Landslides and Engineered Slopes. From the Past to the Future: Proceedings of the 10th International Symposium on Landslides and Engineered Slopes, 30 June - 4 July 2008, Xi'an, China [1 ed.] 0415411963, 9780415411967

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LANDSLIDES AND ENGINEERED SLOPES

PROCEEDINGS OF THE TENTH INTERNATIONAL SYMPOSIUM ON LANDSLIDES AND ENGINEERED SLOPES, 30 JUNE–4 JULY, 2008, XI’AN, CHINA

Landslides and Engineered Slopes From the Past to the Future

Editors

Zuyu Chen China Institute of Water Resources and Hydropower Research, Beijing, China

Jianmin Zhang Department of Hydraulic Engineering, Tsinghua University, Beijing, China

Zhongkui Li Department of Hydraulic Engineering, Tsinghua University, Beijing, China

Faquan Wu Institute of Geology and Geophysics, Chinese Academy of Sciences, Beijing, China

Ken Ho Civil Engineering and Development Department, Hong Kong SAR, China

VOLUME 1

CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business © 2008 Taylor & Francis Group, London, UK Typeset by Vikatan Publishing Solutions (P) Ltd., Chennai, India Printed and bound in Great Britain by Antony Rowe (A CPI-group Company), Chippenham, Wiltshire. All rights reserved. No part of this publication or the information contained herein may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, by photocopying, recording or otherwise, without written prior permission from the publisher. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. Published by: CRC Press/Balkema P.O. Box 447, 2300 AK Leiden, The Netherlands e-mail: [email protected] www.crcpress.com – www.taylorandfrancis.co.uk – www.balkema.nl ISBN set: 978-0-415-41196-7 (2 volumes + CD) ISBN Vol.1: 978-0-415-41194-3 (hbk) ISBN Vol.2: 978-0-415-41195-0 (hbk)

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Table of Contents

Preface

XXIII

VOLUME 1 Keynote lectures Landslides: Seeing the ground N.R. Morgenstern & C.D. Martin

3

Limit equilibrium and finite element analysis – A perspective of recent advances Z. Chen & K. Ugai Improving the interpretation of slope monitoring and early warning data through better understanding of complex deep-seated landslide failure mechanisms E. Eberhardt, A.D. Watson & S. Loew

25

39

Effects of earthquakes on slopes I. Towhata, T. Shimomura & M. Mizuhashi

53

Monitoring and modeling of slope response to climate changes H. Rahardjo, R.B. Rezaur, E.C. Leong, E.E. Alonso, A. Lloret & A. Gens

67

Soil nailing and subsurface drainage for slope stabilization W.K. Pun & G. Urciuoli

85

Special lectures Loess in China and landslides in loess slopes Z.G. Lin, Z.J. Xu & M.S. Zhang

129

Advances in landslide continuum dynamic modelling S. McDougall, M. Pirulli, O. Hungr & C. Scavia

145

Deformation and failure mechanisms of loose and dense fill slopes with and without soil nails C.W.W. Ng

159

Capturing landslide dynamics and hydrologic triggers using near-real-time monitoring M.E. Reid, R.L. Baum, R.G. LaHusen & W.L. Ellis

179

The effects of earthquake on landslides – A case study of Chi-Chi earthquake, 1999 M.L. Lin, K.L. Wang & T.C. Kao

193

The role of suction and its changes on stability of steep slopes in unsaturated granular soils L. Olivares & P. Tommasi

203

Prediction of landslide movements caused by climate change: Modelling the behaviour of a mean elevation large slide in the Alps and assessing its uncertainties Ch. Bonnard, L. Tacher & M. Beniston

217

Geology, geotechnical properties and site characterization Geotechnical appraisal of the Sonapur landslide area, Jainita hills, Meghalya, India R.C. Bhandari, P. Srinivasa Gopalan & V.V.R.S. Krishna Murty

V

231

The viscous component in slow moving landslides: A practical case D.A. González, A. Ledesma & J. Corominas

237

The systematic landslide investigation programme in Hong Kong K.K.S. Ho & T.M.F. Lau

243

General digital camera-based experiments for large-scale landslide physical model measurement X.W. Hu, H.M. Tang & J.S. Li

249

Shear strength of boundaries between soils and rocks in Korea S.G. Lee, B.S. Kim & S.H. Jung

257

Cracks in saturated sand X.B. Lu, S.Y. Wang & P. Cui

263

Some geomorphological techniques used in constraining the likelihood of landsliding – Selected Australian examples A.S. Miner, P. Flentje, C. Mazengarb, J.M. Selkirk-Bell & P.G. Dahlhaus

267

Rock failures in karst M. Parise

275

Geotechnical study at Sirwani landslide site, India V.K. Singh

281

Inferences from morphological differences in deposits of similar large rockslides A.L. Strom

285

Movements of a large urban slope in the town of Santa Cruz do Sul (RGS), Brazil L.A. Bressani, R.J.B. Pinheiro, A.V.D. Bica, C.N. Eisenberger & J.M.D. Soares

293

Geotechnical analysis of a complex slope movement in sedimentary successions of the southern Apennines (Molise, Italy) D. Calcaterra, D. Di Martire, M. Ramondini, F. Calò & M. Parise

299

Application of surface wave and micro-tremor survey in landslide investigation in the Three Gorges reservoir area A. Che, X. Luo, S. Feng & O. Yoshiya

307

A case study for the landslide-induced catastrophic hazards in Taiwan Tuchang Tribute C.Y. Chen & W.C. Lee

313

Pir3D, an easy to use three dimensional block fall simulator Y. Cottaz & R.M. Faure

319

Characterization of the fracture pattern on cliff sites combining geophysical imaging and laser scanning J. Deparis, D. Jongmans, B. Fricout, T. Villemin, O. Meric, A. Mathy & L. Effendiantz

323

In situ characterization of the geomechanical properties of an unstable fractured rock slope C. Dünner, P. Bigarré, F. Cappa, Y. Guglielmi & C. Clément

331

Properties of peat relating to instability of blanket bogs A.P. Dykes

339

Stability problems in slopes of Arenós reservoir (Castellón, Spain) J. Estaire, J.A. Díez & C. Olalla

347

The 22 August, 2006, anomalous rock fall along the Gran Sasso NE wall (Central Apennines, Italy) G.B. Fasani, C. Esposito, G.S. Mugnozza, L. Stedile & M. Pecci

VI

355

New formulae to assess soil permeability through laboratory identification and flow coming out of vertical drains J.C. Gress

361

Structure-controlled earth flows in the Campania Apennines (Southern Italy) F.M. Guadagno, P. Revellino, G. Grelle, G. Lupo & M. Bencardino

365

Geotechnical and mineralogical characterization of fine grained soils affected by soil slips G. Gullà, L. Aceto, S. Critelli & F. Perri

373

Vulnerability of structures impacted by debris flow E.D. Haugen & A.M. Kaynia

381

Engineering geological study on a large-scale toppling deformation at Xiaowan Hydropower Station R. Huang, G. Yang, M. Yan & M. Liu Characterization of the Avignonet landslide (French Alps) with seismic techniques D. Jongmans, F. Renalier, U. Kniess, S. Schwartz, E. Pathier, Y. Orengo, G. Bièvre, T. Villemin & C. Delacourt Deformation characteristics and treatment measures of spillway slope at a reservoir in China N. Ju, J. Zhao & R. Huang Sliding in weathered banded gneiss due to gullying in southern Brazil W.A. Lacerda, A.P. Fonseca & A.L. Coelho Netto

389 395

403 409

Experimental and three-dimensional numerical investigations of the impact of dry granular flow on a barrier R.P.H. Law, G.D. Zhou, C.W.W. Ng & W.H. Tang

415

Temporal survey of fluids by 2D electrical tomography: The ‘‘Vence’’ landslide observatory site (Alpes-Maritimes, SE France) T. Lebourg, S. El Bedoui, M. Hernandez & H. Jomard

421

Characteristics of landslides related to various rock types in Korea S.G. Lee, K.S. Lee, D.C. Park & S. Hencher

427

Two approaches to identifying the slip zones of loess landslides and related issues T. Li & X. Lin

435

Testing study on the strength and deformation characteristics of soil in loess landslides H.J. Liao, L.J. Su, Z.D. Li, Y.B. Pan & H. Fukuoka

443

Failure mechanism of slipping zone soil of the Qiangjiangping landslide in the Three Gorges reservoir area: A study based on Dead Load test X. Luo, A. Che, L. Cao & Y. Lang

449

Post-failure movements of a large slow rock slide in schist near Pos Selim, Malaysia A.W. Malone, A. Hansen, S.R. Hencher & C.J.N. Fletcher

457

Characteristics of rock failure in metamorphic rock areas, Korea W. Park, Y. Han, S. Jeon & B. Roh

463

Shape and size effects of gravel grains on the shear behavior of sandy soils S.N. Salimi, V. Yazdanjou & A. Hamidi

469

Nonlinear failure envelope of a nonplastic compacted silty-sand D.D.B. Seely & A.C. Trandafir

475

An investigation of a structurally-controlled rock cut instability at a metro station shaft in Esfahan, Iran A. Taheri

VII

481

Yield acceleration of soil slopes with nonlinear strength envelope A.C. Trandafir & M.E. Popescu

487

Evaluation of rockfall hazards along part of Karaj-Chaloos road, Iran A. Uromeihy, N. Ghazipoor & I. Entezam

491

Coupled effect of pluviometric regime and soil properties on hydraulic boundary conditions and on slope stability R. Vassallo, C. Di Maio & M. Calvello

495

Mechanical characters of relaxing zone of slopes due to excavation H. Wang & X.P. Liao

501

Deformation characteristics and stability evaluation of Ganhaizi landslide in the Dadu River Y. Wang, Y. Sun, O. Su, Y. Luo, J. Zhang, C. Zhou & S. Zhang

507

Landslide-prone towns in Daunia (Italy): PS interferometry-based investigation J. Wasowski, D. Casarano, F. Bovenga, A. Refice, R. Nutricato & D.O. Nitti

513

Basic types and active characteristics of loess landslide in China W. Wu, D. Wang, X. Su & N. Wang

519

Investigation of a landslide using borehole shear test and ring shear test H. Yang, V.R. Schaefer & D.J. White

525

The importance of geological and geotechnical investigations of landslides occurred at dam reservoirs: Case studies from the Havuzlu and Demirkent landslides (Artvin Dam, Turkey) A.B. Yener, S. Durmaz & B.M. Demir

531

An innovative approach combining geological mapping and drilling process monitoring for quantitative assessment of natural terrain hazards Z.Q. Yue, J. Chen & W. Gao

535

Types of cutslope failures along Shiyan-Manchuanguan expressway through the Liangyun fracture, Hubei Province H. Zhao, R. Wang, J. Fan & W. Lin

543

Advances in analytical methods, modeling and prediction of slope behavior Probability limit equilibrium and distinct element modeling of jointed rock slope at northern abutment of Gotvand dam, Iran M. Aminpoor, A. Noorzad & A.R. Mahboubi

553

Rock block sliding analysis of a highway slope in Portugal P.G.C. Santarém Andrade & A.L. Almeida Saraiva

561

Contribution to the safety evaluation of slopes using long term observation results J. Barradas

567

Delimitation of safety zones by finite element analysis J. Bojorque, G. De Roeck & J. Maertens

573

Laboratory and numerical modelling of the lateral spreading process involving the Orvieto hill (Italy) F. Bozzano, S. Martino, A. Prestininzi & A. Bretschneider

579

Albano Lake coastal rock slide (Roma, Italy): Geological constraints and numerical modelling F. Bozzano, C. Esposito, S. Martino, P. Mazzanti & G. Diano

585

Superposition principle for stability analysis of reinforced slopes and its FE validation F. Cai & K. Ugai

593

Soil suction modelling in weathered gneiss affected by landsliding M. Calvello, L. Cascini, G. Sorbino & G. Gullà

599

VIII

Modelling the transient groundwater regime for the displacements analysis of slow-moving active landslides L. Cascini, M. Calvello & G.M. Grimaldi

607

Numerical modelling of the thermo-mechanical behaviour of soils in catastrophic landslides F. Cecinato, A. Zervos, E. Veveakis & I. Vardoulakis

615

Some notes on the upper-bound and Sarma’s methods with inclined slices for stability analysis Z.Y. Chen

623

Slope stability analysis using graphic acquisitions and spreadsheets L.H. Chen, Z.Y. Chen & P. Sun

631

Efficient evaluation of slope stability reliability subject to soil parameter uncertainties using importance sampling J. Ching, K.K. Phoon & Y.G. Hu

639

Prediction of the flow-like movements of Tessina landslide by SPH model S. Cola, N. Calabrò & M. Pastor

647

Applications of the strength reduction finite element method to a gravity dam stability analysis Q.W. Duan, Z.Y. Chen, Y.J. Wang, J. Yang & Y. Shao

655

Study on deformation parameter reduction technique for the strength reduction finite element method Q.W. Duan, Y.J. Wang & P.W. Zhang Stability and movement analyses of slopes using Generalized Limit Equilibrium Method M. Enoki & B.X. Luong Long-term deformation prediction of Tianhuangpin ‘‘3.29’’ landslide based on neural network with annealing simulation method F. Zhang, C. Xian, J. Song, B. Guo & Z. Kuai

663 671

679

New models linking piezometric levels and displacements in a landslide R.M. Faure, S. Burlon, J.C. Gress & F. Rojat

687

3D slope stability analysis of Rockfill dam in U-shape valley X.Y. Feng, M.T. Luan & Z.P. Xu

693

3-D finite element analysis of landslide prevention piles K. Fujisawa, M. Tohei, Y. Ishii, Y. Nakashima & S. Kuraoka

697

Integrated intelligent method for displacement predication of landslide W. Gao

705

A new approach to in situ characterization of rock slope discontinuities: the ‘‘High-Pulse Poroelasticity Protocol’’ (HPPP) Y. Guglielmi, F. Cappa, S. Gaffet, T. Monfret, J. Virieux, J. Rutqvist & C.F. Tsang Fuzzy prediction and analysis of landslides Y. He, B. Liu, W.J. Liu, F.Q. Liu & Y.J. Luan

711 719

LPC methodology as a tool to create real time cartography of the gravitational hazard: Application in the municipality of Menton (Maritimes Alps, France) M. Hernandez, T. Lebourg, E. Tric, M. Hernandez & V. Risser

725

Back-analyses of a large-scale slope model failure caused by a sudden drawdown of water level G.W. Jia, T.L.T. Zhan & Y.M. Chen

731

Effect of Guangxi Longtan reservoir on the stability of landslide at Badu station of Nankun railway R. Jiang, R. Meng, A. Bai & Y. He

737

Application of SSRM in stability analysis of subgrade embankments over sloped weak ground with FLAC3D X. Jiang, Y. Qiu, Y. Wei & J. Ling

IX

741

Strength parameters from back analysis of slips in two-layer slopes J.-C. Jiang & T. Yamagami Development characteristics and mechanism of the Lianhua Temple landslide, Huaxian county, China J.-Y. Wang, M.-S. Zhang, C.-Y. Sun & Z. Rui

747

755

Modeling landslide triggering in layered soils R. Keersmaekers, J. Maertens, D. Van Gemert & K. Haelterman

761

Numerical modeling of debris flow kinematics using discrete element method combined with GIS H. Lan, C.D. Martin & C.H. Zhou

769

Three dimensional simulation of landslide motion and the determination of geotechnical parameters Y. Lang, X. Luo & H. Nakamura

777

Stability analysis and stabilized works of dip bedded rock slopes J.Y. Leng, Z.D. Jing & X.P. Liao

783

A GIS-supported logistic regression model applied in regional slope stability evaluation X. Li, H. Tang & S. Chen

789

The stability analysis for FaNai landslide in Lubuge hydropower station K. Li, J. Zhang, S. Zhang & S. He

795

Numerical analysis of slope stability influenced by varying water conditions in the reservoir area of the Three Gorges, China S. Li, X. Feng & J.A. Knappett

803

A numerical study of interaction between rock bolt and rock mass X.P. Li & S.M. He

809

Macroscopic effects of rock slopes before and after grouting of joint planes H. Lin, P. Cao, J.T. Li & X.L. Jiang

815

Two- and three-dimensional analysis of a fossil landslide with FLAC X.L. Liu & J.H. Deng

821

Application of the coupled thin-layer element in forecasting the behaviors of landslide with weak intercalated layers Y.L. Luo & H. Peng

827

Numerical modelling of a rock avalanche laboratory experiment in the framework of the ‘‘Rockslidetec’’ alpine project I. Manzella, M. Pirulli, M. Naaim, J.F. Serratrice & V. Labiouse

835

Three-dimensional slope stability analysis by means of limit equilibrium method S. Morimasa & K. Miura

843

Embankment basal stability analysis using shear strength reduction finite element method A. Nakamura, F. Cai & K. Ugai

