes. Agro wastes or agricultural wastes are obtained from the processing of agriculture products which include crop production and harvesting [23]. Urban waste primarily entails organic matter which can be easily decomposed. One of the most cultivated crops all around the world is rice [9]. Being one of the main agricultural crops, its production is enormous. About 1.2 billion tons of rice is produced annually [33]. On an average of about 20%, by weight, rice husk is produced which is a byproduct that is obtained after the rice milling process [27]. These by-products are either burnt or buried because of low levels of protein, which forms ash called rice husk ash (RHA). Apparently, these ashes are rich in silica and lignin content [42] hence cannot be decomposed by the microorganisms. Since waste management and recycling have attenuated a great concern these biomass wastes can be used to help in managing the environment to be clean. So far, much research has been carried out in using RHAs as they have a high content of silica about 5–30% [37]. RHAs have proved their mettle in various engineering applications. The percentage of pozzolanic content is high in RH ashes hence they are fit to be used as in production lightweight concrete [25]. Several experiments have been carried out to use the silica extracted from RHA to produce a catalyst for oxidation of styrene [1] while [6] used catalytic pyrolysis

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for bio-oil production. RH ashes have found their application in subgrade soil of flexible pavement at an optimum percentage of 10 [4]. When compared to other methods these biomasses have low reaction time which makes them suitable for their utilization. The addition of RHA improves the CBR, shear strength parameters, and plasticity of expansive soils [26].

2 RHA as a Sustainable Waste for Stabilization of Problematic Soils The field of civil engineering is very vast and the use of agro-waste has fostered a new era of waste management incorporating their needful properties for beneficial use. Technically soil improvement is either done by alteration or by stabilization. For instance, the addition of chemical binders to the soil so as to change its index properties is called stabilization [3]. These approaches are employed to advance the use of expansive or any problematic soil such that they become apposite for construction. The overdependency on chemical binders such as lime and cement has proved to be costly and has degraded the environment [20]. In recent years, the use of chemical binders has been replaced by solid waste, which not only reduces environmental impact but also proves to be economic. Rice husk is one of the most produced agricultural wastes and a cementitious material has initiated its application in the stabilization of expansive soils [34, 3]. With 82–95% of silica content [36, 21] better results are achieved.

2.1 Effect on Compaction Characteristics Numerous researches have been carried out to investigate the effect of utilizing RHA as a stabilizing material. Hossain [16] presented the characteristics of clayey soil stabilized with rice husk ash and cement kiln dust. This study was carried out in a combination of these stabilizers as well as separately at different contents. It was found that with the increasing percentage of the stabilizers the maximum dry density decreased and optimum moisture content increased which has been confirmed by other researchers also [31, 2, 18, 35]. This increase can be attributed to the fact that RHA has pozzolanic properties whereas CKD has higher specific gravity. Basha et al. [7] observed the potential of rice husk ash in combination with cement as an alternative to common stabilizing materials on residual granite soil. The results demonstrated that with the increase in percentages of RHA and cement the MDD decreased whereas OMC increased. Rahman [28] suggests that the particle size and specific gravity of the stabilizer and soil is the reason for the decrease in dry density value. This means that the compactive energy needed to attain dry density will be

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less and that would result in less cost of compaction. The effect on compaction characteristics with the addition of stabilizers is evident from the figure below.

2.2 Effect on Unconfined Compressive Strength The strength of the stabilized soil can be predicted from the unconfined compressive strength test. It is the basis to decide the quantity of stabilizer to be used in treating the expansive soil [39]. It is found that with the increase in stabilizer content curing age and the value of UCS increases. This can be attributed to various researches that were carried out for various soils. The reason for this increase is the reaction of CaOH of soil and silica in RHA. A similar trend was observed in another experiment in which an expansive soil was treated with RHA and calcium carbide residue (CCR) [20]. At 15% of blending content maximum values of UCS and shear strength were reached with an initial water content of 1.20MC as initial water content has a great effect on UCS. The cohesion and angle of internal friction also improved with the addition of stabilizers. The effectiveness of using these materials utilized for stabilization has gained enormous popularity when compared to other stabilizing materials. Moreover, the higher the content of stabilizing agents, the lesser the crack, hence better integrity of the sample. Hossain [16] confirmed the dependency of UCS on the percentage increase of stabilizer. In this study, a varied percentage of RHA and CKD was used and it was found that compressive strength increases with age and stabilizer content. The results were also checked for the combination also. Another important factor discovered in this research was the effect of RHA and CKD on splitting tensile strength. The results were the same as those for UCS, i.e., the tensile strength also improved.

2.3 Effect on Strength Characteristics The effect on CBR (California bearing ratio) is the same as that of UCS and compaction characteristics in relation to the increase of RHA content. It has been noted in Okafor & Okonkwo [28] that the value of CBR increased from 22.05 to 80.14% with the increase of RHA from 0 to 10%. The aim of this investigation was to determine the effect on geotechnical properties of soil on the addition of RHA. The increase in CBR value corresponds to the increase in the RHA content [7, 16].

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3 RHA as a Source of Replacement in Concrete One of the most reliable construction materials is concrete. However, it has become a major source of environmental degradation. The fact that the use of concrete leads to the immense production of CO2 has become a major concern. It requires 4GJ of energy to produce 1MT of OPC, which enough to produce 1MT of CO2 [24]. Efforts have been made to completely substitute the traditional binder with some agro-waste, which includes RHA [15, 22, 41]. The pozzolanic property of RHA comes from the presence of higher levels of silica content. The table below shows the results of some researchers which depict the percentage of silica in these ashes (Table 1).

3.1 Effect on Concrete Properties (a) Compressive strength With the use of rice husk in concrete porosity, unit weight, water absorption, flowability reduced [17]. The compressive strength of the samples was tested after curing of 1, 7, 28, and 98 days. It was noted that with the replacement of 10% RHA compressive strength increased; however, with the further increase in the content, it resulted in the lessening of strength. The experiment resulted in improved microvoids within the cement particle. Partial inclusion of waste has proved to be economical than 100% replacement and that has been proved in an experiment in which cement was replaced by RHA up to 25% and exhibited good results [19]. The results can be depicted from the table below which illustrates that the highest strength was attained at 25% content of RHA (Table 2). The sustainability of RHA as a replacement in concrete has been testified by various researchers. Compared to normal concrete test results, the addition of rice husk ash upsurges the compressive strength up to certain content. D. R Dara and A. C Bhogayata [16] confirmed that at 25% of ash content the compressive strength is maximum and after this, the strength started decreasing. It is also stated in P. V. Ramani et al [32] experiment that the RHA has a retarding effect on strength beyond 10%. The split tensile strength and flexural strength also showed the same behavior as that of compressive strength. Table 1 Presence of silica in different types of RHA

Author

Silica content %

Sivapullaiah et al. [40]

89.2

Anupam et al. 2016

72.24

Prusty et al. 2016

89.87

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Table 2 Compressive strength of rice husk ash cement [19] Mixture proportion (X) OPC

W/B

Slump (mm)

RHA

Compressive strength (MPa) 3 day

7 day

14 day

28 day

100

0

0.70

65

8.90

13.40

16.08

21.34

75

25

0.77

65

5.97

8.31

10.86

15.38

60

40

0.87

65

4.28

6.57

8.52

11.46

50

50

0.95

65

3.06

4.68

6.72

8.92

(b) Resistance to corrosion and sulfates With the inclusion of RHA, the concrete becomes resistant to corrosion. The results of [32] show that the time for initiation of corrosion was more in concrete having 10– 20% of RHA as compared to the normal concrete. Similarly, the resistance toward sulfates is also improved in the concrete with the incorporation of RHA as compared to the controlled concrete.

4 Application of RHA in Other Fields 4.1 As a Liner Material in Landfill Containments Valorization of agro-waste has proved its beneficial potential in every field. Nowadays, these wastes have found their utilization in the geotechnical field as a barrier material in waste containment. A study investigated the use of RHA in a clayey soil to be used as a liner material. The result testified that the liner had a lower potential to desiccate, which is a required characteristic of the liner and also ascertained adequate strength [38]. Bashar et al. [8] confirmed the potential of rice husk ash to be used as an ideal liner material. RHA was mixed with kapok fiber (KF) in the ratio of 1:1 and 1:2. Results demonstrated that the hydraulic conducted was below 10−9 m/s for the mixes of 1:1 ratio (1RH:1KF), which is an essential factor required for a material to function as a liner material. Also, the mix behaves well as an adsorption medium for the removal of leachate contaminants. The appositeness of any material to be the function as a liner material depends on its hydraulic conductivity or permeability [10, 11]. To evaluate the possibility of using RHA as liner material Eberemu et al. [12] conducted a series of tests on tropical clay treated with RHA. RHA of 0%, 4%, 8%, 12%, and 16% by weight of dry soil was used. Hydraulic conductivity increased as the content of RHA increased to 8%, which ascertains the advantageous use of RHA as liner material.

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4.2 As a Cushion Below the Foundation The problem of constructing any structure in expansive soil is because of expansion and shrinkage. Hence researches have been carried out to overcome this problem. RHA being the best stabilizing agent is used to prevail many problems faced in construction engineering. One of the possible uses of RHA is to use it as a cushion below the foundation in expansive soils to reduce the possible shrinkage and swelling. Sivapullaiah et al. [40] explored the possibility of using RHA as effective cushion material. It was found that the stabilized RHA reduced the bandwidth of differential movements in the soil. The two stabilizers used were lime and cement and it was noted that lime stabilized RHA was more effective The author noticed that vertical movements of LSR (lime stabilized) and CSR (cement stabilized) is 0.15% and 0.13% respectively which indicates that less or negligible movement hence they are designated as non-swelling and non-shrinking material.

4.3 As an Admixture in Subgrade Soils It is a well-known fact that the most crucial structural element of pavement is the soil subgrade on which it is laid. The failure of the pavement is associated with the subgrade beneath it. The alternative to this is the use of by-products like rice husk ash. Anupam et al. 2016 used rice husk ash and fly ash to stabilize the soil of subgrade under repeated loading. The variation of elastic modulus and resilient modulus was greatly improved by the addition of stabilizers. The results obtained were better and hence confirmed the potential of these stabilizers to be used as an admixture in treating the deformation of pavement under repeated loading.

4.4 As a Filler in Epoxy Coatings With the risk of exhaustion of natural resources and environmental pollution, the use of sustainable resources has become important. In recent years, RHA has been used as filler in several polymeric materials. One such use of RHA is in the epoxy coating which is widely used in aerospace, automobile, coating, electronics, and encapsulation of semiconductors [30]. The presence of silica in abundance is considered to be the reason for replacing the commercial silica by fillers made of RHA. An experiment was carried out to compare the effectiveness of RHA to be used as a filler with high purity silica [13]. The results suggested that RHA possesses the same characteristics as that of pure silica and thus can be used as a replacement for it.

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Another experiment was done by Azadi et al. [5] studied the effects of RHA to be used as a cheap filler in epoxy paints at two levels of filler content. The wear resistance, scratch resistance, elongation, and paint plasticity were improved.

5 Conclusion In the awe of meeting the exorbitant demand of human beings, the generation of wastes from industries, mining, domestic and agricultural activities have reached soaring levels and thus have led the environment to the perils of being vulnerable to degradation. With such immense quantities of waste generated, they have a great scope to be utilized in many engineering fields as it is evident from the literature. The use of agro-waste as a sustainable construction material has provided a solution to the exhaustion of natural resources as well as energy. This paper evaluates the researches done so far in utilizing the proficiency of rice husk ash in varied engineering applications and hence safeguarding the environment from deterioration.

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Effect of Sulphate Contamination on Lime-Stabilized Black Cotton Soil Shivanshi, Arvind Kumar Jha , Ankush Kumar Jain, and M. Parwez Akhtar

Abstract Chemical stabilization using lime is considered as a suitable practice to suppress the swelling and shrinkage behaviour of expansive soils. However, several distress encountered throughout the world in lime-stabilized sulphatic soil is mainly due to the generation of heave which leads to the questioning of the effectiveness of lime stabilization. An attempt has been made in the present study to examine the physical and swell behaviour of lime-stabilized expansive soil under sulphate contamination. Detail experimental investigations have been done on untreated and lime-treated soil contaminating with varying concentrations of sodium sulphate (0– 30000 ppm). The results revealed that sulphate contamination influences significantly the swell index of soil. Irrespective of sulphate concentration, the swell index of lime-treated soil is observed to be more than that of the same with untreated soil. The formation of expansive mineral, i.e. ettringite results in swell in lime-treated soil. Keywords Lime stabilization · Sodium sulphate · Swell index

1 Introduction The expansive soil is extended in major parts of the world. It is stretched around 20% of the land in India [1]. Expansive soil is considered as problematic soil as it causes uncertain swelling and shrinkage behaviour upon temporal variation. This uncertain behaviour leads to the occurrence of difficulty in the construction practices, and also causes a huge loss of economy due to the failure of the construction. Hence, different techniques such as mechanical method, chemical method, grouting, reinforcing material, etc. have been adopted to overcome such kind of problematic issues [2]. However, chemical stabilization is one of the most effective and efficient Shivanshi · A. K. Jain · M. P. Akhtar Manipal University Jaipur, Jaipur, Rajasthan, India A. K. Jha (B) Indian Institute of Technology Patna, Patna, Bihar, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_6

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methods of soil stabilization. Hence, various calcium-based stabilizers such as lime and cement are used extensively to improve the engineering properties of expansive soils. Lime treatment improves the plasticity and strength properties of soils [3]. Many factors that affect the application of lime for the stabilization purpose include soil gradation, type and amount of clay minerals present, organic matter, exchangeable ions, pH of soil, moisture content, temperature, type of lime, quantity of lime, etc. [4, 5]. Improvement in the properties of lime-treated soils is due to ionic reactions among silica and aluminium present in soil, and calcium present in the lime with time elapsed. These ionic reactions are classified into the short-term and long-term reactions [6, 7] Short-term reactions such as cation exchange, flocculation, and agglomeration are responsible only for an alteration in physical properties of lime-treated soils [6, 8, 9]. However, long-term reaction, also known as pozzolanic reaction, results in the improvement of engineering properties of soils such as strength, compressibility, permeability, and shear strength [10, 11]. However, several cases are reported about the distress caused due to the treatment of soil with calcium-based stabilizers. This is either due to the presence of sulphate in soils or, the migration of sulphatic water in the lime-stabilized soils [12–14]. Sulphate can be found in many forms such as calcium sulphate dihydrate, sodium sulphate, magnesium sulphate, and potassium sulphate. However, it is reported that 20% of the land throughout the world is covered with gypseous soil [5]. Gypsum is considered as a main source of sulphate in soil but, has a lower dissolution rate whereas, sodium sulphate has a high dissolution rate (408 gm/L) compared to other forms of sulphate. Hence, sodium sulphate can be dissolved quickly with surface or, groundwater, and can lead to the contamination of lime-treated soils [5, 12]. The formation of heave due to calcium-based stabilization of soils containing sulphate is termed generally as a “Sulphate Attack” or, “Manmade Heave”. Sulphate induced heave is caused mainly due to the formation of highly expansive minerals i.e. ettringite or, thaumasite depending upon temperature. The induced heave is dominated by many factors like soil’s physical, chemical, and mineralogical properties, environmental conditions including temperature and humidity [15]. Several researches have been done to understand the behaviour of sulphatic soils stabilized with lime. Further, researches have also been carried out to understand its concept and mechanism of microstructural and mineralogical changes that lead to the formation of heave [13, 16–19]. Relatively very few researches have been carried out to understand the effect of sulphate contamination on the lime-stabilized soils [20–22]. Raja and Thyagaraj [20] studied the physical properties of treated soil with sulphate contamination of 5000–20000 ppm sulphate concentration. Hence, a detailed examination needs to be done to know the potential of lime treatment on varying properties of different locations of soils under a wide range of sulphate contamination for its application in the field. The present work is aimed to examine the effect of sulphate contamination in the behaviour of untreated and treated expansive soil with lime. The objectives are achieved by performing detailed experimental tests such as Atterberg’s limits, free

Effect of Sulphate Contamination on Lime-Stabilized …

53

swell index, and compaction characteristics in both untreated and treated soil contaminating with varying concentrations of sulphatic water which are prepared artificially by using sodium sulphate in the laboratory.

2 Materials Used 2.1 Expansive Soil The Black Cotton Soil (BCS) (common name of expansive soil in India) is collected from Shivdaspura village of Jaipur district, Rajasthan state of India. The physical properties of untreated and lime-treated soil are presented in Table 1. The particle size analysis confirmed that soil is predominated with clay-sized particles (90

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Geotextile Nonwoven geotextiles (Type-2) made of polypropylene fibers having puncture strength higher than 2000 N, elongation greater than or equal to 50% in all directions, inertness to alkali, and with good drainage capability are used below geogrid to enable adequate drainage at the bottom level of lime-soil stabilized pile-cap.

3 Design of Ground Improvement System The maximum height of each section is considered to calculate the pressure and accordingly generalized for the whole section for the Design. Since Section-III is of maximum height among all three sections, maximum pressure will be exerted in this section. The load and pressure calculations are considered at critical points of the subsoil, giving due attention to the position of the RE wall embankment, and the concrete block wall at both sides of the channel in which geogrids and geotextiles are inserted during casting of the wall. The distribution of the embankment load within the subsoil is considered as a strip-load.

3.1 Loads and Pressure Calculations The load data from the RE wall embankment superstructure are considered as supplied from the respective agency and cross-checked based on which the pressure distribution is calculated. The critical nodal points are identified for pressure calculation (Fig. 1), and a sample calculation is shown in Table 3. The pressure bulb for each section is drawn, mentioning pressure at critical points along with the design soil profile (Fig. 2). Critical features of the Design The design of the ground improvement system is displayed in Fig. 3. The various codes of Indian Standards (IS) as well as British Standard (BS) along with the support of research literature are followed wherever found reliable, and appropriate. The design is carried out in such a way that, in the worst case, the primary consolidation will be accomplished during the construction stages. The underlying assumptions of the design are to address the limit state of collapse and serviceability. The flexible bamboo pile with or without vertical lime-sand drain is assumed to meet both requirements of the limit state of collapse and serviceability together by achieving early strength of the subsoil and accelerated primary consolidation settlement as well within acceptable time limit. The high tensile and compressive strength, along with flexibility, is the reason that the bamboo is considered as a flexible pile. The flexible pile-cap of 1.0 m thickness designed with stabilized lime-soil encapsulated by geogrid and geotextile affixed with vertical concrete walls at both sides of the

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B

Z

b

q

B = 2b θ θ

P (a)

(b)

(c)

Δσz = Vertical stress at point P, B = Width of the strip load, Z = Vertical distance of point P from the load surface, x = Horizontal distance of point P from the centre of the strip, b = B/2

Fig. 1 Critical Pressure Points under the embankment within subsoil: a Case-1: Point ‘P’ below the center of the strip at a depth z, b Case-2: Point ‘P’ below depth z in between center and corner of the strip or below the corner, c Case-3: Point ‘P’ below depth z but outside the boundary of the strip

Table 3 Pressure calculation from RE Wall embankment load to the foundation (Section-III) Pressure calculation (Section III)

Unit

Unit weight of soil within RE walls

Symbol value  γ 2 = 1.74t m3  3 γ 1 = 2t m

Height from EGL to top of RE wall

H = 5.796 m

–

Height from founding level to top of RE wall

H1 = 7.181 m

–

Height of pavement fills

H2 = 0.6 m

  P3 = 1.2t m2 P4 = 2.63t m2

Traffic surcharge

H3 = 1.385 m  P = 15.7853 t m2  2 Say 15.8 t m

Unit weight of soil at ground

Height from founding level to EGL Net pressure on the foundation base

 P1 = 2.41t m2  P2 = 14.36t m2

– – –

ground improvement channel will provide sufficient bearing capacity to enable the possible load transfer to the bamboo pile allowing free deformation in all around the ground improvement channel. In the design of the bamboo pile, the reduction factor (α) is taken as 0.35, 0.40, and 0.32 against the ‘N’ values (penetration resistance) of 3, 3, and 2, and the ‘c’ (cohesion) of 3.67 t/m2 , 3.18 t/m2 , and 1.5 t/m2 for Sections III, II and I, respectively. The work of Garg 1988 (book no. ISBN 81–88,237-55–8, pp. 64, Fig. 4.14) regarding skin friction with bamboo strips with soil is also assessed to the support of the assumptions. Furthermore, the improvement of the soil cohesion is considered with a multiplying factor of 2.25 (1.5 × 1.5) to the base cohesion for the calculation of pile capacity because of confinement effects of bamboo piling and lime-soil vertical drain for Section III. The same factor is reduced to 1.5, where vertical sand-lime columns are not introduced (Section-I and II). A load factor of 1.3 and a factor of safety (FOS) of 1.35 is considered to assess the pile capacity

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Fig. 2 Pressures at critical points within pressure bulb

in different manners following the standard procedure of pile mechanics. The pile group settlement is also calculated using the related soil parameters as evaluated from the laboratory and field test data following the same principle. Although the pile mechanics of bamboo–soil interactions are not fully known, and very few works are there as such, the assumptions are made to attain a quantitative result over the qualitative principle.

4 Model Experiment Based on the design, the field models were prepared, applying the techniques of the ground improvement in the respective design of Sections I, II, and III. Load Tests The load tests were performed periodically and differently after installation of the bamboo pile to know the load-carrying capacity of the bamboo pile and sand-lime

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column (vertical drain). The loading increment is maintained based on IS: 1888– 1982 with some realistic assumptions. In total, nine successful tests were performed. The settlement is considered for 40 mm. The objective of the load tests is to validate the designed capacity of the bamboo pile, comparing the pile capacity of a single pile corresponding to a 40 mm settlement from the field experiment. Test Results and Discussion In Section III, three tests (PLT-S1B, PLT-S2B, and PLT-S3B) are performed on bamboo pile out of which PLT-S3B was conducted against 25 mm settlement. The test results are shown in Fig. 4a. Two tests (PLT-SSLC-1 and PLT-SSLC-2) are also conducted on the sand-lime-column, the results of which are shown in Fig. 4b. In Section-II, two tests are performed over the bamboo pile group. The result of this is shown in Fig. 5a. In Section-I, two tests (PLT-S5B and PLT-S6B) are conducted over the bamboo pile group, and the graphical results are shown in Fig. 5b. Pile capacity and settlement In Section III, the recovery of the settlement of the bamboo pile groups is 14.07% with an ultimate pile capacity of 1.73 ton, 68.58% with an ultimate pile capacity of 2.58 ton, and 7.68% with un ultimate pile capacity of 1.44 ton, respectively, for PLT-S1B, PLT-S2B, and PLT-S3B. The highest settlement recovery obtained on the test conducted after three months of the installation, and the lowest is observed on the test that was conducted after fifteen days of the installation against a 25 mm settlement. The mean diameter of the bamboo pile for the Section-III, was 65 mm with a mean spacing of 300 mm c/c, although the numbers of the bamboo piles in PLT-S1B and PLT-S2B was nine, whereas the same for PLT-S3B was 6. The pile length was 3.0 m for the tests in Section III. Also, the load test results on the sand-lime columns with the recovery of the settlement are observed as 5.25% with an ultimate 0

20

40

Settlement (mm)

0

60

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PLT-S1B PLT-S2B PLT-S3B

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-20

-30

-60

20

0

0

-50

(t/m2)

Pressure (a)

-60

Pressure (t/m2) (b)

Fig. 4 a Pressure-Settlement graph from load tests on bamboo piles in Section-III at Ch. 7.00 m (PLT-S1B, 9 bamboo piles), at Ch. 24.00 m (PLT-S3B, 6 bamboo piles), and Ch. 21.90 m (PLT-S2B, 9 bamboo piles) of mean diameter 65 mm, spacing @ 300 mm c/c, b Pressure-Settlement graph from Load Tests on the sand-lime column in Section-III at Ch. 4.50 m (PLT-SSLC-1), and Ch. 23.00 m (PLT-SSLC-2) of diameter 400 mm

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Settlement (mm)

0 0 -5 -10 -15 -20 -25 -30 -35 -40 -45 -50

10

20

Pressure (t/m2)

(a)

30

40 0 -5 PLT-S4B -10 -15 -20 -25 -30 -35 -40 -45 -50

0

10

20

30

40 PLT-S5B PLT-S6B

Pressure (t/m2)

(b)

Fig. 5 a Pressure-Settlement graph from load tests on bamboo piles in Section-II at Ch. 80.50 m (PLT-S4B, 9 bamboo piles) of mean diameter 75 mm, spacing @ 300 mm c/c., b Pressure-Settlement graph from Load Tests on Bamboo piles in Section-I at Ch. 105.00 m (PLT-S5B, 16 bamboo piles), and at Ch. 118.00 m (PLT-S6B, 16 bamboo piles) of mean diameter 55 mm, spacing @ 220 c/c

capacity of 2.58 ton, and 22.61% with an ultimate capacity of 2.54 ton, respectively, for PLT-SSLC-1, and PLT-SSLC-2. In Section-II, the recovery of the settlement of the bamboo-pile groups is 34.33%, with an ultimate pile capacity of 1.97 tons. Also, for the bamboo pile, the mean diameter of the bamboo for Section-II was 75 mm with a mean spacing of 300 mm c/c, and the numbers of the bamboo piles were 9 with a pile length of 5.0 m. In Section I, the recovery of the settlement of the bamboo pile groups is 38.56%, with an ultimate pile capacity of 1.07 ton and 53.24% with an ultimate pile capacity of 1.09 ton, respectively, for PLT-S5B, and PLT-S6B. The highest settlement recovery was obtained on the test conducted after two months of installation of bamboo pile compared to the other as one month, although the pile capacity remains very close for both tests. The mean diameter of the bamboo for Section-I was 55 mm with a mean spacing of 220 mm c/c, and the number of the bamboo piles were16 with a pile length of 5.0 m. The settlement recovery is due to the strength-gain of the subsoil in progress over time. Such interactions allow bamboo to bend (due to flexibility) effectively interacting with the subsoil, probably as a macro-bamboo pile-soil-composite due to which such a high amount of recovery of the settlement was possible. The recovery of the settlement is directly related to the recovery of the energy, and which is unique and advantageous for a bamboo pile foundation system for effectively resisting any cyclic events like an earthquake. The pile spacing was lesser in Section-I, although the diameter of bamboo was less than the other sections.

5 Design Validation The bamboo pile system, as designed, is compared with the experimented load test results. The required pile capacity with the factor of safety 1.0 was calculated on

Ground Improvement of Soft Soil Using Bamboo as Flexible Pile … 3

2.79 2.374

2.5 2 1.5

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1.96

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1.82

1.920 1.760

1.35 1.35 1.35

1.284 0.988

1

1.430 1.080

1.105 0.85 0.654 0.36

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0.480

0.468

0 FOS-attained Designed FOS Required pile capacity (ton)

Section-III

Required factored (1.3)pile capacity (ton) Section-II

Calculated unfactored pile capacity (ton)

Designed allowable pile capacity (ton)

Attained pile capacity (ton)

Section-I

Fig. 6 Designed and experimented load-carrying capacity of bamboo pile

the pressure exerted from the RE wall embankment, which was then multiplied by the load factor of 1.30 to get ‘required factored pile capacity.’ The ‘calculated unfactored pile capacity’ is determined using relevant soil strength parameters and assumptions, which was then divided by a factor of safety (FOS) of 1.35 to get the ‘designed allowable pile capacity.’ The mean pile capacity from the load test against a 40 mm settlement is considered as the basis of the comparison. The outcome of the design and experiment results is given in Fig. 6. The test results, as analyzed in comparison to the design considerations confirm that the designed capacity of the bamboo pile is quite safe, which considerably validates the proposed design approach of the ground improvement system. 90% consolidation settlement was calculated to be within 24 days of the construction of the embankment in stages in Section-III. Section-I and Section-II do not require the consideration of the sand drain as the low pressure at critical points within the pressure bulb, and the corresponding settlement is under permissible limits as calculated. Furthermore, the settlement as experimented from the tests are found reasonably satisfactory to ensure the practical validity of the design, and accordingly, construction is done.

6 Conclusion The present report in this paper related to the Agartala flyover covers a new type of ground improvement system by the combination of four techniques altogether to improve soft soil to take embankment loads with vertical RE wall in safe mode. The address of the problem by introducing bamboo pile and lime-soil stabilized flexible pile-cap encapsulated by geogrid and geotextile affixed to a vertical concrete wall at both sides of the ground improvement channel along with the lime-sand vertical drain is found to be appropriate, optimum, green as well as cost-effective. During the

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staged construction of the RE wall, the maximum settlement may be attained. The lime-soil mix is used based on the laboratory test data of one month only. Practically over the time, the strength of the top layer that is stabilized lime-soil of 1 m thickness and lime-sand vertical drain will attain more strength than what is being calculated and will be an additional boon to the design itself. Bamboo has been ideal in clay. Bamboo is long, and so deep penetration is possible. The innovative techniques, as detailed here in this paper, maybe readily used for such urban conditions where the availability of bamboo is not a concern, but the availability of land is a major concern in open traffic conditions. However, the detail and further in-depth study are required to address the actual mechanism of the flexible bamboo pile and lime-soil stabilized geogrid-geotextile encapsulated flexible pile-cap. Acknowledgements The authors gratefully acknowledge the Nagarjuna Construction Company Limited (NCCL), Agartala, for entrusting the work on behalf of the Public Works Department, Government of Tripura. The authors’ thanks are due to the Jadavpur University, Construction Engineering Department, Kolkata, West Bengal, Indian Institute of Technology, Guwahati and National Institute of Technology, Silchar, Assam to avail the related laboratory facilities of the concerned department to conduct such amount of detail laboratory tests in less time. The appreciations are also due to the field and laboratory testing services as delivered by the CE Testing Company Pvt. Ltd., and Civil Engineering Consultancy Services Pvt. Ltd., Kolkata, West Bengal. Besides, the authors also put their sincere thanks to the VKS Infratech Management Pvt. Ltd. and Reinforced Earth India Pvt. Ltd., New Delhi.

References 1. Lima HC Jr, Willrich FL, Barbosa NP, Rosa MA, Cunha BS (2008) Durability analysis of bamboo as concrete reinforcement. Mater Struct 41:981–989 2. Jha J, Choudhary A, Gill K (2010) Bearing capacity improvement of soil using bamboo micropiles. Int J Plasma Sci Eng 03:167–176 3. Widodo B (2015) Influence of bamboo pile in the pile mattress bamboo construction systems as reinforcement of soft subgrade that support embankment load. In: Fredy Kurniawan SM (ed) Proceedings of narotama international conference on civil engineering. Narotama University Press, Narotama, pp 227–238 4. Mandal JN, Manjunath VR (1995) Bearing capacity of strip footing resting on re-inforced sand subgrades. Constr Build Mater 9:35–38 5. Bergado DT, Bukkanasuta A, Balasubramaniam AS (1987) Laboratory pull-out tests using bamboo and polymer. Geotextiles Geomembranes 153–189 6. NBC (2016) National building code of India. Bureau of India Standard, New Delhi, India

Effect of Reinforcement Form on Bearing Capacity of Sand B. Venkatesh and T. Thyagaraj

Abstract This paper brings out the effect of jute geotextile reinforcement on the load-bearing capacity of sand bed in both planar and geocell forms. A series of laboratory model tests were conducted using a 150 mm diameter (D) circular footing in a steel tank of 900 mm × 900 mm × 800 mm size. For the model tests, the relative density (Rd ) of sand was maintained as 70%. The model tests were carried out using the following optimum geometrical parameters—optimum placement depth of first reinforcement layer (u), optimum spacing between the reinforcement layers (h), optimum number of reinforcement layers (N), optimum pocket size of geocell (d), and optimum size of the reinforcement sheet. The performance of both planar and geocell forms of reinforcement was compared. The quantity of reinforcement was maintained the same in both forms. The test results indicate that the improvement in the load-bearing capacity of the reinforced sand bed with planar form is better than the geocell form. Finally, it is concluded that the stiffness of the reinforcement plays a predominant role in the improvement of the load-bearing capacity of the sand beds. Keywords Jute reinforcement · Planar and geocell · Load-bearing capacity

1 Introduction The improvement of the load-bearing capacity of soils using geosynthetics is a very widely used modern technique. The ground improvement using geosynthetics can be obtained with geotextiles, geogrids, geocomposites in various forms such as planar and geocells. Several researchers studied the effect of geosynthetic reinforcements on load-carrying capacity of soils by conducting numerous experimental and numerical studies, e.g., Binquet and Lee [3, 4], Khing et al. [12], Yetimoglu et al. [19], Shin et al. [15], Boushehrian and Hataf [5], Dash et al. [8], Ghosh et al. [10], Basudhar B. Venkatesh (B) · T. Thyagaraj Department of Civil Engineering, Indian Institute of Technology Madras, Chennai, India e-mail: [email protected] T. Thyagaraj e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_17

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et al. [2], Ghazavi and Lavasan [9], Chakraborty and Kumar [7], Naderi and Hataf [14], Tavangar et al. [17], and Xu et al. [18]. However, the studies comparing both planar and geocell reinforcement forms on the load-bearing capacity of soils are very limited. Therefore, the present laboratory study was carried out to understand the effect of the form of geotextile reinforcement on the load-bearing capacity of soils using both planar and geocell reinforcement forms. In order to achieve this objective, the natural jute geotextile was used as a soil reinforcement layer. The laboratory model studies were conducted using circular footing on both unreinforced and reinforced sand beds. In order to ensure the repeatability of the test results, the tests were repeated three times.

2 Experimental Program 2.1 Materials Sand In the present study, all the laboratory model tests were conducted using clean dry sand collected from Chennai. The properties of sand are summarized in Table 1. Geotextiles The natural woven jute geotextile (JGT) was used for the present model tests. The jute geotextile was procured from the National Jute Board (NJB) approved company— Ballyfabs International Ltd., Chennai, India. The physical and mechanical properties of the jute geotextile were determined as per the ASTM Standards given in Table 2. The load-elongation behavior of the jute geotextile was determined from the standard Table 1 Properties of sand

Property

Value

Specific gravity, G

2.68

Sand (%)

99

Fines (%)

1

D10 (mm)

0.25

D30 (mm)

0.42

D60 (mm)

0.95

Cu

3.80

Cc

0.74

emax

0.666

emin

0.466

Unified soil classification symbol

SP

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Properties

Value

Physical properties Thickness (mm) Mass per unit area

2 (g/m2 )

630

Mechanical property Ultimate tensile strength (kN/m) 28.05 × 25.5

Machine direction (MD) Cross-machine direction (CMD)

5.7 × 6.2

Failure strain (%) Machine direction (MD) Cross-machine direction (CMD) Axial stiffness, EA (kN/m)

428

30 MD CMD

Tensile load (kN/m)

25 20 15 10 5 0 0

1

2

3

4 5 Axial strain (%)

6

7

8

Fig. 1 Tensile load-elongation behavior of jute geotextile

wide-width tension test as per ASTM Standard D 4595–11 [1] and the test results are shown in Fig. 1.

2.2 Laboratory Test Setup The present laboratory model tests were performed on sand beds in a steel tank with inner dimensions of 900 mm × 900 mm × 900 mm. For the model tests, the sand bed was prepared in the test tank up to a depth of 800 mm using an air-pluviation

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technique. In all the model tests, the relative density (Rd ) of sand was maintained as 70%. In the present study, 150 mm diameter (D) model circular footing made with 25 mm thick rigid mild steel was used. The base of the model footing was roughened by affixing a thin layer of sand using Araldite. A hydraulic jack was used to push the footing into the sand bed. The model tests were conducted on both unreinforced sand and sand bed reinforced with jute geotextile in two different forms—planar and geocell forms. The geocell mattress was prepared using jute geotextile with required length and height. A polyester thread was used to stitch the joints of geocell pocket. Irrespective of reinforcement form, the model tests were carried out using optimum geometrical parameters in both planar and geocell reinforcement forms. In the case of planar form of geotextile reinforcement, the optimum geometrical parameters such as depth of first reinforcement layer from the footing base (u), spacing between the reinforcement layers (h), number of geotextile reinforcement layers (N) and size of reinforcement (Br ) were maintained as 0.31 times the diameter (D) and 0.3D, 4 and 3.5D (i.e., 525 mm2 ), respectively, following Buragadda and Thyagaraj [6]. In case of geocell reinforcement form, the optimum placement depth of geocell mattress from the footing base (ug ) was adopted from Latha and Somwanshi [13] as ug = 0.05D. Also, the quantity of reinforcement material used for the preparation of the geocell mattress was maintained the same as that of the quantity of reinforcement material used in case of planar reinforcement form with the optimum diameter (dg ), height of geocell (hg ) and width of the geocell mattress (Bg ) as hg = 0.3D, dg = 0.6D and Bg = 5D, respectively. A precalibrated 50 kN proving ring was used for load application on the footing in accordance with IS 1888–1982 [11]. A schematic diagram of the total test setup of the unreinforced sand bed is shown in Fig. 2. The dial gauges named D1 and D2 are used for measuring the footing settlements Fig. 2.

3 Results and Discussion The improvement in load-carrying capacity of the sand bed with the inclusion of the reinforcement layers is evaluated using the term called Bearing Capacity Ratio (BCR) introduced by Binquet and Lee [3]:   Rein forced soil − bearing capacity qrs   BCR = Unrein forced soil − bearing capacity qus

(1)

where qrs and qus are the soil-bearing capacity of reinforced and unreinforced sand beds at the same settlement (s), respectively. At higher settlements, i.e., beyond the ultimate bearing capacity of the unreinforced soil (qult ), the ultimate bearing capacity (qult ) of footing is used in place of qus .

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Fig. 2 Model test arrangement for the unreinforced sand bed

Figure 3 compares the load-settlement behavior of unreinforced sand and sand reinforced with jute geotextile in both planar and geocell forms. From Fig. 3, it can be observed that the load-settlement behavior of the unreinforced sand and jute geotextile reinforced sand in planar form shows a pronounced peak. However, a clear failure is not observed in case of geocell reinforced sand even at higher footing settlements, i.e., s/D = 30%. In case of planar reinforced sand bed, a pronounced peak has occurred due to rupture failure of the geotextile as shown in Fig. 4. Figure 5 presents the variation of bearing capacity ratio (BCR) with footing settlement ratio (s/D). According to Tafreshi and Dawson [16], the allowable settlement of the footing should be between 10% to 15% of footing diameter (D). Hence, in Fig. 5, the variation of BCR of footings for footing settlement ratios (s/D) is presented up to 12% only. It can be observed from Fig. 5 that at a lower footing settlement ratio (s/D) of 2%, the BCR of geocell reinforced sand bed is far higher in comparison to the sand bed reinforced with planar geotextile. However, with the increase in the footing settlement ratio (s/D) beyond 2%, the BCR of jute geotextile reinforced sand bed in planar form is much greater than the BCR of sand reinforced with the geocell.

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200

400

600

800

1000

1200

1400

1600

0

Foong selement rao, s/D (%)

Unreinforced sand 5

Planar reinforcement form Geocell reinforcement form

10

15

20

25

30

Fig. 3 Comparison of load-settlement curves of unreinforced and geotextile reinforced sand beds in planar and geocell forms

Fig. 4 Rupture of the jute geotextile after plate load test

In case of sand reinforced with planar reinforcements, apart from the factors such as placement, layout, and configuration of reinforcement layers, the improvement of load-carrying capacity depends upon the tensile stiffness of the reinforcement material only. However, the load-carrying capacity of the reinforced sand with jute geotextile geocell mattress not only depends on the tensile stiffness of the reinforcement material but also on the bending stiffness of the material. Further, the visual

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3.5

Bearing Capacity Rao, BCR

Planar reinforcement form 3

Geocell reinforcement form

2.5

2

1.5

1 2

4

6

8

10

12

Foong selement rao, s/D (%)

Fig. 5 Variation of the BCR with footing settlement ratio of jute geotextile reinforced sand beds with planar and geocell forms

observations clearly showed that the geocell pocket walls were folded without rupture of the material joints. From Fig. 3, it can be said that the jute geotextile reinforced sand bed in planar form performed better than the reinforced sand in geocell form, which is attributed to the high tensile stiffness of jute reinforcement material along the axial direction in planar form in comparison to the tensile stiffness in the vertical direction of geocell.

4 Summary and Conclusions The experimental studies demonstrate that the reinforcement material properties— both axial stiffness and bending stiffness—play a major role in the improvement of the load-carrying capacity of sand beds and the improvement in terms of bearing capacity ratio depends on the reinforcement form—planar and geocell. Further, the load-carrying capacity of the jute geotextile reinforced sand bed performed better in planar form in comparison to the geocell form owing to the high tensile stiffness of the reinforcement material in axial direction (planar) as compared to the vertical direction (geocell).

References 1. ASTM Standard D 4595–11. Standard Test Method for Tensile Properties of Geotextiles by the Wide-Width Strip Method. American Society for Testing and Materials, Pennsylvania, USA.

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2. Basudhar PK, Saha S, Deb K (2007) Circular footings resting on geotextile-reinforced sand bed. Geotexiles Geomembranes 25(6):377–384 3. Binquet J, Lee KL (1975) Bearing capacity tests on reinforced earth slabs. J Geotech Eng Div, ASCE 101(12):1241–1255 4. Binquet J, Lee KL (1975) Bearing capacity analysis of reinforced earth slabs. J Geotech Eng Div, ASCE 101(12):1257–1276 5. Boushehrian JH, Hataf N (2003) Experimental and numerical investigation of the bearing capacity of model circular and ring footings on reinforced sand. Geotext Geomembr 21(4):241– 256 6. Buragadda V, Thyagaraj T (2019) Bearing capacity of jute geotextile-reinforced sand bed. Int J Geosynthetics Ground Eng 5:27. https://doi.org/10.1007/s40891-019-0178-6 7. Chakraborty D, Kumar J (2012) Bearing capacity of strip foundations in reinforced soils. Int J Geomech 14(1):45–58 8. Dash SK, Rajagopal K, Krishnaswamy NR (2004) Performance of different geosynthetic reinforcement materials in sand foundations. Geosynthetic Int 11(1):35–42 9. Ghazavi M, Lavasan AA (2008) Interference effect of shallow foundations constructed on sand reinforced with geosynthetics. Geotext Geomembr 26(5):404–415 10. Ghosh A, Ghosh A, Bera AK (2005) Bearing capacity of square footing on pond ash reinforced with jute-geotextile. Geotext Geomembr 23(2):144–173 11. IS 1888–1982. Method of load test on soils. Determination of bearing capacity of soils and its settlement. Bureau of Indian Standards. 12. Khing KH, Das BM, Puri VK, Cook EE, Yen SC (1993) The bearing capacity of a strip foundation on geogrid reinforced sand. Geotext Geomembr 12(4):351–361 13. Latha M, Somwanshi A (2009) Effect of reinforcement form on the bearing capacity of square footings on sand. Geotext Geomembr 27(6):409–422 14. Naderi E, Hataf N (2014) Model testing and numerical investigation of interference effect of closely spaced ring and circular footings on reinforced sand. Geotext Geomembr 42(3):191–200 15. Shin EC, Das BM, Lee ES, Atalar C (2002) Bearing capacity of strip foundation on geogridreinforced sand. J Geotech Geolo Eng 20(2):169–180 16. Tafreshi SM, Dawson AR (2010) Comparison of bearing capacity of a strip footing on sand with geocell and with planar forms of geotextile reinforcement. Geotext Geomembr 28(1):72–84 17. Tavangar Y, Shooshpasha I (2016) Experimental and numerical study of bearing capacity and effect of specimen size on uniform sand with medium density, reinforced with nonwoven geotextile. Arabian J Sci Eng 41(10):4127–4137 18. Xu C, Liang C, Shen P (2019) Experimental and theoretical studies on the ultimate bearing capacity of geogrid-reinforced sand. Geotext Geomembr 47(3):417–428 19. Yetimoglu T, Wu JTH, Saglamer A (1994) Bearing capacity of rectangular footings on geogridreinforced sand. J Geotech Eng Div, ASCE 120(12):2083–2099

Deformation of Stone Column Subjected to Earthquake Loading by Numerical Analysis Maniam Rajan Priyadharshini

and Premalatha Krishnamurthy

Abstract Stone columns have been widely used as one of the economical ground improvement techniques in soft soils. Stone columns improve the settlement characteristics of soft soil deposits. Besides, stone columns are capable of dissipating excess pore water pressure when subjected to earthquake loading, by acting as a drainage component. The mechanism of stone column under static loading conditions is well known, but the study on the behaviour of stone column subjected to earthquake loading is meagre. The objective of the study is to compare the bulging pattern of stone column and stress concentration factor under earthquake loading using numerical analysis software, PLAXIS 2D. For validation, numerical analyses are carried out from field test results, for static loading and earthquake loading. These cases are simulated in software and the load–settlement curves are found to be comparable with the field test results. For the present study, static analyses are carried out for loading on stone column alone and equivalent area of stone column. Earthquake analyses are carried out by loading stone column alone. The variation in the bulging pattern and stress concentration factor for the above said conditions are observed. Keywords Stone column · Settlement · Earthquake loading · PLAXIS 2D · Deformation pattern

1 Introduction Stone column installation as a ground improvement technique developed over many decades paved the way to utilize weaker soil deposits all around the world. This technique thereby led to infrastructure development even in unsuitable ground profile.

M. R. Priyadharshini (B) · K. Premalatha Department of Civil Engineering, College of Engineering Guindy, Anna University, Chennai, India e-mail: [email protected] K. Premalatha e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_18

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Stone columns improve the strength of the soil deposit by acting as a reinforcement and drainage component. Stone columns act as a vertical drainage and increase the drainage of clay deposits and reduce post-construction settlement of the structure. Stone columns were studied extensively and still it is a topic of interest with enhancement in field applications such as encased stone columns. Also, development in technologies led to study of the seismic behaviour of the stone column with the help of laboratory tests and numerical modelling.

2 Literature Review Stone columns were introduced decades back and kinds of literature started getting published from the 1960s. They were studied extensively both experimentally and numerically. Research on static loading is abundant. In recent years with innovative technologies and numerical models, dynamic studies are being carried out. With such an overview, works of literature are presented here for static and dynamic loading.

2.1 Static Loading The design of the stone column was developed for treating cohesive soil based on the theory of passive resistance by Greenwood [1] through field observations and Hughes and Withers [2] using a pressure meter test. Datye [3] proposed a theoretical solution for stone columns based on cavity expansion theory. Balaam [4] presented an analytical solution based upon the theory of elasticity to determine the settlement magnitude and settlement rate. Alamgir [5] proposed an analytical solution for predicting stresses and settlement based on the unit cell concept. Lee [6] proposed a numerical model for the elastic and elasto-plastic analysis of stone columns. The solutions were able to predict the stone column behaviour based on experimental results. Even in recent years also many solutions are being proposed based on numerical and experimental observations. Extensive studies by experimental investigations reported the mechanism of stone columns. The behaviour of stone columns, when subjected to widespread load, is by bulging of the column, effect of surcharge and bearing support from the intervening soil [7]. The critical length of the stone column is the depth beyond which the column fails by bulging. The depth at which maximum bulging occurs was reported by many researchers through experimental studies and varies between 0.5D and 2D [1, 2, 8, 9]. This variation is due to the presence of very soft cohesive soil which may cause bulging even at a shallow depth of the soil deposit [7]. Also, such a reduced depth of bulging can be observed if the applied load is confined to the stone column alone. The maximum critical length of the stone column is reported as 4D–6D [1, 2, 10, 11].

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Parametric studies using numerical models were also conducted widely based on many factors. Some of the findings are as follows. The spacing between stone columns is 2D–3D [7, 8]. An increase in the load-carrying capacity of the stone column increases with an increase in diameter. Settlement of the stone column reduces with the increase in the length of the column [12]. Stress concentration factor (n) value depends on the length of the stone column and shear strength of the surrounding soil. With time, stress concentration ratio increases as the consolidation process occurs. For a higher L/D ratio, stress concentration factor increases [13, 14]. Stress concentration ratio was found to be in the order of 2–3 computed on the level of the stone column [14, 15]. This value tends to increase when the load on the stone column is applied through rigid foundations [7]. For encased stone columns (ESC), the provision of encasement improved the loadcarrying capacity of the column. It also reduced the lateral bulging of the column. The effective length for providing encasement is reported as 2D [16]. The performance of ESC is independent of the strength of the surrounding soil [8]. The stress concentration factor of ESC is higher than the conventional stone columns because of the increased stiffness due to encasement. Small diameter ESC can perform well due to the mobilization of confining stress when compared with larger diameter ESC [10].

2.2 Earthquake Loading Studies on stone column subjected to earthquake loading involve both experimental testing and numerical modelling. Some of them are discussed here. Experimental Investigation. Kim [17] conducted a 1-g shake table test to study the seismic behaviour of stone column reinforced ground. The experimental setup was instrumented using accelerometers. Earthquake time history of Hachinohe (1994) and Northridge (1994) [18] were given as input motion. From the observations, it was reported that stone columns can reduce shear deformation of the improved ground to a great extent when compared to unimproved ground. The frequency content of the earthquake motion influenced the seismic performance of the stone column. Cengiz [19] experimentally investigated the seismic performance of encased and ordinary stone columns. Shake table tests were conducted in the laboratory using large steel tanks. The experiment was well instrumented using strain gauges and laser displacement sensors. El-Centro (1940) and Kobe (1995) [18] earthquake time history were given as the input motion. From the observations, it was reported that end bearing stone columns performed well when compared with floating columns. An encased stone column with higher stiffness deformed more uniformly. When seismically loaded, encasement in stone columns showed higher deformation at a depth of 5D–7D which was more than the usual encasement depth of 2D. Numerical Modelling. Erol [20] numerically studied the behaviour of encased stone column using TNO DIANA software. A field study of the construction of an airport is modelled. Izmit, Turkey (1999) [18] earthquake time history was given as input

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motion. From the analyses, it was observed that the encasement of stone columns prevent bulging failure during seismic action. Also, the confinement provided by the encasement helps to keep the encased stone column as a single unit during seismic excitation. It was also observed that in the upper portion of the encased stone column, tensile stresses developed was higher when compared to the lower part of the stone column. This was observed due to the prevention of column bulging in the upper layers of the stone column. Chamala [21] studied the behaviour of stone column in the sloping ground with layered deposits by numerical modelling. OpenSeesPL software was used for the analysis. For the analysis, Nepal 2015 [18] earthquake time history was given as the input motion. Single stone column was analyzed numerically. Numerical results were compared for sloping ground with and without stone columns. Peak ground acceleration, lateral displacement, excess pore water pressure and shear strain values were compared for both cases. For the treated slope the observed values were found to be lower than the untreated ground by 0.7%, 6%, 2% and 8%, respectively. Overall improvement by the use of stone column in sloping ground was observed. Ferhat [22] numerically studied the seismic behaviour of floating stone columns. PLAXIS 2D software was used for the numerical analysis. A parametric study was carried out to analyse the bearing capacity improvement of stone column reinforced ground. Van Muradiye (2011) [18] earthquake time history was given as input motion. From the analysis results it was observed that bearing capacity of the stone column reinforced ground improved by 38% under static condition and by 36% when subjected to earthquake loading. Therefore as a net effect, earthquake loading reduces the performance of the stone column. From the above literature, it is clear that the behaviour of a stone column was studied using experimental techniques and numerical modelling. The results were reliable and found to be comparable with the postulated theories. Consequently, for the present study numerical analysis is chosen to examine the behaviour of stone columns. For stone columns subjected to earthquake loading, understanding the bulging behaviour, stress concentration factor and influence of other parameters on load-settlement behaviour also required to be studied as the literature available are meagre. Therefore based on the identified gap in study, the objective coined for this paper is to compare the deformation pattern, load–settlement curve and stress concentration factor of stone column when subjected to static and earthquake loading using numerical analysis software, PLAXIS.

3 Numerical Modelling Using PLAXIS 2D PLAXIS 2D is a finite element software for two-dimensional geotechnical analysis. In software, the stone column is modelled in axisymmetric condition. Fourth-order 15-noded elements are used for deformation study in soil. The model dimensions are chosen based on the extent of bulging of the stone column. The model geometry is 3 m and 10 m in horizontal and vertical direction respectively. The mesh quality is set

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as a medium with a mesh refinement in stone column region for accuracy in output results. In the framework of this study, a homogenous clay deposit of 10 m thickness, reinforced with stone column is considered. The length of the stone column is 6 m with a triangular pattern of arrangement and spacing of two times the diameter of the stone column. The groundwater table is assumed to be at the existing ground level. This field condition is simulated in PLAXIS software and is elucidated in the following section. Static analysis is carried out by loading on the stone column alone and loading on an entire area of stone column [8]. Though loading on stone column alone is not practically applicable, this condition is chosen to understand the effective transfer of load by the stone column. Earthquake analysis is performed by loading the stone column alone. The calculation module includes five construction stages. In the first phase, initial stresses are generated. The initial phase is followed by stone column installation, consolidation of the ground, loading of the stone column. For earthquake loading, in addition to the above-mentioned phases, simulation of the earthquake is included.

3.1 Validation of the Numerical Model Validation for static analysis. The field behaviour of clay and stone column is simulated using the Mohr–Coulomb model. The validity of the Mohr–Coulomb model in the analysis was verified using field study [23]. The stone column diameter and length is 600 mm and 10.5 m respectively. Unit weight of stone column material is 22 kN/m3 , angle of internal friction is 47° and Young’s modulus is 15 MPa. The properties of soil for the bore log profile are presented in Table 1. The properties of the soil were not given in the bore log report; therefore correlations for N value as recommended by Bowels [24] were used for the analysis. The settlement of the stone column from field test and numerical analysis were 10.8 mm and 12.12 mm respectively. The percentage variation in the results is about 11%. The load– ettlement curves from the field test report and the numerical analysis were comparable as shown in Fig. 1. Validation for earthquake analysis. For verifying the validation of the numerical model under earthquake loading condition, a field case study involving the occurrence of earthquake in Guatemala was selected [25]. The soil properties of the field profile are given in Table 2. The site occupies a thermal power plant. As the region is earthquake-prone, power plant foundations were instrumented to study the seismic effect. Power plant was constructed on the ground improved with stone columns of diameter 0.95 m and length 20 m, placed in a square grid of 2.70 m. After the completion of the construction of power plant, an earthquake of magnitude 7.4 hit Guatemala on 12 November 2012. The Peak Ground Acceleration recorded was 0.04 g at Guatemala City [26]. The settlement of foundations before and after the earthquake was monitored as 14 mm and 4 mm, respectively. The field details are simulated in PLAXIS and the settlement values are observed as 16.31 mm and

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Table 1 Properties of the soil profile for static analysis Layer

Soft clay

Loose sand

Medium dense sand

Loose sand

Medium stiff clay

Loose sand

Very stiff clay

Soil layer thickness (m)

0−1.5

1.5−4.5

4.5−6

6−7.5

7.5−9

9−10.5

>10.5

Average SPT N value

4

7

14

8

6

6

22

Saturated 14 unit weight (kN/m3 )

16

18

16

17

16

19

Cohesion (kPa)

–

–

–

37.5

–

100

Angle of – friction, ϕº

28

30

28

–

30

–

Young’s 4500 modulus, E (kPa)

15,000

20,000

15,000

25,000

15,000

100,000

18.75

Pressure (kPa) 0

100

200

300

Settlement (mm)

0 Numerical analysis result

5

field test result 10 15

Fig. 1 Comparison of load–settlement curve with the field test result Table 2 Properties of the soil profile for earthquake analysis

Layer

Clayey sandy gravel

Sandy layer

Silty sand and sandy layers

Soil layer thickness (m)

0−6

6−10

10−15

Unit weight (kN/m3 )

14.95

14.68

12.49

Angle of friction, ϕº

44

38

30

Young’s modulus, E (kPa)

150,000

45,000

20,000

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4.17 mm, respectively. The percentage difference is 16.5% and 4% for settlement before and after the earthquake. As the numerical model was able to bring out the behaviour of the soil system both in static and earthquake loading system, it is considered to be valid. Input parameters for the present study. Numerical analyses were executed by varying the diameter of the stone column and magnitude of earthquake loading. The parameters considered for the analyses are shown in Table 3. The properties of clay and stone column used for numerical modelling are listed in Table 4. The safe load on the stone column is calculated as per IS 15284 (Part 1): 2003 [7]. The load capacity calculation as per the code is given for widespread loads based on the contribution of resistance against bulging of column, presence of surcharge and capacity of virgin soil. For the current study, it is observed from the numerical analysis that the load capacity of the stone column is purely by resisting force against bulging and capacity of virgin soil to some extent, as the clay deposit considered here has lesser shear Table 3 Parameter details Type of soil deposit

Clay

Thickness of soil deposit

10 m

Stone column diameter

300, 450 and 600 mm

Stone column length

6m

Type of stone column

Floating type

Loading

Stone column alone loaded, equivalent area of stone column loaded

Magnitude of earthquake [19]

5.5–PGA–14.1 cm/s2 (Nagarkot-Nepal 2015), 6.5–PGA–23.1 cm/s2 (Jaco-Costa Rica 2017) and 7.5–PGA–0.66 cm/s2 (Papua, New Guinea 2018)

Safe load as per IS 15284 (Part 1): 2003

14 kN, 33 kN and 63 kN for 300 mm, 450 mm and 600 mm respectively

Applied load on stone column

10 kN, 22 kN and 36 kN for 300 mm, 450 mm and 600 mm respectively

Table 4 Soil and stone column properties used in the present study [27]

Material (depth)

Clay (10 m thick) Stone column

Model

Mohr–Coulomb

Mohr–Coulomb

Unit weight, γsat (kN/m3 ) 14

18.4

Young’s modulus, E (kPa) 1000

10,000

Poisson’s ratio, υ

0.45

0.25

Cohesion, Cu (kPa)

6

–

Friction angle, ϕ (°)

–

48°

Permeability, kx (m/day)

8.1e-4

1

ky (m/day)

8.1e-4

1

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strength. There is no surcharge acting over the stone column, so the contribution by surcharge effect is neglected.