851

Back analysis based on SOM-RST system H. Owladeghaffari & H. Aghababaei

857

Temporal prediction in landslides – Understanding the Saito effect D.N. Petley, D.J. Petley & R.J. Allison

865

3D landslide run out modelling using the Particle Flow Code PFC3D R. Poisel & A. Preh

873

Double-row anti-sliding piles: Analysis based on a spatial framework structure T. Qian & H. Tang

881

X

Centrifuge modeling of rainfall-induced failure process of soil slope J.Y. Qian, A.X. Wang, G. Zhang & J.-M. Zhang

887

A GIS-based method for predicting the location, magnitude and occurrence time of landslides using a three-dimensional deterministic model C. Qiu, T. Esaki, Y. Mitani & M. Xie

893

Application of a rockfall hazard rating system in rock slope cuts along a mountain road of South Western Saudi Arabia B.H. Sadagah

901

Model tests of collapse of unsaturated slopes in rainfall N. Sakai & S. Sakajo

907

Calibration of a rheological model for debris flow hazard mitigation in the Campania region A. Scotto di Santolo & A. Evangelista

913

Optical fiber sensing technology used in landslide monitoring Y.X. Shi, Q. Zhang & X.W. Meng

921

Finite element analysis of flow failure of Tailings dam and embankments R. Singh, D. Mitra & D. Roy

927

Landslide model test to investigate the spreading range of debris according to rainfall intensity Y.S. Song, B.G. Chae, Y.C. Cho & Y.S. Seo

933

Occurrence mechanism of rockslide at the time of the Chuetsu earthquake in 2004 – A dynamic response analysis by using a simple cyclic loading model N. Tanaka, S. Abe, A. Wakai, H. Kawabata, M. Genda & H. Yoshimatsu Analysis for progressive failure of the Senise landslide based on Cosserat continuum model H.X. Tang Large-scale deformation of the La Clapière landslide and its numerical modelling (S.-E. de Tinée, France). E. Tric, T. Lebourg & H. Jomard

939 945

951

A novel complex valued neuron model for landslide assessment K. Tyagi, V. Jindal & V. Kumar

957

Prediction of slope behavior for deforming railway embankments V.V. Vinogradov, Y.K. Frolovsky, A. Al. Zaitsev & I.V. Ivanchenko

963

Finite element simulation for the collapse of a dip slope during 2004 Mid Niigata Prefecture earthquake in Japan A. Wakai, K. Ugai, A. Onoue, K. Higuchi & S. Kuroda

971

Sensitivity of stability parameters for soil slopes: An analysis based on the shear strength reduction method R. Wang, X.Z. Wang, Q.S. Meng & B. Hu

979

Back analysis of unsaturated parameters and numerical seepage simulation of the Shuping landslide in Three Gorges reservoir area S. Wang, H. Zhang, Y. Zhang & J. Zheng

985

Slope failure criterion: A modification based on strength reduction technique Y.G. Wang, R. Jing, W.Z. Ren & Z.C. Wang

991

Unsaturated seepage analysis for a reservoir landslide during impounding J.B. Wei, J.H. Deng, L.G. Tham & C.F. Lee

999

A simple compaction control method for slope construction L.D. Wesley

XI

1005

Numerical analysis of soil-arch effect of anti-slide piles Y. Xia, X. Zheng & R. Rui

1011

Determination of the critical slip surface based on stress distributions from FEM D. Xiao, C. Wu & H. Yang

1017

Effect of drainage facilities using 3D seepage flow analysis reflecting hydro-geological structure with aspect cracks in a landslide – Example of analysis in OODAIRA Landslide area M. Yamada & K. Ugai

1023

Back analysis of soil parameters: A case study on monitored displacement of foundation pits B. Yan, X.T. Peng & X.S. Xu

1031

3D finite element analysis on progressive failure of slope due to rainfall G.L. Ye, F. Zhang & A. Yashima

1035

Block-group method for rock slope stability analysis Z. Zhang, Y. Xu & H. Wu

1043

Quantitative study on the classification of unloading zones of high slope D. Zheng & R.Q. Huang

1051

Investigations on the accuracy of the simplified Bishop method D.Y. Zhu

1055

Author index

1059

VOLUME 2 Landslide mechanism, monitoring and warning GIS-based landslide susceptibility mapping in the Three Gorges area – Comparisons of mapping results obtained by two methods: Analytical hierarchy process and logistic regression S. Bai, J. Wang, G. Lu, P. Zhou, S. Hou & F. Zhang

1067

Importance of study of creep sliding mechanism to prevention and treatment of reservoir landslide J. Bai, S. Lu, J. Han

1071

Stability prediction of landsides before and after impoundment for Lijiaxia hydropower station J. Bai, S. Lu & J. Han

1077

The technical concept within the Integrative Landslide Early Warning System (ILEWS) R. Bell, B. Thiebes, T. Glade, R. Becker, H. Kuhlmann, W. Schauerte, S. Burghaus, H. Krummel, M. Janik & H. Paulsen

1083

The Åknes rockslide: Monitoring, threshold values and early-warning L.H. Blikra

1089

DInSAR techniques for monitoring slow-moving landslides D. Calcaterra, M. Ramondini, F. Calò, V. Longobardi, M. Parise, C.M. Galzerano & C. Terranova

1095

Multitemporal DInSAR data and damages to facilities as indicators for the activity of slow-moving landslides L. Cascini, S. Ferlisi, D. Peduto, G. Pisciotta, S. Di Nocera & G. Fornaro

1103

The Serre La Voute Landslide (North-West Italy): Results from ten years of monitoring M. Ceccucci, G. Maranto & G. Mastroviti

1111

Onset of rockslide by the peak-residual strength drop Q.G. Cheng & G.T. Hu

1119

Analysis of mechanism of the K31 landslide of Changzhi-Jincheng express highway Y. Cheng

1127

XII

A plane-torsion rockslide with a locked flank: A case study Q. Cheng

1133

Monitoring of natural thermal strains using hollow cylinder strain cells: The case of a large rock slope prone to rockfalls C. Clément, Y. Gunzburger, V. Merrien-Soukatchoff & C. Dünner

1143

Landslide hazards mapping and permafrost slope InSAR monitoring, Mackenzie valley, Northwest Territories, Canada R. Couture & S. Riopel

1151

Advanced monitoring criteria for precocious alerting of rainfall-induced flowslides E. Damiano, L. Olivares, A. Minardo, R. Greco, L. Zeni & L. Picarelli

1157

Investigation of slope failure mechanisms caused by discontinuous large scale geological structures at the Cadia Hill Open Pit J. Franz & Y. Cai

1165

Two approaches for public landslide awareness in the United States – U.S. geological survey warning systems and a landslide film documentary L.M. Highland & P.L. Gori

1173

Formation and mechanical analysis of Tiantai landslide of Xuanhan county, Sichuan province R.Q. Huang, S. Zhao & X. Song

1177

Development of wireless sensor node for landslide detection H.W. Kim

1183

Redox condition and landslide development Y.H. Lang, S.Y. Liang & G.D. Zheng

1189

Prepa displacement mechanism and its treatment measures for Hancheng landslide T.F. Li & L.C. Dang

1195

Investigation of the stability of colluvial landslide deposits X. Li & L.M. Zhang

1205

Choice of surveying methods for landslides monitoring S.T. Liu & Z.W. Wang

1211

No. 1 landslide on the eastern approach road to ErLang Mountain tunnel: Inference factors and controlling measures H.M. Ma & Z.P. Zhang Estimation of landslide load on multi-tier pile constructions with the help of a combined method S.I. Matsiy & Ph.N. Derevenets

1217 1225

The use of PSInSAR™ data in landslide detection and monitoring: The example of the Piemonte region (Northern Italy) C. Meisina, F. Zucca, D. Notti, A. Colombo, A. Cucchi, G. Savio, C. Giannico & M. Bianchi

1233

Fill slopes: Stability assessment based on monitoring during both heavy rainfall and earthquake motion T. Mori, M. Kazama, R. Uzuoka & N. Sento

1241

The mechanism of movement of mud flows in loess soils, successful and unsuccessful cases of forecast R.A. Niyazov, Sh.B. Bazarov & A.M. Akhundjanov

1247

Influence of fine soil particles on excess of basal pore-fluid pressure generation in granular mass flows Y. Okada & H. Ochiai

1253

An early warning system to predict flowslides in pyroclastic deposits L. Pagano, G. Rianna, M.C. Zingariello, G. Urciuoli & F. Vinale

XIII

1259

Monitoring and modeling of slope movement on rock cliffs prior to failure N.J. Rosser & D.N. Petley

1265

Active tectonic control of a large landslide: Results from Panagopoula landslide multi parametric analyses S. El Bedoui, T. Lebourg & Y. Guglielmi

1273

A warning system using chemical sensors and telecommunication technologies to protect railroad operation from landslide disaster H. Sakai

1277

Distributive monitoring of the slope engineering B. Shi, H. Sui, D. Zhang, B. Wang, G. Wei & C. Piao

1283

Observational method in the design of high cutting slope around bridge S. Sun, B. Zhu, B. Zheng & J. Zhang

1289

Ultrasonic monitoring of lab-scaled underwater landslides Q.H. Truong, C. Lee, H.K. Yoon, Y.H. Eom, J.H. Kim & J.S. Lee

1297

Interaction between landslides and man-made works G. Urciuoli & L. Picarelli

1301

Desiccation fissuring induced failure mechanisms for clay levees S. Utili, M. Dyer, M. Redaelli & M. Zielinski

1309

Stability analysis by strength reduction finite element method and monitoring of unstable slope during reinforcement Z.Q. Wang, H.F. Li & L.M. Zhang

1315

Displacement monitoring on Shuping landslide in the Three Gorges Dam reservoir area, China from August 2004 to July 2007 F.W. Wang, G. Wang, Y.M. Zhang, Z.T. Huo, X.M. Peng, K. Araiba & A. Takeuchi

1321

Deformation mechanism and prevention measure for strongly expansive soft-rock slope in the Yanji basin X. Wu, N. Xu, H. Tian, Y. Sun & M. He

1329

Twenty years of safety monitoring for the landslide of Hancheng PowerStation M.J. Wu, Z.C. Li, P.J. Yuan & Y.H. Jiang

1335

A time-spatial deterministic approach to assessment of rainfall-induced shallow landslide M.W. Xie, C. Qiu & Z.F. Wang

1343

Introduction of web-based remote-monitoring system and its application to landslide disaster prevention M. Yamada & S. Tosa

1349

Deformation mechanism for the front slope of the left bank deposits in Xiluodu hydro-electrical power station, China M. Yan, Z. Wu, R. Huang, Y. Zhang & S. Wang

1355

Monitoring of soil nailed slopes and dams using innovative technologies J.-H. Yin, H.-H. Zhu & W. Jin

1361

Application of multi-antenna GPS technique in the stability monitoring of roadside slopes Q. Zhang, L. Wang, X.Y. Zhang, G.W. Huang, X.L. Ding, W.J. Dai & W.T. Yang

1367

Effects of earthquakes on slopes Influences of earthquake motion on slopes in a hilly area during the Mid-Niigata Prefecture Earthquake, 2004 S. Asano & H. Ochiai

XIV

1375

The 1783 Scilla rock-avalanche (Calabria, southern Italy) F. Bozzano, S. Martino, A. Prestininzi, M. Gaeta, P. Mazzanti & A. Montagna Self-excitation process due to local seismic amplification and earthquake-induced reactivations of large landslides F. Bozzano, S. Martino, G. Scarascia Mugnozza, A. Paciello & L. Lenti

1381

1389

Geological constraints to the urban shape evolution of Ariano Irpino (Avellino province, Italy) D. Calcaterra, C. Dima & E. Grasso

1397

Landslide zones and their relation with seismoactive fault systems in Azerbaijan, Iran E. Ghanbari

1405

Ground movements caused by lateral spread during an earthquake S.C. Hsu, B.L. Chu & C.C. Lin

1409

2-D analysis of slope stability of an infinite slope during earthquake J. Liu, J. Liu & J. Wang

1415

High-cutting slopes at Qingshuichuan electric power plant in the North of Shaanxi: Deformation and failure modes and treatment scheme H. Liu, Z. Liu & Z. Yan

1421

GIS-based real time prediction of Arias intensity and earthquake-induced landslide hazards in Alborz and Central Iran M. Mahdavifar, M.K. Jafari & M.R. Zolfaghari

1427

Geomorphology of old earthquake-induced landslides in southeastern Sicily P.G. Nicoletti & E. Catalano

1433

Coseismic movement of an active landslide resulting from the Mid-Niigata Prefecture Earthquake, Japan T. Okamoto, S. Matsuura & S. Asano

1439

Characteristics of large rock avalanches triggered by the November 3, 2002 Denali Fault earthquake, Alaska, USA W.H. Schulz, E.L. Harp & R.W. Jibson

1447

FE analysis of performance of the Lower and Upper San Fernando Dams under the 1971 San Fernando earthquake C. Takahashi, F. Cai & K. Ugai

1455

Reduction of the stability of pre-existing landslides during earthquake B. Tiwari, I. Dhungana & C.F. Garcia

1463

Probabilistic hazard mapping of earthquake-induced landslides H.B. Wang, S.R. Wu, G.H. Wang & F.W. Wang

1469

Investigation on stability of landfill slopes in seismically active regions in Central Asia A.W. Wu, B.G. Tensay, S. Webb, B.T. Doanth, C.M. Ritzkowski, D.Z. Muhidinov & E.M. Anarbaev

1475

Mechanism for loess seismic landslides in Northwest China L. Yuan, X. Cui, Y. Hu & L. Jiang

1481

Climate, hydrology and landslides Evaluation of the landslide potential in Chahr Chay dam reservoir slopes K. Badv & K. Emami Effect of well pumping on groundwater level and slope stability in the Taiwan Woo-Wan-Chai landslide area M. Chang, B.R. Li, Y.S. Zhang, H.S. Wang, Y.H. Chou & H.C. Liu

XV

1489

1493

Case study: Embankment failure of Cable-Ski Lake development in Cairns K. Chen

1501

Analysis method for slope stability under rainfall action X.D. Chen, H.X. Guo & E.X. Song

1507

Hydrological modelling of the Vallcebre landslide J. Corominas, R. Martín & E. Vázquez-Suñé

1517

Landslides in stiff clay slopes along the Adriatic coast (Central Italy) F. Cotecchia, O. Bottiglieri, L. Monterisi & F. Santaloia

1525

Research on the effect of atomized rain on underground water distribution in Dayantang landslide J. Ding

1533

Landslide hydrogeological susceptibility in the Crati valley (Italy) P. Gattinoni

1539

Sustainable landslide stabilisation using deep wells installed with siphon drains and electro-pneumatic pumps A. Gillarduzzi

1547

Biological and engineering impacts of climate on slopes – learning from full-scale S. Glendinning, P.N. Hughes, D.A.B. Hughes, D. Clarke, J. Smethurst, W. Powrie, N. Dixon, T.A. Dijkstra, D.G. Toll & J. Mendes

1553

Some attributes of road-slope failure caused by typhoons M.W. Gui, C.H. Chang & S.F. Chen

1559

A small rock avalanche in toppled schist, Lake Wanaka, New Zealand G.S. Halliday

1565

NRCS-based groundwater level analysis of sloping ground L.I. Ju, O.T. Suk, M.Y. Il & L.S. Gon

1571

A numerical case study on load developments along soil nails installed in cut slope subjected to high groundwater table A.K.L. Kwong & C.F. Lee

1575

Landslides at active construction sites in Hong Kong T.M.F. Lau, H.W. Sun, H.M. Tsui & K.K.S. Ho

1581

Landslide ‘‘Granice’’ in Zagreb (Croatia) ˇ Ortolan Z. Mihalinec & Z.