4 Results and Discussion Based on numerical analyses load–settlement curves, bulging of a stone column and stress concentration factors are discussed here for static and earthquake loading.

4.1 Static Analysis Stone column alone loaded. Analyses were performed by loading the stone column alone. From the output results, some typical figures of deformed mesh, settlement and vertical stress are shown in Fig. 2. Load–settlement curves. Load–settlement curve for diameter 300, 450 and 600 mm are shown in Fig. 3. Stone columns are analyzed for loading till the verge of

Fig. 2 Typical results for 300 mm diameter for stone column alone loaded condition. a Deformed mesh b Vertical settlement c Radial displacement d Vertical stress

Fig. 3 Load—settlement curves for different diameters

Load (kN)

Selement(mm)

0 0 5 10 15 20 25 30 35 40 45

10

20

30

40 300 mm dia SC 450 mm dia SC 600 mm dia SC

Deformation of Stone Column Subjected to Earthquake Loading … Fig. 4 Bulging of stone column for different diameters

195

x/D 0.000 0

0.005

0.010

0.015

z/D

2 4

D=300 mm

6

D=450 mm

8

D=600 mm

10

failure. Though the load-settlement curve is non-linear, the linearity portion increases with an increase in the diameter of the stone column. The ultimate load for 300, 450 and 600 mm diameters are 8 kN, 19 kN and 30 kN, respectively. For the permissible settlement of 12 mm, corresponding load capacities noted are 7, 14 and 21 kN for 300 mm, 450 mm and 600 mm diameters respectively. Bulging pattern. Prominent bulging pattern can be seen from Fig. 2. Radial displacement or bulging of a stone column is plotted for 300, 450 and 600 mm diameters and shown in Fig. 4. For comparison, bulging and depth are expressed in terms of the diameter of the stone column. The value of maximum bulging was 3.7 mm, 5.5 mm and 7.5 mm for 300 mm, 450 mm and 600 mm respectively. Maximum bulging is observed at a depth of 0.5D and the radial distance up to which bulging is observed is up to a distance of 2D from the periphery of the stone column. Total depth of bulging is 4D. Stress concentration factor (SCF). SCF is defined as the ratio of average stress in the stone column to the stress in the soil within a unit cell. SCF is computed for failure load. Based on the analysis results, vertical effective stresses are read as an average value over the head of the stone column and clay within 1.05 times spacing of stone column (equivalent diameter for triangular pattern). For 300 mm diameter, stress on a stone column is observed to be 135.8 kN/m2 , whereas in clay it was 2.12 kN/m2 . SCF computed is 63. Similarly, for 450 and 600 mm, SCF is 61 and 59. The capacity of the stone column increases with diameter, but there is a reduction in SCF by a percentage of 3%. Such a large value of SCF is due to the loading of the stone column alone. This is because the maximum extent of the load is taken by the stone column. Equivalent area of stone column loaded The equivalent diameter of the stone column is loaded and analysed. The equivalent diameter for the triangular pattern of stone column is assumed. For a spacing of two times, the diameter of the stone column the equivalent diameter is computed and utilized for the numerical analysis. From the output results, some typical figures of loading condition and deformed mesh are shown in Fig. 5. Load-settlement curve. The load–settlement curve of stone columns loaded over the equivalent area for 300 mm, 450 mm and 600 mm diameters are shown in Fig. 6.

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Fig. 5 Equivalent area of 300 mm diameter stone column loaded a typical loading b deformed mesh

Fig. 6 Load–settlement curves for different diameters

Load (kN) 0

5

10

15

20

selement (mm)

0 10 20

300 mm 450 mm 600 mm

30 40 50

The ultimate load for 300 mm, 450 mm and 600 mm diameters are 3 kN, 6.5 kN and 12.5 kN, respectively. For the permissible settlement of 12 mm, corresponding load capacities observed are 2.8 kN, 5.4 kN and 8.3 kN for 300 mm, 450 mm and 600 mm diameters, respectively.

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Bulging pattern. From Fig. 7, it can be pointed out that bulging is in maximum value at a depth of 1.5D and bulging is observed up to a radial distance of 2.7D from the periphery of the stone column. Total depth of bulging observed in 5D. Bulging values are 49 mm, 60 mm and 62 mm for 300 mm, 450 mm and 600 mm diameter, respectively. In this case, clay is also subjected to loading which offers less resistance to bulging. This increases bulging with an increase in the diameter of the stone column. Stress concentration factor (SCF). When the equivalent area is loaded up to failure, both column and clay bear the load, thereby reducing the SCF value. Similar to the previous section, SCF values are computed. Irrespective of the diameter of the stone column, SCF value remains the same. SCF value is lesser than the values calculated for stone column alone loaded condition.

4.2 Earthquake Analysis When the stone column alone is loaded and subjected to an earthquake of different magnitudes, there will be some change in the behaviour of stone column in terms of settlement, bulging pattern and stress concentration factor. Settlement of the stone column. Earthquake is simulated using time history input motion using a prescribed displacement option. For simplification, instead of generating load-settlement curves, settlement values alone are reported and compared in Table 5. Settlement depends on the diameter of the stone column and magnitude of earthquake. The decrease in settlement value for a 7.5 magnitude earthquake could be due to the lesser peak acceleration when compared to 5.5 and 6.5 magnitude earthquake. Bulging pattern of stone column. From the analyses results, it was observed that irrespective of the magnitude and diameter of the stone column, the maximum depth of bulging is 0.5D and total depth of bulging is 4D as observed in static condition.

z/D

Fig. 7 Bulging of stone column for different diameters

0.000 0 1 2 3 4 5 6 7

0.020

x/D 0.040

0.060

0.080

300mm 450mm

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Table 5 Comparison of settlement for Static and Earthquake Analysis Diameter of stone column (mm)

Settlement of stone column (mm) Magnitude of earthquake

Static loading condition

5.5

6.5

7.5

300

37.96

35.81

37.01

24.0

450

49.98

50.8

50.3

31.0

600

51.52

50.57

51.98

39.0

Table 6 Comparison of stress concentration factors Diameter of stone column (mm)

Stress Concentration Factor Earthquake Magnitude

Static loading condition

5.5

6.5

7.5

300

103

101

104

63

450

132

130

133

61

600

127

125

127

59

Bulging can be observed up to a radial distance of 3D from the periphery of the stone column, which is higher than the static condition. Stress Concentration Factor (SCF). Stress concentration factor for earthquake loading is of higher-order when compared with the static condition and is presented in Table 6. SCF depends on the diameter of the stone column and the magnitude of an earthquake. From the observations, for a stone column diameter and different magnitude, SCF values are almost uniform. For 450 mm diameter and varying magnitude, SCF value is slightly higher which possibly is influenced by the amplitude, time duration and frequency content of the earthquake.

5 Conclusions The results described here are from numerical analyses. The conclusions made are presented below: For static loading, the load-carrying capacity improves with an increase in the diameter of the stone column. Maximum bulging occurs at a depth of 0.5D when the stone column alone is loaded. Bulging is observed up to a radial distance of 2D from the periphery of the stone column. Total depth of bulging is observed as 4D. Stress concentration factor is of higher order when stone column alone is loaded. For static loading, when the equivalent area of the stone column is loaded, bulging occurs at a depth of 1.5D. The total depth of bulging is 5D. Bulging increases with an increase in the diameter of the stone column. Irrespective of the column diameter, stress concentration factor remains the same. Bulging is seen up to a radial distance of 2.7D from periphery of the stone column.

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For earthquake loading, when stone column alone is loaded the depth at which maximum bulging occurs is 0.5D and total depth of bulging is 4D, which are observed irrespective of the magnitude and diameter of the stone column. Stress concentration is much higher when compared to the static condition. This indicates the efficiency of the stone columns in mobilizing the stresses when subjected to earthquake loading. All the views are made from the results of numerical analyses. Though the results confirm with the established theories, they need to be verified with the experimental investigations and field-based data for a better understanding of the static and seismic behaviour of the stone column. This forms the future scope of the study.

References 1. Greenwood DA (1970) Mechanical Improvement of soils below ground surface. In: Proceedings of ground engineering conferebnce instn civil engineers 2. Hughes JMO, Withers NJ (1974) Reinforcing of soft cohesive soils with stone columns. Ground Eng 7(3):42–49 3. Datye KR (1981) Design approach and field control for stone columns. In: Proceedings of 10th international conference on SMFE, vol 3, pp 637–640 4. Balaam NP, Booker JR (1981) Analysis of rigid rafts supported by granular piles. Int J Numer Anal Methods Geomech 5(4):379–403 5. Alamgir M, Miuran N, Poorooshab HB, Madhav MR (1996) Deformation analysis of soft ground reinforced by columnar inclusions. Comput Geotech 18(4):267–290 6. Lee JS, Pande GN (1998) Analysis of stone-column reinforced foundations. Int J Numer Anal Meth Geomech 22(12):1001–1020 7. IS: 15284 (Part 1) (2003) Indian Standard code of practice for design and construction for ground improvement–guidelines, Stone columns, Compendium of Indian Standard on Soil Engineering 8. Ambily AP, Gandhi SR (2007) Behaviour of stone columns based on experimental and FEM analysis. J Geotech Geoenviron Eng 133(4):405–415 9. Mohanty P, Samanta M (2015) Experimental and numerical studies on response of the stone column in layered soil. Int J Geosynthetics Ground Eng 1(3):1–14 10. Malarvizhi SN (2007) Comparative study on the behaviour of the encased stone column and conventional stone column. Soils Found 47(5):873–885 11. Sivakumar V, Jeludine DKNM, Bell A, Glynn DT, Mackinnon P (2011) The pressure distribution along stone columns in soft clay under consolidation and foundation loading. Géotechnique 61(7):613–620 12. Madun A, Meghzili SA, Tajudin SAA, Yusof MF, Zainalabidin MH, Al-Gheethi AA, Md Dan MF, Ismail MAM (2018) Mathematical solution of the stone column effect on the load-bearing capacity and settlement using numerical analysis. J Phys: Conf Ser 995(1):1–9. IOP Publishing 13. Fattah MY, Shlash KT, Mohammed Al-Waily MJ (2011) Stress concentration ratio of model stone columns in soft clays. Geotech Testing J 34(1):50–60 14. Al-Ani W, Wanatowski D, Chan SH, Serridge CJ (2015) Numerical modelling of stone columns in soft soils 15. Mitchell JK, Huber TR (1985) Performance of a stone column foundation. J Geotech Eng 111(2):205–223 16. Murugesan S, Rajagopal K (2006) Geosynthetic-encased stone columns: numerical evaluation. Geotext Geomembr 24(6):349–358 17. Kim J, Son S, Mahmood K, Ryu J (2012) Site response and shear behaviour of stone column improved ground under seismic loading. In: Proceeding of the 15th worlds conference on earthquake engineering

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18. https://strongmotioncenter.org/. Last accessed 28 Dec 2019 19. Cengiz C, Güler E (2018) Seismic behaviour of geosynthetic encased columns and ordinary stone columns. Geotext Geomembr 46(1):40–51 20. Guler E, Alexiew D, Abbaspour A, Koc M (2014) Seismic performance of geosynthetic-encased stone columns. Transp Res Record 2462(1):77–88 21. Reddy CS, Mohanty S (2017) 87 Seismic behaviour of stone column on a sloping layered soil, 6IYGEC2017, pp1–4 22. Sahinkaya ¸ F, Vekli M, Çadır CC (2017) Numerical analysis under seismic loads of soils improvement with floating stone columns. Nat Hazards 88(2):891–917 23. Consultancy work (2018) Stone column design for secondary clarifier tank for Manali Petro Chemicals Pvt. Ltd. 24. Bowles JE (1997) Foundation analysis and design. McGraw-Hill Companies Inc., Singapore 25. Callejas F, Luna R (2017) Seismic ground improvement: stone columns performance for a power plant in the southern alluvial plains of Guatemala. In: Proceedings of the 19th international conference on soil mechanics and geotechnical engineering. Seoul 26. https://www.eeri.org/cohost/Special-Earthquake-Reports/M7.4_Guatemala_Earthquake/ EERI-AGIES_Guatemala_Report.pdf. Last ccessed 19 Jan 2020 27. Han J, Ye S-L (2001) Simplified method for consolidation rate of stone column reinforced foundations. J Geotech Geoenviron Eng 127(7):597–603

Ground Improvement Technique By Psuedo-Binghamian Grout for Fly Ash and Black Cotton Soil Nirali B. Hasilkar, Lalit S. Thakur, and Atul Panchal

Abstract Various processes of ground improvement are available to increase the strength, reduce compressibility and permeability or improve groundwater condition. The techniques involved for the attainment of the required improvement properties are referred to as a geotechnical process. While considering any foundation problem, a civil engineer will probably examine the use of grouting as one of the possible solutions. The selection of grout material and its technique is of utmost importance for the success of grout. Cement, cement and admixture, chemicals, fly ash, pozzolans, etc., are the commonly used grouting materials. The process of injecting a suitable fluid under pressure was the invention of a French engineer Charles Berginy, who in 1802 prevented scouring of a sluice gate by injecting clay-based grout. The present project intends to design a grout using sodium silicate as the base and calcium chloride as a hardener. Fly ash and black cotton soil are used as a filler material to enhance the property of the designed grout. An experimental study is carried out to study the physical properties and engineering properties of raw and grouted sand for the designed grout. The experimental study conducted gave the following broad inferences that gel time reduces as hardener concentration increases, while specific gravity, time of afflux, needle penetration strength, and syneresis increases. It has been observed that UCS and triaxial strength of raw and grouted (dry and saturated sand) increases as the time of curing increases. Cohesion and angle of internal also increases as curing time increases. Keywords Flyash · Black cotton soil · Psuedo-binghamian grout · Syneresis

N. B. Hasilkar (B) · L. S. Thakur Civil Engineering Department, BITS Edu Campus, Varnama, Vadodara 391240, Gujarat, India e-mail: [email protected] A. Panchal Afcons Infrastructure Limited, Shapoorji Pallonji Group, Ahmedabad 380018, Gujarat, India © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_19

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1 Introduction The engineering treatment of soils involves improving their geotechnical character for construction purposes. Problems regarding construction in soils arise from their lack of strength, which manifests itself in their deformation which beneath the foundation takes the form of settlement, or in some exceptional circumstances gives rise to ground failure. Ground improvement by grouting is used in various geotechnical engineering applications like in the construction of tunnels, shafts, and dams for the purpose of either reducing permeability and/or increasing the mechanical stability in water-bearing soil or rock. The conventional OPC grouts pose problems of penetration into medium-fine sands. To overcome this problem, various chemical grouts have been developed and implemented during the past few years to penetrate into finely fissured rocks or fine sands tending to silt irrespective of their cost. The present-day scenario to look into the environmental compatibility of grouts has made sodium silicate grouts to be more commonly used. For cost-effective solution, black cotton soil and fly ash were mixed with the sodium silicate based grouts. An attempt is made to study the physical and engineering properties of grouted sand.

2 Literature Review Sodium silicate grouts are the most popular grouts because of their safety and environmental compatibility. Sodium silicates have been developed into a variety of different grout systems. Singh et al. [1] used formamide for evolving silica gel and studied the efficacy of the grout for dune sand by measuring the unconfined strength up to a value of 38 kg/cm2 for a moist cured sample. Tokoro and Kashimo (1982) initiated research on flash setting grout to be used for special geotechnical problems dealing with drilling in fractured rock and grouting in interbedded sand under ground-water flow. Flash-setting grout is developed employing sodium silicate and a hardener containing a combination of bisulfate, sulfate, and bicarbonate. The time-strength relationships of silica gel developed by using phosphoric acid and sodium aluminate were studied by Shroff and Moghe (1980). Rhone Prongil (1972) described material motion changes in unconfined compressive strength of silica gel-sand test samples immersed in water, in terms of the sodium silicate’s rate of destabilization (Ratio reagent/sodium silica in the gel). Warner (1970) used the water cure method to test silica gels for durability. After setting, the silica gel is subject to various alterations which may affect its durability in varying degrees. It should be pointed out, that finer the sand, the less intense will be the disturbances caused by syneresis [2]. In order to identify the possible effect of these different phenomena in specific terms, the behavior of the gel needs to be examined using various curing methods such as curing in an airtight medium, dry curing, wet curing, and/or curing in water under pressure.

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Cambefort and Caron [3] analyzed the strength of silica gel by the permanence aspect of silica gel produced by acid, sulfate, bicarbonate, and aluminates as precipitants, by performing, washout test of grouted sand. Bicarbonate and phosphate gels have an almost insoluble framework that prevents the passage of water, while in other gels the framework diminishes in importance more or less rapidly.

3 Work Outline The experimental work has been carried out to determine the effect of various percentages of hardeners keeping the percentage of sodium silicate constant in the grout. As an additional material, fly ash and black cotton soil were used so that the grouts can be in both categories of Newtonian and Binghamian fluids.

3.1 Scope of Work • To design and develop an optimal Pseudo-Binghamian grout using sodium silicate, fly ash, black cotton soil, water, and CaCl2 as hardener into it. • To compare the Pseudo-Binghamian grout with a pure Binghamian grout and to study the comparison of properties to formulate a cheaper option to cement grout with the advantage of properties of Newtonian grouts.

3.2 Scheme of Investigation The basic grout materials used in the investigation are the following: • Sodium Silicate (SS), Fly ash (F), Black Cotton Soil (BC), and Calcium chloride (Cl) for the Pseudo-Bingham grout. • Sodium Silicate (SS), Calcium chloride (Cl), and water (W) for the Newtonian grout (Table 1).

4 Experimental Work 4.1 Mixing of Grout The term mixing is a very important process of manufacturing the grout. Mixing of hardener with water is almost essential before sodium silicate is added for preparing grout mix to obtain the gel. It has been seen from this work that sequence of adding

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Table 1 Schedule of experiments Grout type (SS + W = 25% + 75%)

Grout type SS + F + BC + W = 25% + 4% + 6% + 65%

H = 2% √

H = 2.5% √

H = 2% √

H = 2.5% √

√

√

√

√

14 days √

14 days √

14 days √

14 days √

Needle penetration 1, 3 days test

1, 3 days

1, 3 days

1, 3 days

UCS and Triaxial 3, 7 and 28 days test (raw grout and grouted sand)

3, 7 and 28 days 3, 7 and 28 days

Test

Specific gravity pH value Syneresis Marsh cone viscosity

3, 7 and 28 days

Where, T = Curing time, B = Bentonite part by weight of cement, w/c = Water ce-ment ratio, Test performed

√ =

Table 2 Specification of Mixer Motor details

1/16 HP, 220/230 V, 50 cls. A.C. Power supply

Types of rotor

12.5 mm diameter, 55 mm long with three-bladed propeller of 65 mm sweep made up stainless steel

Maximum speed 0–6000 rpm Stirring capacity

25–30 L

first hardener then water and then sodium silicate gave gel for pure chemical grout. In case of Pseudo-Binghamian grout, sodium silicate is mixed with a small amount of water, thereafter block cotton soil and fly ash are added and the mixture is mixed thoroughly. After thorough mixing diluted solution of hardener in water is added and again mixed, this forms the grout. The mixing time of 3 min and maximum speed was kept constant throughout the experimental work. The specification and arrangement of mixer is described in Table 2.

4.2 Grout Mix Design Approaches There are three methods of grout mix design listed as follows: • Framework Approach, • Equivalent Weight Approach, and • Conceptual Framework Approach.

Ground Improvement Technique By Psuedo-Binghamian Grout …

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The latter two are used for chemical grout while the framework approach is used for particulate cement grouts. In the present investigation, conceptual framework approach is adopted.

4.3 Determination of Engineering Properties Measurement of specific gravity, fluidity, viscosity, water retentively, compressive strength, flexure strength, tensile strength, penetrability, and washout resistance/critical Hydraulic Gradient are essential for the assessment of properties of grouts.

5 Analysis and Discussion Figure 1 shows the evaluation of Specific gravity for different types of Grouts. It has been observed that as the hardener concentration increases specific gravity of grout increases. Figure 2 shows the effect of afflux time characteristics for different types of raw grouts. Afflux time varies as the hardener concentration increases. For Newtonian grout with hardener concentration 2.0 and 2.5% were 34.6 and 37.9 s, while for Pseudo-Binghamian grout 41.09 and 42.91 s. Figure 3 shows the variation in pH for different raw grouts. The pH is 11.4 and 11.2 for Pseudo-Binghamian grout, 10.8 and 11.2 for pure chemical grout. It has been observed that as the hardener concentration increases the pH increases for pure and Pseudo-Binghamian grout. Figure 4 shows the syneresis versus grout type with a variation of hardener. The percentage of syneresis is 61.35 and 56.58% for Pseudo-Binghamian grout, 55.95 and 53.49% for Newtonian. In wet cured condition, all grouts do not show syneresis.

Fig. 1 Evaluation of specific gravity for different types of raw grouts

206

Fig. 2 Afflux time characteristics for different types of raw grouts

Fig. 3 Variation in pH for different raw grouts

Fig. 4 Syneresis versus grout type with variation of hardener

N. B. Hasilkar et al.

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NPR Strength in Kg/cm2

Fig. 5 Effect of hardener concentration on NPR strength for pure chemical grout

207

0.7

1Day

3 Day

0.6 0.5 0.4 0.3 0.2 0.1 0 1.5

2.0

2.5

3.0

Hardener Concentration (%)

It has been observed from the graph that for pure chemical and Pseudo-Binghamian grouts syneresis decreases as the concentration of hardener (CaCl2 ) increases.

5.1 Needle Penetration Resistance for Grouts The early gel strength of grouts was determined using Needle Penetration test. Figures 5, 6, 7 and 8 show the effect of hardener concentration on NPR strength with a variation of hardener from 1.5% to 3.0% after 1 and 3 days for pure chemical and Pseudo-Binghamian grouts. The needle penetration resistance for Newtonian grout increases from 0.018 kg/cm2 to 0.553 Kg/cm2 at 24 h and 0.044 Kg/cm2 to 0.753 kg/cm2 at 72 h after gellification with hardener concentration with variation from 1.5% to 3.0%. For Pseudo-Binghamian grout, it is increases from 0.156 kg/cm2 to 1.79 kg/cm2 at 24 h and 0.379 kg/cm2 to 2.213 kg/cm2 at 72 h after gellification with hardener concentration vary from 1.5% to 3.0%. The graph shows that as the needle penetration resistance increases immediately after gelification. It also indicates that as the hardener concentration increases the needle penetration resistance increases, while for cement bentonite grout as bentonite increases needle penetration resistance decreases.

5.2 Time-Strength Characteristics of Raw Grouts Unconfined compression and triaxial compression tests were performed to determine the strength of hardened raw Newtonian and Pseudo-Binghamian. Figures 9 and 10 shows the peak strength versus curing time for Unconfined and Triaxial compression test after 3, 7, and 28 days for raw grouts. It has been observed that the for pure chemical grout with hardener concentration 2.0 and 2.5% peak strength

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NPR Strength in Kg/cm 2

3 2.5

1Day

3 Day

2 1.5 1 0.5 0 1.5

2.0

2.5

3.0

Hardener Concentration (%)

Fig. 6 Effect of hardener concentration on NPR strength for Pseudo-Binghamian

NPR strenght in Kg/cm2

0.8 1.5% CaCl2 2.0% CaCl2 2.5% CaCl2 3.0%CaCl2

0.7 0.6 0.5 0.4 0.3 0.2 0.1 0 1 DAY

Time in days

3 DAY

Fig. 7 Effect of hardener concentration on NPR strength for pure chemical grout

NPR strenght in Kg/cm2

3 2.5

1.5%Cacl2

2%CaCl2

2.5% CaCl2

3% CaCl2

2 1.5 1 0.5 0 1 DAY

Time in days

3 DAY

Fig. 8 Effect of hardener concentration on NPR strength for Pseudo-Binghamian grout with time

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1.8 SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2

SS-25,F-4,BC-6,H-2.5

W/C-5,B-60

W/C-5,B-40

1.600

2

UCS Strength (kg/cm )

1.6 1.4 1.2 1 0.8

0.673

0.6

0.473

0.457

0.4 0.146

0.216

0.2

0.034 0.026 0.039 0.029

0.214

0.157

0.139

0.087

0.086

0.142

0.066

0.015

0 3 day

7 day

28 day

Curing Time

Fig. 9 Unconfined compressive strength versus Curing time for Raw grouts 3.50

Triaxial Strength (Kg/cm 2)

3.00

SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2

SS-25,F-4,BC-6,H-2.5

W/C-5,B-60

W/C-5,B-40

3.06

2.50 2.00

2.00

1.50 1.20

1.00 0.63

0.51

0.50 0.11 0.140 0.10

0.17 0.33

0.13 0.10

0.15 0.128

0.19 0.12

0.161

0.25

0.00

3 da ys

7 da y s

Curing Time

Fig. 10 Peak triaxial strength versus curing time for raw grouts

2 8 da y s

210

N. B. Hasilkar et al. 12

UCS Strain (%)

10 8

9.813

10.033

SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2

SS-25,F-4,BC-6,H-2.5

W/C-5,B-60

W/C-5,B-40

7.787 7.185

6.853

6.78

6.975 6.579 6.415

6.305

6 4.770 3.947

4 2.632 2 . 6 3 2

2.385

2.721

2 .6 3 2

2.303

2 0

3 day

7 da y

2 8 da y

CuringTime

Fig. 11 Peak strains versus curing time for raw grout in UCS test

increases from 0.026 kg/cm2 to 0.015 kg/cm2 and 0.029 kg/cm2 to 0.086 kg/cm2 . For Pseudo-Binghamian grout with hardener concentration 2.0% and 2.5% peak strength increases from 0.034 to 0.142 kg/cm2 and 0.039 to 0.066 kg/cm2 . Figure 5.35 shows the peak triaxial strength versus curing time for raw grouts tested after 3, 7 and 28 days. It has been observed that the for Newtonian grout with hardener concentration 2.0 and 2.5% peak strength increases from 0.140 to 0.161 kg/cm2 and 0.166 to 0.251 kg/cm2 . For Pseudo-Binghamian grout with hardener concentration 2.0 and 2.5% peak strength increases from 0.095 to 0.122 kg/cm2 and 0.108 to 0.186 kg/cm2. Figure 11 shows the peak UCS strain versus curing time for raw grouts tested after 3, 7, and 28 days. It has been observed that for Newtonian grout with hardener concentration of 2.0 and 2.5% peak strain decreases from 9.813% to 6.579% and 7.185% to 6.415%. For Pseudo-Binghamian grout with hardener concentration of 2.0 and 2.5% peak strain increases from 10.033% to 4.77% and 6.853% to 3.947%. Figure 12 shows the peak triaxial strain versus curing time for raw grouts tested after 3, 7, and 28 days. It has been observed that for Newtonian grout with hardener concentration of 2.0% and 2.5% peak strain increases from 2.122% to 5.127% and 2.897% to 6.124%. For Pseudo-Binghamian grout with hardener concentration of 2.0% and 2.5% peak strain increases from 4.824% to 2.851% and 4.715% to 3.399%. Figure 13 shows the modulus of elasticity versus curing time for raw grouts tested after 3, 7, and 28 days. It has been observed that for Newtonian grout with hardener concentration 2.0% and 2.5% modulus of elasticity increases from 0.71 kg/cm2 to 2.45 kg/cm2 and 0.79 kg/cm2 to 2.54 kg/cm2 . For Pseudo-Binghamian grout with hardener concentration, 2.0 and 2.5% modulus of elasticity increases from 0.834 kg/cm2 to 5.0 kg/cm2 and 1.604 kg/cm2 to 4.162 kg/cm2 .

Ground Improvement Technique By Psuedo-Binghamian Grout … SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2.5

W/C-5,B-60

8 7.347

211

SS-25,F-4,BC-6,H-2 W/C-5,B-40 7.566 7.017

Triaxial Strain (%)

7 6.124

6.031

6

5.482 4.824 4.715

5

5.127

5.044 4.934

4.934

4 3.125

2.897

3

3.399

3.287 2.851

2.122

2 1 0 3 days

7 days

28 days

Curing Time

Fig. 12 Peak strains versus curing time for raw grouts in triaxial test

90 85 80 75

84.88

SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2

SS-25,F-4,BC-6,H-2.5

W/C-5,B-60

W/C-5,B-40

2

E Value (Kg/cm )

70 65 60 55 50 45 40

35.06

35

29.37

30.38

30 25 17.39

20 9.65

15 10 5

0.71

0.83 0.79

1.61

2.69

0.86 1.461.86

5.00 4.16

2.45 2.54

0 3 days

7 days

28 days

Curing Time

Fig. 13 Modulus of elasticity versus curing time for raw grouts in UCS test

Figure 14 shows the modulus of elasticity versus curing time for raw grouts tested after 3, 7, and 28 days in triaxial test. It has been observed that for Newtonian grout with hardener concentration 2.0% and 2.5% modulus of elasticity increases from 1.98 kg/cm2 to 6.21 kg/cm2 and 3.12 kg/cm2 to 8.67 kg/cm2 . For PseudoBinghamian grout with hardener concentration, 2.0 and 2.5% modulus of elasticity increases from 3.841 kg/cm2 to 9.931 kg/cm2 and 4.296 kg/cm2 to 20.267 kg/cm2 .

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N. B. Hasilkar et al. 80

SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2

SS-25,F-4,BC-6,H-2.5

W/C-5,B-60

W/C-5,B-40

70

66.78

2

E Value (Kg/cm )

60 50 40

34.55 29.75

30

20.27

20

13.80

12.37 8.43

10

3.12 1.98

10.93 4.93

4.30 3.84

3.47

5.63

6.21

8.67

9.93

0 3 days

7 days

28 days

Curing Time

Fig. 14 Modulus of elasticity versus curing time for raw grouts in triaxial test

Figure 15 shows the cohesion and angle of internal friction versus curing time for raw grouts tested after 3, 7, and 28 days in UCS test. It has been observed that for the Newtonian grout with hardener concentration of 2.0% and 2.5%, cohesion is 0.032 kg/cm2 to 0.022 kg/cm2 and 0.034 kg/cm2 to 0.012 kg/cm2 . For PseudoBinghamian grout with hardener concentration 2.0% and 2.5% cohesion increases from 0.037 kg/cm2 to 0.026 kg/cm2 and 0.036 kg/cm2 to 0.010 kg/cm2 . Cohesion increases as the hardener concentration increases for PseudoBinghamian grout, while for Binghamian grout cohesion decreases as curing time increases. In addition to this, the angle of internal friction increases with the curing time.

6 Conclusion Pseudo-Binghamian grout consists of sodium silicate, black cotton soil, fly ash water, and hardener, i.e., CaCl2 and is compared with Newtonian grout consisting of sodium silicate, water, and CaCl2 as hardener and then compared. • Specific gravity for Newtonian grout and Pseudo-Binghamian grout increase with the increase of hardener concentration. • Time of afflux for the fluidity measurement by Marsh cone test increases with increase of hardener concentration for both Newtonian and Pseudo-Binghamian grouts.

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1

213

SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2.5

W/C-5,B-40

SS-25,F-4,BC-6,H-2 W/C-5,B-60

SS-25,H-2

SS-25,H-2.5

SS-25,F-4,BC-6,H-2

SS-25,F-4,BC-6,H-2.5

W/C-5,B-40

W/C-5,B-60 0.931

0.9

16 14

0.8 12 10

0.6 0.5

8

0.432

0.4

6

0.343

0.3 0.204

0.2 0.1

Ø (Degrees)

C (Kg/cm2)

0.7

4

0.152

0.034 0.036 0.037 0.032

0.039

0.063

0.058 0.045 0.058

0.022

0.026 0.01

2

0.012

0

0

3 days

7 days

28 days

Curing Time

Fig. 15 Cohesion and angle of internal friction versus curing time for raw grouts

• The pH value of Pseudo-Binghamian and Newtonian grout increases with hardener concentration and for Binghamian grout it increases with bentonite percentage and it is more than 7. • The percentage syneresis increases with the hardener concentration for both Newtonian and Pseudo-Binghamian grout. • Needle penetration resistance for Newtonian and Pseudo-Binghamian grout increases with the increase of hardener concentration after gelification. • The unconfined compressive strength of raw grouts (Newtonian and PseudoBinghamian) increases with curing time and hardener concentration. But it has been observed that UCS strength for Pseudo-Binghamian grout is slightly more than as compared to pure chemical grout. • The unconfined compressive strength and strains of grouted dry sand is more than grouted saturated sand with Newtonian and Pseudo-Binghamian grout under wet cure condition for curing time of 3, 7, and 28 days. • The triaxial strength of grouted sand is more than the raw grouts after wet curing for a period of 3, 7, and 28 days. • The value of cohesion for raw Newtonian and Pseudo-Binghamian grouts decreases with the increase of curing period. It has been also observed that the cohesion reduces with increases of bentonite proportion in the grout mix. The value of the angle of internal friction for raw Binghamian grout with 40% bentonite

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increases and then decreases as the curing period increases. But for Newtonian and Pseudo-Binghamian, grouts decreases as the curing period increases. • The value of cohesion for dry grouted sand is more than the saturated grouted sand and increases with an increase in the curing period. The value of angle of internal friction for grouted dry sand increases as the curing period increases and also it is more than the grouted saturated sand.

References 1. Singh A (1983) Behaviour of silicate stabilized dune sand. In: Proceeding on Indian Geotechnical conference, Madras, pp 7–12 2. Caron C (1967) Synthetic rasins, Laboratory tests and application of grouting, Rilem collageue, pp 183–186 3. Cambefort H, Caron C (1957) The erosion of sodium silicate gels. In: Proceeding of 4th international conference on SMFE. Butter worths, London, pp 324–337

Ground Improvement for Liquefaction Mitigation of Sand Deposits in Southern Dubai B. Soundara and S. Bhuvaneshwari

Abstract The present study involves the case study of the geotechnical investigation carried out in the commercial area of Jebel Ali region in southern Dubai, United Arab Emirates. The place is a coastal city and a commercial hub with many infrastructural developments. As a part of an expansion project, it was proposed to construct a double-span steel portal building, a steel drum factory, and 100 m high tower, with a superimposed load of 35 kN/m2 and column loads varying from 200 to 700 kN. Nearly 14 boreholes were drilled with varying depths from 10 to 50 m maximum and 24 static cone penetration tests with pore water pressure measurements were performed to ascertain the sub-surface profile. The undisturbed and disturbed soils were also collected for further laboratory analysis. The soil investigation program revealed fine silt as the upper soil layers and medium dense silty sand and calcareous sandstone prevailing at larger depths. Shallow foundations were recommended for the proposed structures. Isolated strip footings for the auxiliary structures, raft foundations for the tall tower, and drilled pile foundations for every main structure were recommended. However, the upper fill soil is loose and highly susceptible to liquefaction effects, which include, loss of bearing strength, surface settlement, negative skin friction on piles, and uplift pressures on the lightweight structures. The region is also grouped under zone 2A with the seismic zone factor (Z) of 0.15. The phenomenon of liquefaction can cause large total and differential settlements, hence ground replacement with the stone column is recommended. Stone columns of diameter 500 mm and length 3 m were adopted to densify the upper soil fill. A total of 1621 stone columns were installed in a rectangular grid pattern with 1.50 m c/c. The overall increase in the bearing capacity and the improvement is verified through a number of cone penetration tests, plate load tests, and zone load tests. Keywords Liquefaction · Stone column · Cone penetration · Bearing strength B. Soundara (B) Bannari Amman Institute of Technology, Sathyamangalam 638401, India e-mail: [email protected] S. Bhuvaneshwari SRM Institute of Science and Technology, Kattankulathur 603203, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_20

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1 Introduction 1.1 Liquefaction Phenomenon The majority of the failures of structures like embankment, dams, slopes, etc., are triggered by the liquefaction phenomenon during the event of an earthquake. According to Castro and Poulos [1], “the liquefaction phenomenon could be defined as the phenomenon, wherein the saturated sand loses its shear strength due to monotonic/ cyclic loading and flows like a liquid. The buildings on the liquefied soil float until the shear stresses are countered by the reduced shear resistance of the soil”. During an earthquake, under the dynamic loading conditions, loose saturated sand deposits, compress and the stresses are transferred to the pore water. The high pore water pressure generated, is unlikely to drain, resulting in the reduction of the effective confining pressure. This leads to loss of strength, stiffness, and excessive deformation of the soil deposits leading to sinking and floating of lighter structures [2]. The soil softens allowing large cyclic deformations. This leads to ground oscillation and lateral spread as a consequence of either flow deformation or cyclic mobility [3]. Liquefaction poses a great threat to the failure of earth structures such as the earthen dams, embankments, pavements, etc. [4, 5]. Another important aspect of liquefaction is that it is a progressive phenomenon, as the liquefaction of the top layers in the first few cycles of earthquake motion reduces the overburden of the lower layers further initiating the liquefaction of the deep layers [2]. The phenomenon of liquefaction gained worldwide momentum after the most famous Niigata earthquake, 1964. Extreme damages were observed in highways, railways, and airports. The liquefaction of the loose saturated sand along the rivers of Niigata city was the cause of these major damages incurred [6]. The other major earthquakes, San Fernando (1971), Loma Prieta (1989), Kobe earthquake (1995) are some of the major earthquakes which contributed to the significant observation and information toward liquefaction studies [2]. Indian subcontinent also witnessed liquefaction damages caused by earthquake events. During 1988, earthquake-triggered liquefaction caused severe damages to embankments, bridges, and foundation structures in the Bihar–Nepal Border region.

1.2 Factors Affecting Liquefaction The soil liquefaction phenomenon depends on a multitude of factors ranging from the earthquake features such as the intensity, magnitude, and duration of the ground motion and distance of the earthquake to the soil properties such as the soil type, depth of soil layer, grain size, fines, plasticity content, etc., [1, 7–10]. The liquefaction induced failures are also a function of the thickness of the liquefiable soils and non-liquefiable soil layers [11]. The sinking of the ground, flow behavior, excessive settlements of the structures, which are associated with the liquefaction phenomenon

Ground Improvement for Liquefaction Mitigation of Sand Deposits …

217

could be potentially mitigated through remediation, and appropriate ground improvement techniques. However, the liquefaction mitigation methods require appropriate liquefaction potential assessment of the soil layers. The evaluation of liquefaction potential could be carried out by laboratory tests and in situ tests. The in situ test methods of liquefaction potential assessment involve reliable samples with smaller disturbances which could represent the liquefaction potential of the soil more accurately [5]. A number of studies have provided practical methods on the liquefaction assessment using field tests such as cone penetration test (CPT), standard penetration test (SPT), shear wave velocity (Vs), Backer penetration test (BPT) [3]. The field methods such as the SPT cannot provide continuous parameters of the blow count of the SPT as a measurement made at intervals. Nevertheless, the evaluation using CPT provides a more continuous profile of the penetration resistance and reliable measurements [3, 12]. The liquefaction of a particular site depends on the factor of safety against liquefaction. This is obtained as the ratio of the capacity of the soil to resist liquefaction–– CRR (cyclic resistance ratio) to that of the seismic demand imposed on it, calculated as the CSR (cyclic stress ratio). The CRR/CSR could be obtained from the in situ test data such SPT, CPT, and other field tests, earthquake parameters such as the peak ground acceleration (PGA), moment magnitude (Mw), and soil properties related to the particle size distribution and fines content [10, 13].

1.3 Mitigation Methods The excessive damages caused by the liquefaction phenomenon in loose sand deposits are attributed to the loss of strength and stiffness of the soil, associated with excess pore water pressure generation due to the cyclic loading. Hence most of the mitigation and remediation methods are focused on soil densification and pore water pressure dissipation. Some of the usual methods adopted are soil densification through vibration techniques, blasting, soil reinforcement techniques, and compaction grouting measures [14]. The method of soil densification through granular columns or the stone columns is a more sought-after technique as it lowers the potential for excessive pore water pressure due to the dilatancy effect. The stone columns increase the soil density during installation and offer higher effective lateral confining stress. The inclusions also provide an additional drainage path for excess pore water pressure mitigation [15–17]. The present study involves the discussion on the field improvement technique adopted for the densification of the loose sand layers as a measure to mitigate liquefaction. The in situ strength is evaluated through CPT carried out upon post/ pretreatment. The paper deliberates on the granular columns inclusion to densify soil, adopted at the Jebel Ali area of Dubai for probable expansion of the infrastructure.

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2 Case Study for Granular Inclusions for Liquefaction Mitigation 2.1 Ducab Jebel Ali Expansion Project, Jebel Ali, Dubai: Vibro Stone Columns Around Bored Piles (2009) The expansion project at the Jebel Ali Area, Dubai consisted of 5–6 major building constructions. The project includes the construction of a special cable plant building, HV/EHV plant building, steel drum factory, 100 m high tower, warehouse building, and five single-story accommodations. The column loads varied from 200 to 6000 kN. The site situated in Dubai was characterized by a hot and arid climate, strong drying winds, and high solar radiation. For the subsoil exploration, 14 boreholes were drilled at the site for depths ranging from 10 to 50 m. Standard penetration tests were carried out. Based on the bore log reports, the following observations were made. 1. The Water table lies at a shallow depth (0.5–1 m) below EGL in all locations. 2. The borehole profile shows that there are general similarities and continuities of the sub-surface materials, in spite of some local variations. Very loose to dense sand/sandy silt with gypsum crystals were found for a thickness of 2.3–5.2 m and very week sandstone/calcareous siltstone were present in the lower strata (Table1).

2.2 Foundation Details For the proposed structures, it was recommended to adopted shallow foundations, raft foundations and for heavier columned structures, pile foundation was recommended. However, it was proposed after the required densification or excavation up to the top 2–3 m. The presence of loose upper soil is susceptible to high settlement and liquefaction, which includes loss of bearing strength, surface settlement, lateral spreading, and negative friction on piles. This could also cause huge damages to water pipes, electrical installations, and pavement structures. Based on the SPT data, it was understood that the upper soil would be susceptible to liquefaction under cyclic loading during earthquakes. Ground densification is recommended for the site which was verified by the CPTU tests to confirm that the engineering properties meet the anticipated design requirements. According to the seismic history from previous records available and instrumental data, the local authorities, have grouped Dubai under zone 2A with a seismic factor of 0.15. In order to avoid ground liquefaction of the topsoil layers in seismic conditions, ground improvement became necessary. Since the upper layers of subsoil predominantly consist of very loose sand/sandy silts, the vibro replacement method (vibro stone columns) was recommended. The major advantages of vibro stone columns being the improved shear resistance, increased load-bearing capacity and soil density, high permeability of the treated soil. In case of an earthquake event with excess pore

2.22

5.50

6.02

5.58

5.55

4.63

6.87

7.12

0.70

-0.86

1.38

1.80

1.39

1.45

1.58

2.54

1.97

2.62

2.14

−1.00

−5.02

−5.22

−4.85

−5.49

−5.49

−5.32

5.05

4.41

4.63

4.58

4.74

4.73

4.85

4.84

4.82

4.99

4.36

4.50

1.00

0.36

0.70

−0.86

1.38

1.80

2.37

2.85

2.3

3.27

3.28

3.35

2.78

3.25

5.27

4.35

4.32

0.36

4.68

3.88

1.00

4.88

Thickness (m)

To

From

Elevation DMD

6

5

4

3

2

1

14

13

12

11

10

9

8

7

6

5

4

3

2

1

BH No

Summary of Tests Results

Light brown, pinkish brown, reddish brown, offwhite, mottled greenish gray, calcareous SANDSTONE / CALCARENITE / CONGLOMERATE / CALCIRUDITE / calcareous SILSTONE/CALCISILTITE, embedded with shell fragments, gravel and voids, partially to distinctly weathered, fractures very close to widely spaced and interbedded with cemented calcareous sand/gravel/silt

RQD (%) 0 - 95 UCS (MPa) 0.08–6.32

Brown, light brown, pinkish brown, reddish SPT’N’:1–50 brown, slightly silty very silty fine/medium to fine / medium to coarse SAND / sandy SILT with gypsum crystals and occasional to many gravel size sandstone fragments

Geologic Description

Table 1 Borelog for the soil at Jebel Ali site, Dubai. (with permission from ACES)

(continued)

Very week to moderat-ely week

Very loose to very dense

Material Class

Ground Improvement for Liquefaction Mitigation of Sand Deposits … 219

6.62

21.9

47.6

22.15

7.53

7.68

−5.17

−20.32

−45.06

−20.18

−4.91

−5.54

−5.60

: Rock Quality Designation

: Standard Penetration Test

1.45

1.58

2.54

1.97

2.62

2.14

−1.00

RQD

SPT

4.60

6.55

−5.16

1.39

Thickness (m)

To

From

Elevation DMD

Table 1 (continued)

14

13

12

11

10

9

8

7

BH No

EOB

UCS

Geologic Description

: End of Boring

: Unconfined Compressive Strength

Summary of Tests Results

Material Class

220 B. Soundara and S. Bhuvaneshwari

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Fig. 1 Typical arrangement of the vibro stone columns

water pressure generation, it could be assumed that a significantly faster reduction of the water pressure would be possible with granular column inclusions. Based on the site investigations, the diameter of the vibro stone columns was set to 0.50 m, the spacing to 1.5 m, and the depth to 3 m or up to refusal. Pretreatment cone penetration tests (CPTs), post-treatment CPTs, plate load tests, and large-scale load tests were executed in order to confirm the design requirements. A total of 1621 stone columns were installed in a rectangular grid pattern. The typical stone column grouping and load test markings were shown in Fig. 1.

2.3 Cone Penetration Test Twenty-four (24) static cone penetration tests were carried out at site up to refusal, at the specified locations. A Self-Anchoring Penetrometer (Self Propelled) 200 kN Capacity—TG73-200, manufactured by Pagani (Italy) was used. An Electric Piezocone with a projected area of 10 sq. cm., the apex angle of 60°, and friction sleeve area of 150 sq. cm complying with all requirements of ISSMFE – International Reference Test Procedure. The piezocone is equipped with a pore water pressure measuring facility and manufactured by Van Den Berg (the Netherlands). The test was carried out by penetrometer tip with electric sensors permitting simultaneous data acquisition and graphical presentation of Tip resistance, qc (MPa), sleeve friction, fs (kPa), pore water pressure, U2 (kPa). The piezocone was continuously pushed into the ground at the standard rate of 20 mm/s. A Standard saturation procedure was used of the cone for pore water pressure measurement. Tilt values were also continuously monitored during the test. The obtained test results were presented in two

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standard formats. The graphical presentations show tip resistance, sleeve friction, friction ratio, pore water pressure versus depth. In Fig. 2, the cone penetration resistance of the unimproved soil and the calculated penetration resistance required for the safety against ground liquefaction are presented both with (Post CPT) and also without stone columns (Pre CPT). As shown, the risk of ground liquefaction in the depth between 1.0 m and 2.0 m can be avoided by the installation of stone columns. The Load test was done to monitor the load versus settlement performance of the surface below the footing during the load test of up to 150% of the design bearing pressure. The large-scale plate load test was performed using a 2.05 m × 2.05 m firm footing. A sufficient working area was allowed to place the test plate and test loads. The reaction of the test was provided by means of a steel kentledge system using concrete blocks as the counterweight. Four dial gauges were placed on reference beams on the four corners of the footing for monitoring the settlement. Figure 3 shows the results of the zone load test. The specified allowable load was 66 tons and the maximum load applied was 90 tons. For a maximum applied load of 99 ton, the observed maximum settlement was 11.46 mm.

3 Summary and Conclusions The Jebel Ali site of Dubai was identified as the liquefaction susceptible zone based on the in situ tests using SPT. The top 2–4 m soil was of loose-fill which required densification or removal depending on the foundation recommendation. However, the shallow foundation was recommended for the site after the densification of the top layers. Granular column inclusions were adopted as a densification method. Nevertheless, this method also increases the strength and stability of the soil and also acts as a drain for the dissipation of the excess pore water pressure generated during the event of earthquakes. Around 1600 stone columns were installed in the entire site to a depth of 3 m, to improve the entire stretch of the loose soil profile. The diameter of 500 mm stone columns was adopted. The improvement in the strength of the soil layer was assessed through CPT carried out in the aftermath of the stone column installation. There was a considerable increase in the tip resistance in the range of 25 MPa to 30 MPa in the 3–4 m depth of soil. These were further assessed by the plate load tests and the load settlement relationship is assessed. Thus, the present case study delineates the effectiveness of the stone column technique for the effective mitigation of liquefaction in seismic zones.

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Fig. 2 a and b Typical Cone penetration test including required tip resistance with and without vibro stone columns

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Pressure Selement Graph 125 Load (Tons)

100 75 50 25 0 0

2

4

6 Selement (mm)

8

10

12

Time-Selement Graph

120 100 80 60 40 20 0

Selement (mm)

Load (Tons)

Time-Load Graph

0

5 10 Time (hrs)

15

0 2 4 6 8 10 12 14

0

Time (hrs) 5 10

15

Fig. 3 Load test Results

References 1. Castro G, Poulos JS (1977) Factors affecting liquefaction and cyclic mobility. J Geotech Eng Div 103.GT6 2. Idriss IM, Boulanger RW (2008) Soil liquefaction during earthquakes. Earthquake engineering research institute 3. Youd TL, Idriss IM, Andrus RD, Arango I, Castro G, Christian JT, Dobry R, Finn WDL, Harder LF Jr, Hynes ME, Ishihara K, Koester JP, Liao SSC, Marcuson WF III, Martin GR, Mitchell JK, Moriwaki Y, Power MS, Robertson PK, Seed RB, Stokoe KH II (2001) Liquefaction resistance of soils summary report from 1996 NCEERand 1998 NCEER/NSF workshops on Evaluation of Liquefaction Resistance of Soil. J Geotech Geoenviron Eng 127:817–833 4. Lee KL, Fitton JA (1969) Factors affecting the cuclic loading strength of soil. In: Vibration effects on soils and Foundations, Special Technical Publication 450, Americal Society for testing and Materials, Philadelphia, Pa 5. Huang Y, Yu M (2017) Hazard analysis of seismic soil liquefaction. Springer Natural Hazards 6. Iwasaki T (1986) Soil liquefaction studies in Japan: state-of-the-art. Soil Dyn Earthq Eng 5:1 7. Youd TL, Perkins DM (1978) Mapping of liquefaction induced ground failure potential. J Geotech Eng Div ASCE 104(4):433–446 8. Kramer SL (1996) Geotechnical earthquake engineering. Prentice Hall, New Jersey

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9. Tuttle M, Chester J, Lafferty R, Dyer-Williams K, Cande B (1999) Paleo seismology study northwest of the New Madrid Seismic Zone, US Nuclear Regulatory Commission, NUREG/CR-5730,Washington, DC 10. Dixit J, Dewaikar DM, Jangid RS (2012) Assessmnet of liquefaction potential index for Mumbai City. Nat Hazards Earth Syst Sci 12:2759–2768 11. Ishihara K, Yamazaki A, Haga, K (1985) Liquefaction of Ko consolidated sand under cyclic rotation of principal stress direction with lateral constraint. Soils Found Jpn Soc Soil Mech Found Eng 5(4):63–74 12. Liu CN, Chen CH (2006) Mapping liquefaction potential considering spatial correlations of CPT measurements. J Geotech Geoenviron 132(9):1178–1187 13. Seed HB, Idriss IM (1971) Simplified procedure forvaluating soil liquefaction potential. J Soil Mech Found Div ASCE 97(SM9):1249–273 14. Mitchell JK (2008) Mitigation of liquefaction potential of silty sands. In: Laier JE, Crapps DK, Hussein MH (eds) From research to practice in geotechnical engineering, geotechnical special publication, vol 180, ASCE, Reston, VA, pp 453–451 15. Baez JI (1995) A design model for the reduction of soil liquefaction by vibro-stone columns. PhD dissertation, Univ. of Southern California, Los Angeles 16. Adalier K, Elgamal A (2004) Mitigation of liquefaction and associated ground deformations by stone columns. Eng Geol 72(3–4):275–291 17. Green RA, Olgun CG, Wissmann KJ (2008) Shear stress redistribution as a mechanism to mitigate the risk of liquefaction. In: Zeng D, Manzari MT, Hiltunen DR (eds) Geotechnical earthquake engineering and soil dynamics IV. ASCE, Reston, VA, pp 1–10

Shake Table Studies to Assess the Effect of Reinforced Backfill Parameters on Dynamic Response of MSE Walls Tirtha Sathi Bandyopadhyay, Pradipta Chakrabortty, and Amarnath Hegde

Abstract The mechanically stabilized earth retaining wall offers several benefits in various areas of geotechnical engineering. Due to the interaction between backfill soil and reinforcement, the MSE wall behaves like a coherent structure. Thus, suitable to sustain substantial loading. The purpose of this study is to analyze the seismic behavior of geogrid reinforced soil retaining walls under three different input motions. The influence of relative densities of fine backfill sand and the number of reinforcing layers was studied using shaking table tests.1-D shake table tests were performed on small scale retaining wall models with two and three layers of geogrid. Six models were constructed and instrumented with LVDTs and accelerometers to study the performance of the wall and backfill. Base excitations with three different peak ground acceleration (PGA) values were provided to all these models. The lateral displacement and acceleration amplification factors (AAF) were assessed at different elevations for every model. It was witnessed that the acceleration amplification factor (AAF) was more at the wall panels as compared to sand backfill for the same elevation. The AAF was higher in dense soil backfill as compared to the loose state. The experimental results also indicated that, with the increase in the relative density of backfill, the horizontal face displacement decrease. Hence, it can be concluded that retaining walls with fine sand backfill in the dense condition performs better in terms of displacement under dynamic loads. Keywords Mechanically stabilized earth (MSE) wall · Shake table test · Geogrid · Peak ground acceleration · Acceleration amplification factor

1 Introduction A mechanically stabilized earth (MSE) wall is a composite structure consisting of alternating layers of compacted backfill and soil reinforcement elements, fixed to a wall facing. The failure of retaining walls in the seismic prone zones due to severe T. S. Bandyopadhyay (B) · P. Chakrabortty · A. Hegde Department of Civil and Environmental Engineering, IIT Patna, 801106 Patna, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_21

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earthquakes cause devastating loss to life and property. Past records show that some dreadful earthquakes have occurred with ground accelerations reaching up to 0.7 g. However, effective use of the geosynthetics reinforced earth walls can be useful in resisting these strong motions to a greater extent. Analyzing the dynamic behavior of retaining structures under seismic ground motions helps in a better understanding of their performance during an earthquake and to design these structures efficiently [10]. Mechanically stabilized earth walls have become a good alternative to the traditional retaining walls. These walls consist of compacted soil backfill, layered with extensible or inextensible reinforcements fixed to wall panels. The soil backfill provides friction and the reinforcements provide tensile strength which combinedly provides stability to the structure. These structures also provide improved seismic performance and aesthetic benefits over conventional retaining walls [5, 8, 14, 15]. Soil reinforcements can be steel strips or geosynthetics, having unique characteristics for the tensile strength and pullout capacity, corrosion, and durability. Studies conducted on small scale model tests of geosynthetic reinforced soil structures with different types of reinforcements revealed the decrease in the lateral displacement with the increase in the reinforcing length [3, 7, 9]. Facing walls basically avoids the erosion of soil and tolerate some differential movement [1, 6]. Several models with discrete and continuous facing with changing the slope of facing wall have been studied [4]. Wall with discrete facings had less deformation than continuous facing [13]. Lateral wall displacement and reinforcement loadings depend on the number of reinforcing layers and some other factors such as stiffness and length of the reinforcements [2]. The relative density of the backfill also plays an important role in the overall designing and the stability of the retaining walls [11, 12]. It was observed from the literature that the combined effects of relative density with the simultaneous change in number of reinforcement layers in the dynamic response of the MSE wall not been studied. The purpose of the present study is to study the effect of different layers of geogrid reinforcement and the density of the backfill soil on the dynamic response of the MSE wall. A series of six numbers of 540 mm high models with fine sand as backfill, reinforced with geogrids and rigid panels as wall panels were tested on a shake table designed for horizontal 1-D shaking. The tests were conducted by applying sinusoidal motions. The effect of backfill densities, simultaneously varying the number of geogrid layers was investigated for three different peak ground accelerations (PGA), viz., low (0.090 g), medium (0.234 g), and high (0.594 g).