1587

Improvement of subsurface drainage provisions for recompacted soil fill slopes in Hong Kong K.K. Pang, J.M. Shen, K.K.S. Ho & T.M.F. Lau

1595

Biotechnical slope stabilization and using Spyder Hoe to control steep slope failure P. Raymond

1603

Rapid landslides threatening roads: Three case histories of risk mitigation in the Umbria region of Central Italy D. Salciarini, P. Conversini, E. Martini, P. Tamburi & L. Tortoioli

1609

Assessment of the slope stabilisation measures at the Cadas Pangeran road section, Sumedang, West Java D. Sarah, A. Tohari & M.R. Daryono

1615

Analysis of control factors on landslides in the Taiwan area K. Shou, B. Wu & H. Hsu

1621

Inclined free face riverbank collapse by river scouring J.C. Sun & G.Q. Wang

1627

Drainage control and slope stability at an open pit mine: A GIS-based hydrological modeling C. Sunwoo, Y.S. Choi, H.D. Park & Y.B. Jung

1633

XVI

Assessment of regional rainfall-induced landslides using 3S-based hydro-geological model C.H. Tan, C.Y. Ku, S.Y. Chi, Y.H. Chen, L.Y. Fei, J.F. Lee & T.W. Su

1639

Investigation of a landslide along a natural gas pipeline (Karacabey-Turkey) T. Topal & M. Akin

1647

Influence of extreme rainfall on the stability of spoil heaps I. Vanicek & S. Chamra

1653

Behavior of expansive soil slope reinforced with geo-grids M.Y. Wang, X.N. Gong, M.Y. Wang, J.T. Cai & H. Xu

1659

Geotechnical properties for a rainstorm-triggered landslide in Kisawa village, Tokushima Prefecture, Japan G. Wang & A. Suemine Yigong rock avalanche-flow landslide event, Tibet, China Q. Xu, S.T. Wang, H.J. Chai, Z.Y. Zhang & S.M. Dong

1667 1675

Key issues of emergency measures and comprehensive remediation projects to control the Danba landslide, Sichuan province, China Q. Xu, X.-M. Fan, L.-W. Jiang & P. Liu

1681

Enhanced slope seepage resulting from localized torrential precipitation during a flood discharge event at the Nuozhadu hydroelectric station M. Xu, Y. Ma, X.B. Kang & G.P. Lu

1689

An issue in conventional approach for drainage design on slopes in mountainous regions Z.Q. Yue

1697

Analysis of geo-hazards caused by climate changes L.M. Zhang

1703

Slope stabilization and protection Back experience of deep drainage for landslide stabilization through lines of siphon drains and electro-pneumatics drains: A French railway slope stabilization example S. Bomont

1713

Experimental geo-synthetic-reinforced segmental wall as bridge abutment R.M. Faure, D. Rossi, A. Nancey & G. Auray

1721

Rock slope stability analysis for a slope in the vicinity of Take-off Yard of Karun-3 Dam M. Gharouni-Nik

1725

Stabilization of a large paleo-landslide reactivated because of the works to install a new ski lift in Formigal skiing resort J. González-Gallego, J. Moreno Robles, J.L. Garc´ı a de la Oliva & F. Pardo de Santayana

1731

A case study on rainfall infiltration effect on the stability of two slopes M.W. Gui & K.K. Han

1737

Consolidation mechanism of fully grouted anchor bolts S. He, Y. Wu & X. Li

1745

Stability analysis for cut slopes reinforced by an earth retention system by considering the reinforcement stages W.P. Hong, Y.S. Song & T.H. Kim

1751

Landslide stabilization for residential development I. Jworchan, A. O’Brien & E. Rizakalla

1757

Influence of load transfer on anchored slope stability S.K. Kim, N.K. Kim, Y.S. Joo, J.S. Park, T.H. Kim & K.S. Cha

1763

XVII

Review of slope surface drainage with reference to landslide studies and current practice in Hong Kong T.M.F. Lau, H.W. Sun, T.H.H. Hui & K.K.S. Ho Analysis of dynamic stability about prestressed anchor retaining structure H. Li, X. Yang, H. Liu & L. Du Safety analysis of high engineering slopes along the west approach road of ZheGu mountain tunnel T.B. Li, Y. Du & X.B. Wang

1769 1775

1781

Landslide stabilizing piles: A design based on the results of slope failure back analysis M.E. Popescu & V.R. Schaefer

1787

Landslides on the left abutment and engineering measures for Manwan hydropower project X. Tang & Q. Gao

1795

Factors resulting in the instability of a 57.5 m high cut slope J.J. Wang, H.J. Chai, H.P. Li & J.G. Zhu

1799

Orthogonal analysis and applications on anchorage parameters of rock slopes E.C. Yan, H.G. Li, M.J. Lv & D.L. Li

1805

Waste rock dump slope stability for a gold mine in California H. Yang, G.C. Rollins & M. Kim

1811

Properties of the high rock slope of Hongjiadu hydropower project and its engineering treatment measures Z. Yang, W. Xiao & D. Cai

1817

Typical harbor bank slopes in the Three Gorges reservoir: Landslide and collapse and their stability control A. Yao, C. Heng, Z. Zhang & R. Xiang

1825

Weighting predisposing factors for shallow slides susceptibility assessment on the regional scale J.L. Zêzere, S.C. Oliveira, R.A.C. Garcia & E. Reis

1831

Analyses of mechanism of landslides in Tongchuan-Huangling highway L. Zhang & H. He

1839

Treatments of Loess-Bedrock landslides at Chuankou in Tongchuan-Huangling expressway J.B. Zhao

1847

Types, characteristics and application conditions of anti-slide retaining structures J. Zheng & G. Wang

1855

The stabilization of the huge alluvial deposit on the left bank and the high rock slope on the right bank of the XiaoWan Hydropower Project L. Zou, X. Tang, H. Feng, G. Wang & H. Xu

1863

Risk assessment and management Malaysian National Slope Master Plan – Challenges to producing an effective plan C.H. Abdullah & A. Mohamed

1873

Spatial landslide risk assessment in Guantánamo Province, Cuba E. Castellanos Abella & C.J. van Westen

1879

Landslide risk management: Experiences in the metropolitan area of Recife – Pernambuco, Brazil A.P. Nunes Bandeira, & R. Quental Coutinho

XVIII

1887

Societal risk due to landslides in the Campania region (Southern Italy) L. Cascini, S. Ferlisi & E. Vitolo

1893

Landslide risk in the San Francisco Bay region J.A. Coe & R.A. Crovelli

1899

A first attempt to extend a subaerial landslide susceptibility analysis to submerged slopes: The case of the Albano Lake (Rome, Italy) G.B. Fasani, C. Esposito, F. Bozzano, P. Mazzanti & M. Floris

1905

Landslide susceptibility zonation of the Qazvin-Rasht-Anzali railway track, North Iran H. Hassani & M. Ghazanfari

1911

Assessment of landslide hazard of a cut-slope using linear regression analysis S. Jamaludin, B.B.K. Huat & H. Omar

1919

Global monitoring strategy applied to ground failure hazards E. Klein, C. Nadim, P. Bigarré & C. Dünner

1925

Regional slope stability zonation based on the factor overlapping method J.F. Liu, G.Q. Ou, Y. You & J.F. Lui

1933

Landslide hazard and risk assessment in the areas of dams and reservoirs of Serbia P. Lokin & B. Abolmasov

1939

The evaluation of failure probability for rock slope based on fuzzy set theory and Monte Carlo simulation H.J. Park, J.G. Um & I. Woo

1943

Macro-zoning of areas susceptible to flowslide in pyroclastic soils in the Campania region L. Picarelli, A. Santo, G. Di Crescenzo & L. Olivares

1951

Zoning methods for landslide hazard degree J. Qiao & L.L. Shi

1959

A proposal for a reliability rating system for fluvial flood defence embankments in the United Kingdom M. Redaelli, S. Utili & M. Dyer

1965

Simplified risk analysis chart to prevent slope failure of highway embankment on soft Bangkok clays A. Sawatparnich & J. Sunitsakul

1971

Determining landslide susceptibility along natural gas pipelines in Northwest Oregon, USA J.I. Theule, S.F. Burns & H.J. Meyer

1979

Landslide susceptibility assessment using fuzzy logic Z. Wang, D. Li & Q. Cheng

1985

Prediction of the spatiotemporal distribution of landslides: Integrated landslide susceptibility zoning techniques and real-time satellite rainfall H. Yang, R.F. Adler, G.J. Huffman & D. Bach Entropy based typical landslide hazard degree assessment in Three Gorges Z. Yang & J. Qiao The optimal hydraulic cross-section design of the ‘‘Trapezoid-V’’ shaped drainage canal of debris flow Y. You, H.L. Pan, J.F. Liu, G.Q. Ou & H.L. Pan Practice of establishing China’s Geo-Hazard Survey Information System K. Zhang, Y. Yin & H. Chen A XML-supported database for landslides and engineered slopes related to China’s water resources development Y. Zhao & Z. Chen

XIX

1991 1995

2001 2005

2011

Landslide and engineered slopes in China Failure and treatment technique of a canal in expansive soil in South to North Water Diversion project Y.J. Cai, X.R. Xie, L. Luo, S.F. Chen & M. Zhao

2019

High slope engineering for Three Gorges ship locks G.J. Cao & H.B. Zhu

2027

Large-scale landslides in China: Case studies R.Q. Huang

2037

Early warning for Geo-Hazards based on the weather condition in China C.Z. Liu, Y.H. Liu, M.S. Wen, C. Tang, T.F. Li & J.F. Lian

2055

Slope engineering in railway and highway construction in China G. Wang, H. Ma, M. Feng & Y. Wang

2061

Mining slope engineering in China S. Wang, Q. Gao & S. Sun

2075

Structure and failure patterns of engineered slopes at the Three Gorges reservoir Y.P. Yin

2089

Slope engineering in hydropower projects in China J.P. Zhou & G.F. Chen

2101

A thunder at the beginning of the 21st century – The giant Yigong Landslide Z.H. Wang

2111

Author index

2119

XX

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Preface

The city of Xi’an, China is privileged to have the honor of hosting the 10th International Symposium on Landslides and Engineered Slopes, following its predecessors: Rio de Janeiro, Brazil, 2004; Cardiff, U.K., 2000; Trondheim, Norway, 1996; Christchurch, New Zealand, 1992; Lausanne, Switzerland, 1988; Toronto, Canada, 1984; New Delhi, India, 1980; Tokyo, Japan, 1977; and Kyoto, Japan, 1972. China is one of the countries in the world that suffer severely from landslide hazards. Statistics have shown that every year 700 to 900 people are killed by landslides. With the large scale infrastructure construction, failures of engineered slopes are increasing, and have become a serious concern of the government, various enterprises and technical societies. The Chinese geological and geotechnical communities look forward to this unique opportunity of exchanging and sharing technical know-how and experience of combating landslides disasters with our international peers. Xi’an is a historical city of China. It has been the capital for China’s twelve dynasties, spanning over 1200 years, and was a starting point of the famous Silk Road. It is a nice place for participants from all over the world to meet and review our past experience, and in the meanwhile, look forward to a productive future against landslides and slope failures. From this book, readers will find that there are 7 Symposium Themes, as allocated by the Steering Committee, including 13 keynote and special papers. At the JTC1 meeting held at the 9th Symposium, it was decided that a special session entitled ‘China Afternoon’ would be organized, whose papers are arranged under a separate theme ’Landslides and Engineered Slopes in China’ in this Proceedings. This symposium was jointly organized by the Chinese Institution of Soil Mechanics and Geotechnical Engineering-CCES, Chinese National Commission on Engineering Geology, Chinese Society of Rock Mechanics and Engineering, and the Geotechnical Division of the Hong Kong Institution of Engineers. The Organizing Committee is grateful to the reviewers for reviewing more than 300 submitted papers. It is practically impossible to list the large number of these volunteers here. However, their contributions must be fully acknowledged, without which the quality of this book would not have been maintained. Special thanks also go to the members of JTC1 for their constant attention and useful comments during the preparation of this symposium. Zuyu Chen

XXI

Keynote lectures

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Landslides: Seeing the ground Norbert R. Morgenstern & C. Derek Martin Dept. Civil & Environmental Engineering, University of Alberta, Edmonton, Canada

ABSTRACT: Landslide engineering requires the consideration of a number of complex processes ranging from geological and hydrogeological characterization to geomechanical characterization, analyses and risk management. This paper concentrates on recent advances that improve site characterization applied to landslide problems. It presents the view that one of the most exciting developments is the growing potential for application of Geographical Information Systems (GIS) and that making GIS goetechnically smart is a transformative development. Examples are given of integrating remote sensing data in GIS to improve visualization, mapping and movement characterization. Application of analysis of rockfall within GIS and complex slope stability evaluation with the aid of GIS are presented to illustrate recent developments and provide direction for future enhancements. 1

• geotechnical site characterization • the geotechnical model, seepage, stability and deformation analyses • risk assessment and risk mitigation

INTRODUCTION

A landslide, whether it occurs in a natural or an engineered slope, is a complex process. When Laurits Bjerrum, at the end of his Terzaghi Lecture (Bjerrrum, 1967), reminded us of the recognition in Japan of ‘‘a landslide devil who seems to laugh at human incompetence’’, he was reminding us of the complexity of the landslide process. Managing complexity invariably requires simplification into a Process Model. A Process Model captures the essentials required to meet the objectives of using the model, without including details that are extraneous to these objectives. In geotechnical engineering these objectives can range from ensuring that an engineered structure will perform as intended to managing the risk associated with natural hazards over a larger scale. Establishing the appropriate process is both site and project dependent. It underpins the value associated with the practice of geotechnical engineering. Understanding the landslide process and being able to simplify it effectively calls on interpreting a number of contributory processes and activities. The main ones are as follows:

There has been very substantial progress in all of these areas in recent years. New tools are applied to site characterization. The range of geomechanical models that can be usefully applied in practice has grown substantially. The capacity to analyse often exceeds the capacity to characterize. Risk assessment and management of slopes is maturing quickly as a valuable tool for dealing with landslides both locally and regionally. Yet much uncertainty persists in geotechnical practice. The intrinsic presence of uncertainty in geotechnical practice was emphasized by Morgenstern (2000) who provided numerous examples of unanticipated behaviour of geotechnically engineered facilities, often with unfavourable results. In developing the theme for this paper, we have drawn on our experience to conclude that the greatest uncertainties in the process modeling of landslides arise from inadequacies in site characterization, in the broadest sense, and therefore we concentrate on a discussion of recent advances that improve site characterization applied to landslide problems.

• geomorphology—the multiplicity of physical and chemical processes that have affected the surface and near-surface of the site • hydrology—the role of surface water in infiltration, erosion, etc. • geology—the sequence and characteristics of the soils and rocks • hydrogeology—the factors affecting the groundwater distribution

2

VIEWING THE GROUND SURFACE

2.1 Geographical Information Systems (GIS) It is our opments potential is that it

3

view that one of the most exciting develfor landslide engineering is the growing for application of GIS. The power of GIS enables us to ask questions of a database,

perform spatial operations on databases and generate graphic output that would be laborious or impossible to do manually. Rhind (1992) observes that a GIS can answer five generic questions: Question

Type of Task

1. What is at . . . ..? 2. Where is. . ...? 3. What has changed since. . .?

Inventory Monitoring Inventory and monitoring Spatial analysis Modelling

4. What spatial pattern exists..? 5. What if. . ...?

The first three questions are simple queries, while the last two are more analytical. GIS on its own adds enormously to our capacity to see and interpret surface geospatial information which is essential for landslide engineering. A first fly-by experience in GIS soon convinces the landslide engineer of its potential. However, GIS has limitations in presenting three-dimensional (3D) geologic and geotechnical data since it was originally developed to deal with two dimensional plane problems. Some GIS systems, like ArcGIS, provide a functional developer kit which can be used to develop the 3D capability for geotechnical engineering problems. As pointed out by Lan & Martin (2007), the many current developments in 3D GIS are still not sufficient to meet the needs of the geotechnical engineer. Mining software such as Surpac Vision provides a comprehensive system for geological modeling, but not geotechnical modeling. For example, while three-dimensional solid modeling and two-dimensional sections can be easily created in Surpac, querying inclinometer or piezometer data is not easily accomplished. However an integrated system, which is illustrated in Figure 1, can be developed. Even within Stage 1, limited ground behaviour can be modelled. The aim of Stage 2 is to establish a comprehensive ground model of the site. Stage 3 links geotechnical numerical analysis tools to conduct geotechnical analyses and assist in decision-making. Making GIS geotechnically smart is a transformative development for geotechnical engineering. Examples to illustrate this will follow in subsequent sections of the paper which will return to demonstrate the role of GIS in a number of slope related problems.

Figure 1. The architecture for an integrated system. It is composed of three different stages which required the implementation of specific tasks.

the landslide engineer for many years, although often neglected in many geotechnical curricula. The latter has usually been the preserve of specialists. To recognize landslides, API relies on characteristic morphology, vegetation and drainage. Parise (2003) provides an example of how diagnostic surface features can be related to certain types of movement, the degree of activity and the depth of movement. The study of sequential photographs can provide information on the progressive evolution of landslides and can lead to a better understanding of their causes (Chandler & Brunsden, 1995; Van Westen & Getahun, 2003). GIS facilitates the application of API, the archiving of the photos and the production of geomorphological maps that arise from the interpretation. While the application of API is common, more quantitative studies are rare, probably due to limited availability of good quality photographs, adequately fixed control points and cost. Modern photogrammetric software has been developed that should encourage greater use of photogrammetry for the construction of high quality digital elevation models (DEM). Differential DEMs will quantify landslide movements.

2.2 Aerial and terrestrial photographs Aerial photographs are a well established resource for landslide studies and this is well-understood (Soeters & Van Westen 1996). Aerial photographs can be used for interpretation (API) for qualitative analysis and photogrammetry for extracting quantitative information. The former has been an essential tool for

4

slopes are bare of vegetation and, under good conditions, can produce DEM’s that are comparable to those produced by terrestrial based LiDAR systems (Martin et al. 2007). 2.3 Satellite sensing Van Westen (2007) has recently summarized the use of remote sensing imagery in creating landslide inventories and notes that medium resolution satellite imagery such as LANDSAT, SPOT, ASTER, IRIS-D etc. are used routinely to create landuse maps and inventories of landslides. At a broader scale, areas of global landslide susceptibility have been determined by correlating information on land surface features from the NASA Shuttle Radar Topography Mission (SRTM) which can resolve features up to 30 m in size. Correlations with other satellite-based information, such as that related to precipitation, results in a global landslide susceptibility map that is broadly supported empirically (Hong, Adler & Huffman 2007). While technically fascinating, the restricted resolution of the DEM and the lack of sub-surface information limits the value of this very remote sensing information for the landslide engineer. While optical images with resolutions larger than 3 m (see medium resolution list above) have proven useful for interpretation of landslides in some individual cases (Singhroy 2005), Van Westen (op. cit) advises that very high resolution imagery (QuickBird, IKONOS, CARTOSAT-1, etc.) has become the best option for landslide mapping from satellite imagery and the number of earth observation satellites with stereo capabilities and resolution of 3 m or better is increasing rapidly. Van Westen (op cit) also draws attention to the high resolution imagery that is available within Google Earth and notes that the 3-D capabilities and zooming functions that are available in Google Earth, together with the possibility of drawing polygons on the image, greatly facilitates the interpretation and mapping of slopes and landslides. These images can be transported into GIS for storage and mensuration.

Figure 2. Flowchart of the working procedures used by Walstra et al (2007) for using historical aerial photographs in landslide assessment.

Walstra, Dixon & Chandler (2007) provide a flow chart for this process, reproduced here as Figure 2, and then illustrate its application to the Mam Tor landslide, looking back over photographs of adequate quality to 1953, and to the East Penwyn landslide over a comparable period. The need for accurate orthophotos are common to both the studies discussed in this paper as well as any spatial quantification and referencing within GIS. As the authors observe, the two case histories have demonstrated the value of the aerial photographic archive to extract spatial data necessary for assessing landslide dynamics. Terrestrial-based digital photography is also gaining acceptance as an efficient method for creating DEMs as well as capturing geological structure of rock slopes (Pötsch et al. 2007). When designing remedial measures, the DEM’s can be used to produce sections as well as estimate volumes. Findley (2007) describes the application of Sirovision technology using terrestrial-based digital photography to characterise a rock slope along Interstate 90 USA. He notes that the while the technology saves time and effort it is not a substitute for field mapping. It should also be noted that the technique works best when the

2.4 LiDAR technology LiDAR (Light Detection and Ranging) has become such an essential tool in the mapping and characterizing of landslides, that it is becoming difficult to imagine assessing all but the most local problems without it. Airborne LiDAR uses a powerful laser to map the ground surface in swathes. The literature on the application of LiDAR to landslide studies is growing rapidly and the procedures are becoming almost routine in professional practice in some countries. In our view, the definitive studies published so far are those carried out by the U.S. Geological Survey to assess

5

earth DEM’s. Examples of a bare earth DEM with a 1 m resolution applied to landslide studies and a bare-earth DEM with 25–50 cm resolution applied to faulting studies are citied by Carter et al. (2007). Airborne LiDAR is increasingly being applied to map landslides and contribute to infrastructure locations such as pipelines and to develop landslide maps which contribute to risk analysis. The ability to penetrate forest cover, even with reduced DEM accuracy, is of enormous value. The increased accuracy of DEM’s foresees the increasing use of differential LiDAR to measure ground deformations with time. As an illustration of the use of LiDAR in current practice, Figure 3 shows the bare earth projection side by side with conventional aerial photography of a potential pipeline crossing of a river in central British Columbia. While most crossings are by horizontal directional drilling, design for conventional crossings by excavation methods are needed as a stand-by. Hence landslide identification is an important consideration in route selection. In this case the river has downcut into a deep deposit (∼200 m) of glacio-lacustrine clays. The contrast between the information revealed by the bare earth imagery and conventional airphotos is striking. While airborne LiDAR is commonly used for larger scale studies, it is also of value to enhance detailed geological studies at specific sites. Jaboyedoff et al. (2007) has suggested that the high resolution LiDAR DEM can be used to extract both regional and local scale geological structures. The advantages of such techniques are obvious when dealing with steep mountainous terrain. Figure 4 shows a portion of the famous Turtle Mountain-Frank Slide in Canada and the use of shading relief of a high resolution LiDAR DEM to portray the extent of tension cracks that still exist beyond the scarp of the slide (Sturznegger et al. 2007). The potential instability associated with these cracks is a matter of concern.

landslide susceptibility in Seattle, Washington (Schulz 2004, Schulz 2007). In this case, airborne laser pulses were uniformly spaced within a 600 m wide swath with an average pulse density of 1/m2 . Up to four laser returns were collected for each pulse resulting in a vertical profile of ground features for each pulse location. Each pulse generates multiple returns due to reflections from features such as powerlines, buildings, trees, undergrowth and the ground surface. Simultaneous acquisitions of aircraft position and laser direction located laser returns with absolute vertical and horizontal accuracy of 15 cm and less than 1 m, respectively (Schulz 2004). Swathes are stitched together into a seemless DEM during processing. All ground features that produce returns are represented in the laser survey, including buildings, trees and boulders. One of the most valuable developments is that the trees can be stripped away because some pulses penetrate the tree canopy and others are reflected off the forest floors. The latter can be separated from reflections from the trees to produce bald-earth DEM’s. This processing for deforestation is a remarkable contribution but as pointed out by Haugerud & Harding (2001), there are some limitations in the algorithms that need to be recognized in interpreting the bare-earth DEM’s. The technique has even been applied to faulting studies in highrelief Alpine landscapes, with spectacular results (Cunningham et al. 2006). In the case of the Seattle bare-earth DEM, the vertical accuracy is typically about 30 cm, but is significantly less in areas of high vegetation. The data in the DEM have a grid cell size of 1.8 m. This DEM was entered into a GIS to produce a landslide map using derivatives of the DEM such as shaded relief maps (hill shades), a slope map, a topographic contour map and numerous ground surface profiles with a 2 m contour interval. This was supplemented by historical information and ground mapping. The strength and weaknesses of LiDAR mapping are discussed in detail by Schulz (2007). It is of interest to note his conclusions that aerial photographs appeared to be more effective than LiDAR in the Seattle area for discerning boundaries of recently active landslides within landslide complexes. The resolution of the LiDAR data appeared inadequate to resolve landslide boundaries within landslide complexes. However, LiDAR was much more effective for identifying presumably older landslides and the boundaries of complexes in which recently active landslides occurred. Another recent example illustrating the value of high resolution DEM’s provided by LiDAR for mapping landslides has been provided by Ardizzone et al. (2007). Improvements in LiDAR technology, including processing, are rapidly leading to even more accurate bare

Figure 3. Example of a LiDAR bare-earth projection compared to the conventional aerial photograph.

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broader Engineering Geology or Geomorphology to Geotechnical Engineering. The development of a landslide map, an essential for any hazard on risk assessment tool, relies primarily on the visualization techniques summarized previously. Aerial photo interpretation remains widely used and is increasingly enhanced by LiDAR imagery. GIS and image processing software facilitate the process. Soeters & Van Westen (1996) have summarized the geomorphic features that are diagnostic of landslides both recent and relict. The assessment of landslide susceptibility goes beyond the cataloguing of past and current landslides by including areas that are susceptible to sliding. Ideally a susceptibility assessment is based on field reconnaissance to determine factors contributing to instability, utilizing the landslide inventory as a first step. Landslide susceptibility maps have been published for many decades (e.g., Radbruch & Crowther 1973). However the ability to manipulate geomorphic data within GIS has proliferated the number of landslide susceptibility studies and their associated methodology. Even prior to the use of GIS based techniques, relative landslide susceptibility in terms of simple bivariate analyses or more complex multivariate analyses had been developed. Early zonation methods based on these developments have been discussed by Varnes (1984). More recent GIS-based developments are listed in Chacón et al (op cit). Some highlights cited are:

Figure 4. Structural and tension crack mapping of the Frank Slide using LiDAR DEM.

In addition to airborne LiDAR, terrestrial-based LiDAR is also finding applications in slope stability studies. Examples of rock slope assessment, where LiDAR has been used to evaluate rock structure are given by Kemeny et al. (2006) & Martin et al. (2007). Using terrestrial based LiDAR portable scanners that can operate in the range of 50 m to 800 m greatly enhance our ability to map the slope discontinuities. Sturzenegger et al. (2007) combined both airborne and terrestrial-based LiDAR to map the structural features associated with Frank Slide (Fig. 4). Terrestrial-based LiDAR is also being used for direct monitoring of the process of hard rock coastal cliff erosion (Rosser et al. 2005).

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• Franks et al. (1998) prepared detailed 1:1000 thematic maps in GIS for landslide hazards on Hong Kong Island, based on a very rich database. • Wachal & Huduk (2000) used GIS to assess landsliding in a 1,500–2,000 km2 area in the USA based on four factors—slope angle, geology, vegetation and distance to faults. • Dai & Lee (2004) developed probabilistic measures of landslide susceptibility for Lantau Island using multivariate logistic regression of presence-absence of dependent variables relating landslides and contributing factors such as lithology, slope angle, slope aspect, elevation, soil cover, and distance to stream channels.

GIS AND LANDSLIDE SUSCEPTIBILITY

As stated by Van Westen (op cit), GIS has determined, to a large degree, the current state of the art in landslide hazard and risk assessment, particularly for landslide studies that cover large areas. Chacón et al. (2006) have recently conducted a comprehensive general review of GIS landslide mapping techniques and basic concepts of landslide mapping. From this extensive investigation they identify three main groups of maps that have been propagated by means of GIS:

There is a tendency to incorporate increasingly complex statistical methods in these landslide susceptibility analyses. Spatial validation is essential for any practical application. Temporal considerations most commonly enter into landslide susceptibility forecasting by coupling rainfall probability assessment as an important triggering factor (Lan et al. 2005). This can be undertaken empirically or on a more process-based consideration. The work of Mejia-Navarvo et al (1994) provides an example of the former while that of Dietrich et al. (1995) is an early example of the latter.

1. Spatial incidence of landslides 2. Spatial-temporal incidence and forecasting of landslides (hazard susceptibility) 3. Consequence of landslides. Regional studies might characteristically have scales of 1:50,000 and smaller, while site specific studies will have larger scales ranging from 1:1000–1:25,000 depending on the project. At these larger scales one is characteristically merging from

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slowly been incorporated into geo-engineering practice for mapping the rates and extents of ground deformations associated with landslides. As the potential applications and limitations of this tool are gradually being understood, the range of terrains and situations to which it may be applied are expanding. The strength of this technique is that either available archives of data can be utilized to better understand historical movements or new data can be acquired for go-forward monitoring for large areas (up to 2500 km2 ) using a remote platform that can acquire data at night or through clouds. Synthetic aperture radar (SAR) is an active sensor that can be used to measure the distance between the sensor and a point on the earth’s surface. A SAR satellite typically orbits the earth at an altitude of approximately 800 km. The satellite constantly emits electromagnetic radiation to the earth’s surface in the form of a sine wave, which reflects off the earth’s surface and returns back to the satellite. The back-scattered microwave signal is used to create a SAR satellite image (a black and white representation of ground reflectivity) using SAR signal processing methods. SAR radar images are made up of pixels, with the specific size influenced by the SAR sensor resolution; the higher the resolution the smaller the pixel size. To measure differential ground movements over a specified time period, InSAR requires two SAR images of the same area taken from the same flight path, within typically 500 m laterally. During InSAR processing the phase of the corresponding pixels of both images are subtracted. The phase difference between the two SAR images can be used to determine the ground movement in the line-of-sight of the satellite. Froese et al. (2004) discussed some of the limitations and applications of differential InSAR (D-InSAR) in mapping ground deformations associated with landslides. These included data availability, rate of motion, direction of movement, steep slope distortions and loss of coherence due to a variety of factors such as vegetation, ground moisture and atmospheric effects. Therefore the potential application of InSAR to landslide mapping and monitoring requires consideration on a case-by-case basis to determine the suitability of this method to a particular set of site conditions. Over the past few years a number of advances have lead to an increased reliability of InSAR for measuring ground motion in an ever increasing number of ground conditions. PS-InSAR: In the last fifteen years, the available number of spaceborne SAR sensors (ERS 1/2, Radarsat 1, JERS, ALOS), has increased significantly. The capability of InSAR has been considerably improved by using large stacks of SAR images acquired over the same area, instead of the classical two images used in the standard configurations.

The inclusion of geomechanical and hydrological process considerations within GIS based modeling and landslide hazard analysis marks a convergence between the techniques for regionalbased studies developed by Engineering Geomorphologists and Geologists and the inputs of the Geotechnical Engineer. Here the example offered by Delmonaco et al. (2003) is of interest. In this case the infinite slope analysis was applied at a river basin scale with basin scale characterization of all of the inputs to this classical equation. In order to calculate the likely pore pressure development, Green-Ampt infiltration analyses were also carried out over the basin scale, reflecting the variation of rainfall with different return periods. The relation between potential instability and return period was determined and the predicted scenarios of instability were found to correspond sensibly with observations made after an extreme rainfall event in 1966. Examples like this encourage the integration of process-based considerations into GIS-based hazard and risk analyses. The coupling of landslide susceptibility forecasts with earthquake effects have already been investigated in a GIS environment (Refice & Capolongo 2002). The centrality of GIS-based processing has greatly advanced regional landslide hazard and risk analysis as summarized by Chacón et al. (2006). There has been some convergence between the tools used in regional studies and those used by geotechnical engineers in more site specific problems. As stated in Section 2, current GIS technology has significant limitations in truly three-dimensional problems. Chacón et al. (2006) concluded that ‘‘the use of three-dimensional GIS for large scale, detailed hazard or risk maps will be one of the significant developments in the near future’’. This, applied to landslide engineering, is the fundamental theme of this paper. Günther et al (2004) illustrate the kind of progress that is being made in their extension to GIS, designated RSS-GIS, that incorporates the deterministic evaluation of rock slope stability and is particularly useful for regional stability assessment. It incorporates grid-based data on rock structures, kinematic analyses for hard rock failure modes, some pore pressure effects and stability evaluations on a pixel basis. Other examples of integration of geotechnical considerations with GIS follow later in this paper, see Section 5, 6 and 7.

4 4.1

MAPPING GROUND MOVEMENT InSar

Since the late 1990’s the application of spaceborne Interferometric Synthetic Aperture Radar (InSAR) has

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stabilizing the existing road versus a re-route away from the area of most significant instability. As there is limited point source geotechnical instrumentation available in the areas that are easily accessible from the highway, larger portions of the valley slope do not have quantitative monitoring information. In the fall of 2006, a series of 18 corner reflectors were installed on both the southwest and northeast valley walls in order to characterize the differential movements of the various portions of each valley walls (Figure 5). Between November 2006 and November 2007, scenes of Radarsat-1 ascending F2N scenes were obtained and processed by the Canadian Centre for Remote sensing using IPTA software (Froese et al. 2008). The preliminary results available at the time of the preparation of this paper indicate that for the reflectors that are situated on landslide blocks moving with the line-of-sight of the satellite, the movements observed from the CR-InSAR are greater than those found over the same time period as those observed on conventional slope inclinometers. As these slides are moving in colluvium, likely with a rotational component, the CR-InSAR results may be more representative of the actual deformations that are only represented in the horizontal plane by slope inclinometers. Evaluation of this data is currently on going. Future Development: While the available resolution of the SAR sensors and the number of satellites has in the past been a limitation to the technique, the launching of new, higher resolution satellites provides the opportunities to overcome some of these limitations. With the launch of Radarsat-2 in December 2007, the ability of the satellite to look both right and left and obtain 3 m pixel resolution data will likely increase the directions of slope movements that can be measured and increase the amount of data that can be obtained.