2 Shake Table Tests 2.1 Experimental Setup MSE wall models with the dimensions 655 mm long × 580 mm wide × 540 mm tall were constructed within a rigid wooden box. The box was having dimensions of

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800 mm long × 580 mm wide × 770 mm tall that was fixed to the shaking table. The wooden box was open at the front side. Two sets of rigid panels with thickness of 25 mm were used as facing wall. The size of the base of the box was slightly smaller than that of the shaking table to ensure complete contact between them. The table was connected to a motor having a power of 2 HP and 1200 RPM to provide stepwise increased horizontal sinusoidal motions by changing its frequency. Figure 1a-d shows a schematic view of the model set up. The sand raining technique was adopted to pour the backfill sand into the box, maintaining specific heights to achieve the required relative densities using a hopper.

2.2 Materials Used Soil Fine sand was used as backfill for all the MSE wall models with different relative densities. The specific gravity of the sand was 2.69. Other physical properties are tabulated in Table 1. Reinforcement Biaxial geogrids with a square aperture of 35 mm × 35 mm were used as reinforcing elements. Geogrids were made up of firm polyester coated with the bituminous solvent providing high grabbing and friction resistance. The length of the geogrid layers was maintained 400 mm, thus the reinforcement to wall height ratio (L/H) was 0.74. The L/H ratio was chosen as per the guidelines of FHWA [5], which recommends a minimum value of 0.7 H. Nominal stiffness (EA) and tensile strength of the geogrid reinforcement were determined as 90 kN/m and 13 MPa, respectively.

2.3 Preparation of Test Bed and Instrumentation A series of six model tests were conducted to check the influence of the number of reinforcements and relative density of backfill soil on the dynamic response of the geogrid reinforced soil walls. Table 2 shows the details for reinforcement parameters and backfill soil used in the tests. The sand was poured into the box using the sand raining technique maintaining a certain height to achieve the required relative density. Three target relative densities of 22%, 45%, and 70% representing loose, medium, and dense states, respectively were considered. In these model walls, two and three layers of geogrid were used with a vertical spacing of 270 mm and 190 mm, respectively. Data acquisition system was used to capture the response of the MSE wall after the base excitations. The models were instrumented with accelerometers and LVDTs at different positions of the facing wall and soil backfill. Three LVDTs were placed to measure the wall displacement at the bottom, middle, and top of the facing panels. Accelerometers were placed at different elevations of facing as well as

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Fig. 1 Schematic representations of the shake table showing the layout of the MSE wall models a the seismic box mounted on the shake table, b two geogrids model (side view), c three geogrids model (side view), d top view (for both type of models cut from the section of geogrid) (All dimensions are in mm) Table 1 Backfill soil physical properties

Properties

Values

Particle size

0.075 mm–0.425 mm

emax

0.849

emin

0.59

 (at relative density = 22%)

29°

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Table 2 Test model reinforcement and backfill parameters Model no.

Backfill property

Relative density (%)

Reinforcement No. of geogrid layers

Spacing (mm)

M1

Loose state

22

2

270

M2

Medium state

45

2

270

M3

Dense state

70

2

270

M4

Loose state

22

3

190

M5

Medium state

45

3

190

M6

Dense state

70

3

190

at the soil backfill to measure the acceleration response of the wall and the backfill. For acquiring data from the accelerometers, Bruel and Kjaer set-up was used with PULSE software. For data processing, and filtering the PULSE reflex was used. Three different sinusoidal base input motions, namely, Low PGA, medium PGA, and high PGA varying from lowest to highest were applied to the model. Intensities of accelerations changing from 0.09 to 0.594 g were applied to the model by changing the frequency of the motor attached to the shake table. The input acceleration time histories are shown in Fig. 2. 5.0 2.5

(a) PGA = 0.594g

2

Input acceleration (m/s )

0.0 -2.5 -5.0 2 1

(b) PGA = 0.234g

0 -1 -2 1.2 0.8 0.4 0.0 -0.4 -0.8 -1.2

(c) PGA = 0.090g

0

1

2

3

4

5

6

7

8

9

10

Time (sec)

Fig. 2 Recorded input acceleration time histories a High PGA, b Medium PGA, c Low PGA

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3 Test Results 3.1 Acceleration Response The effect of base excitation on the retaining structure was presented in the form of Acceleration Amplification Factors (AAF) at different positions of the facing wall and backfill material. The acceleration amplification factor is defined as the ratio of the response acceleration value at any definite point to the corresponding peak input acceleration value. Three accelerometers were placed on the backfill soil reinforced with geogrids, one at the base, at the middle, and at the top to measure the acceleration variations along the elevations of the backfill. Similarly, three accelerometers were placed one each at the bottom, at the middle, and at the top of the facing panels. AAF variations with the height of the facing wall were compared for three different backfill densities and two different layering configurations of reinforcement under three input motions. AAF depends on the backfill properties and frequency of the input motions, hence it varies for different PGA. Figure 3 shows the vertical distributions of acceleration amplification factors for three different motions (i.e., high PGA, medium PGA, and low PGA) at the wall. The results indicated that horizontal accelerations were amplified while transmitted through the facing panels. Nonlinear curve was obtained for all the cases. With the increase in density of the backfill, AAF was also found to increase. At the top of the facing, AAF increased by over 61% with an increase in the density from loose to dense for two layers of geogrid at high PGA. For three layers of geogrids, AAF increased by 35.1% for the same PGA at the top of the wall. The results shown in Fig. 4 gives a comparison of the acceleration amplification factors with an increase in the elevation in soil backfill for all the three input motions. The variation is found to be non-linear and increasing with the height of the backfill soil. From the model study, it was found that AAF in backfill soil also varied in nonlinear pattern with the height of the wall. AAF was maximum at the top. Due to more rigidity in the facing panels, the magnitudes of AAF was found to be greater than the backfill at the same elevation. AAF was found to be 85% greater at a dense state as compared to the loose state for two layers of geogrids at high PGA. For three layers of geogrid, AAF was 54.6% more for dense condition than that of loose state for the same input motion. Similar results and trends were observed for the medium and low PGA as well. Comparing AAF for change in the reinforcement layers, it was found that AAF for three layers was reduced by 17.4% at the top for high PGA as compared to two layers. For medium and low PGA, AAF was reduced by 13.2% and 6.3%, respectively, at the top of reinforced backfill sand.

Height of MSE wall (mm)

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M1 M2 M3 M4 M5 M6

600 500 400 300 200 100 0 0.5

1

1.5

2

2.5

3

AAF

M1 M2 M3 M4 M5 M6

600 500

Height of MSE wall (mm)

Height of MSE wall (mm)

(a)

400 300 200 100 0

0.5

1.5 AAF

(b)

2.5

600 500

M1 M2 M3 M4 M5 M6

400 300 200 100 0 0.5

1.5

2.5

AAF

(c)

Fig. 3 Variation of amplification of base input acceleration with height of facing panel a Low PGA, b Medium PGA, c High PGA; inset figures showing the position of the accelerometers, which is common for all the three results

3.2 Displacement Response The displacement response of the MSE wall models was compared in terms of the horizontal movement of the facing panels measured at three different elevations. LVDTs were placed at three different elevations of the facing wall to measure the displacements of the wall after the base accelerations were provided for 10 s. Horizontal displacement response of the facing walls at three different input motion values are plotted in Fig. 5. From the test results, it was noted that for all the cases, the maximum displacement was found at the top of the facing panel. The displacement for high PGA was comparatively high as compared to the medium and low PGA. With an increase in the relative density of the backfill, the friction between the soil grains increased thus strengthening the stability of the structure, causing the decrease in facing displacement. For high PGA, the maximum displacement at the top in the dense state was

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Height of backfill (mm)

234 M1 M2 M3 M4 M5 M6

600 500 400 300 200 100 0 0.5

1

1.5

2

2.5

3

3.5

AAF

M1 M2 M3 M4 M5 M6

600 500

Height of backfill (mm)

Height of backfill (mm)

(a)

400 300 200 100 0

M1 M2 M3 M4 M5 M6

600 500 400 300 200 100 0

0.5

1.5

2.5

0.5

1.5

AAF

AAF

(b)

(c)

2.5

Fig. 4 Variation of amplification of base input acceleration with height of backfill soil a Low PGA, b Medium PGA, c High PGA; inset figures showing the position of accelerometers which is common for all the three results

reduced by 72% and 19.7% as compared to loose and medium states respectively in two-layer geogrid case. For three layers of geogrids, the reduction at the same height and backfill state was about 74% and 2%, respectively. Similar observations were found for the other two input motions with the marginal reductions in displacement value. The displacements for two layers of geogrids was higher than three layers for all the three base excitations. The horizontal displacement at the top for high PGA was 80% higher for two layers as compared to the three layers of reinforcement. For medium and low PGA, the maximum displacement at the top for two layers of geogrids was 58.5% and 55.2% more than three layers. Thus, as the number of layers of reinforcement increased, which decreases the vertical spacing of the geogrids, the grabbing tendency between the reinforcement and the wall increased, hence lateral displacement was reduced. It was also observed through the model tests that the

Height of MSE wall (mm)

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M1 M2 M3 M4 M5 M6

600 500 400 300 200 100 0 0

0.2

0.4

0.6

0.8

Dispalcement (mm)

Height of MSE wall (mm)

600 500 400 M1 M2 M3 M4 M5 M6

300 200 100

Height of MSE wall (mm)

(a) 600 500 400 300

M1 M2 M3 M4 M5 M6

200 100 0

0 0

0.5

1

0

5

10

Displacement (mm)

Displacement (mm)

(b)

(c)

15

Fig. 5 Variation of facing wall displacement with the height of the wall a Low PGA, b Medium PGA, c High PGA; inset figures showing the position of LVDTs which is common for all three results

predominant mode of deformation was a combination of bulging of the facing and rotation about the wall base with minor sliding in the base.

4 Conclusions A series of shake table tests were performed to investigate the effects of change in the number of reinforcement layers and backfill density on the performance of the MSE walls. The acceleration and deflection response for three different peak ground acceleration values were studied. The following conclusions were drawn:

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• The type of deformation was found to be a combined form of bulging of the facing and rotation about the wall base with slight base sliding. • The input accelerations generated at the base were amplified along with the height of the facing wall as well as in the backfill soil. The trend followed in the graphical representation of acceleration amplification factors with elevation was non-linear. • The acceleration amplification factors for backfill soil were less than that in facing wall amplification factors. • The AAF was higher for two layers of reinforcement as compared to three numbers for almost the same input motions and at the same density and elevation. • Using fine sand in the dense state as fill material in the retaining wall produces lower deformations in the facing wall as compared to the loose state of the sand. • Increase in the number of layers of geogrid decreases the displacement of the facing wall. Maximum displacement reduced by almost 80% for three layers of geogrid as compared to two layers for high PGA at the top of the wall.

References 1. Bathurst RJ, Cai Z, Alfaro M, Pelletier MJ (1997) Seismic Design Issues for Geosynthetic Reinforced Segmental Retaining Walls. In: Wu JTH (ed) Balkema, Proceedings of the International Symposium on Mechanically Stabilized Backfill. Denver, Colorado, USA, pp 79–97 2. Bathurst RJ, Hatami K (1998) Seismic response analysis of a geosynthetic-reinforced soil retaining wall. Geosynth Int:127–166 3. El-Emam MM, Bathurst RJ (2007) Influence of reinforcement parameters on the seismic response of reduced-scale reinforced soil retaining walls. Geotext Geomembr:33–49 4. El-Emam MM, Bathurst RJ (2005) Facing contribution to seismic response of reduced-scale reinforced soil walls. Geosynth Int:215–238 5. FHWA (2009) Design and construction of mechanically stabilized earth walls and reinforced soil slopes, vol 1. U.S. Department of Transportation and Federal Highway Administration, pp 1–332 6. Huang CC, Chou LH, Tatsuoka F (2003) Seismic displacements of geosynthetic-reinforced soil modular block walls. Geosynth Int:2–23 7. Jackson P, Bowman ET (2012) Cubrinovski M.: Seismic testing of model-scale geosyntheticreinforced soil walls. Bull NZ Soc Earthq Eng:1–13 8. Komak Panah A, Yazdi M, Ghalandarzadeh A (2015) Shaking table tests on soil retaining walls reinforced by polymeric strips. Geotext Geomembr:148–161 9. Koseki J, Munaf Y, Tatsuoka F, Tateyama M, Kojima K, Sato T (1998) Shaking and tilt table tests of geosynthetic-reinforced soil and conventional-type retaining walls. Geosynth Int 5(1– 2):73–96 10. Kramer SL (1996) Geotechnical earthquake engineering. Pearson 11. Madhavi Latha G, Murali Krishna A (2008) Seismic response of reinforced soil retaining wall models: influence of backfill relative density. Geotext Geomembr:335–349 12. Madhavi Latha G, Santhanakumar P (2015) Seismic response of reduced-scale modular block and rigid faced reinforced walls through shaking table tests. Geotext Geomembr:307–316 13. Matsuo O, Tsutsumi T, Yokoyama K, Saito Y (1998) Shaking table tests and analysis of geosynthetic-reinforced soil retaining walls. Geosynth Int:97–127

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14. Ramakrishnan K, Budhu M, Britto A (1998) Laboratory seismic tests on geotextile wrap-faced and geotextile-reinforced segmental retaining walls. Geosynth Int 5(1–2):55–71 15. Tatsuoka F, Tateyama M, Uchimura T, Koseki J (1996–1997) Geosynthetic-reinforced soil retaining walls as important permanent structures. Geosynth Int:81–136

Efficacy of Consolidation Grouting in Improving Dynamic Characteristics of a Nuclear Facility Foundation Strata G. Padmanabhan, R. Mano, Sudipta Chattopadhyay, L. Davy Herbert, V. Manoharn, and BPC Rao

Abstract Consolidation grouting is often carried out to improve the permeability characteristics and deformation properties of the foundation medium. The improvement is often measured in terms of permeability and represented by the Lugeon value. In this paper, the efficacy of consolidation grouting carried out at foundation medium of a nuclear facility site was evaluated in terms of Lugeon value and surface wave velocity obtained from advanced Multi-Channel Analysis of surface wave test, which was carried out before and after the consolidation grouting to evaluate the dynamic properties of the foundation medium. This study indicated an increase of shear wave velocity and dynamic shear modulus of the foundation medium after consolidation grouting and improvement of the engineering property of foundation medium. Keywords Consolidation grouting · Lugeon value · MASW test · Shear wave velocity

1 Introduction Nuclear power plant (NPP) structures are designed and constructed in such a way that the structures are safe during the entire life of the plant. These structures are generally founded on competent rock medium which has sufficient bearing capacity to resist the forces acting on the foundation. However, after exposing the rock surface surprises like fissures, fractures, joints, weak zones are often noticed in the foundation medium which needs to be treated and improved. This will improve the deformability characteristics of the foundation medium and also reduce the possible uplift pressure acting on foundations. The study area (Fig. 1) is located along the East Coast Peninsula of India and comprises a Crystalline/Archean complex overlain by beach deposits. The site comprises sand followed by silty sand and clay. Hard rock is available at a depth of 15–20 m below the existing ground level. Geologically, the area comprises G. Padmanabhan (B) · R. Mano · S. Chattopadhyay · L. D. Herbert · V. Manoharn · B. Rao Fast Reactor Fuel Cycle Facility, Indira Gandhi Center for Atomic Research, Kalpakkam 603102, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_22

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Fig. 1 Location of the study area

of two distinct formations, namely, the charnockite rock and more recent sediments [1]. The proposed nuclear safety related structure is to be founded on moderately weathered grade rock available at a depth of 18.6 m below the existing ground levels. After reaching the required founding level, geological mapping was carried out according to the principles of [2] and weak zones in the foundation medium were identified. Subsequently, prior to the consolidation grouting, pre-permeability tests were carried as per [3] to establish the permeability of the foundation medium and consolidation grouting was taken up as specified by the standard [4]. The current practice of evaluation of the efficacy of the consolidation grouting is carried out by estimating post grout permeability of foundation medium. Often the improvements of rock mass in terms of dynamic properties after consolidation grouting were seldom studied due to the absence of incorporation of such advanced test in the investigation programme. The effect of consolidation grouting on the dynamic properties of the foundation medium is evaluated by conducting Multi-Channel analysis of the surface wave test and addressed in this study.

2 Properties of Foundation Medium The geological mapping carried out at the site indicated that the foundation strata primarily comprises of charnockite-type rock. Apart from this, the region also has some patches of granite and chlorite schist. These rocks are generally classified into fresh to moderately weathered types with visible signs of weathering with slight

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Fig. 2 Geological Map of the study area

discoloration. The rocks are in general classified into weathering grade II to III. However, the foundation strata are characterized by discontinuities of length ranging from a few centimeters to several meters. These rock discontinuities were altered and weathered and filled with finer materials. The thicknesses of the altered zone along these discontinuities vary from 1 to 20 cm. The Rock Mass Rating (RMR) of the foundation strata ranges from 48 to 77 as per [5]. The RMR in some locations is comparatively low ranging from 48 to 58. A typical geological map of the study area along with significant features is shown in Fig. 2. The geological mapping suggested that as the foundation rock mass is characterized by numerous discontinuities grouting is be required to improve the overall strength of the rocks depending upon the bearing capacity.

3 Plan and Execution of Consolidation Grouting Consolidation grouting consists of drilling of grout holes in rock and injecting cement grout under pressure, which eventually sets in the openings and voids present in the rock. The criterion for groutability of stratum is determined by pre permeability values which are generally expressed in terms of Lugeon and the holes, which absorb water greater than 1 Lugeons is, considered as groutable holes. In this project, drilling of grout holes was carried out with rotary drilling rigs. The grout holes of diameter 90 mm were drilled vertically up to a depth of 6.0 m. After cleaning of grout holes, the pre-grout percolation tests were performed as per standard [4] to determine the permeability and extent of further grout requirement. Figure 3 shows the single packer permeability test assembly used for grouting operations at site.

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Fig. 3 Single packer assembly

Subsequent to the determination of pre-grout permeability values of the area which is to be treated with consolidation grouting, neat cement grout with initial Water– Cement ratio 5:1 (by weight) is injected by pump. The water–cement ratio was gradually increased up to 1:1 in steps when pressure started to rise after continuous injection over a period of 10 min for a given Water Cement ratio. For a selected zone, grouting is carried out from the outer region to the inner region. Grouting of outer holes was carried out initially for the confinement of grout and to ensure effective subsequent treatment of inner cracks. When grout intake is constantly less than 2 l/min. averaged over a period of 10 min, grouting operation is terminated. A typical photograph of grouting operation at the site is shown in Fig. 4. Upon completion of grouting in respective areas, test holes were drilled for performing confirmatory permeability test. If the observed permeability does not

Fig. 4 Photograph of grouting operation

Efficacy of Consolidation Grouting in Improving … Fig. 5 Comparison of Lugeon Values

243

Lugeon Value 1

2.37

3 4 5

Pre Permeability Lugeon Values

2.98

2

Post Permeability Lugeon Values

2.07 3.02 4.19

meet the acceptance criteria, of Lugeon 1.5 Mg/m3 . 2. The time taken for saturation in case of the sample with higher density was more than that of others, i.e., 2 > 1.75 > 1.5 Mg/m3 .

References 1. Chen YG, Cui YJ, Tang AM, Wang Q, Ye WM (2014) A preliminary study on hydraulic resistance of bentonite/host-rock seal interface. Géotechnique 64(12):997–1002 2. He Y, Cui YJ, Ye WM, Conil N (2017) Effects of wetting-drying cycles on the air permeability of compacted Téguline clay”. Eng Geol 228:173–179 3. Kale RC, Ravi K (2019) Influence of thermal history on swell pressures of compacted bentonite. Process Saf Environ Prot 123:199–205 4. Komine H (2004) Simplified evolution on hydraulic conductivities of sand-bentonite mixture backfill. Appl Clay Sci 26:13–19 5. Lee JO, Kang IM, Cho WJ (2010) Smectite alteration and its influence on the barrier properties of smectite clay for a repository. Appl Clay Sci 47:99–104 6. Lee JO, Lim JG, Kang IM, Kwon S (2012) Swelling pressures of compacted Ca- bentonite. Eng Geol 129–130:20–26 7. Pusch (1980) Swelling pressure of highly compacted bentonite. SKB Technical Reports TR80–13 8. Rao S, Ravi K (2015) Influence of initial degree of saturation on swell pressures of compacted Barmer bentonite specimens. Ann of Nuc Energy 80:303–311 9. Yong RN (1999) Overview of modeling of clay microstructure and interactions for prediction of waste isolation barrier performance. Eng Geol 54:83–91 10. Zhu CM, Ye WM, Chen YG, Chen B, Cui YJ (2013) Influence of salt solutions on the swelling pressure and hydraulic conductivity of compacted GMZ01 bentonite. Eng Geol 166:74–80

Lateral Load Carrying Mechanism of Fibre Reinforced Concrete Pile in Sandy Soil Koushik Sukla Das, Plaban Deb, and Sujit Kumar Pal

Abstract In the past few decades, pile foundation has become a sustainable solution for the offshore and onshore structure that is usually subjected to lateral loads. In order to sustain these high amounts of lateral loads, the tensile strength and the ductility of the pile need to be improved by adding some supplementary materials. In this study, behaviour of laterally loaded fibre reinforced concrete pile in sandy soil was investigated to study the lateral load carrying mechanism of fibre reinforced concrete pile through small-scaled laboratory model tests. The model study involves a monotonic loading system, a testing tank and an instrumented model concrete pile. To measure the bending moment and shaft resistance along the pile shaft, the pile was instrumented with a number of foil strain gauges. Five different fibre to concrete ratios by volume were used to fabricate the model piles. Then the performance of the fibre concrete pile was compared with the conventional concrete pile under static lateral loading. The results of the laboratory model test reveal that the fibre reinforced with concrete pile shows more ductility and provides higher lateral load carrying capacity compared to the conventional model concrete pile. Fibre reinforced concrete pile also reduces the bending moment of the model pile and provides more structural stability as compared to the conventional concrete pile. Keywords Fibre reinforced concrete pile · Sandy soil · Lateral load carrying mechanism · Bending moment

K. S. Das (B) · P. Deb · S. K. Pal National Institute of Technology Agartala, Agartala, Tripura, India e-mail: [email protected] P. Deb e-mail: [email protected] S. K. Pal e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_29

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1 Introduction Pile foundation is a type of deep foundation that takes the whole amount of load directly or indirectly coming from the structure. The primary benefit of this pile foundation is to transfer the vertical load into the deep bearing layer. But in case of offshore and onshore structure, lateral loads also appear in the form of wave, wing, impact loading, etc. Lateral loads are nothing but the horizontal forces acting on the structure and that may cause the deflection of the foundation. In many offshore structures, the lateral loads are developed due to the impact of seawater. The lateral loads coming from earthquakes and tsunamis are the most dangerous and can be a reason for foundation failure. That is why the impact of lateral loading is an important factor in the case of offshore and onshore structure. There are several loading cases in the piles subjected to static lateral loading such as the piles near natural or man-made slopes for bridge abutment foundations or as retaining walls, piles near slopes carrying lateral loads from power towers or wind turbines, piles for stabilizing slopes, piles for the foundation of high-rise buildings near slopes, etc. Due to this lateral load, reinforced concrete piles are damaged and tensile cracks are formed [1, 2]. Previous researches illustrate that there is a necessity to check the durability of reinforced concrete pile against the tensile crack which is developed due to the high amount of lateral loads. Therefore, to sustain these high amounts of lateral loads, it is very much important to improve the durability of the concrete pile by adding some supplementary materials like fly ash, fibre, etc. It has been well known that the flexural strength, compressive strength, tensile strength and the crack resistance ability of concrete pile can be improved by adding the fibre content on concrete [3]. Ozcan et al. [4] conducted the three-point bending test on steel fibre reinforced concrete beam and conventional reinforced concrete beam and observed that the load carrying capacity of fibre reinforced concrete is almost 18% larger than conventional one. [5, 6] performed a comparison study between the flexural behaviour of conventional concrete and fibre mixed concrete. An experimental and finite element analysis of fibre reinforced concrete pile and conventional pile was conducted by [7] and concluded that the addition of fibre in concrete could improve the lateral load carrying capacity [7, 8]. Many researchers have studied the lateral load transfer curves for many cases including loading conditions, soil types, geometry conditions, pore water pressure conditions in soil and many other different conditions. But research associated to sustain the high amounts of lateral loads by adding fibres and the lateral response of fibre reinforced concrete piles are very scarce. Therefore, in this paper, the lateral response of fly ash mixed concrete pile and fibre reinforced concrete pile in sandy soil are compared with the conventional concrete pile.

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319

Fig. 1 Experimental set-up

2 Model Test Set-Up and Loading System To perform the experimental work, a testing tank made of plexiglass and steel sheet of size 90 cm × 90 cm × 65 cm is prepared. Model set-up and loading system is shown in Fig. 1. To study the lateral response of fibre reinforced concrete pile, a pile head connector is prepared which is made of iron steel as shown in the figure and a calibrated tensile load cell is connected with the pile head connector by means of a wire rope which is associated with the winch. Lateral load is applied in the pile through pulley, winch and a wire rope. To measure the lateral displacement a LVDT is connected with the pile head. LVDT and load cell are further connected with the data logger.

3 Materials and Methodology 3.1 Foundation Soil The soil used for this test is sandy soil which is collected from the river bed. To find the properties of the sandy soil in the laboratory, specific gravity test, compaction test, direct shear test, grain size analysis test are performed. The sandy soil is having specific gravity of 2.64, and 96.2% of sand proportion, 3.8% of silt proportion. The coefficient of uniformity is 1.85 and the coefficient of curvature is 1.17. The optimum moisture content corresponding to the maximum dry density of 1.67 gm /cc is 9.5%.