This multi-image InSAR technique was introduced as Permanent/Persistent Scatterer Interferometric Synthetic Aperture Radar (PS-InSAR) (Ferretti et al., 1999, 2000, 2001). With these advances the InSAR techniques are becoming more and more quantitative geodetic tools for deformation monitoring, rather than simple qualitative tools. Numerous recent projects in Europe (Farina et al. 2006, Colesanti & Waskowkski, 2006, Meisina et al. 2007) have shown good correlation between results obtained from PS-InSAR and traditional geotechnical instrumentation in urban areas impacted by landslide movements. CR-InSAR: While the PS-InSAR technique is ideally suited to urban environments where buildings can be used as artificial reflectors or where suitable natural exposures exist, the application of this technique is more limited in northern boreal regions with sparse development and more dense vegetation cover. It is often in these remote northern areas where large slowly moving landslides whose size and rates of deformation are ideally suited for the InSAR technology are located. In order to overcome the issues associated with loss of coherence in vegetated and moist ground conditions the introduction of artificial, phase stable reflectors is emerging. This technique has been called either Corner Reflector InSAR (CR-InSAR) or Interferometric Point Target Analysis (IPTA). One of the first documented case histories of the use of artificial reflectors for monitoring of landslides was by Rizkalla & Randall (1999) where five corner reflectors were installed on the Simmonette River pipeline crossing as a trial to monitor slope movements. More recently Petrobras has utilized this technique along a pipeline crossing in Brazil (McCardle et al. 2007). Both of these applications have focused on the application of artificial reflectors along linear corridors in vegetated terrain. Perhaps the most complex landslide monitoring attempted utilizing CR-InSAR is for the Little Smoky River crossing of Highway 49 in northern Alberta, Canada. The application of D-InSAR to this site was first attempted in 2003 (Froese et al. 2004) but the heavy vegetation and ground moisture conditions limited the success of this application. Both valley walls at the Highway 49 crossing of the Little Smoky River are subject to ongoing movements of deep seated, retrogressive slides in glacial materials and bedrock. The movements of each valley wall are very complex as there are a variety of zones of movement that differ in aspect and level of activity based on their proximity to the present day river. Since the completion of the bridge across the Little Smoky in 1957, there have been significant ongoing maintenance issues due to slope instability impacting on the highway. In order to provide a more stable long term solution to mitigate the impacts of slope movements on the highway, options were considered for

Figure 5. Layout of the corner reflector array in relation to recently installed instrumentation and profile locations (from Froese et al. 2008).

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The introduction of 1 m pixel resolution available from the recently launched TerraSAR-X will also continue to increase the density of data that is available for target detection and monitoring. As the quality and density of this data improves, three dimensional deformation information for landslides may become a reality. Recent studies by Farina et al. (2007) for the Ciro Marina village in Calabria, Italy have shown the potential for using data from different radar platforms to estimate the geometry of movement patterns, an essential step for defining the geometry of the three dimensional nature of the rupture surface.

Figure 6. Example of rock fall hazard along a section of railway in British Columbia.

4.2 Surface Radar (SSR) The application of differential interferometry using synethic aperature radar has been recently applied to the monitoring of rock slopes. The technique is called Slope Stability Radar (SSR) and instead of using synthetic aperature radar from a moving radar platform, the SSR uses a real-aperture on a stationary platform positioned 50 to 1000 metres away from the foot of the slope (Harries & Roberts 2007). A major advantage of the technique is that it provides full coverage of the rock slope without the need to install reflectors. According to Harries & Roberts (2007) the technique offers sub-millimetre precision of slope wall movements without being affected by environmental conditions such as rain, dust, etc. The accuracy of this technique diminishes in areas of vegetative cover and hence the technique has been primarily used in open pit mines. 5

rock fall analyses at regular intervals and often the hazards come from inaccessible natural rock slopes well upslope of the track with previously undeterminable flow paths (Figure 6). GIS has been used as an effective tool in hazard delineation, but seldom is GIS used for rock fall process modeling (Dorren & Seijmonsbergen 2003). Stand alone computer software to assess rock falls have been developed to analyze trajectories, run-out distance, kinetic energies, and the effect of remedial measures (Pfeiffer & Bowen 1989, Guzzetti et al 2002, Jones et al. 2000). This software typically does not interact directly with existing GIS software. As a result to use these programs, one must first extract the digital elevation model (DEM) and then recompile it in a form that is suitable for the rock fall software. RockFall Analyst, a three dimensional rock fall program that was developed as an extension to ArcGIS, is used to illustrate the added value GIS technology provides for hazard assessment for rock falls (Lan et al. 2007). Rock fall hazard assessment for engineering purposes must capture as many variables as possible in relation to the rock fall process, kinetic characteristics and their spatial distribution (Dorren & Seijmonsbergen 2003). As a geomorphologic slope process, rock falls are characterized by high energy and mobility despite their limited volume. The dynamics of the rock fall process is dominated by spatially distributed attributes such as: detachment conditions, geometry features and mechanical properties of both rock blocks and slopes (Agliardi & Crosta 2003). Today accurate three dimensional morphology can be obtained from LiDAR data but the geotechnical parameters (the coefficient of restitution and friction) must be calibrated using historical rock fall events. The historical rock fall database records provides the information of past rock fall events including location of source and deposition, timing of events, size, influence on the railway operations and the effectiveness of existing barriers, should such barriers exist.

ROCK FALL PROCESS MODEL

Rock fall is the simplest of landslide processes and it is a surface phenomenon. If GIS can be made geotechnically smart, the development of rock fall simulation within GIS, a Stage 3 development in Figure 1, should form a starting point. The Canadian railway industry has been exposed to various ground hazards since the first transcontinental line was constructed in the 1800s. One of the frequently occurring ground hazards is rock fall. These events in mountain regions occur as the result of ongoing natural geomorphologic processes (Figure 6). The slope and rock properties controlling the initiation and behaviour of these rock falls can vary widely and it is not practical to eliminate these rock fall hazards due to the extent and area of potential rock fall source zones. Nonetheless, reducing this hazard to an acceptable level of safety requires proactive risk management strategies. Due to the linear corridor occupied by railways there is often a need to conduct a large number of

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Lim (2008) showed that high resolution airphotos can also be used to aid in the assigning the geotechnical parameters to various regions of the slope and in the source zone characterisation process. 5.1

Rock fall modelling

Spatial modeling of rock fall hazards along a section of railway in the Rocky Mountains was carried out using RockFall Analyst, a GIS extension, which combines 3 dimensional dynamic modeling of the rock fall physical process with distributed raster-image modeling of rock fall spatial characteristics. The rock fall modeling process involves the following steps: 1. The potential instabilities areas (source areas) are evaluated using the DEM from LiDAR data and the spatial database of historical rock fall events. The available LiDAR DEM provided a spatial resolution of 0.15 m. 2. The rock fall hazard was assessed using RockFall Analyst by taking into account the distributed geometry and mechanical parameters, and spatial pattern of rock fall characteristics. This step included two parts: (1) rock fall process modelling and (2) raster image modelling. 3. Raster image modelling of rock fall spatial frequency, flying height (potential energy) and kinetic energy was used to produce the rock fall hazard map. Once the DEM was created from the LiDAR data and the spatial attributes assigned, potential rock falls were simulated from all of the possible source areas. Figure 7a shows the results from the simulation where the source is simulated as a horizontal line source. Using the ArcGIS tools, the number of rock fall trajectories occurring was calculated for each grid cell with an area of one square meter. Spatial geostatistical techniques within the GIS software were used to analyze the trajectories and determine the rock fall spatial frequency for the whole study area (Fig. 7b). The rock fall frequency occurring in the region of railway track was classified into 10 classes (0–9) and plotted in ArcGIS in three-dimensional view (Fig. 7c). Figure 7 demonstrates the advantage of using GIS technology to simulate the rock fall process model. One of the concerns that frequently arises when conducting rock fall analysis is the level of resolution required for the DEM to provide reliable results. The resolution of the DEM controls the geometry of the rock fall impacts and trajectories and may control the physics of the impact. In British Columbia, where the case history is located, a DEM at a 10-m-grid spacing is freely available. For comparison purposes a 1-m-grid LiDAR survey was obtained to provide more accurate geometry of the slope. The rock fall simulations described above using the DEM from the

Figure 7. Spatial frequency analysis of rock falls using RockFall Analyst. All modeling and processing of the information was carried out in ArcGIS.

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Figure 8. Comparison of the historical rock fall frequency impacting the railway tracks with the results from the RockFall Analyst simulations using the 1-m and 10-m digital elevation models.

LiDAR survey were repeated using the 10-m DEM. Figure 8 compares the historical rock fall data base to the results from RockFall Analysts for both the 1-m and 10-m-grid. It is evident from Figure 8, that the rock fall simulations from the 1-m provides better agreement with the historical rock fall events (Lim 2008).

Figure 9. Rock fall hazard assessment based on rock fall frequency and kinetic and potential energy.

as well as historical data for calibration purposes. Once calibrated, the energy output from RockFall Analyst may be used in the design of protective measures.

5.2 Hazard zoning Once the spatial distribution of the rock fall has been computed the energy from such events is required to complete the hazard assessment. Two raster layers were created in ArcGIS to assess the spatial distribution of rock fall potential-energy (flying/bouncing height relative to ground) and the kinetic energy (velocity). The energy raster layers combined with the rock fall frequency is used to produce the rock fall hazard assessment shown in Figure 8. The rock fall hazard map clearly identifies the section of railway with the greatest risk, consistent with the historical evidence. Once the hazard has been identified, the energy raster layers can also be used to provide input to the design of protective barriers. Rock fall protection requires an assessment of both the height (bouncing/flying) and velocity of the rock falls. Without such information the design of protective measures is usually based on single 2 dimensional analysis or qualitative methods.

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INTEGRATING GIS INTO GEOTECHNICAL PRACTICE

Assessment of slope movement and associated hazards demands an understanding of the site characteristics and their spatial and temporal variability. Current geotechnical modelling tools are focused on numerical analyses and are not generally designed to facilitate the requirements of site investigation and characterization. Site characterisation must address key geospatial issues, e.g., complex geology, highly irregular porewater pressure, complex surface geometry and slip surface definition as appropriate. There is little doubt that capturing more complete geomorphological, geological and geotechnical information improves the quality of geotechnical site investigation, particularly when the site is geologically and geotechnically complex (Luna & Frost 1998, Tsai & Frost 1999, Parsons & Frost 2002, Jaboyedoff et al. 2004, Kunapo et al. 2005). Culshaw (2005) in the fifth Glossop lecture suggested that ‘‘the rapid development in technology over the last twenty years and the digitization of increasing amounts of geological data has brought engineering geology to a situation in which the production of meaningful three-dimensional spatial models of the shallow subsurface is feasible’’. Despite these advances there are very few spatial tools that help the geotechnical engineer achieve this goal.

5.3 Summary Rock falls are a significant hazard to Canadian railways. The assessment of such hazards over long sections of railway requires an efficient means for storing historical data and conducting rock fall simulations. The development of the three dimensional RockFall Analyst as an extension to ArcGIS, provides the framework for rapid assessment of rock fall hazards. The use of such tools requires a detailed DEM

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The main aim of Stage 2 in Figure 1 is to establish a comprehensive ground model for the site. The construction of the ground model is enhanced using three dimensional geological modelling tools commonly available in the mining industry. Finally, geotechnical analyses and engineering decisions are performed in Stage 3 (see Figure 1). In this step the ground behaviour is analysed using commercially available geotechnical numerical tools. It is essential that the tools used in Stage 2 communicate with the tools used in Stage 3 so that the data integration is maintained across all stages. One of major issues in this integrated approach involves data input and output. In order to develop an appropriate easy-to-use input and output function, some industrial standard file formats are employed for data conversion and communication throughout the three Stages. Shape file (.SHP) from ESRI and Data Exchange File (.DXF) from AutoDesk are both industrial standard formats supported by almost all PC-based CAD and GIS products. The communication between different Stages in the system developed by Lan & Martin (2007) was implemented using these two file formats. The integration of the tools described above offers effective digital tools to model heterogeneous geology, complex stratigraphy and slip surface geometry, and variable pore pressure conditions which are critical to complex slope stability problems or other analyses. The tools also provide for incorporating findings from monitoring data. In the following section the tools are demonstrated using a translational bedrock slide in Edmonton Alberta.

While GIS is increasingly viewed as a key tool for managing spatial distribution of data (Nathanail & Rosenbaum 1998, Parsons & Frost 2000, Kunapo et al. 2005) it has significant limitations in presenting three-dimensional (3D) geologic and geotechnical data. Some GIS systems, like ArcGIS, developed by the Environmental Systems Research Institute, Inc. (ESRI), provide a functional developer kit which can be used to create 3D capability. However the current developments in 3D GIS are still not sufficient to meet the needs of the geotechnical engineer. Mining software such as Surpac Vision developed by Gemcom Software International Incorporated provide a comprehensive system for geological modelling but not geotechnical modelling. For example, while three-dimensional solid modelling and twodimensional sections can be easily created in Surpac, querying inclinometer or piezometer data is not readily accomplished. In the following section we describe an integrated approach using ArcGIS, Surpac Vision and numerical modeling, to develop a three-dimensional spatial model of a shallow subsurface slide locally referred to as the Keillor Road Slide. This integrated approach illustrates the added value obtained when data and analyses are tightly integrated. 6.1 Development of an integrated approach Nearly all slope site characterization efforts deal with surface mapping, geological information from borehole data and monitoring data. The work flow from data collection through to engineering analyses was outlined by Lan & Martin (2007) and can be summarized in three stages (see Figure 1). Stage 1 involves the data collection, management and geosynthesis of the data. Modern GIS software provides effective tools for the handling, integrating and visualizing diverse spatial data sets (Brimicombe 2003). Therefore, in Stage 1, the functionality of GIS provides an essential role in collecting, storing, analyzing, visualizing and disseminating geospatial information. This information could be basic site investigation data, such as geomorphology and geology conditions, and diverse, continually evolving geotechnical parameters, such as displacement and pore pressure readings from geotechnical instruments. Most GIS tools have limitations in representing time series data such as the displacement data from inclinometer or pore pressures from piezometers. Therefore additional functional tools were required for the standard ArcGIS software. These tools have been implemented using ArcObject, an ArcGIS developer kit, and Visual studio.net, a software developing package by Microsoft. These development tools provide capability for users to interact and communicate with various data sets.

6.2

Case study: Keillor Road slide

A complete description of the Keillor Road bedrock slide was given by Soe Moe et al. (2005). The failure of the slope occurred over a number of years with the largest deformations occurring in 2002. The slide took place along the bank of the North Saskatchewan River valley in Edmonton, Alberta Canada (Fig. 10). The site investigations for the slide were conducted over a period of 15 years using traditional boreholes and monitoring systems. Figure 11 shows a plan view of the site created in ArcGIS indicating the outline of the slide, the topography of the area, location of the tension cracks and location of the main scarp. Figure 11 also shows the location of the boreholes that had been used in the site investigations over the 15 year period. The major benefit of assembling the data in ArcGIS is that the borehole symbols are dynamically linked to the data base and instrumentation data installed in the boreholes. To reconstruct the dynamics and kinematics of the processes acting on the slope and to determine their spatial and temporal distribution, Lan & Martin

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deceleration of slope movement. In addition to the displacement versus time-plots, plots of displacement vector directions and displacement rate offer the ability to identify and evaluate the spatial and temporal character of the deformations, all in a user-friendly environment. In addition to being able to process the data quickly in both a visual manner as well as conduct specific depth queries, the user can quickly assess the kinematics of the slide. Multiple movement zones at different depth are often detected in slope inclinometer plots. In this case, two movement zones were identified in the inclinometer readings for borehole B02-2 (Fig. 12). One zone was from depth 0.61 m to depth 2.44 m (zone 1) and the other zone was from depth 7.32 m to depth 8.53 m (zone 2). Their deformation histories can be rapidly shown on the plan map (Fig. 12). It can be seen clearly that zones 1 and 2 show different movement characteristics. The moving direction of the shallow zone 1 changed direction frequently while the direction of zone 2 was essentially unchanged. Once the major rupture surface is identified, the spatial deformation pattern from all the inclinometer

Figure 10. Photograph of the Keillor Road Slide, from Soe Moe et al (2005).

Figure 11. Plan view of Keillor Road slope showing location of the site investigation boreholes, tension cracks and outline of the slide. Contour elevations have been removed for clarity.

(2007) developed add-on tools for the processing of borehole information, plotting of time-series displacement data from slope inclinometers, pore pressure data from piezometers and relative geomorphological features, such as tension cracks. The add-on tools provide all of the standard types of plots for analysing slope inclinometer data. From these standard plots discrete movement zones can be defined by specifying the from-to-depths. The resultant time-displacement plots for these discrete zones show acceleration or

Figure 12. Displacement history at different shear zones of site B02-2. Shearing zone 1 and zone 2 are characterized by obviously different displacement vectors.

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data can be shown (Fig. 13). This provides a consistency check within the data sets as well highlights the more active portions of the slide as both total displacements as well as displacement rate can be shown. Geotechnical parameters are usually measured at points during site investigation by in-situ tests or by laboratory tests. Geostatistical kriging and simulation techniques in GIS offer powerful spatial modeling tools for visualising the spatial variability of these parameters (Nathanail & Rosenbaum 1998). Parsons & Frost (2002) argued that such statistical approaches improve the quality of site investigation data. Pore pressure is an essential parameter in slope stability studies. Lan & Martin (2007) used geostatistical techniques to interpret the point pore pressure data into a spatial pore pressure surface. When conducting such geostatistical analysis it is important to ensure that the pore pressures are being measured on the same geological unit which can be readily verified by comparing the borehole logs and piezometer installation locations. Lan & Martin (2007) also attached the three dimensional displacement curves obtained from the inclinometer data to the boreholes as lines to show a spatial relationship between displacement locations and pore pressure.

As mentioned earlier, Surpac Vision provides advanced tools for viewing and interpretation of geology data. Connecting to the same geological database as used by ArcGIS, the three-dimensional geology model for the Keillor Road Slide was created in Surpac Vision. Together with the other data imported from ArcGIS, such as the geomorphological surface, displacement and pore pressure readings, and tension crack planes, a comprehensive ground model for Keillor Road slope was created in Surpac Vision. From the model, the spatial extent of the slope which is at risk from instability can be immediately defined by the displacement data and the surface mapping information. Critical profile sections can be extracted in Surpac Vision along the section lines parallel to the displacement vectors. These sections now include all information managed and produced in GIS and Surpac. These section profiles can be exported to DXF files which can then be optimized for slope stability analysis. Nearly all modern slope stability software such as Slope/W or Slide can readily import DXF files. However, it is important that the user examine these DXF files to ensure that the relevant information is captured. Slope/W is widely used in geotechnical engineering practice for analyzing the stability of slopes. It uses limit equilibrium theory to compute the critical factor of safety (Krahn 2003). The essential geometry elements in Slope/W include the ground surface, complex geological regions, and pore pressure line and tension crack lines. In many slope stability problems it is very important to establish an accurate representation of the slope surface geometry because small changes in the slope profile can have a significant impact on the calculated factor of safety especially when the rupture surface is relatively flat such as the translational slide at Keillor Road. Creating the section profile from LiDAR survey ensures that the most accurate surface geometry is captured. Figure 14 shows the final geometry and geology modeled using Slope/W. The integration of the

Figure 13. Spatial distribution of slope deformation from all the inclinometer data at the same discrete movement zone. Two major slope portions with different displacement evolution are divided by Keillor Road.

Figure 14. Slope stability analysis using Limit equilibrium and/or Finite-element analysis.

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has cut through about 150 metres of glacial sediments (Porter et al. 2002). Quaternary sediments occur within the major valleys where deep valley fills have been dissected and terraced by postglacial downcutting of the trunk rivers. The landslides occurred on the steep walls of an inner valley that formed during the Holocene when Quaternary sediments filling the broader Thompson River valley were incised. The valley fill consists dominantly of permeable sediments, the exception being a unit of rhythmically-bedded silt and clay in the Pleistocene sequence (Clague & Evans 2003). The surficial materials in the area are tills, fluvial, fluvioglacial, lacustrine and colluvial deposits (Ryder 1976). Individual investigations had been carried out for the six most active of these earth slides in the Thompson Valley since the early 1980s. A major effort was initiated in 2003 to re-analyse the data that had been collected over the past 20 years using the spatial capabilities inherent in GIS tools. Eshraghian et al (2007) completed a comprehensive study of the slides and concluded that the rupture surfaces that had been detected in the individual slides followed the highly plastic, overconsolidated clays within a Pleistocene stratigraphic unit consisting of up to 45 metres of rhythmically-bedded glaciolacustrine deposit of silt and clay couplets, ranging from less than 1 cm to several tens of centimetres thick (Fig. 16). These sediments may be several hundred thousand years old and

GeoStudio Software means that this model can also be used for conducting deformation or stress analyses.

7

ANALYSING COMPLEX LANDSLIDES

In the previous section we showed the benefit of integrating technologies when analyzing a single slide. In this section we demonstrate the added value when considering multiple complex landslides. 7.1 Background Large translational landslides with rupture surfaces through glacial lake sediments in preglacial valleys are common hazards within river valleys of Western Canada (Evans et al. 2005). Eleven, retrogressive, multiple, translational earth slides have occurred along 10 kilometres of the Thompson River valley between the communities of Ashcroft and Spences Bridge in south-central British Columbia, Canada. The Canadian Pacific Railway (CPR) and Canadian National Railway (CN) main rail lines were constructed through the Thompson River valley in 1885 and 1905 respectively. Both have had recurring slope stability problems along this valley (Fig. 15). Given that the two national railroads traverse the same landslide prone area, the evaluation of risk at this location is a matter of considerable significance. The Ashcroft area is part of the Thompson Plateau, a subdivision of the Interior Plateau of British Columbia. The Thompson River flows south and

Figure 16. Geological units in the earth slides and highland terraces in Thompson Valley. The arrows indicate the rupture surfaces. Modified from Eshraghian et al 2007.

Figure 15. Major landslides south of Ashcroft, BC, (modified from Eshraghian et al 2007).

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Figure 17. Example of an aerial photograph draped over a digital elevation model produced from an airborne Lidar survey (modified from Eshraghian et al 2007).

thus of Middle or Early Pleistocene age (Clague & Evans 2003). Samples of this unit from boreholes in South Slide show layers of brown, high-plastic clay 1 to 20 cm thick between thicker layers of olive silt (Figure 16). The deposits of the three glacial sequences are separated by unconformities. Figure 16 shows the geological succession synthesized from borehole logs and outcrops in scarps and terraces in the Ashcroft area based on units proposed by Clague & Evans (2003). Eshraghian et al. (2007) used GIS technology to estimate the slide volumes which varied from 1.8 to 21.4 Mm3 and spatially correlate the rupture surface. They concluded that the stratigraphic boundaries have tilts of 1.7 m/km similar to the glacial lake bottom and that sliding was occurring along essentially the same layers within the glaciolacustrine sediments (Figure 16). They then examined the surface exposure of the larger slides using LiDAR technology. The airborne LiDAR provided a vertical resolution of ±150 mm and when combined with the aerial photographs illustrated the multiple blocks associated with retrogressive slides of this nature (Figure 17). 7.2

Figure 18. Simplified multiblock model illustrating the sliding process since deglaciation at the Slide CN50.9 (modified from Eshraghian et al 2007).

horizontal movement by block A and horizontal and vertical movement by block B. This sliding and also the Thompson River erosion caused progressive failure on the deeper rupture surface and the slide was ready for another retrogression (Stage 4, Figure 18). The most recent retrogression of the Slide CN50.9 happened in September 1897. During this stage, block D moved down on the main scarp and rest of slide material moved horizontally toward the river (Stage 5, Fig. 4). During this retrogression in the early morning of September 22, 1897, residents of Ashcroft were awakened by loud, thunder-like rumblings. The landslide constricted the Thompson River without completely blocking it (Clague & Evans 2003). From these explanations, the movement rate during this retrogression is estimated to be rapid. Following the slide, Thompson River removed part of the toe of the slide, mainly within block A (stage 6, Figure 18). Eshraghian et al 2007 used the conceptual model developed in Figure 18 to develop the geological multiblock model for the retrogressive slide shown

Multiblock retrogressive model

Eshraghian et al 2007 examined Slide CN50.9 using traditional site investigation boreholes and the combined Lidar DEM and aerial photograph model to develop the geological and movement history since deglaciation (Fig. 18). During the first stage a braided Thompson River started cutting through the glacial sediments after deglaciation. Thompson River continued its down-cutting until it reached the first weak layer and potential rupture surface (stage 2, Fig. 18). Progressive failure within this weak layer caused sliding of blocks A and B on the shallower rupture surface. More down-cutting by Thompson River encountered a deeper weak layer (Stage 3, Fig. 18). This time, movement happened without retrogression on the deeper rupture surface. It caused more

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the Thompson River levels and slide movements from 1970 to 2000. Data prior to 1970 are sparse. They presented a correlation between the cumulative river level difference from the average river level (CRLD) and active years. They concluded that the main trigger for reactivation of these slides was the discharge of Thompson River that produced above average river levels for prolonged periods (Figure 20). Clague & Evans (2003) suggested ‘‘irrigation of the bench lands above the river, especially in the late 1800s, introduced large amounts of water into the valley fill. High pore pressures probably developed locally at the top of the rhythmically bedded siltclay unit, triggering large landslides’’. They added ‘‘although high pore pressures generated by irrigation related groundwater discharge probably triggered most of the historical landslides in the Ashcroft area, the fundamental causes are geological’’. Morgenstern (1986) also concluded that the primary trigger for the movement in these complex slides was related to their retrogressive nature characterised by toeinitiated movements. Eshraghian (2007) examined the following possible trigger mechanisms: (1) rainfall, (2) irrigation, (3) Thompson River Level and (4) toe erosion. He concluded that the changes in the river level had the largest impact on the stability of the slides and that this occurred when the river remained high for a long period and then retreated to cause a drawdown effect on the slope toe blocks. Those slides which when combined with toe-erosion showed the greatest potential for rapid movements.

Figure 19. Slide CN50.9 cross-section showing the rupture surfaces, stratification, and borehole locations (see Figure 15 for location, modified from Eshraghian et al 2007).

in Figure 19. This slide is now moving on two rupture surfaces as a multiple translational earth slide. The positions of the rupture surfaces were determined by inclinometers and the rate of movement differs not only on the main rupture surfaces but between inclinometer measurements on the shallower rupture surface, suggesting possible more small blocks at the toe within R-3 block. Developing a single factor of safety for such a complex slide does not convey the geological complexity of the slide nor communicate the risk to the railways. 7.3 Trigger mechanisms The location of the piezometers at the toe of Slide CN50.9 indicate upward gradients near the toe of the slope. All piezometers respond to changes in the river level but the shallower the piezometer and the closer it is located to the river, the greater the response. The piezometers also show a 7 to 10 days delay between the river level changes and the piezometer on the deeper rupture surface. The piezometers indicate that the slide portion near to the scarp, i.e. the slide head, is generally a recharge zone and the toe is a discharge zone when the Thompson River level is low. On the other hand, when the Thompson River starts rising, the water from the river seeps towards the slide mass and offsets the upward gradient condition at the toe. However, the river may not stay at these high levels for sufficient time to allow the flow system to reach equilibrium. Therefore, the top part of the rhythmically bedded silt and clay layer (unit 2) is more affected by river level changes than the lower part. In the years that the Thompson River stays at high levels for longer periods, the piezometers show the greatest increase in pore water pressures because the water has more time to seep through the soil mass. The average rainfall in the area has been increasing since the 1920s from 150 mm/year to 240 mm/year (Porter et al. 2002). Despite this rainfall increase in the area, Eshraghian et al. 2005a did not find a correlation between slide movements and short term or long term rainfall. Eshraghian et al. (2005a, 2005b) examined

Figure 20. Illustration of the average Thompson River Level compared to 1981 when typical slow movements were recorded and 1982 when the Goddard Slide was reactivated, (modified from Eshraghian et al 2007).

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7.4

well as uncertainties from the groundwater modeling and toe erosion. The probabilistic rates of movement were calculated using the frequency of the trigger (the Thompson River flood, Figure 21), the historic movement rates, for each reactivation block. The result of calculating the probability of movement for reactivation blocks are a movement probability distribution (Figure 22) which shows the probability distribution of different movement rates which may happen during the design life time of the project. They also reported the results for each reactivation block in the form of probability of different movement rates using the movement rate class suggested by Cruden & Varnes (1996) calculated for the designed life time of 100 years (Fig. 23).

Movement and risk

It is obvious from the discussion above that attempting to capture the risk from the movement associated with such landslides with a single number is not practical. Eshraghian et al 2007 used a quantitative hazard analysis in a framework that considered the different post-failure movement rates. They demonstrated the approach using probabilistic stability analyses that included material and trigger uncertainties as

7.5 Summary

Figure 21. Thompson River level for different yearly discharge return periods.

The Introduction for this paper drew attention to the large number of contributory processes that have to be considered in developing an effective process model of a landslide and its consequences. All of these contributory processes have had to be considered in the example just presented; from geology and geomorphology through hydrological and geotechnical characterization and finally geotechnical and risk

Figure 22. Histogram frequency distribution of movement rate for two translational blocks on shallower and deeper rupture surfaces.

Figure 23. Frequency of different movement rate classes for reactivation blocks defined within Slide CN50.9 for a 100 year return period.

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iv. The Gateway Pipeline Project for providing Figure 3.

analyses, utilizing both deterministic and probabilistic considerations. The capacity for undertaking such complex landslide analyses was enhanced by the availability of recently developed tools such as LiDAR imagery. However, it is unlikely that the end product could have been achieved without utilizing GIS for spatial data management, correlations and analysis. 8

The development of the ArcGIS tools and the case histories described in this paper was supported by the Canadian Railway Ground Hazard Research Program, a collaborative research program between Canadian National Railway, Canadian Pacific Railway, Transport Canada, Geological Survey of Canada, University of Alberta, Queen’s University and the Natural Sciences and Engineering Council of Canada.

CONCLUDING REMARKS

A number of recent technical advances are leading to dramatic improvements in the study of landslides and the evaluation of appropriate risk mitigation measures. This paper draws attention to some, such as the application of LiDAR to delineate landslides more clearly than aerial photographs, and the role of InSAR to monitor ground movements over large areas with increasing accuracy. A number of other tools are entering practice that merit discussion but were beyond the scope of this paper. It has been the central premise of this paper that the most important advances have been associated with improved visualization of landslides and related processes, both through surface and sub-surface features. To this end, our experience leads us to the view that GIS is capable of making transformative contributions. Examples of use of GIS in geotechnical assessment, beyond its routine application of archiving surface information, have been provided. A general rockfall simulation model has been developed with GIS. Other examples illustrate the integration of GIS with subsurface modelling capability with interfaces to any kind of geotechnical analysis software. Landslide data management and analysis of all kinds in GIS will be essential for future progress in landslide engineering as three-dimensional visualization and modelling capabilities improve.

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ACKNOWLEDGMENTS The authors wish to acknowledge the assistance of the following: i. Dr. Hengxing Lan, Research Engineer, for his leadership in developing GIS based tools at the University of Alberta. ii. Mr. Cory Froese, Team Leader—Geological Hazards, Alberta Geological Survey/Energy and Utilities Board, for assisting us with understanding the current status of InSAR applied to landslide studies. iii. The graduate students who have collaborated with us in these and related studies over the past few years.

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Varnes, D.J. 1984. International Association of Engineering Geology Commission on Landslides and Other Mass Movements on Slopes. Landslide Hazard Zonation. International Association of Engineering Geology. UNESCO Natural Hazard Series 3(63) pp. Wachal, D.J. & Hudak. P.F. 2000. Mapping landslide susceptibility in Travis County, Texas, USA. GeoJournal 51: 245–253. Walstra, J., Dixon, N. & Chandler, J.H. 2007. Historical aerial photographs for landslide assessment: two case histories. Quarterly Journal of Engineering Geology and Hydrogeology 40: 315–332.

23

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Limit equilibrium and finite element analysis – A perspective of recent advances Zuyu Chen China Institute of Water Resources and Hydropower Research, China

Keizo Ugai Department of Civil Engineering, Guma University, Japan

ABSTRACT: This paper gives a general review of the recent advances in the applications of limit equilibrium (LEM) and Finite Element Methods (FEM) for slope stability analysis. Accuracies of various LEM including Sarma’s have been reviewed. Special attentions have been given to the strength reduction finite element method regarding its applicability, criteria for failure indications, and the treatment for modulus, Poisson ratio and dilation angles.

1

2

INTRODUCTION

Analyses play an important role in assessing the risks involved in a potential landslide and the design of an engineered slope. The traditional approach based on the limit equilibrium methods (LEM) has found wide applications in slope stability analysis. The bench-mark reports given by Morgenstern (1992, 1991) and Duncan (1996) have covered almost all important aspects with regard to the use of LEM. Latest development in this area may not be substantial, but the improvement in enhancing its efficiency and widening its applicability deserve a special review. On the other hand, 3D LEM is an area that has received wide attentions whose review is also worthwhile. The finite element method (FEM) offers an alternative that is more rigorous, being free of the assumptions regarding static indeterminacy, and descriptive, being able to provide both stress and deformation information. FEM normally does not give an explicit result for factor of safety, which limits its applicability in engineering judgments. The strength reduction finite element method (SRF), originally advocated by Zienkiewicz, et al (1975), has received warm response recently as it has similar theoretical background and the associated factor of safety to those of the conventional approach using LEM.

THE LIMIT EQUILIBRIUM ANALYSIS METHODS

2.1 General When commenting the further efforts in updating the general limit equilibrium approaches, Morgenstern (1992) said that various LEM, at least for twodimensional analyses, are well understood. The impact of new studies can be slight. Indeed, new papers dealing with the improvement of the analytical aspects of LEM have not been many, compared to the early stages of the 1960s or 1970s. Some conclusive remarks on the accuracies of various LEM have been made (Duncan, 1996). More papers have focused on the search techniques for locating the critical slip surfaces, especially on the stochastic approaches of optimization methods. 2.2 The generalized formulations A number of general formulations have been proposed in the literatures. The authors tried to fix various LEMs in a unified framework. Among these research outcomes (Zhu, et al 2003, Espinoza, et al 1994, Li, 1992), we wish to briefly review the work by Chen & Morgenstern (1983), whose analytical forms will greatly facilitate the calculations by spread sheets (Chen. et al., 2008).