320 Fig. 2 Pile sampler

Fig. 3 Strain gauge attachment in a pile

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Lateral Load Carrying Mechanism of Fibre Reinforced Concrete Pile in Sandy Soil Table 1 Soil properties

321

Properties

Sandy soil

Specific gravity, G

2.66

Sand (%)

96.2

Silt (%)

3.8

Maximum dry density (MDD) gm/cc

1.67

Optimum moisture content (OMC) (%)

9.5

Angle of internal friction, (degree)

29.77

Cohesion, C (kN/m2 )

Cohesion less

Coefficient of curvature

1.17

Coefficient of uniformity

1.85

Table 2 Model test pile notation Designation

Percentage of fly ash and polypropylene fibre

Notation

Conventional plain concrete pile

0

CP

Conventional reinforced concrete pile

0

RCP

Fly ash mixed conventional plain concrete pile

20

FCP-20

30

FCP-30

Fly ash mixed conventional reinforced concrete pile

20

FRCP-20

30

FRCP-30

Polypropylene fibre mixed conventional plain concrete pile

0.5

PCP-0.5

Polypropylene fibre mixed conventional reinforced concrete pile

1

PCP-1

1.5

PCP-1.5

0.5

PRCP-0.5

1

PRCP-1

1.5

PRCP-1.5

From the direct shear test, it is observed that the sand is cohesionless and the angle of internal friction is 29.77°. All the properties of sandy soil are tabulated in Table 1.

3.2 Model Pile and Instrumentation To prepare piles of different mixes, three pile samplers made of steel pipe is prepared in an industrial shop at Agartala. The diameter of the steel pipes is 8 cm and the height is 55 cm. After preparing the samplers, the mix design for different fibre content is calculated and according to the different mixes, model piles are prepared. Model piles are then kept in the water tank for 28 days of curing. All the model piles are instrumented with five foil type strain gauges to measure the bending moment.

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Table 3 Mix proportions for the pile Pile/ cube no & detail

Water (kg)

Cement (kg)

Sand (kg)

Coarse aggregate (kg) 10 mm

Admixture (gm)

20 mm

Fly ash (kg)

Polypropylene (gm)

CP

0.816

1.77

2.56

1.62

2.43

35.4

–

–

FCP-20

0.735

1.28

2.48

1.56

2.35

31.9

0.320

–

FCP-30

0.700

1.07

2.86

1.64

2.45

30.5

0.458

–

PCP-0.5

0.8

1.68

2.50

1.61

2.42

35

–

46

PCP-1

0.8

1.68

2.50

1.61

2.42

35

–

92

PCP-1.5

0.8

1.68

2.50

1.61

2.42

35

–

138

Pile samplers and instrumented concrete pile is shown in Figs. 2, 3. In the same proportions, concrete cubes of size 15 × 15 × 15 cm3 are also prepared to check the compressive strength of the concrete mixes. The variation of different mix proportion and their respective notations are represented in Tables 2 and 3.

3.3 Preparation of Sand Bed As the height of the soil tank is 65 cm and the height of the pile is 55 cm, the soil bed need to be prepared considering the fact that some part of the pile must be over the height of the soil bed so that the pile head can be connected with the wire of the trench through a pulley. The top 5 cm of the testing tank is kept free so that the overflow of the sandy soil during the loading condition can be minimized. For that, the lower part of the tank is filled with sand and the filling is done with rainfall technique maintaining a uniform relative density of 60%. Then the Pile is placed in the middle of the tank and the tank is filled with sand using the same technique, and thus the bed preparation is completed. The trench wire is then connected to the pile head. 50 cm of the pile is under the sandy soil and 5 cm of the pile is over the sandy soil.

4 Results and Discussion 4.1 Analysis of Compressive Strength Compressive strength test is carried out in a compressive testing machine with a cube specimen of 15 cm × 15 cm × 15 cm. The loading rate was 4 t/min. For each mix, the compressive strength test is conducted and the variation of compressive strength is shown in Fig. 4. It is seen that the compressive strength has been increased

Compressive strength (MPa)

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323

40 35 30 25 20 15 10 5 0

CP

FCP-20 FCP-30 PCP-0.5 PCP-1 PCP-1.5 Type of concrete mix

Fig. 4 Variation of compressive strength for different mixes

when fly ash and polypropylene fibres are added to the concrete mixes. However, the improvement in compressive strength for polypropylene fibre mixed concrete is higher than the fly ash mixed concrete. There is almost 9% and 16% increase in the compressive strength when 20% and 30% fly ash are added with the concrete, respectively. On the other hand, compressive strength has increased up to 26% when 1.5% polypropylene fibre is added to the concrete.

4.2 Analysis of Tensile Strength The tensile strength of the concrete specimens is carried out with cylinders of 15 cm diameter and 30 cm height. The compressive load was applied to the diameter of the specimen and the tensile strength was evaluated from Eq. 1. T ensilestr ength =

P πd L

(1)

where P is the applied load, d is the diameter of the sample and L is the height of the sample. Figure 5 is showing the tensile strength of each type of concrete mixes. The tensile strength of concrete inherently increases with the addition of fly ash, as well as the polypropylene fibre, but the increase in the tensile strength in case of fibre addition is comparatively higher than that of fly ash addition. When fibres are added to the concrete mix, it acts as a reinforcing agent, thereby increasing the tensile behaviour of the concrete.

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Tensile strength (MPa)

6 5 4 3 2 1 0

CP

FCP-20

FCP-30

PCP-0.5

PCP-1

PCP-1.5

Types of concrete mix

Fig. 5 Tensile strength of different type of concrete mixes

4.3 Analysis of Lateral Response Normalized lateral load on the pile head is represented in Fig. 6. Each lateral load (Pi ) obtained from the load cell are normalized by dividing L2 dγ, where L, d and γ are the embedded length of the pile, diameter of the pile and unit weight of soil, respectively, and the pile head lateral displacement is normalized by dividing the pile diameter. PN =

Pi 2 L dγ

and y N =

yi d

(2)

From the figure, it is observed that the normalized lateral load–displacement profile is non-linear. It is also observed from the test results that polypropylene fibre and fly ash mixed reinforced/plain concrete pile can sustain more lateral load as compared to the conventional reinforced concrete pile and conventional plain concrete pile, respectively. It is also seen from the figure that the first creaking load is gradually increasing with the increase of fly ash and fibre content. This can be attributed to the fact that the increase in the fly ash and the fibre content enhances the tensile behaviour of the pile. The first creaking load of PRCP-1.5 is enhanced by about 47% as compared to the conventional reinforced concrete pile.

4.4 Moment Distribution Bending moment profile of the pile can be obtained by using the formula suggested by Deb and Pal [9]. The recorded bending moment (Mi ) is normalized by dividing L3 dγ, where L, d and γ are the embedded length of the pile, diameter of the pile and

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325

Fig. 6 a, b Normalized lateral load versus normalized lateral displacement comparison curves

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unit weight of soil, respectively, the depth is normalized by the embedded length of the pile. MN =

Zi Mi and Z = L 3 dγ L

(3)

The variation of the normalized moment (M N ) with the normalized depth (Z) is represented in Fig. 7. It is clearly seen from the figure that the addition of fly ash and polypropylene fibre have improved the creaking moment capacity of the pile subjected to lateral loading. This is due to the fact that the addition of fly ash and polypropylene fibre enhanced the flexural behaviour of RC pile. From the figure, it is also observed that the maximum creaking moment is generated at a distance of 0.3 to 0.4 times the length of the pile. The maximum creaking moment capacity for pile configuration of PRCP-1.5 is improved by about 61% as compared to the conventional reinforced concrete pile.

Fig. 7 Moment distribution curve

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327

5 Conclusion Based on the experimental results and its analysis, the following conclusions are drawn: 1. From the compressive strength test, it is found that fly ash mixed concrete and polypropylene fibre mixed concrete showed more compressive strength as compared to plain concrete, and the compressive strength have increased up to 26% when 1.5% of polypropylene fibre is added to the concrete. 2. From the tensile strength test, it is observed that the tensile strength of concrete intrinsically increases with the addition of fly ash and the polypropylene fibre to the conventional concrete. 3. Lateral load–displacement test reveals that polypropylene fibre and fly ash mixed reinforced/plain concrete pile can sustain more lateral load as compared to the conventional reinforced concrete pile and conventional plain concrete pile. 4. From the moment distribution profile of the pile, it is observed that the maximum creaking moment is generated at a distance of 0.3 to 0.4 times the length of the pile and the maximum creaking moment capacity for pile configuration of PRCP1.5 is improved by about 61% as compared to conventional reinforced concrete pile.

References 1. Matsui T, Oda K (1996) Foundation damage of structures. Soils and Foundations 36 (Special), 189–200 2. Tokimatsu K, Mizuno H, Kakurai M (1996) Building damage associated with geotechnical problems. Soils and foundations 36(Special), 219–234 3. Nataraja MC, Dhang N, Gupta AP (1999) Stress–strain curves for steel-fiber reinforced concrete under compression. Cement Concr Compos 21(5–6):383–390 4. Özcan DM, Bayraktar A, Sahin ¸ A, Haktanir T, Türker T (2009) Experimental and finite element analysis on the steel fiber-reinforced concrete (SFRC) beams ultimate behavior. Constr Build Mater 23(2):1064–1077 5. Chunxiang Q, Patnaikuni I (1999) Properties of high-strength steel fiber-reinforced concrete beams in bending. Cement Concr Compos 21(1):73–81 6. Dancygier AN, Savir Z (2006) Flexural behavior of HSFRC with low reinforcement ratios. Eng Struct 28(11):1503–1512 7. Ozden G, Akdag CT (2009) Lateral load response of steel fiber reinforced concrete model piles in cohesionless soil. Constr Build Mater 23(2):785–794 8. Akdag CT, Özden G (2013) Nonlinear behavior of reinforced concrete (RC) and steel fiber added RC (WS-SFRC) model piles in medium dense sand. Constr Build Mater 48:464–472 9. Deb P, Pal SK (2019) Numerical analysis of piled raft foundation under combined vertical and lateral loading. Ocean Eng 190:106431

Centrifuge Modelling of Marginal Soil Slopes Under Rainfall with Hybrid Geosynthetic Inclusions Dipankana Bhattacherjee

and B. V. S. Viswanadham

Abstract The present study focuses on investigating the effect of inclusion of an innovative hybrid geosynthetic on the seepage, deformation and stability aspects of marginal soil slopes subjected to rainfall. Model hybrid geosynthetics were prepared in the study by integrating the drainage potential of nonwoven geotextile with the reinforcement function of woven geogrid. The model soil was a blend of fine sand and kaolin in the ratio of 4:1 by dry weight. The silty sand exhibited a percentage of fines equal to 20% and a saturated permeability of 1.54 × 10–6 m/s, thereby representing the properties of marginal soils found in major portions of India and other parts of the world. Centrifuge-based physical modelling was adopted at 30 gravities on slopes of 7.2 m height and crest width of 7.5 m using the 4.5 m radius beam centrifuge facility available at IIT Bombay, India. Rainfall was simulated using a custom-designed rainfall simulating assembly for a prototype rainfall intensity of 20 mm/h. It was observed that the unreinforced slope model experienced a catastrophic failure, while the hybrid geosynthetic reinforced slope experienced negligible deformation throughout the rainfall event. The surface settlements and slope face movements decreased substantially by about 94% and 71%, respectively, owing to the geogrid component. Further, the inclusion of geotextile component of hybrid geosynthetics resulted in a reduction of pore water pressures by almost 66%, thereby indicating the importance of hybrid geosynthetics in alleviating the instability of marginal soil slopes subjected to rainfall. Use of hybrid geosynthetics thus facilitate the use of marginal soils in reinforced earth construction, thereby economizing the project. Keywords Slope stability · Hybrid geosynthetics · Rainfall · Centrifuge test · Marginal soils D. Bhattacherjee (B) Department of Civil Engineering, Indian Institute of Engineering Science and Technology, Shibpur, Howrah 711103, India e-mail: [email protected]; [email protected] B. V. S. Viswanadham Department of Civil Engineering, Indian Institute of Technology Bombay, Powai, Mumbai 400076, India e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_30

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1 Introduction In recent times, global warming and associated climatic changes have adversely affected the environmental balance and triggered the frequency of hydrometeorological events. As a consequence, instability of natural and engineered soil slopes and retaining walls induced by rainfall have come to the forefront, resulting in significant economic damage and loss of life. The annual statistical review report of the Centre for Research on the Epidemiology of Disasters [1] reveals that, in the year 2016, almost 75% of the natural disasters that occurred in the country may be attributed to rainfall, and the estimated average losses due to rainfall-triggered landslides (especially in the Himalayan regions) exceed Rs. 550 crores/year (about 77 Million US $ per year), as reported by Dahal and Hasegawa (2008) [2]. The instability may be associated primarily with the loss of soil matric suction under rainfall, leading to the build-up of positive pore water pressure within slopes. The situation aggravates if the soil used in reinforced earth construction exhibits low permeability, and cannot dissipate the pore water pressures generated during meteorological events. However, due to increasing scarcity of good quality permeable granular soil, onsite low-permeable soils (or marginal soils) are being widely utilized in the field in recent times. This has led to increased incidents of reinforced slope/wall failures, and landslides being reported owing to the reduced strength, considerable fines content, and low permeability associated with such marginal soils. An alternative mitigation methodology is to ensure freely draining condition within marginal soil slopes during rainwater infiltration for enhanced stability. The present study focuses on investigating the effect of inclusion of a special variety of geosynthetic material, referred to as hybrid geosynthetic (or geogrid based geocomposite) on the seepage, deformation, and stability aspects of marginal soil slopes subjected to rainfall. Hybrid geosynthetic is an assembled material possessing both in-plane drainage and reinforcement characteristics derived from a nonwoven geotextile and geogrid, respectively, as investigated by Bhattacherjee and Viswanadham (2016) [3] and Viswanadham and Bhattacherjee (2015) [4]. In the literature, the use of permeable inclusions within natural and engineered slopes have been investigated by numerous researchers including Tatsuoka and Yamauchi (1986) [5], Zornberg et al. (1998) [6], Akay et al. (2014) [7], Thuo et al. (2015) [8] and Cotecchia et al. (2016) [9], whereas the importance of reinforcement function was reported by Iryo and Rowe (2005) [10], Wu and Chou (2013) [11] and Abd and Utili (2017) [12]. However, till date, studies on the potential coupling of reinforcement and drainage functions are limited, especially with respect to slope stability under rainfall condition tested in a geotechnical centrifuge. Hence, this forms a topic of major research interest. An innovative rainfall simulator was designed for inducing rainfall at high gravities, which is a novel aspect of the present research.

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331

2 Model Materials and Scale Factors The model soil used in slope preparation was formulated in the laboratory by blending locally available fine sand and commercially available kaolin in the ratio of 4:1 by dry weight. The model silty sand was formulated such that it has a percentage of fines equal to 20% and a saturated permeability (k sat ) of 1.54 × 10–6 m/s, thereby representing the properties of locally available marginal soils. In this regard, it should be mentioned that marginal soils are defined as that containing fines in excess of 15%, as per Christopher and Stuglis (2005) [13], and having a saturated coefficient of permeability (k sat ) of the order of 1 × 10–6 m/sec or less, as per Holtz and Kovacs (1981) [14]. Model hybrid geosynthetics (G1N1) were prepared by integrating the drainage potential of nonwoven geotextile (N1) with the reinforcement function of woven geogrid (G1). In order to minimize particle size scale effects in a centrifuge, a ratio of S t /D50 > 10 as suggested by Izawa and Kuwano (2010) [15] was adopted during model geogrid selection, where S t is the spacing between transverse ribs of the reinforcement and D50 is the average particle size of model soil. The ultimate tensile capacity (T gu ) and ultimate tensile strain (εgu ) of model hybrid geosynthetic was evaluated as 2.1 kN/m and 22.8%, respectively, along the machine direction as per wide-width tensile test procedure outlined in ASTM D 4595 (2005) [16]. Further, the drainage potential of hybrid geosynthetic was ascertained based on the radial flow principle outlined in ASTM D 6574 (2006) [17], and a transmissivity of 1.9 × 10–6 m2 /s was determined in the laboratory. Centrifuge-based physical modelling technique was adopted in the present study to replicate similar stress history and retain identical state of stresses in the model as that of the full-scale prototype. During geotechnical centrifuge testing, a centrifugal acceleration of high gravities (Ng) is applied relative to that of earth’s normal gravity (g). Standard scaling relationships are employed to link the model behaviour with the corresponding prototype. The parent geotextile was scaled based on identical transmissivity requirements outlined in Raisinghani and Viswanadham (2011) [18], whereas the parent geogrid was modelled based on scaling considerations proposed by Viswanadham and König (2004) [19]. Modelling of rainfall at high gravities was performed based on standard scaling laws outlined in Tamate et al. (2010) [20] and Bhattacherjee and Viswanadham (2018a) [21].

3 Model Preparation and Test Procedure Centrifuge-based physical modelling was performed at 30 gravities on a representative silty sand slope of 240 mm height, 60 mm base layer and 2 V:1H inclination, having a crest width of 250 mm. Tests were conducted using the 4.5 m radius beam centrifuge facility available at IIT Bombay, INDIA. The above gravity level (30 g) and radius of centrifuge ensured minimum scale effects due to variation of gravity level with model depth and horizontal distance in the model, as outlined in Taylor (1995)

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[22]. The model slope corresponded to a prototype height of 7.2 m and a crest width of 7.5 m in the field. Rainfall was simulated using a custom-designed rainfall simulating assembly consisting of spray nozzles, nozzle hanging rods, nozzle assembly attaching plate, a water container assembly with support system and additional components involving a solenoid valve, a seepage tank and run-off collector. Adequate measures were taken to minimize the effects of Coriolis force generated at high gravities by enabling a shift in the position of nozzles depending upon the direction of centrifugal rotation. The nozzles are specially designed pneumatic nozzles capable of producing fine mist at a uniform rate in-flight condition for intensities ranging from 2 mm/h to as high as 80 mm/h. A set of four nozzles were placed at the slope crest, while another four were placed at the inclined face to ensure uniform distribution of rainfall over the slope surface. The slopes were instrumented with four pore pressure transducers (PPTs) placed above the base layer at distances of 20 mm (PPT4), 125 mm (PPT3), 250 mm (PPT2), and 350 mm (PPT1) from the perforated face of the seepage tank in model dimensions. Further, L-shaped plastic markers made from thin transparency sheets of 20 mm × 10 mm dimensions were embedded within the slope front elevation to track slope displacements with the progress of rainfall. Additional L-shaped plastic markers were glued on to the slope face to facilitate computation of slope face movements with rainfall. The model hybrid geosynthetic (G1N1) was cut to a total length of (LA + LF + LR ) and to a width of 200 mm. The anchorage length (LA ) was equivalent to 0.25 times the model slope height h, LF represented the length along the slope face and the reinforcement length (LR ) was 0.85 h. The various stages involved in the construction of unreinforced and reinforced slope models in the centrifuge are discussed elaborately in Bhattacherjee and Viswanadham (2018b) [23]. The response of unreinforced and reinforced slope models was monitored for a prototype rainfall intensity of 20 mm/h, which corresponds to a heavy rainfall event as per standard global thresholds of Llasat (2001) [24].

4 Results and Discussion The surface settlements, slope face movements, and pore water pressure profiles developed during rainfall at various time intervals were investigated in the present study based on data recorded by pore water pressure transducers and through image analysis [25] of selected images captured during centrifuge tests. The results of two centrifuge model tests (T1 and T2) are discussed in this section, wherein Model T1 represents an unreinforced slope and Model T2 corresponds to a reinforced slope with six layers of hybrid geosynthetic (G1N1) inclusions. In both the cases, an initial water table was maintained up to the slope toe at the onset of rainfall by means of horizontal seepage induced by a seepage tank. The duration of centrifuge tests was maintained as 30 min (18.75 days in prototype dimensions) from the period of starting rainfall for the reinforced slope (Model T2), or until failure in case of the unreinforced slope (Model T1).

Centrifuge Modelling of Marginal Soil Slopes …

333 Stable slope (no visible movement)

Toe failure

(b)

(a)

Unreinforced slope

Reinforced slope

Fig. 1 Front elevation of slopes a Model T1 [t = 9.375 days] b Model T2 [t = 18.75 days]

4.1 Deformed Slope Profiles Observed Post Rainfall The front elevation of unreinforced and hybrid geosynthetic reinforced slopes captured during the ultimate stage of centrifuge tests at the end of 9.375 days and 18.75 days of rainfall (in prototype dimensions) are presented in Fig. 1a, b respectively. It can be observed from Fig. 1a that the unreinforced slope (Model T1) experienced a catastrophic toe failure due to rainwater infiltration. However, the hybrid geosynthetic reinforced slope (Model T2) was stable throughout the rainfall event, with no visible slope movements captured until the end of the rainfall event shown in Fig. 1b.

4.2 Pore Water Pressure Generation with Rainfall The pore water pressure generation during rainfall for unreinforced slope and hybrid geosynthetic reinforced slope are shown in Fig. 2. The values measured by PPT3 (uPPT3 /γ h) placed vertically below the mid-point of the crest of the slope on the base layer are herein presented. The pore pressures (u) are normalized with respect to the unit weight of model soil (γ ) multiplied by the slope height (h), and are expressed in prototype dimensions, starting from the time of occurrence of rainfall. As visible from Fig. 2, the unreinforced slope Model T1 exhibited increasing uPPT3 /γ h values with rainfall, the peak value being 0.438. On the contrary, the slope model reinforced with hybrid geosynthetic layers, namely Model T2 (G1N1) exhibited lower uPPT3 /γ h values for the entire duration of tests with a peak value of 0.028. The maximum normalized pore pressure (umax /γ h) recorded by PPT4 for both the models are presented in Table 1, which indicates that the presence of hybrid geosynthetics resulted in a reduction of pore water pressure values by almost 66%. The above implies the effectiveness of the geotextile component of hybrid geosynthetics in

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0.5

T2 [G1N1]

PPT 3 is located at 7.35 m distance from slope toe

u PPT 3 /γh

0.4

Rainfall started at t = 0 days

0.3 0.2 0.1 0.0

0.0

2.5

5.0

7.5

10.0

12.5

15.0

17.5

20.0

22.5

25.0

Time of occurence of rainfall in prototype scale, t(days)

Fig. 2 Variation of normalized excess pore water pressure with rainfall

Table 1 Summary of test results

Parameter

Model T1 [Unreinforced]

Model T2 [Reinforced]

Time elapsed during rainfall (days)

a 9.375

b 18.75

0.554

0.185

cu

max /γ h

S c,max /h

0.492

0.020

S f,max /h

0.166

0.048

Note All values are reported in prototype dimensions; h: Height of slope; u/γh: Normalized pore water pressure; Sc,max : Max. crest settlement; Sf,max : Max. deformation along slope face; a Time corresponding to failure; b During ultimate stage of test‚ beyond which no significant variations in pore water pressure magnitudes or slope deformation was observed; c For PPT4 placed at 350 mm from slope toe

dissipating the excess pore water pressures generated during a rainfall event, thereby ensuring the stability of the slope under rainfall conditions.

4.3 Variation of Surface Settlements with Rainfall Figure 3 presents the variation of surface settlements measured from the slope crest at the ultimate stage of the tests. As evident from Fig. 3, the unreinforced slope (Model T1) showed a gradual increase in surface settlement with rainfall of 20 mm/h, the maximum value being 3.54 m (in prototype dimensions) at the crest at the ultimate

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Slope crest 0 0.0

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Horizontal distance from crest of the slope (m) 1

2

3

4

5

6

7

8

Surface settlement (m)

0.5 1.0 1.5 2.0

T1 [unreinforced]

2.5 3.0 3.5

T2 [G1N1] Surface settlement

4.0 Fig. 3 Variation of surface settlements from slope crest with rainfall

stage. This high magnitude of deformation is attributed to the absence of reinforcement function within the marginal soil slope subjected to rainfall. On the contrary, the slope reinforced with hybrid geosynthetic layers (Model T2) recorded negligible settlement of the order of 0.092 m (in prototype dimensions) under the same rainfall intensity of 20 mm/h, as can be seen from Fig. 3. Hence, the inclusion of hybrid geosynthetics resulted in reduced vertical settlements at slope surface by about 94%.

4.4 Slope Face Movements with Rainfall Figure 4 presents the displacements observed at the slope face with the progress of rainfall, obtained by tracking the co-ordinates of inclined markers at the slope face in case of unreinforced slope (Model T1), and that of plastic markers stuck to hybrid geosynthetic layers facing towards the slope face for reinforced slope (Model T2). The face movements have been plotted considering the slope face to be vertical, and coinciding with the vertical axis and origin at the toe. As evident from Fig. 4, the unreinforced slope (Model T1) recorded a sudden displacement at the toe in the order of about 1.20 m in prototype dimensions at the time of failure (t = 9.375 days). This may be attributed to the building up of pore water pressures within the slope due to rainfall [observed previously in Fig. 2], giving rise to positive seepage forces. On the contrary, the slope reinforced with hybrid geosynthetic layers (Model T2) depicted a negligible increase in lateral displacements at slope face with rainfall,

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6.6 6.0

Height of slope (m)

5.4 4.8 4.2

Slope face

3.6 3.0 2.4 1.8 1.2 T2 [G1N1]

T1 [unreinforced]

0.6 0.0 0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

2

Movement along slope face in prototype dimension (m) Fig. 4 Slope face movements observed during rainfall

the maximum value being 0.36 m in prototype scale (t = 18.75 days). Hence, the geogrid component of hybrid geosynthetics resulted in reduced slope face movements by about 71% as compared to unreinforced marginal soil slopes. The results derived from the centrifuge model tests conducted in the present study are hereby summarized in Table 1.

5 Conclusions The present study highlights the importance of coupling the two functions of drainage and reinforcement offered by geogrid and nonwoven geotextile, respectively, into one integral material referred to as hybrid geosynthetic for improving the performance of marginal soil slopes subjected to rainfall. Modelling of hybrid geosynthetics in a geotechnical centrifuge and use of the same in alleviating the problems associated with marginal soil slopes under rainfall using an innovative in-flight rainfall simulator developed for this purpose may be considered as the novel aspects of the present study. Based on the centrifuge model tests conducted in the study, it is inferred that the unreinforced slope model experienced a catastrophic failure and increasing phreatic levels with rainfall. On the contrary, the hybrid geosynthetic reinforced slope experienced negligible deformation throughout the rainfall event, and the surface settlements and slope face movements decreased substantially by about

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94% and 71%, respectively, owing to the geogrid component. Further, the inclusion of geotextile component of hybrid geosynthetics resulted in a reduction of pore water pressure values by almost 66%. The above finding facilitates the use of soils available locally at the construction site in infrastructural projects, thereby economizing project costs, and preventing to a large extent the unsustainable over-mining of natural sand deposits for construction purposes.