25

The governing force and moment equilibrium equations provided by Chen and Morgenstern are 

b

a



a

p(x)s(x)dx = 0

(1)

p(x)s(x)t(x)dx − Me = 0

(2)

b

The conventional definition for factor of safety reduces the shear strength parameters c′ and φ ′ by the following equations. ce = c/F

(3)

tanφe = tan φ/F

(4)

Eqs. (1) and (2) involve the following definitions: dW sin(φe − α) + q sin(φe − α) dx dW · sec α sin φe + ce sec α cos φe − ru dx dW +η cos(φe − α) dx

p(x) =

s(x) = sec(φe − α + β)    x dβ × exp − tan(φe − α + β) dζ dζ a  x t(x) = (sin β − cos β tan α)

(5)

(6) Figure 1. The generalized method of slices. (a) The failure mass, (b) Spencer method, (c) tanβ to be fixed at both ends.

a

× exp Me =



a

b

η



a

ξ

tan(φe − α + β)

dW he dx dx

 dβ dζ dξ dζ

(7)

that is zero at x = a and x = b. Figure 1(c) is an example that adopts a sine function for f (x). It is possible to find F and λ from Eqs. (1) and (2) by iterations.

(8)

where α = inclination of the slice base; β = the inclination of the inter-slice force. dW/dx = weight of the slice per unit width; q = vertical surface load; η = the coefficient of horizontal seismic force, he = the distance between the horizontal seismic force and base of the slice, ru = pore pressure coefficient. In order to avoid violating the Principle of complementary shear stresses, Chen and Morgenstern (1983) argued that β must be fixed at both ends of the sliding mass. They suggested [Figure 1(b) and (c)]. tan β = fo (x) + λf (x)

2.3 The simplified methods The various simplified methods in common use can be derived from Eqs. (1) and (2) as follows. 1. Spencer, 1966 This method takes fo (x) = 0 and f (x) = 1 in Eq. (9), which means [refer to Figure 1(b)], dβ/dx = 0

(10)

Eqs. (1) and (2) thus respectively reduce to

(9)

in which f (x) is a linear function that allows the values fo (a) and fo (b) to be equal to specified values of tan β at x = a and x = b respectively. f (x) is another function

26

s(x) = sec(φe − α + β)

(11)

t(x) = sin β(x − xa ) − cos β(y − ya )

(12)

The force and moment equilibrium equations (1) and (2) are simplified as: 

b



b

p(x) sec(φe − α + β)dx = 0

a



(13)

× (x sin β − y cos β)dx = Me

2.4.1 General remarks The accuracies of various limit equilibrium methods have been conclusively summarized by Duncan as follows: • The Ordinary (Fellunius) Method is highly inaccurate for effective stress analyses of flat slopes with high pore-pressure; • Bishop’s simplified method is accurate for all conditions; • Factors of safety calculated by force equilibrium methods are sensitive with the assumed inclinations of the side forces between slices; • Methods that satisfy all conditions of equilibrium are accurate for any conditions.

(16)

4. Janbu, 1954 The simplified Janbu’s method assumes β=0

For years our profession has been puzzled by the fact that Bishop’s simplified method always gives factors of safety in good agreement with those that satisfy complete equilibrium conditions. Zhu (2008) found that the omitted term on the summation of unbalanced shear forces on the interfaces in Bishop’s simplified approach can be set to zero if a particular shear force distribution for inter-slices is assigned, which in the meanwhile allows the force equilibrium condition to be satisfied.

(17)

By specifying a particular value of β for each slice, it is possible to solve for F in Eq. (1) for the methods 2, 3 and 4. 5. Bishop 1952 This method is concerned with circular slip surfaces whose center is taken to establish the moment equilibrium equation. The Bishop’s simplified method assumes β = 0 that makes s = sec(φe − α)

2.4.2 Illustrative examples To illustrate the statements regarding the accuracies of the LEM methods in Section 2.3, we present the following two examples. Example 1 Figure 2 shows a slope with simple geometry and material properties: c = 5 × 9.8 kN/m2 , φ = 35◦ , γ = 1.7 × 9.8 kN/m3 . Table 1 summarizes the factors of safety associated with different arc angle α of the circle and pore pressure coefficient ru . It can be found that the Bishop’s simplified method in all cases gives basically the same results of Spencer’s method. The discrepancies of Sweden’s method, compared to Spencer’s, increases rapidly as α and ru enlarges. At α = 117.6 and ru = 0.6, the relative error is: (FS − FF )/FF = (1.381 − 0.769)/1.381 = 44.3%.

(18)

and t=− =−



x

tan αdξ a



a

x

dy dξ = −y = −R cos α dξ

(20)

2.4 Applicability of various simplified methods

3. Low, J. III and Katafiath, 1960. The authors assumed that the inclination of the inter-slice force of each slice is equal to the average of the slopings of the top and base of the slice. (α + γ ) 2

 dW Rd dx = 0 dx

which can be demonstrated to be identical to the original formulation given by Bishop.

(14)

(15)

β = β′ =

dW dW cos α tan(φe − α) − ru sin φe dx dx   b × sec(φe − α) dx + ce sec(φe − α) × cos φe − η

2. U. S. Army, Corps of Engineers, 1967. This method assumes that the inclination of the inter-slice force of each slice is parallel to the average sloping of the slope surface, whose inclination is designated γa , β = γa

a



a

p(x) sec(φe − α + β)

a

b

(19)

where R is the radius of the circle. Substituting Eqs. (18) and (19) into Eq. (2), we have

27

Figure 2. An illustrative example explaining the accuracies of Bishop, Fellunius and Spencer methods. Table 1. Factors of safety associated with various arc angle α of the circle and pore pressure coefficient ru .

Figure 3.

Back analysis of the Huaihexin Dike.

Table 2.

Geotechnical parameters for example 2.

Soil layer number

γ (kN/m3 )

c (kPa)

φ(◦ )

1 2 3

19.11 13.03 17.12

11.0 2.67 27

26.0 0 2.84

Table 3. of β.

Factors of safety associated with different values

ru

α(◦ )

Fb

FF

Fs

β(◦ )

0.0

1.55*

5.0

10.0

15.0

20.0

0.0

117.6 95.2 81.1 70.9 63.2 117.6 95.2 81.1 70.9 63.2 117.6 95.2 81.1 70.9 63.2 117.6 95.2 81.1 70.9 63.2

3.020 2.614 2.451 2.371 2.332 2.444 2.121 1.994 1.936 1.910 1.876 1.634 1.542 1.504 1.490 1.325 1.157 1.098 1.078 1.076

3.009 2.608 2.446 2.368 2.329 2.444 2.122 1.995 1.937 1.910 1.895 1.648 1.552 1.511 1.496 1.381 1.195 1.126 1.098 1.091

2.544 2.322 2.245 2.216 2.209 1.953 1.820 1.782 1.775 1.783 1.361 1.317 1.318 1.334 1.354 0.769 0.814 0.855 0.893 0.929

F

1.013

1.027

1.070

1.147

1.232

1.341

0.2

0.4

0.6

compared with that obtained by Spencer’s method that gives F = 1.027 and β = 1.55◦ . It can be found that F varies with β considerably and the additional moment equilibrium method is indeed necessary to find a reasonable solution for F. 3

THE UPPER BOUND ANALYSIS

3.1 Sarma’s (the upper-bound) method Sarma presented a method that divides the failure mass into a number of slices with inclined interfaces. The limit equilibrium condition has been applied to both the base and inter-slice faces. This method is particularly applicable to rock slopes as advocated by Hoek (1983, 1987). The original approach by Sarma (1979) is based on the force equilibrium conditions (Figure 4), which has complex recurrence formulations. Donald and Chen (1997) presented an identical approach which is theoretically supported by the upper bound theorem and practically easy to handle. This method starts with a kinematically admissible velocity field, in which the slice moves in a direction that inclined at a friction angle relative to its neighboring slice or the base (detailed discussion has been given in Chen (2008).

NOTE: Fb , FF , Fs are factors of safety obtained by methods of Bishop, Fellunius and Spencer respectively.

Example 2 Back analysis of the Huaihexin Dike This example, shown in Figure 3, takes from a dike case in which the authors tried to use different methods to back analysis the failure (Chen, 1999). The strength parameters are shown in Table 2. The factors of safety given by the approach of satisfying the force equilibrium method only, associated with different input of β, are shown in Table 3,

28

Figure 5. Sketch for the analyses by the energy approach of Sarma’s method in finite difference forms. Figure 4. Schematic illustrations for Sarma’s method.

1. Formulations based on the finite differences The velocity of a slice numbered i, designated Vi can be determined by (refer to Figure 5) (21)

Vi = κV1

Figure 6. Sketch for the analyses by the energy approach of Sarma’s method in integral forms.

where V1 is the velocity of the first slice. κ is defined as κ=

i j  sin(αil − φeil − θi ) j

j=1

sin(αir − φeir − θi )

  V = κ exp −

(22)

x0

θ is the angle of the velocity with reference to the positive x axis. The superscript j refers to the variable on the interfaces, and l and r refer to the left and right sides of the interfaces. The factor of safety, based on Eqs. (3) and (4) is obtained by the work-energy balance equation, n  i=1



−

 dα dζ V1 dζ

(24)

Eq. (23) then becomes xn  (ce cos φe − u sin φe ) sec α − −

− W sin(α − φe )]i

i=1

cot(α − φe − θj )

x0

κ[(ce cos φe − u sin φe ) sec α x

n−1 

x

×

κ(cej cos φej − uj sin φej )i



 dW sin(α − φe ) E(x)dx dx

xn x0

(cej cos φej − uj sin φej )L csc(α − φe − θj )

dα E(x)dx + Ki = 0 dx

(25)

where Ki is a coefficient accounting for possible discontinuities in α, φe and ce .

× csc(α r − φer − θj )i sin( α − φe )i Li = 0

Ki = −

(23)

n  i=1

(cej cos φej − uj sin φej )i

× Li csc(α r − φer − θj ) sin( α − φe )i E l (xi )

The first term of the left-hand side of Eq. (23) refers to the work done by the external loads and the energy dissipation on the slip surface, while the second term is the energy dissipation on the interfaces between two contiguous slices. 2. Formulations based on integrals Eq. (21) can be transformed to an integral if the width of the slice approaches to infinitesimally small (Figure 6),

(26)

3.2 The optimization process The conventional limit equilibrium methods and the upper-bound method (Donald and Chen, 1997) include an optimization process that finds the critical failure mode associated with the minimum factor of safety.

29

Early research work, such as Chen and Shao’s (1988), has been continued recently by a number of researchers (Goh, 1999, Pham and Fredlund, 2003, Cheng, et al., 2003, Sarma and Tan, 2006). The slip surface is discretized into a number of nodal points that are connected by either smooth curves or straight lines designated A1 , A2 , . . . , A6 (Figure 7). The variables defining the failure mode includes the co-ordinates of the nodal points and the inclinations of the interfaces if the upper-bound method is adopted. The optimization method will find these variables that give the minimum factor of safety designated B1 , B2 , . . . , B6 . 3.3

problem and demonstrated that Eq. (25) is reducible to the closed-form solution. Figure 9 shows an example with the material property parameters c = 98 kPa, φ = 30◦ . The inclination of the slope surface is γ ′ = 45◦ , the bearing capacity q calculated by the closed form solution (refer to Chen, 2008) is 10921.1 kPa. In Figure 9(a) the initially guessed slip surface is represented by 5 nodal points connected by straight lines. The inclinations of the interfaces are set arbitrarily. The factor of safety given by Eq. (23) is 1.047. Figure 9(b) shows the critical mode associated with F = 1.013. Figure 9(c) shows a more accurate result that employs 16 nodal points with F = 1.006. It can be seen that the upper-bound method gives an accurate result both in terms of the factor of safety and the critical failure mode, compared to the slip-line field method.

Test examples

3.3.1 Theoretical verifications A series of test problems based on the closedform solutions provided by the slip-field method (Sokolovski, 1960) has been performed using the numerical approaches described in this Section (Donald and Chen, 1997; Chen, 1999). The results showed good agreements, demonstrating that the upper-bound method approach is more rigorous than the conventional method of vertical slices. As an extension, this method has been successfully applied to the calculation for bearing capacity analysis, in which the conventional method is generally not applicable (Wang et al, 2001). This means that various empirical coefficients involved in the conventional approaches accounting for the effect of soil weight, embedment of footing, complicated ground heterogeneities and water conditions are no longer necessary. Example 3 Comparisons with the closed-form solution Figure 8 shows a uniform slope subjected to a vertical surface load q. The weight of the soil mass is neglected. The closed-form solution for the ultimate vertical surface load has been provided by Sokolovski (1960). Chen (2008) gives a detailed description of the

3.3.2 Comparisons with the conventional methods Test examples have also shown that the upper-bound method is also able to give comparative results of factor of safety to those obtained by the conventional methods. The following two examples are taken from the ACADS slope stability programs review by Donald and Giam (1992) from which one may find the details including the material and geometry parameters. Example 4 The ACADS test example EX1(a) For a simple test example shown in Figure 10 the ‘referee answer’ based on the simplified Bishop’s

Figure 8. Verifications of Eq. (24) compared to the theoretical solution provided by the slip-line field method, example 3.

Weak seam

Figure 7. The optimization process for locating the critical failure mode.

Figure 9. Example 3, an example describing the upper bound approach.

30

3.4 Practical considerations with Sarma’s method

method for the critical slip surface is 1.00. The upper bound method defined 4 nodal points designated A, B, C, D with arbitrary interface inclinations as shown in the Figure 10(a). Factor of safety for this initial failure mode is 1.304. Figure 10(b) shows the critical failure mode associated with a minimum F of 0.997, which is very close to the ‘referee answer’ Example 5 The ACADS test example EX1(c) Using the similar algorithms the factor of safety for the initial failure mode was 1.630 as shown in Figure 11(a), and the minimum F for the critical failure mode shown in Figure 11(b) was 1.401. The upper bound results can be compared with those given by the conventional methods shown in Figure 12. The slip surfaces 1, 2, 3 are related to the methods of Spencer, Bishop, and Sarma respectively. Chen (2008) illustrated that the limit equilibrium methods with the vertical and inclined slices can be approximately fixed in the theoretical framework of the lower and upper theorems of Plasticity.

The experience of using Sarma’s method shows that the following two issues are frequently encountered, which require proper treatments. 1. The alternative directions of shear force or relative velocity on the inter-slice surface It has been shown that there are two possible directions for a relative velocity between two contiguous slices. The conventional method only considers the condition that the left slice moves upward relative to the right one, as shown in Fig 7. However, it is sometimes likely that the left slice moves down ward relative to the right one. Failure to identify this alternative may occasionally yield wrong results. 2. Treatment when tension develops on the interfaces and/or the base of a slice Sarma’s method assumes that shear failure develops along the slip surface and the interfaces. However the calculated results may show some tensile internal forces, which is contradictive to the original assumptions. In his program SARMA, Hoek gives a warning but no solution is offered. An approximate treatment is proposed by Chen (2008). Details working on the two issues deserve a special paper which is contained in this Proceedings (Chen, 2008).

Figure 10. The ACADS test example EX1(a). (a) The initial failure mode; (b) the critical one.

4

THE 3D LIMIT EQUILIBRIUM AND UPPER BOUND ANALYSIS

4.1 3D analysis based on the ‘method of columns’ In their keynote and state-of-the-art reports, Morgenstern (1992) and Ducan (1996) advocated the importance of the development of 3D limit equilibrium methods for slope stability analyses. A great number of papers on 3D slope stability analysis methods have emerged during the subsequent 10 years (e.g., Stark and Eid 1998, Chen et al. 2001, Huang and Tsai 2000, Jiang and Yamagami 2004). All these papers have dealt with the ‘method of column’ that can be considered to an extension of the ‘‘method of slices’’ in the two-dimensional area (Figure 13). On the other hand, the method employing columns or blocks with inclined interfaces have been developed (Michalowski, 1989, Chen et al. 2001a,b, Farzaneh and Askari 2003), which can be regarded as an extension of the 2D Sarma’s method. Review of the development of 3D limit equilibrium and upper bound analysis deserves a full paper, and indeed has been tried by Chen et al. (2006) in

Figure 11. The upper bound solutions for ACADS test example EX1(c). (a) The initial failure mode; (b) the critical one.