References 1. CRED (Centre for Research on the Epidemiology of Disaster) (2012) Annual Disaster Statistical Review 2011: The Numbers and Trends. Research Unit, Université Catholique de Louvain, Louvain-la-Neuve, Belgium, CRED 2. Dahal RK, Hasegawa S (2008) Representative rainfall thresholds for landslides in the Nepal Himalaya. Geomorphology 100(3–4):429–443 3. Bhattacherjee D, Viswanadham BVS (2016) Effect of hybrid geosynthetic layers on soil walls with marginal backfill subjected to rainfall. In: De A, Reddy KR, Yesiller N, Zekkos D, Farid A (eds.) Proceedings of Geo-Chicago 2016, Geotechnical Special Publication No 269, ASCE (Pubs.), pp 362–371 4. Viswanadham BVS, Bhattacherjee D (2015) Studies on the performance of Geocomposite reinforced low-permeable slopes subjected to rainfall. Japn Geotech Soc Spe Pub 2(69):2362– 2367 5. Tatsuoka F, Yamauchi H (1986) A Reinforcing Method for Steep Clay Slopes using Non-woven Geotextile. Geotext Geomembr 4(3–4):241–268 6. Zornberg JG, Sitar N, Mitchell JK (1998) Performance of geosynthetic reinforced slopes at failure. J Geotech Geoenviron Eng 124(8):670–683 7. Akay O, O¨zer AT, Fox GA, (2014) Assessment of EPS block geofoam with internal drainage for sandy slopes subjected to seepage flow. Geosyn Int 21(6):364–376 8. Thuo JN, Yang KH, Huang CC (2015) Infiltration into unsaturated reinforced slopes with nonwoven geotextile drains sandwiched in sand layers. Geosyn Int 22(6):457–474 9. Cotecchia F, Lollino P, Petti R (2016) Efficacy of drainage trenches to stabilise deep slow landslides in clay slopes. Géotechnique Let 6(1):1–6 10. Iryo T, Rowe RK (2005) Hydraulic behaviour of soil geocomposite layers in slopes. Geosyn Int 12(3):145–155 11. Wu JY, Chou NN (2013) Forensic studies of geosynthetic reinforced structure failures. J Perform Construct Facil ASCE 27(5):604–613 12. Abd AH, Utili S (2017) Design of geosynthetic-reinforced slopes in cohesive backfills. Geotext Geomembr 45(6):627–641 13. Christopher BR, Stuglis RS (2005) Low permeable backfill soils in geosynthetic reinforced soil wall: State of the practice in North America. In: Proceedings of North American Geosynthetics conference (NAGS2005), GRI-19, Las Vegas, USA, pp 14–16 14. Holtz RD, Kovacs WD (1981) An introduction to geotechnical engineering Prentice Hall. Englewood Cliffs, New Jersey 15. Izawa J, Kuwano J (2010) Centrifuge modelling of geogrid reinforced soil walls subjected to pseudo-static loading. Int J Phys Modell Geotech 10(1):1–18 16. ASTM D 4595 (2005) Standard test method for tensile properties of geotextile by the widewidth strip method, Annual Book of ASTM Standards, Section 4, Volume 04.13, Geosynthetics, American Society for Testing and Materials, West Conshohocken, Pennsylvania, USA 17. ASTM D 6574 (2006) Standard test method for determining the (in-plane) hydraulic transmissivity of a geosynthetic by radial flow, American Society for Testing and Materials, West Conshohocken, Pennsylvania, USA

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18. Raisinghani DV, Viswanadham BVS (2011) Centrifuge model study on low permeable slope reinforced by hybrid geosynthetics. Geotext Geomembr 29(6):567–580 19. Viswanadham BVS, König D (2004) Studies on scaling and instrumentation of a geogrid. Geotext Geomembr 22(5):307–328 20. Tamate S, Suemasa N, Katada T (2010) Simulating shallow failure in slopes due to heavy precipitation. In: Springman S, Laue J, Seward L (eds.) Proceedings of the 7th international conference in physical modelling in geotechnics – 7th ICPMG, Switzerland Taylor and Francis group (Pubs.), vol 2, pp 1143–1149 21. Bhattacherjee D, Viswanadham BVS (2018) Design and Performance of an Inflight Rainfall Simulator in a Geotechnical Centrifuge. Geotech Test J 41(1):72–91 22. Taylor RN (1995) Centrifuges in modelling: principles and scale effects. Geotechnical Centrifuge Technology, Blackie Academic and Professional, Glasgow, U.K. 23. Bhattacherjee D, Viswanadham BVS (2018) Effect of geocomposite layers on slope stability under rainfall condition. Indian Geotech J 48(2):316–326 24. Llasat MC (2001) An objective classification of rainfall events on the basis of their convective features. Int J Climatol 21(1):1385–1400 25. Image-Pro Plus (2004) Image-Pro Plus Manual. Ver. 5.1. Media Cybernetics, Inc., USA

Study of Geosynthetic Reinforced Retaining Wall under Various Loading Ratnesh Ojha, Ananya Srivastava, and Vinay Bhushan Chauhan

Abstract Reinforcement in the soil can be used to enhance the behavior of retaining walls under seismic loading in terms of improved overall stability of the structure. Mechanically Stabilized Earth (MSE) retaining walls, referred to as “segmental” retaining walls owing to their use of segmental blocks as facing elements for reinforcement, have been extensively used in recent years as permanent retaining structures. Nevertheless, in case of high retaining wall, tiered walls are preferred over a single height retaining wall due to their cost-effectiveness and more stability than conventional MSE walls. However, the behavior of tiered MSE retaining wall is not yet studied thoroughly. In the present paper, an attempt has been made to examine the behavior of tiered MSE walls having a height of 12 m under static and seismic loading using the numerical analysis. The present paper compares the stability of tiered MSE walls with a conventional MSE wall in terms of factor of safety (FOS) and possible mode of failure. This study discusses the effect of the length of geogrid in multiple layers on the stability of the reinforced retaining wall due to various seismic excitation levels and also investigates the generated failure modes. Keywords MSE walls · FEM · Numerical modeling · Two-tier wall

1 Introduction Soil Reinforcement is a technique to increase the shear strength of soil to resist the load. It is used to increase the stability of the structure using geosynthetics as a tensile reinforcement. MSE wall is a combination of compacted soil, reinforced materials R. Ojha · A. Srivastava · V. B. Chauhan (B) Civil Engineering Department, Madan Mohan Malaviya University of Technology, Gorakhpur 273010, India e-mail: [email protected] R. Ojha e-mail: [email protected] A. Srivastava e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_31

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in the form of layers which are directly associated with the facing of the structure. Modular blocks, concrete panels, and geosynthetics are used for the facing of MSE walls. In the past decades, reinforced soil walls have been widely adopted due to their cost-effectiveness and easy construction. Out of many alternatives to retain high backfill, retaining wall with relief shelves can be a potential alternative solution for most places, where reinforced concrete or gravity type walls have traditionally been used to retain soil [1–3]. However, the flexible nature of MSE walls is found to the best alternative for seismic loading. In the past decades, reinforced soil walls have been widely adopted due to their cost-effectiveness and easy construction. MSE retaining structures are cost-effective alternatives, where reinforced concrete or gravity type walls have traditionally been used to retain the soil. These structures include bridge abutments and wing walls, as well as areas where the right-of-way is restricted, such that an embankment or excavation with stable side slopes cannot be constructed. These are the cost-effective structures in hilly areas, at a location of slope instability or the region having soil strata with poor bearing capacity to lay down a foundation for the rigid retaining walls [4, 5]. A tiered retaining wall is a system in which two or more short walls or tiers are constructed at different levels, where each tier is placed on the backfill surface of the preceding tier and spaced apart at a certain offset distance [6]. These walls provide a beautiful appearance, space for the plantations, and a better load bearing capacity. Spacing between the successive reinforcement layers plays a vital role as the decrease in spacing between the successive reinforcing layers improves the performance of the structure under seismic loading. And also, seismic performances of the reinforced walls can be improved by using multi-tiered arrangements [7–10]. The design and performance of the tiered MSE wall are majorly governed by the offset distance between the two consecutive tier walls. It is important to know that in some conditions if the walls are tiered, the lower wall may fail due to the extra loading of the upper wall. It is found from the previous studies that whenever the upper wall is located at the distance less than twice the height of the lower wall, the upper wall can apply extra load to the lower wall. So to avoid this condition, the spacing between the walls should be appropriate such that the system can get overall stability [11– 13]. Analysis, design, and the behavior of tiered MSE walls are often more complex than conventional monolithic walls. As the tiered wall is a complex structure, so the estimation of geogrid for such a wall cannot be done by standard design charts. However, for the monolithic retaining wall, reinforcement length for the desired factor of safety is much higher because reinforcement length is directly proportional to the height of the wall. In case of the same length and spacing of reinforcement for the monolithic and tiered wall, the backfill settlement in the tiered wall is lower than that of the monolithic wall [7]. Furthermore, the width of excavation behind the wall to lay the reinforcement will be high, which subsequently increases the overall cost of the high monolithic MSE retaining structures, i.e., increases construction, as well as reinforcing material cost. In such cases, tiered walls are preferred where a great height of soil retention is required.

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The study of tiered MSE walls is complex, therefore, the behavior under seismic loading has not been well investigated. To have an insight into the behavior of tiered MSE walls in terms of relationship among soil, reinforcement, and walls, a study has been carried out using a finite element modeling of a 12 m high wall. Previous studies give an idea about the length of geogrid in the MSE wall such as it has a massive effect to control the failure of the structure. In the present study, the performance of a 2 tier wall, each having a height of 6 m is compared with a 12 m high monolithic wall with varying reinforcement length under static and seismic loading while evaluating their stability in terms of the factor of safety and possible mode of failure.

2 Numerical modeling of MSE wall In the present study, FHWA recommendations have been followed to design the walls (spacing and length of reinforcement) and to examine the behavior under gravity and seismic loading using finite element modeling for various lengths of reinforcement. For walls of greater heights, two-tiered or multi-tiered walls are preferred for designing. Here in this study, we have considered a two-tiered wall for numerical modeling, the guidelines for which suggests that the individual walls in the tiered system can be treated as separate, independent walls when the offset distance, D, exceeds the following: D = Hm tan(90 − ϕr )

(1)

where, H m : the height of the lower of the two tiers and ϕ r : the friction angle for the reinforced soil backfill [12]. For the present study, a two-tiered superimposed geogrid reinforced soil retaining wall and one monolithic MSE wall having a total height of 12 m with a rigid facing were considered to investigate their stability under gravity and seismic loading. A leveling pad with dimensions 2 m × 0.2 m is constructed on the foundation soil over which a layer of backfill soil with a lift of 0.6 m is deposited. A leveling pad with dimensions 2 m × 0.2 m is constructed on the foundation soil over which a rigid concrete facing with dimensions 1.5 m × 0.6 m is placed and a layer of backfill soil with a lift of 0.6 m is deposited behind the wall facing. In between the concrete facings, a geosynthetic layer (L) of the desired length is placed. Again, above it, the backfill layer is deposited with concrete facing and geosynthetic layer and this procedure is repeated till the desired height of the wall is achieved. The geosynthetic material is taken as an elastic material which can bear the tensile load. The backfill model is the Mohr–Coulomb model. The block geosynthetic interface value is taken as 0.8 and that for soil geosynthetic interface is 0.65. For the numerical analysis of geogrid reinforced soil retaining wall, a series of organized simulations were executed using a finite element based computational tool Optum G2 [14] in the two-dimensional analysis.

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Fig. 1 Typical mesh for the conventional MSE wall having a height of 12 m considered in the present study

The stability of the overall structure is evaluated using the strength reduction method with the consideration of reduced strength parameters of the backfill and the factor of safety is calculated. The total height of the monolithic wall is 12 m (H) and the height of the two-tiered wall is divided into a lower and upper tier (H 1 = H 2 = 6 m). Complete details of the geometrical configuration of monolithic and tiered geogrid reinforced MSE walls are shown in Figs. 1 and 2, respectively. Material properties used for the numerical simulation of walls are given in Table 1. As per the suggestions laid by FHWA for tiered walls, the range of geogrid reinforcement length in the lower and upper walls should not be less than 0.7 times of the individual wall height for static case [12, 13, 15]. In the present study, the length of reinforcement is varied from L/H ratio 0.5 to L/H ratio 1.1 at an interval of 0.2 units, where L/H is the ratio of the length of the geogrid to the height of the wall. Although the L/H ratio of 0.5 is not recommended according to FHWA, still it has been considered to analyze the failure plane that occurs on such a ratio. The stability assessment of reinforced retaining wall under seismic loading is carried out for a range of horizontal seismic acceleration coefficient (k h = 0.12, 0.24, and 0.36) and compared with the non-seismic case (k h = 0). Failure modes of the wall and their transition from one mode to another mode of failure due to geogrid length configuration has been studied and analyzed. The results obtained from the analysis have been discussed in the next section.

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Fig. 2 Typical mesh for a two-tier MSE wall having a height of 12 m considered in the present study

Table 1 Material properties considered in the present study

Property

Geogrid Backfill material

Stiffness of geosynthetic (kN/m)

880

–

Yield force of geosynthetic (kN/m) 88

–

Modulus of elasticity (MPa)

–

35

Poisson’s ratio

–

0.25

Cohesion (kN/m2 )

–

0.1

Dry unit weight (kN/m3 )

–

16

Internal friction angle (degrees)

–

36.5

Dilation angle (degrees)

–

5

3 Results and Discussion The failure plane diagram for L/H ratio = 0.5 in two-tiered wall and monolithic wall is shown, respectively, in Figs. 3 and 4. As we see in the figure above, the failure planes for two-tiered walls begin at the toe below the leveling pad and propagate towards the backfill in a rough wedge like pattern. The soil mass beneath the offset distance does not intersect the failure path, thus justifying the fact that after a calculated offset distance, the walls exert no influence over each other. In monolithic walls, it is observed that as the seismicity increases, the soil below the leveling pad sinks which causes failure.

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Fig. 3 Mode of failure for two-tiered wall having L/H=0.5 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

Fig. 4 Mode of failure for Monolithic wall having L/H = 0.5 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

The failure plane diagram for L/H ratio = 0.6 in two-tiered wall and monolithic wall is shown, respectively, in Figs. 5 and 6. In two-tiered walls, we see that as the seismicity increases the failure pattern shows a more definite wedge like geometry. The failure begins at the toe and propagates towards the backfill. In monolithic wall, failure also occurs at the end of the geosynthetic layer, further propagating towards the backfill.

Fig. 5 Mode of failure for two-tiered wall having L/H = 0.6 (a) kh = 0, (b) kh = 0.12, (c) kh = 0.24, (d) kh = 0.36

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Fig. 6 Mode of failure for Monolithic wall having L/H = 0.6 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

The failure plane diagram for L/H ratio=0.8 in two-tiered wall and monolithic wall is shown, respectively, in Figs. 7 and 8. In the figure above, for two-tiered wall at k h = 0.24, the failure propagates from the toe of the wall and at this place, maximum stresses are developed. In monolithic walls, as the seismicity increases the force on the geosynthetic layer is also increased. In case of two-tiered walls, the geosynthetic layer is not much affected.

Fig. 7 Mode of failure for two-tiered wall having L/H = 0.8 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

Fig. 8 Mode of failure for Monolithic wall having L/H = 0.8 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

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The failure plane diagram for L/H ratio=1.0 in two-tiered wall and monolithic wall is shown, respectively, in Figs. 9 and 10. Here we observe that for monolithic walls the geosynthetic layer gets affected as the seismicity increases, but in two-tiered walls, the failure plane propagates from toe to the backfill in a wedge like structure without much affecting the geosynthetic layer. The failure plane diagram for L/H ratio = 1.2 in two-tiered wall and monolithic wall is shown, respectively, in Figs. 11 and 12.

Fig. 9 Mode of failure for two-tiered wall having L/H = 1.0 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

Fig. 10 Mode of failure for Monolithic wall having L/H = 1.0 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

Fig. 11 Mode of failure for two-tiered wall having L/H = 1.2 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

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Fig. 12 Mode of failure for monolithic wall having L/H=1.2 a k h = 0, b k h = 0.12, c k h = 0.24, d k h = 0.36

In two-tiered walls, stresses are generated in soil mass behind the geosynthetic layer and at the toe of both the walls. In monolithic walls, failure propagates in between the geosynthetic layer and also in the backfill. By comparing the FOS values in both cases of walls, one can see that at increased seismicity, two-tiered wall does not show much reduction in FOS valuescompared to the monolithic wall, which is favorable as the two-tiered walls reduce the cost of construction (previously discussed) and also show the promising values of FOS. The graph for FOS versus L/H ratio in two-tiered wall and monolithic wall at Kh = 0 is shown, respectively, in Fig. 13. For two-tiered wall, the FOS is maximum at L/H ratio = 1.2 and is minimum at L/H = 0.5. In monolithic walls, the L/H ratio increases, FOS increases almost linearly for increased horizontal ground acceleration coefficients (K h ). It is also observed that, with increased L/H ratio, minutely higher difference is noted in lower bound and

Fig. 13 Variation of FOS with the L/H with error bars indicating the worst case between upper and lower bound solutions for non-seismic case (K h = 0) a two-tiered wall b monolithic wall

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upper bound values than the same at a lower ratio of L/H, where the points are very convergent in nature. The graph for FOS versus L/H ratio in two-tiered wall and monolithic wall at K h = 0.12 is shown, respectively, in Fig. 14. From the above graph, the monolithic wall shows a very constant rise in the values of FOS with increased L/H ratio with upper and lower bound values almost convergent. The two-tiered wall shows a minimum change in FOS with the increment of L/H ratio giving maximum FOS of 1.35. The graph for FOS versus L/H ratio in two-tiered wall and monolithic wall at K h = 0.24 is shown, respectively, in Fig. 15. Here the difference in lower and upper bound values is higher as compared to the other two cases discussed before. The two-tiered wall shows a gradual and small

Fig. 14 Variation of FOS with the L/H with error bars indicating the worst case between upper and lower bound solutions for seismic case (K h = 0.12) a two-tiered wall b monolithic wall

Fig. 15 Variation of FOS with the L/H with error bars indicating the worst case between upper and lower bound solutions for seismic case (K h = 0.24) a two-tiered wall b monolithic wall

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increment in FOS value, whereas in monolithic wall, the value of FOS is far more at increased L/H ratio than it was in the beginning, also the graph is a straight line with a steep slope. The graph for FOS versus L/H ratio in two-tiered wall and monolithic wall at Kh = 0.36 is shown, respectively, in Fig. 16. In this case, both the walls show unsatisfactory results as their FOS value is less than 1.0. The two-tier wall shows a very minute difference in FOS value with varying geosynthetic length, whereas in monolithic wall, a higher difference is noticed. Also, the two-tiered wall shows a greater difference in lower and upper bound values than the monolithic wall. Based on the graphs shown above, optimum lengths of the geogrid for both cases can be summarized in Table 2.

Fig. 16 Variation of FOS with the L/H with error bars indicating the worst case between upper and lower bound solutions for seismic case (K h = 0.36) a two-tiered wall b monolithic wall

Table 2 Optimum length of Reinforcement for monolithic and two-tiered wall for different K h value Horizontal acceleration coefficient (K h )

Reinforcement length (L) for monolithic wall (H = 12 m)

Reinforcement lengths (L 1 , L 2 ) for two-tiered wall (H 1 = H 2 = 6 m) (m)

Kh = 0

L = 14.4

L1 = L2 = 4.8

Kh = 0.12

L = 14.4

L1 = L2 = 4.8

Kh = 0.24

L = 14.4

L1 = L2 = 5.4

Kh = 0.36

L = 14.4

L1 = L2 = 6.0

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4 Conclusion The present study observes the stability of the reinforced soil retaining wall and its modes of failure with the variation of geogrid length. From the results obtained, it can be concluded that the provision of geogrid has a major effect on improving the overall stability of reinforced retaining walls of greater height in both reinforced single wall and two-tiered wall. The length of geogrid significantly controls the stability of the overall structure, as well as the failure modes of the reinforced soil wall system. For all cases (seismic as well as non-seismic), possible modes of failure at different L/H ratios have been observed for two-tiered reinforced retaining wall with rigid facing at different K h values. The FOS values show a very minor difference during the comparison of monolithic and two-tiered wall. This suggests that when FOS is satisfactory, it is much convenient to construct two-tiered walls owing to their benefits discussed in the present study. Also, in places where backfill material has to be transported, the two-tiered walls are a better alternative as the excavation area is lesser in the latter case. The FOS increases with the increase in L/H ratio and decreases when K h increases. Moreover, it is observed that after a particular L/H ratio there is hardly any difference in FOS values in both cases of walls.

References 1. Chauhan VB, Dasaka SM (2016) Behaviour of rigid retaining wall with relief shelves with cohesive backfill. In: 5th international conference on forensic geotechnical engineering, Bangalore, India, pp. 350–357 2. Chauhan VB, Dasaka SM (2018) Performance of a rigid retaining wall with relief shelves. J Perform Constr Facil 32(3):04018021. https://doi.org/10.1061/(ASCE)CF.1943-5509.0001161 3. Chauhan VB, Khan R, Dasaka SM (2019) Reduction of surcharge induced earth pressure on rigid non-yielding retaining wall using relief shelves, In: Anirudhan IV, Maji V (eds) Geotechnical applications. Lecture notes in civil engineering, vol 13. Springer, Singapore, pp. 209–217. https://doi.org/10.1007/978-981-13-0368-5_23 4. Han J (2015) Principles and practice of ground improvement. Wiley, New Jersey 5. Ojha R, Chauhan VB (2019) Performance of geosynthetic reinforced segmental retaining wall. In: Shehata H, Brandl H, Boussida M, Sorour T (eds) Sustainable thoughts in ground improvement and soil stability GeoMEast. Sustainable civil infrastructures. https://doi.org/10.1007/ 978-3-030-34184-8 6. Tiered walls, VERSA-LOK standard design & installation guidelines. http://www.versa-lok. com 7. Liu H, Yang G, Ling HI (2014) Seismic response of multi-tiered reinforced soil retaining walls. Soil Dyn Earthq Eng 61–62:1–12 8. Liu H (2016) Comparing the seismic responses of single and multi-tiered geosynthetic reinforced soil walls. In: Geo-frontiers congress 2011, Dallas, Texas, United State, 13–16 March 2011. https://doi.org/10.1061/41165(397)356 9. Ling HI, Leshchinsky D, Mohri Y, Wang JP (2012) Earthquake response of reinforced segmental retaining walls backfilled with substantial percentage of fines. J Geotech Geoenviron Eng 934–944. https://doi.org/10.1061/(ASCE)GT.1943-5606.0000669 10. Richardson GN, Lee KL (1975) Seismic design of reinforced earth walls. ASCE J Geotech Eng Div 101(2):167–188

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11. Bathurst RJ, Hatemi J, Alfaro MC (2002) Geosynthetic reinforced soil walls and slopes: seismic aspects, In: Shukla SK (ed) Geosynthetic and their application. Thomas Telford, pp 327–392 12. FHWA (2009) Influence of soil reinforcement on horizontal displacement of MSE wall. Design and construction guidelines of mechanically stabilized earth walls and reinforced soil slopes 13. Holtz RD (1995) FHWA, geosynthetic design and construction guidelines. Federal Highway Administration Publication No. FHWA-HI-95-038 14. OptumG2 (2017) Optum Computational Engineering. Copenhagen NV, Denmark 15. Federal Highway Administration Publication No. FHWA-NHI-10-024, vol I. US Department of Transportation, Washington, DC, USA (2009)

Numerical Analysis on the Effect of Compaction Induced Stresses on the Performance of MSE Walls Akhil B. Alex and Renjitha Mary Varghese

Abstract A Mechanically Stabilized Earth (MSE) wall is a composite structure consisting of compacted backfill and reinforcement. When an earthfill is subjected to compaction, it results in the development of residual lateral stresses known as Compaction Induced stresses (CIS). The conventional design of MSE wall mainly considers the friction developed at the soil-reinforcement interface and the influence of compaction induced stresses on the reinforcements during the construction phase is commonly neglected. This study is focused to find the effectiveness of compaction provided during the construction of the retaining wall on its post construction behaviour. Initially, a numerical model was developed using the finite element method and validated with literature. Parametric studies were conducted with varying reinforcement spacing, reinforcement length and reinforcement strength under different surcharge loads. The effects of different compaction efforts on the development of CIS were also noted and compared. From the results, it was evident that reinforcement spacing had a significant influence on the performance of MSE walls compared to other parameters. Results also showed that lateral deformation will be higher during the construction phase when CIS is taken into account. However, the post construction lateral deformations were less at different surcharge loads. Higher the compaction effort, lesser will be the post construction deformation. Keywords MSE walls · Compaction induced stress · Reinforcement spacing

1 Introduction Mechanically Stabilized Earth (MSE) walls have gained popularity across the world during the recent decades and are now being widely used in slope stabilization and the field of transportation as such in case of bridge abutments, carriageway widening A. B. Alex (B) · R. M. Varghese Department of Civil Engineering, NIT Calicut, Kozhikode, Kerala, India e-mail: [email protected] R. M. Varghese e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_32

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across a slope, etc. This pronounced acceptance of MSE walls over conventional retaining walls is due to its rapid and straightforward construction procedures, less space required for construction, cost-effectiveness and technical feasibility to achieve heights above 30 m. An MSE wall is considered as a composite structure consisting of alternate layers of compacted soil and reinforcement. The soil-reinforcement interaction taking place between these elements contribute to the stability of the system [1]. During the construction of MSE walls, backfilling is done in stages along with the placement of reinforcement. Generally, after the placement of each lift, compaction is applied to each layer. In some cases, failures of MSE walls have been reported as inadequate compaction of the backfill [1]. Hence, it is essential to take into account and analyze the influence of compaction and the corresponding stresses developed, in the performance of MSE walls. When an earthfill is subjected to compaction, which is similar to repeated loading, results in the development of both vertical and lateral stresses. Even though vertical stress reduces to zero during the unloading phase, lateral stress developed does not get completely relieved and there exists residual lateral stress, termed as Compaction Induced Stresses (CIS). The magnitude of CIS developed depends upon the level of confinement existing in the soil [2]. Many studies have shown that a well compacted granular fill will be able to carry more loads while undergoing comparatively less lateral movement [3, 4]. CIS in a reinforced soil mass is expected to be more significant than CIS induced in unreinforced soil, as reinforced soil structure tends to restrict lateral movement through the interface friction developed between soil and reinforcement [5]. The objectives of this study were to understand the effect of surcharge, reinforcement spacing and reinforcement strength on the behaviour of MSE walls, by numerical modelling. While modelling, each layer of soil with corresponding reinforcement was subjected to different compaction efforts. The lateral deformations at the end of construction and post construction were monitored.

2 Numerical Analysis of Compaction Induced Stresses The numerical analysis of the MSE wall was done using the finite element software called PLAXIS 2D. This software enables the end users to analyze complex problems in geotechnical engineering with the help of 15 nodded elements. The results presented from the experimental model set up at the Geotechnical laboratory of COPPE, Federal University of Rio de Janeiro [6], were taken for numerical validation. This laboratory model of 1.4 m wall height was able to simulate the behaviour of the prototype, which was 6.8 m high and having a wrapped facing. The geometry of the model developed in Plaxis represented the prototype. The height of the wall taken was 6.8 m. Length of reinforcement adopted was 4.8 m and was placed at a vertical spacing of 0.4 m. The reinforcement length adopted was based on the standard specifications which suggested that minimum reinforcement length should be 70% of the wall height [5]. The wrapped facing was modelled with an inclination of

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6° to the vertical. Figure 1 shows the geometry of the model adopted for validation. The constitutive model adopted for the study is the Hardening Soil model, which is a hyperbolic soil model and can simulate the soil stiffness variations more realistically. Table 1 shows the input parameters used in this validation. Staged construction was considered; soil layers were placed every 0.2 m and compacted until the final wall height was achieved. Horizontal displacements were restricted at a distance of 11 m from the facing and both horizontal and vertical displacements were restricted at the bottom. The results obtained in plaxis were in good agreement with the results obtained by Mirmoradi and Ehrlich (2013). Various methods for simulating compaction operation have been put forward by many researchers [6]. A commonly adopted procedure was the application of a uniformly distributed load above each soil layer. Maximum stress due to roller compaction operation will be developed at roller-soil contact and it decreases with depth. It is assumed that every point in the soil layer is directed to an equal stress level during compaction. As stated earlier, since compaction stresses decrease with depth, the stresses acting beyond the soil lift thickness are neglected and the stresses will be considered acting throughout the depth, which leads to an overestimation of the stresses. A more reliable and accurate method was proposed by Mirmoradi and Ehrlich [6]. In this procedure, compaction is simulated by the application of uniformly distributed loads above and below each soil layer (a single load-unload cycle). Here, only the soil elements present in the corresponding layer experience compaction stresses, and at points below the layer, only geostatic stresses are present. Figure 2 shows the variations of lateral deformations at different levels of MSE walls with and without considering the compaction effort. Post construction

Fig. 1 Model geometry adopted for validation [6]

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Table 1 Input parameters for validation [6]

Parameter

Value

Backfill soil Peak plane strain friction angle, ϕ(0 )

50

Cohesion, C (kPa)

1

Dilation angle, ψ(0)

0

Unit weight, γ (kN/m3 )

21

ref E 50 ref E oed ref E ur

(kPa)

42,500

(kPa)

31,800

(kPa)

127,500

Stress dependence exponent (m)

0.5

Failure ratio Rf

0.7

Poisson’s ratio, ν

0.25

Reinforcement Elastic axial stiffness (kN/m)

600

Facing Elastic axial stiffness (kN/m) Elastic bending stiffness

60

(kNm2 /m)

1

8

Wall Height (m)

7

without compaction

6

without compaction+30kPa surcharge

5 4

60kPa compaction

3

60kPa compaction+30kPa surcharge

2 1 0

0

100

200

300

Horizontal deformation (mm) Fig. 2 Horizontal deformation produced in compacted and uncompacted model at the end of construction and when a surcharge was applied

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surcharge of 30 kPa was applied to the model and the lateral deformations soon after construction and the application of surcharge load were monitored. The effects of CIS were observed after providing a compaction of 60 kPa to each soil layer. It has been noted that the deformation at the end of construction is higher for the compacted retaining wall. However, the post construction deformation with a surcharge of 30 kPa is less for the compacted model than the uncompacted one (Fig. 2). While the uncompacted model produced a lateral deformation of 19 mm when a surcharge of 30 kPa was applied, the compacted model had undergone only a lateral deformation of 13.7 mm for the same conditions (Fig. 2). It is expected that as the compaction effort increases, lesser will be the post constructional displacement.

3 Parametric Study of MSE Walls Considering CIS Parametric studies were conducted with varying compaction efforts, reinforcement spacing, and reinforcement strength. Lateral deformations produced in the MSE wall for different parameters were analyzed at different surcharges. The geometry of the wall for the parametric study was the same as that adopted for numerical validation. The other input parameters were kept the same as that taken for validation.

3.1 Effect of Surcharge The behaviour of the MSE wall under static loading conditions was analyzed by applying different surcharge loads. The maximum lateral deformations were analyzed under various surcharge loads. Figure 3 shows extreme lateral displacements with respect to the height of the wall, represented as normalized facing

Lateral deformation / Height (x 10-3)

16 14 12 10 8

Without compaction With compaction (60kPa)

6 4 2 0 0

10

20

30

40

50

60

Surcharge (kPa)

Fig. 3 Ratio of extreme lateral deformation to wall height under different surcharges

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Post construction lateral deformation (mm)

35 30

without compaction with compaction (60kPa)

25 20 15 10 5 0

10

20

30

40

50

Surcharge (kPa) Fig. 4 Post construction lateral deformation at different surcharge loads

displacement (denoted as x/H * 10−3 ), at different surcharge loads for the models with and without compaction and a comparison of the post construction deformation between the compacted and uncompacted model is shown in Fig. 4. The x/H value at the end of construction for compacted and uncompacted MSE wall models were 10.3 and 6.7, respectively. During the post construction phase, x/H increased up to 14.6 for compacted wall and 11.2 for uncompacted wall until 50 kPa surcharge load. Though the range of x/H is higher for compacted wall, the rate of increase in x/H value during the post construction phase is 15% less compared with that of uncompacted model. While a post construction lateral deformation of 34.8 mm was noticed at a surcharge of 50 kPa in compacted wall, only 29.7 mm was noticed in uncompacted wall at the same surcharge condition, i.e. a reduction of 5 mm in post construction lateral deformation was observed (Fig. 4). It was observed that lateral displacement increases linearly with an increase in surcharge loads in both cases. However, the rate of increase in lateral displacement for compacted wall is reduced after the end of construction.

3.2 Effect of Different Compaction Effort A parametric study involving different compaction efforts were studied. Compaction efforts of 20, 40 and 60 kPa were simulated and analyzed along with the uncompacted MSE wall model. Figure 5 shows the normalized facing displacement produced for different compactions at various surcharge loads. The x/H value varied in the range of 6.1–10.2 for compaction efforts varying from 0 to 60 kPa at the end of the construction phase. During the post construction phase, when a surcharge of 50 kPa was applied, x/H value increased to a range of 11.2–14.6 for the compaction efforts ranging from 0 to 60 kPa. It has been observed that more than fifty per cent

Lateral deformation/ Height( x 10-3)

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15

Compaction 13

60 kPa 40 kPa

11

20 kPa

9

0 kPa

7 5

0

10

20

30

40

50

60

Surcharge (kPa) Fig. 5 Normalized facing displacement for different compaction efforts at various surcharge loads

Percentage Reduction in post consrtuction deformation (%)

reduction in postconstruction deformation was obtained for 60 kPa compaction at a surcharge load of 10 kPa (Fig. 6). The percentage reduction in post construction lateral displacement at 10 kPa surcharge load for 40 kPa and 20 kPa compactions were 36% and 19%, respectively. The lateral deformation due to compaction induced stresses are relatively more noticeable at a higher magnitude of compaction. The percentage reduction is prominent at lower surcharges and decreases with an increase in surcharge loads. For 60 kPa compaction, percentage reduction value declined to 23% at a surcharge of 40 kPa. Higher the compaction, more significant is the percentage reduction in postconstruction deformation. For lower compaction efforts of 20 and 40 kPa, the percentage reduction values decline significantly after a specific limit of surcharge load. 60 50

Compaction

40

60 kPa

30

40 kPa

20

20 kPa

10 0

0

10

20

30

40

50

Surcharge (kPa) Fig. 6 Percentage reduction in lateral deformation versus surcharge for different compaction efforts

A. B. Alex and R. M. Varghese

Lateral deformation/Height (x10-3)

360 18 15 12 9 6

without compaction 3

with compaction

0 0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

Reinforcement spacing(m) Fig. 7 Ratio of lateral deformation to wall height at various reinforcement spacing for models without and with compaction(60 kPa)

3.3 Effect of Reinforcement Spacing Figure 7 shows the lateral deformations experienced by the wall at various reinforcement spacings for compacted and uncompacted conditions. The x/H value varied from 8.3 to 17.4 when the reinforcement spacing was varied from 0.2 m to 0.6 m in the case of compacted wall. For uncompacted wall, x/H value increased from 3.7 to 9.5 for the same variations in reinforcement spacing. The normalized facing displacement seems to vary linearly with reinforcement spacing in the uncompacted wall, while there was an immediate significant increase in normalized facing displacement in compacted wall when the reinforcement spacing was increased further from 0.4 m to 0.6 m. There was a decrease in lateral deformation when spacing was reduced to 0.2 m and an increase of 68% for a spacing of 0.6 m when compaction induced stresses were taken into account. The stability of the system is primarily achieved through soil-reinforcement interaction consisting of friction and tension. Many research studies have shown that this interaction is limited to a certain depth [1]. Hence, the vertical spacing of the reinforcement plays a much prominent role in the load-carrying capacity of the wall system. This composite effect is achieved generally when the spacing is less than 0.4 m.

3.4 Effect of Reinforcement Strength A comparative study was done by varying the reinforcement stiffness from 600 to 1200 kN/m keeping other parameters the same in the compacted model. The MSE wall with weaker reinforcement of stiffness 600 kN/m had shown an increase in x/H values from 10.6 to 14.6 during the post construction. At the same time, the wall with reinforcement stiffness of 1200 kN/m had x/H in the range of 8–10.7 (Fig. 8). The

Lateral deformation/Height (x10-3)

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16 14 12 10 8

Reinforcement strength

600 kN/m

6

1200 kN/m 4

0

10

20

30

40

50

60

Surcharge(kPa) Fig. 8 Normalized facing displacement versus surcharge for two different reinforcement strength

rate of increase in x/H during the post construction is significantly reduced in the wall having higher stiffness. A decrease of 22% in lateral deformation at the end of construction was observed when the reinforcement strength was doubled. Figure 9 showed that there was a considerable reduction in post construction lateral displacement at various surcharges for compacted wall compared with the uncompacted case. A 37% decrease in post construction lateral displacement was observed at a surcharge of 50 kPa. This reduction in post construction lateral deformation tends to increase with increasing surcharge.

Post construction lateral deformation (mm)

30 25 20

Reinforcement strength 600 kN/m 1200 kN/m

15 10 5 0 10

20

30

40

50

Surcharge (kPa) Fig. 9 Post construction lateral deformation for different reinforcement strengths at various surcharges

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4 Summary and Conclusions A series of parametric studies were performed to analyze the influence of different reinforcement parameters and also to examine the effect of compaction induced stresses in the performance of MSE walls by simulating a series of various compaction efforts. The following conclusions were drawn from the study: • Modelling of compaction induced stresses was done, and the deformations were determined corresponding to various compaction efforts. Higher the compaction effort; higher is the lateral deformation at the end of construction. However, the post construction deformation tends to decrease with greater compaction efforts. This indicates that compaction induced stresses provide a prestressing effect to MSE wall system. • Lateral deformation tends to vary linearly with an increase in reinforcement spacing in the case without compaction, however, there is a significant increase of 68% when spacing was increased from 0.4 to 0.6 m in compacted model. This can be due to the loss of effective soil-reinforcement interaction existing in the soil composite. • A decrease of 22% in lateral deformation at the end of construction was observed when the reinforcement strength was doubled and a considerable reduction in the rate of increase in lateral deformations in post construction phase was noticed.

References 1. Wu JTH (2019) Characteristics of Geosynthetic Reinforced Soil (GRS) walls: an overview of field-scale experiments and analytical studies. Transp Infrastruct Geotechnol 6(2):138–163. https://doi.org/10.1007/s40515-019-00074-x 2. Wu JTH, Pham TQ (2010) An analytical model for evaluation of compaction-induced stresses in a reinforced soil mass. Int J Geotech Eng 4(4):549–556. https://doi.org/10.3328/IJGE.2010. 04.04.549-556 3. Guler E, Hamderi M, Demirkan MM (2007) Numerical analysis of reinforced soil-retaining wall structures with cohesive and granular backfills. Geosynth Int 14(6):330–345. https://doi.org/10. 1680/gein.2007.14.6.330 4. Sukmak K, Han J, Sukmak P, Horpibulsuk S (2016) Numerical parametric study on behavior of bearing reinforcement earth walls with different backfill material properties. Geosynth Int 23(6):435–451. https://doi.org/10.1680/jgein.16.00008 5. Berg R, Christopher B, Samtani N (2009) Design and construction of mechanically stabilized earth walls and reinforced soil slopes. In: Fed. High W. Adm., vol. I, November 2009. FHWANHI-10-024 & FHWA-NHI-10-025 6. Mirmoradi SH, Ehrlich M (2014) Modeling of the compaction-induced stresses in numerical analyses of GRS walls. Int J Comput Methods 11(2):1–14. https://doi.org/10.1142/S02198762 13420024

Evaluation of Moisture Susceptibility of Pyro-Oil Modified Bitumen by Surface Free Energy Approach Shubham D. Suryawanshi, Hemantkumar P. Hadole, and M. S. Ranadive

Abstract Municipal solid waste (MSW) is one of the vital environmental problems of Indian cities. Improper management of municipal solid waste (MSW) causes hazards to inhabitants. Various studies reveal that about 90–95% of MSW is disposed of unscientifically in open dumps and landfills, creating problems to public health and the environment. The surface free energy (SFE) is the essential property of materials. SFE parameters can be used to explain moisture damage, rutting, fracture, healing, etc. problems of bituminous mixes. In this paper, a comparison of surface free energy for VG30 (viscosity grade) and pyro-oil modified bitumen (POMB) was done in terms of adhesion and cohesion. The basic test like viscosity test, flash point test, penetration test, and softening point test was done on the VG 30, POMB and combination of both materials. For this research, the sessile drop method was used to calculate the surface free energy of VG 30 and pyro-oil modified bitumen. The surface free energy is used to evaluate cohesion bond energy of the material and adhesion bond energy between two conditions. Pyrolysis is one of the latest technologies and the best option for disposal. The bio oil generated from the pyrolysis process is proposed to use as a substitute to asphalt binder. The pyrolysis is one of the promising technology for the total disposal of waste. The pyro-oil (pyrolysis of plastic) derived from the pyrolysis of municipal solid waste can be best utilized as a bitumen modifier or bitumen extender in flexible pavement. The pyro-oil obtained by pyrolysis is of high-density polyethylene (HDPE) at about 600 °C. Pyro-oil modified bitumen was prepared with 1, 3 and 5% of HDPE by total weight of bitumen and mixed with 3000 rpm at 120 min by using a homogenizer instrument. The Pyrolysis of HDPE was done at 600 °C using a pilot pyrolysis plant developed in the transportation laboratory of College of Engineering, Pune, Maharashtra. Both materials were aged by short term ageing using Rolling thin film oven (RTFO). The comparison between S. D. Suryawanshi (B) · H. P. Hadole · M. S. Ranadive Department of Civil Engineering, College of Engineering, Pune, India e-mail: [email protected] H. P. Hadole e-mail: [email protected] M. S. Ranadive e-mail: [email protected] © The Author(s), under exclusive license to Springer Nature Singapore Pte Ltd. 2021 T. G. Sitharam et al. (eds.), Ground Improvement Techniques, Lecture Notes in Civil Engineering 118, https://doi.org/10.1007/978-981-15-9988-0_33

363

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surface free energy without ageing and with ageing material. Here SFE measurements were reported to compute work of adhesion, work of debonding, work of cohesion, wettability, energy ratio. Keywords SFE · POMB · HDPE · VG 30

1 Introduction Moisture damage is one of the major failures of asphalt pavement along with rutting, cracking failures, etc. Moisture damage failure of asphalt pavement can be solved by surface free energy (SFE) parameter. Moisture damage is defined as bonding failure at asphalt binder-aggregate interface (adhesion failure) or loss of asphalt binders tensile strength (cohesive failure). There are some adhesion theories like mechanical adhesion theories, chemical reaction theories, surface free energy theory, surface structure theory, molecular orientation theory, electrostatic theory. But most of the researchers follow SFE theory to predict stripping in asphalt mixture. According to the surface physical chemistry theory, SFE can be defined as the energy separating liquid or solid to produce a new interface in a vacuum. If the separated material is not homogeneous and two new different surfaces are formed such energy is called adhesion. If the separated material is homogeneous, then the energy is called as cohesion. The surface free energy (SFE) of aggregates and different types of asphalt binders can be used to calculate the work of adhesion between aggregate and asphalt binder and work of cohesion of homogeneous material, i.e. asphalt binder. Good adhesion between aggregates and asphalt binder improves mixture properties and provide a more durable mixture structure. So that, the evaluation of surface free energy of aggregate and asphalt is very important. There are different methods to determine SFE of asphalt and aggregates, which are sessile drop method, wilhelmy plate method, pendant drop method, de Noug ring method, atomic force microscopy, etc. Among the above-mentioned method, sessile drop method and wilhelmy plate method was performed at room temperature and most of the researchers had used these methods to calculate SFE. The sessile drop method represents a simple and easy way to calculate SFE. Hence, in the present study, the sessile drop technique was used to measure contact angle between different probe liquids and aggregates or asphalt binders, from which SFE of asphalt binder and aggregates were calculated using the Owens–Wendt theory. From SFE, the work of adhesion between asphalt and aggregate was accordingly calculated. In this paper, firstly the SFE for unmodified asphalt binder (VG30) and pyrooil modified bitumen (modified bitumen) were determined using the contact angle (sessile drop) technique, and SFE components of gravel, sandstone, granite aggregates were obtained from the literature. Consequently, a wide range of binder type, percentage of pyro-oil and aggregate were covered in this study. Finally, work of

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365

adhesion, work of debonding, work of cohesion, wettability, energy ratio were estimated to mechanistically discuss the effect of asphalt binder, percentage of pyro-oil and aggregates on the moisture damage of asphalt mixes containing pyro-oil. The results presented herein are expected to be helpful in the evaluation of aggregates– asphalt binder-pyro-oil combination, during the material selection for asphalt mixes, in order to minimize the possibility of moisture damage in pavement.

2 Objectives This study aims to evaluate the effect of pyro-oil modified bitumen and aggregate type on the moisture damage of mixes using the SFE approach. The specific objectives of this study were to: • Determine the SFE component of unmodified bitumen (VG 30) with and without the addition of different amounts of pyro-oil (i.e. 0, 1, 3, 5%) using the sessile drop method. • Determine the SFE of base binder and modified binder with and without ageing, i.e. by short term ageing (RTFO). • Evaluation of the unmodified bitumen coating quality with and without the addition of pyro-oil on different types of aggregates using the wettability parameter (spreading coefficient). • Evaluation of moisture damage of mixes containing pyro-oil with different types of aggregate and asphalt binders based on the energy ratio parameter calculation based on work of adhesion, work of debonding and wettability.

3 Background on Surface Free Energy 3.1 Surface Free Energy Components For the calculation of SFE for bitumen and aggregates, several methods are proposed in the literature. In this paper, we have used the Van Oss-Chaudhury-Good method (acid-base) [1]. In this approach, the SFE is calculated using γ L (1 + cos θ) = 2

√ L W L W  √ b a  √ a b  + γL γS + γL γS γL γS

(1)

a Since the Van Oss-Chaudhury-Good method has three unknowns, γ LW S , γ S and three contact angle measurements with three liquids are taken: one dispersive liquid (Formamide), and two polar liquids (water and glycerol). The SFE of probe liquid is given by Table 1.

γ bS ,

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Table 1 SFE components of probe liquids Liquid

SFE component γ

γ LW

γ +

γ−

γ AB

Water

72.6

21.6

25.5

25.5

51

Glycerol

64

34

3.6

57.4

30

Formamide

58

39

2.28

39.6

10

The SFE of a material is defined as the amount of external work required to create a unit area of a new surface in a vacuum [2]. The SFE of any material is divided into three components as follows: Lifshitz-van der waal component (non-polar components) (γ LW ), Lewis acid component (γ +) and Lewis base component (γ −) both components combine called as polar component (γ AB ). These components are used to calculate the total SFE (γ T ) as per Eqs. (2) and (3) [2].      T γ = γ LW + γ AB γ AB = 2

√

γ +γ −



(2) (3)

3.2 Performance Parameter Evaluated Using SFE Work of cohesion: (W cohesion ) The bonding within the bitumen is known as cohesive bonding. The work of cohesion is twice the total SFE of bitumen calculated using Eq. (4). Wcohesion = 2γ T

(4)

where γ T = total SFE of bitumen Work of Adhesion in dry condition: (W AB ) The energy required to create two interfaces from one interface, consisting of two different phases in contact is defined as the work of adhesion between an asphalt binder and aggregate can be determined from Eq. (5) W AB = 2

√ LW LW  √ + −  √ + −  + γL γ S + γL γ S γL γ S

(5)

where suffix “L” is used for probe liquids and suffix “S” is used for unknown (i.e. asphalt binder or aggregate).

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Work of Adhesion in wet condition (work of debonding): (W wet ) ABW When moisture contact with aggregate and asphalt binder (i.e. presence of water), there is a reduction in the energy, such energy is called as wet adhesion energy or work of adhesion in wet condition. wet = γ AW + γ BW − γ AB W ABW

(6)

where γ BW , γ AW , γ AB is the interfacial energy between asphalt binder and water, aggregate–water and aggregate–asphalt binder, respectively. Using SFE, the determination of interfacial energy between any two material “m” and “n” Γmn = γm + γn + 2

√ L W L W  √ + −  √ + −  γm γ n − γm γ n − γm γ n

(7)

Wettability: The spreading coefficient (S A/S ) is a significant index of the wettability of the binder over the aggregate. The wettability is defined as the reduction in the SFE of solid surface and forming a new liquid–solid and vapour–liquid interface. The spreading coefficient is given in Eq. (8). S A/S = γ S − γ AS − γ A

(8)

where, S A/S = spreading coefficient of asphalt binder over the aggregate, γ S = total SFE of aggregate, γ AS = interfacial energy between aggregates and asphalt binders and γ A = total SFE of asphalt binder. Energy ratio (ER): From the above performance parameter was used to evaluation of energy ratio. There are two types of energy ratios ER1 and ER2, respectively. ER1 is the ratio work of adhesion to work of debonding. As given in Eq. (9) wet ER1 = W AB ÷ W ABW

(9)

ER2 is the ratio of wettability to work of debonding given by Eq. (10). Work of adhesion and wettability should be more so that the energy ratio will be increased. High energy ratio shows more resistance to moisture damage [3]. wet ER2 = S A/S ÷ W ABW

(10)

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4 Materials 4.1 Pyro-Oil In this study, VG30 was used as base bitumen for research work. HDPE plastic waste was used for pyrolysis to get Pyro-oil for modification of base bitumen. HDPE plastic waste was collected from MSW of Pune City, Maharashtra, India. Pilot Pyrolysis plant was developed in transportation laboratory, College of Engineering, Pune, Maharashtra, India and used for pyrolysis at about 750 °C. Pyro-oil production percentage was shown in Table 2. For the modification of base binder, blending of bitumen with pyro-oil was performed at temperature of about 160–170 °C at about 5000 rpm for first 15 min and then at a reduced speed of 3000 rpm for last 5 min using a high shear mixer of IKA instruments. 1, 3 and 5% of HDPE pyro-oil by weight of binder was used to modify virgin bitumen and abbreviated as POMB1, POMB3 and POMB5, respectively. Here, the type of aggregates selected for SFE calculation were basalt, Gravel and limestone. The SFE components of these aggregates were taken from the literature [4].

4.1.1

Test Methods

Virgin Binder (VG30) and HDPE pyro-oil modified binder samples were checked for various physical tests like penetration, viscosity and ductility and results are tabulated in Tables 3 and 4. Table 2 Output of HDPE pyro-oil Sr. No.

Types of waste

Quantity of waste (kg) Quantity of pyro-oil (%)

Reactor temperature (°C)

1

HDPE

3

720

69.6

Table 3 Properties of VG30 and POMB Test

VG30

POMB1

POMB3

POMB5

Penetration at 25 °C (1/10 mm)

59

79

83

101

Softening Point (°C)

48

54

48

32

Ductility (cm)

78

88

90

104

Viscosity at 60 °C (poise)

2856

2110

1540

1047

Kinematic viscosity at 135 °C (cSt)

345

264

248

175

Viscosity at 150 °C (poise)

355

635

464

295

Loss in mass (%)

72 h

Sample discarded

12

01:02 PM

>5 h

>72 h

Sample discarded

setting time of geopolymer grout. For the compressive strength of geopolymer grout, the mortar samples were prepared as per the IS: 2250-1981. Table 3 shows the details of material proportion and compressive strength of geopolymer mortar samples prepared with zone 2 uniform sand. Here the geopolymer mortar was prepared with alkaline solution/binder ratio as 1, and NaOH to Na2 Sio3 ratio was taken as 2 with different molarity 8 M, 10 M, and 12 M.

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Fig. 1 Mixing of material for geopolymer grout

Fig. 2 Initial setting of geopolymer grout material

4 Results and Discussion The Geopolymer grout material was prepared from the fly ash as a binder material and a mixture of NaOH and Na2 SiO3 as an alkaline activator. Figure 4 shows the composition of the samples. Figure 5 shows the setting time of Geopolymer Grout samples. The Fig. 5, shows that from sample 1 to sample 9, the setting time increases. While samples 10–12, were not formed properly and the setting time was not reported. The increase in setting time may depend on various factors like molarity, alkaline solution to binder ratio, NaOH to Na2 SiO3 ratio. Figure 6 indicates the effect of NaOH: Na2 SiO3 ratio on the setting time of geopolymer grout material. The trend shows that an increase in the ratio decreases the setting time of geopolymer grout material. This response may be due to an increase in alkaline solution, reacts with a silica content of fly ash (binder), and forms the gel network because of geopolymerisation process with less time. Figure 7 indicates

Feasibility of Geopolymer Grout for Granular Soil

393

Table 3 Compressive strength of geopolymer mortar sample Alkaline solution to binder ratio

Molarity

Materials

Quantity (gm)

8M

NaOH Solution

265

Sodium Silicate

109

Fly Ash

374

1

10 M

12 M

Sand

2242

NaOH Solution

253

Sodium Silicate

125

Fly Ash

378

Sand

2267

NaOH Solution

242

Sodium Silicate

139

Fly Ash

381

Sand

2289

Test results (MPa)

Sample 1

Sample 2

Sample 3

Average

6.2

6.7

7.8

6.9

2.8

3.5

3.1

3.13

2.9

3.1

2.5

2.83

Fig. 3 Final Setting of geopolymer grout material

the effect of the alkaline solution to fly ash ratio on setting time. It shows that as the ratio increases, the setting time also increases. It may be due to the increase in the amount of solution that reacts slowly with a binder and which leads to an increase in the setting time of geopolymer grout. Figure 8 indicates the compressive strength of Geopolymer mortar made with uniform sand. The results show that the compressive strength after 28 days with ambient curing higher with 10 M solution geopolymer grout design. However, literature shows that the geopolymer mortar sample gave strength more than the reported one.

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200 Weight (grms)

NaOH

Na2SIO3

Flyash

150 100 50 0

1

2

3

4

5

6 7 Samples

8

9

10

11

12

Samples

Fig. 4 Composition of geopolymer grout material

12 11 10 9 8 7 6 5 4 3 2 1 0

50

100

150

200

250

300

350

Final Setting time(Minutes) Fig. 5 Final Setting time of geopolymer grout

Setting Time(Minutes)

350 300 250 200 150 100 50 0 0

0.5

1

1.5

2

NaOH:NA2SIO3 ratio Fig. 6 Effect of NaOH: Na2 SiO3 ratio on setting time

2.5

3

Feasibility of Geopolymer Grout for Granular Soil

395

Setting time(Minutes)

350 300 250 200 150 100 50 0 0

1

2

3

4

5

Alkaline solution to flyash(binder) ratio

Compressive strength (Mpa)

Fig. 7 Effect of alkaline to fly ash ratio on setting time 8 7 6 5 4 3 2 1 0 8M

10M

12M

Molarity(M) Fig. 8 Compressive strength of geopolymer mortar

5 Conclusion From the feasibility study of geopolymer grout material and available literature, the following conclusions are derived. • Using the industrial waste material like fly ash, the geopolymer grout material is feasible to design. • The Geopolymer grout material was feasible but there are number of trials required to form the optimum design of geopolymer grout material considering other design parameters. • The setting time is a very important control parameter because when such grout material is applied to the field, the control on setting time is required as per the type of application. • The results of setting time show the range between 40 and 298 min.

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• The results of compressive strength of geopolymer mortar with uniform sand were 6.9 MPa after 28 days with ambient curing for 8 M of alkaline solution, alkaline solution/binder ration = 1 and NaOH/NA2 SiO3 ratio = 2. However, there are number of trials required for the compressive strength of geopolymer mortar. Acknowledgements The research work was conducted under the Minor Research Project funded by Nirma University. Authors would like to express their gratitude to the Civil Engineering Department, Nirma University for providing the research facilities. Authors also thank the students who helped while conducting the research activities.

References 1. Skvara F, Jilek T, Kopecky L (2005) Geopolymer materials based on fly ash Ceram.-Silik 49(3):195–204 2. Mirza J, Saleh K, Roy V, Mirza MS (1997) Use of high volume fly ash in grouting applications. American Concrete Institute 281–298 3. Wardhono A, Law DW, Strano A (2015) The strength of alkali-activated slag/fly ash mortar blends at ambient temperature. Proc Eng (Science Direct) 125:650–656. The 5th International Conference of Euro Asia Civil Engineering Forum (EACEF-5) 4. Khayat KH, Yahia A, Sayed M (2008) Effect of supplementary cementitious materials on rheological properties, bleeding, and strength of structural grout. ACI Mater J 5. Trimurtiningrum R, Ekaputri J (2016) Geopolymer grout material. Mater Sci Forum 841:40–47. https://doi.org/10.4028/www.scientific.net/MSF.841.40 6. Subhashree S, Singh SP, Palsule P (2017) Synthesis and characterization of geo-polymer grout material Indian geotechnical conference, GeoNEst, IIT Guwahati, India 7. Subektil S, Bayuaji R, Darmawan MS, Husin NA, Wibowo B, Anugraha B, Irawan S, Dibiantara D (2017) Review: potential strength of fly ash-based geopolymer paste with substitution of local waste materials with high-temperature effect. In: IOP conference series: materials science and engineering 267. https://doi.org/10.1088/1757-899x/267/1/012001 8. Departmental Material Specification 4675 (2014) Cementitious grouts and mortars for miscellaneous applications. Texas Department of Transportation