Figure 12. Comparisons of the critical slip surfaces obtained various methods. (1) Spencer, F = 1.366; (2) Bishop, F = 1.378; (3) Sarma, F = 1.401.

31

Figure 13. Sliding mass consisting of prisms with vertical interfaces.

Figure 14. The rock wedge failure ρl , ρr = the dilatant angles. The subscripts ‘l’ and ‘r’ stand for the left and right planes respectively. For other parameters, refer to Chen (2004).

their keynote paper of GeoShangai. Therefore no more elaboration will be made in this Paper. 4.2 The generalized solution to rock wedge analysis Among a variety of three-dimensional stability problems of slopes, wedge is a special and also the simplest case that requires a special study. It has been found that the limit equilibrium approach commonly used for tetrahedral rock wedge stability analysis actually involves an assumption that the shear forces applied on the failure planes are parallel to the line of intersection. It is because of this assumption that makes the solutions for normal forces applied on the left and right failure planes possible, as depicted in Figure 14. To illustrate the impact of this finding, Chen took an example that has a symmetric geometry and material properties with respect to the line of intersection. The cohesion of the two failure surfaces is set to zero. The angle between the line of intersection and the shear force applied on the failure surface is denoted by γ . For this symmetric wedge with simple geometry, it is possible to establish a formulation to calculate F associated with different values of γ . The case γ = 0◦ corresponds to the conventional method and gives a value F = 0.727. However F increases as γ becomes larger and eventually reaches a maximum of 1.002 at γ = 42.5◦ , as shown in Figure 15. A new method that allows an input of various shear force directions has been presented by Chen (2004). The controlling equation is reducible to the conventional solution and permits a formal demonstration to confirm that when ρl = φel and ρr = φer factor of safety will obtain its maximum. Chen also discussed the theoretical implications of these findings regarding some fundamental understanding in Plasticity.

Figure 15. Factors of safety associated with various shear force directions on the failure planes.

5

FACTORS OF SAFETY BY THE FINITE ELEMENT METHODS

5.1 Definition of the factor of safety The finite element method normally gives information of stress and strain fields. Various approaches have been proposed to transfer them to the factor of safety that is a common concern in engineering practice. 1. Based on the stress levels Keep σ3′ of an element unchanged, draw a circle that is tangent to the Mohr-Coulomb’s failure envelope with a corresponding diameter (σ1′ − σ3′ )f and the stress level (σ1′ − σ3′ )/(σ1′ − σ3′ )f . The factor of safety is defined as FFE1 = 

32

dl σ1′ − σ3′ dl (σ1′ − σ3′ )f 

(27)

where the integral represents the scalar summations along a potential slip surface 2. Based on the shear stress on an element For a stress state σx , σy , τxy , the normal and shear stresses σn′ and τ on the slip surface can be determinated by: τ=

1 (σy − σx ) sin 2α + τxy cos 2α 2

σn′ = σx sin2 α + σy cos2 α − τxy sin 2α

(28) (29) Figure 16. Factors of safety associated with various shear force directions on the failure planes.

where α the inclination of the slip surface to the x axis. The shear strength that can be developed along the slip surface is

Table 4. Comparisons of factors of safety obtained by different definitions. Factor of safety

′

τf = c +

σn′

tan φ

′

(30)

The factor of safety along the entire slip surface can be defined as FFE2 =



(c′ + σn′ tan φ ′ )dl  τ dl

(31)



(c′ + σn′ tan φ ′ )dl

σ1′ − σ3′ (σ1′ − σ3′ )f

Types

FFE1

FFE2

FFE3

1 2 3

Embankment by layers One-layer embankment Excavation

1.001 1.000 1.082

1.001 1.003 1.044

1.001 1.000 1.044

6

3. Based on the weighted stress levels This approach defines the factor of safety by the following equation: FFE3 = 

No.

THE STRENGTH REDUCTION FINITE ELEMENT METHOD

6.1 The advances SRF has made SRF was used for slope stability analysis as early as 1975 by Zienkiewicz et al. In this method, the shear strength parameters are reduced by Eqs. (3) and (4), allowing the evolution of large area of plastic yielding. Firstly, a gravity turn-on is implemented under elastic state to determine the initial stress distribution inside the slope. Then, stresses and strains are calculated by the elasto-plastic finite element method. The shear strength reduction factor, F, is then increased incrementally until the global failure of the slope reaches, which means that the finite element calculation diverges under a physically realistic convergence criterion (Refer to 6.3.1). In a benchmark paper, Griiffiths and Lane (1999) provided a series of test examples that show good agreements both in terms of factor of safety and plastic zones with the conventional LE method. SRF can be a powerful alternative to the traditional limit equilibrium methods. This technique has also been adopted in some well-known commercial software, such as FLAC, for practical applications. The main advantages of the SRM can be summarized as follows:

(32)

(c′ + σn′ tan φ ′ )dl

5.2 Search for the critical slip surface Having given the definition of the factor of safety, a search technique, similar to that commonly used in the limit equilibrium analysis area, can be employed to find the critical slip surface associated with the minimum factor of safety (Zhou et al., 1995). Donald et al. (1985) used an algorithm called CRISS to calculate the factor of safety of a slope taken from ex1(a) of the test problems issued by ACADS. Example 5 Reevaluations of Example 2 by FEM The referee answer based on the conventional limit equilibrium method by the ACADS review program Ex1(a) is 1.00 associated with a critical slip surface passing through the toe of the slope as shown in Figure 16. Table 4 summarizes the associated minimum factors of safety. It can be found that the results are close to one other.

• It requires no assumptions which have been commonly involved in LEM. • The critical failure surface is found automatically.

33

• It offers much more detailed information such as the plastic zone, stress and deformation field, etc., compared to LEM. • It is possible for SRF to include piles and anchors that produce the coupled stress fields for soil and structure simultaneously (Cai et al. 1998; Cai & Ugai 2000).

and Lane, 1999; Cheng et al, 2007) has confirmed good agreements between LEM and SRF in terms of factor of safety based on the definitions given by Eqs. (1) and (2). In this Proceedings, Duan et al. (2008) reported the calculated results by SRF for the gravity dam stability problems with weak seams, which are briefly summarized here. Example 6 An example that compares the results of FEM and Sarma’s method (Duan, et al., 2008). Figure 17 shows the geometry of the example with material properties listed in Table 5. For the analysis along the weak seam ABC, Sarma’s method gave a factor of safety F = 2.12. Figure 18 shows the plastic zones from which one may find that the plastic zones along the weak seam extend as FOS increases. At F = 2.20, yielding dominates throughout the seam with a plastic zone near the dam toe shown in Figure 17, compared to the result F = 2.12 by Sarma’s method. Duan et al (2008) further investigated a case where section BC no longer exists as shown in Figure 17(b). The critical location is determined by an automatic search process in Sarma’s method. The results obtained by Sarma and SRF were also in good agreement.

Perhaps, the most important contribution of SRF is that it makes geotechnical calculations by FEM selfcheckable and reproducible. Since a large-scale nonlinear finite element analysis involves complicated constitutive equations and iterations, it always happens that different computer programs cannot give same analytical results. Lack of unique and widely accepted solutions has discouraged the extensive use of FEM in geotechnical practice. Now SRF can be a tool to test the applicability of an EEM program which, as reliable software, should provide comparative results with LEM if SRF is performed using this program. 6.2 An illustrative example A number of research work (Naylor, 1981; Donald and Giam, 1992; Matsui and San, 1992; Ugai and Leshchinsky, 1995; Dawson et al, 1999; Griffiths

6.3 General issues with SRF 6.3.1 The failure criteria There have been a number of criteria that define failure at which the calculation by SRF terminates: (1) nonconvergence of the numerical process; (2) rapid increase of displacement at some critical points; or (3) development of basically continuous plastic zones. The value of F at this moment is believed to be the solution for factor of safety of this problem. Experience has shown that these criteria do not lead to substantially different values of F.

Dam

A

C

B Seam

Bedrock

D

(a) Dam

6.3.2 The constitutive laws In SRF, both associated and non-associated elastoplastic constitutive models can be adopted. The MohrCoulomb yield criterion is used to define the yield function if non-associated flow law is used.

A Bedrock

Seam

D

1 f = −c′ cos φ ′ − I1 sin φ ′ 3

 1 + J2 cos  − sin  sin φ ′ 3

Figure 17. An example that compares the results of FEM and Sarma’s method, the calculation by Sarma.

(33)

Table 5. Geotechnical properties for example 6. Material

Density (kN/m3 )

Modulus (GPa)

Poisson ratio

Friction angle

Cohesion (MPa)

Tensile strength (MPa)

Dam Bedrock Seam

24.0 25.6 18.0

20 10 2.5

0.17 0.26 0.35

35 35 19.8

2.0 1.3 0.115

1.85 0 0

34

Eq. (37) was derived on the assumption that c is negligible and the overburden h is very big. Duan et al. (2008) gave a more generalized criterion,

where =

 √ 3 3 J3 1 −1 , − sin 3 2 J23/2

π π − ≤≤ 6 6

sin(φ + α)/ cos α ≥ (1 − 2μ)

(34)

where α is defined by

and the Drucker-Prager criterion is normally adopted to define the plastic potential function g = −αI1 + where



tan 

, α= √ 9 + 12 tan2 

2c/γ h 2 2c (1 + K)2 + γh

sin α = 

(35)

J2 − κ

κ= √

3c′ 9 + 12 tan2 

(39)

K is approximately taken to be coefficient of earth pressure at rest, which can be taken as:

(36)

K=

In the above equations c, φ, and  are the effective cohesion, friction angle, and dilatant angle, respectively. I1 , J2 , and J3 are the first invariant of the effective stress, and the second and third invariants of the deviatoric stress, respectively.

μ 1−μ

(40)

The detailed work is documented in a paper of this Proceedings (Duan et al., 2008). 2. Young’s modulus From physical point of view, it is obviously advantageous to reduce both E and μ. Duan et al. (2008) also suggested a hyperbolic stress strain relationship, similar to that proposed by Duncan and Chang (1970) to derive the criterion for the reduced Young’s modulus. Figure 19 illustrates how the modulus can be reduced based on the reduction of the strength envelop. As a matter of fact, the results shown in Example 4 and Figure 20 are based on the reduced values of μ. If μ keeps unreduced, a large area of

6.3.3 Treatment of other parameters The normal practice of SRF reduces the shear strength parameters during calculations while keeping other parameters constant. The necessity of treating these unchanged parameters has been discussed. 1. The Poisson ratio Zheng et al. (2002) found that if c and tan φ are reduced considerably while Poisson ratio is still kept unchanged, it is likely that an element will inevitably yield, which is unrealistic. An approximate condition was suggested: sin φ ≥ 1 − 2μ

(38)

(37)

where μ is the Poisson ratio. Therefore a better approach can be reducing the Poisson ratio simultaneously with the reduction of c and tan φ.

Figure 19. The concept of reducing Young’s modulus E based on the hyperbolic stress strain relationship, the subscript ‘o’ and ‘r’ stand for the original and reduced variables respectively.

Figure 18. An example that compares the results of FEM and Sarma’s method, the calculation by SRF.

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This paper also reviews the use of finite element method for slope stability analysis with particular attentions to the strength reduction approach (SRF). Main findings are: • In the 2D areas, good agreements between LEM and SRF can be found. SRF can thus provide a useful approach for stability analysis; • It would be advantageous to make appropriate reductions for other material parameters, such as Poisson ratio, Young’s modulus, etc. Some associated criteria have been proposed.

Figure 20. The large area of plasticity at the bottom by using unreduced values of μ by SRF.

plastic zone would develop at the bottom of the foundation, as shown in Figure 20. Although the factor of safety obtained was still F = 2.12, this unrealistic stress distribution would limit the credit and applicability of SRF. 3. The dilatant angles A certain elasto-plastic constitutive law will be employed in the nonlinear finite element calculations, which can be associative or non-associative. The latter means a dilatant angle other than the friction angle can be assigned to an element. A number of papers (e.g. Cheng et al., 2007) investigated the influence of different values of dilatant angles to the final calculated results. The general conclusion is that adoptions of different values do not affect the final solution to factor of safety substantially in the 2D analysis.

7

REFERENCES Bishop, A.W. 1955. The use of the slip circle in the stability analysis of slopes. Geotechnique 5(1):7–17. Cai, F., Ugai, K., Wakai, A. & Li, Q. 1998. Effects of horizontal drains under rainfall by three-dimensional finite element analysis. Computers and Geotechnics 23:255–275. Cai, F. & Ugai, K. 2000. Numerical analysis of the stability of a slope reinforced with piles. Soils and Foundations 40(1):73–84. Chen, L.H., Chen, Z.Y. & Sun, P. 2008. Slope stability analysis using graphic acquisitions and spreadsheets. Proceedings of the 10th International Symposium on landslide and engineered slopes. Xi’an. Chen, Z.Y. & Morgenstern, N.R. 1983. Extensions to the generalized method of slices for stability analysis. Canadian Geotechnical Journal 20(1):104–119. Chen, Z.Y. & Shao, C.M. 1988. Evaluation of minimum factor of safety in slope stability analysis. Canadian Geotechnical Journal 25(4):735–748. Chen, Z.Y. 1999. Discussions: Prior and back stability analysis of the Huaihongxin Dike. Chinese Journal of Geotechnical Engineering 21(4):518–519 (in Chinese). Chen Z.Y., Wang xX.G., Haberfield C., Yin, J.H. & Wang, Y.J. 2001. A three-dimensional slope stability analysis method using the upper bound theorem-part I: theory and methods. International Journal of Rock Mechanics & Mining Sciences 38(3):369–378. Chen Z.Y., Wang J., Wang Y.J., Yin, J.H. & Haberfield C. 2001. A three-dimensional slope stability analysis method using upper bound theorem-part II: numerical approaches, applications and extensions. International Journal of Rock Mechanics & Mining Sciences 38(3):379–397. Chen, Z.Y. 2004. A generalized solution for tetrahedral rock wedge stability analysis. International Journal of Rock Mechanics & Mining Sciences 41:613–628. Chen, Z.Y., Yin, J.H. & Wang, Y.J. 2006. Keynote lecture: The Three-Dimensional Slope Stability Analysis: Recent Advances and a Forward Look. Advances in Earth Structures, Research to Practice, Proceedings of Sessions of Geoshanghai, ASCE Special Publication No. 151:1–42. Chen, Z.Y. 2007. The limit analysis in soil and rock: a mature discipline of geomechanics. Journal of Zhejiang University SCIENCE 8(11):1712–1724.

CONCLUSIONS

This paper summarizes the recent advances in the traditional limit equilibrium and upper-bound methods, referred to as the method of slices with vertical and inclined interfaces respectively. Main findings in this area are: • The analytical forms, such as Eqs. (1), (2) and Eqs. (13) and (14), can be derived for the generalized method of slices. They are particular useful for calculation by a spread sheet. • Theoretical studies and test examples have shown that various LEM basically yield factors of safety close to one another, all lie on the lower bound side. • Sarma’s method with the slices of inclined interfaces can be formulated by an upper-bound approach represented by Eq. (23) or (25). This method enjoys a sound mechanical background. Some treatments for the alternative inter-slice shear force directions and internal tensions have been proposed. • The three-dimensional limit equilibrium and upperbound methods with vertical and inclined columns have been made possible. As a special case, the traditional rock wedge stability analysis method has been generalized allowing different input of shear force directions in the failure surfaces.

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Chen, Z.Y. 2008. Some notes on the upper-bound and Sarma’s methods with inclined slices for stability analysis. Proceedings of the 10th International Symposium on landslide and engineered slopes. Xi’an. Cheng, Y.M. 2003. Locations of critical failure surface and some further studies on slope stability analysis. Computers and Geotechnics 30:255–267. Cheng, Y.M., Lansivaara, T.B. & Wei, W.B. 2007. Two-dimensional slope stability analysis by limit equilibrium and strength reduction methods. Computers and Geotechnics 34:137–150. Dawson, E.M., Roth, W.H. & Drescher A. 1999. Slope stability analysis by strength reduction. Geotechnique 49(6):835–840. Donald, I.B., Tan, C.P. & Goh, T.C.A. 1985. Stability of geomechanical structures assessed by finite element method. Proc. 2nd Int. Conf. In Civil Engr. Hangzhou, 845–856. Beijing: Science Press. Donald I.B. & Giam, S.K. 1988. Application of the nodal displacement method to slope stability analysis. In: Proceedings of the fifth Australia—New Zealand conference on geomechanics. 456–460. Sydney, Australia. Donald, I.B. & Chen, Z.Y. 1997. Slope stability analysis by the upper bound approach: fundamentals and methods. Canadian Geotechnical Journal 34:853–862. Donald, I.B. & Giam, P. 1992. The ACADS slope stability programs review. Proc. 6th International Symposium on Landslides. 3:1665–1670. Duan, Q.W., Chen, Z.Y., Wang, Y., Yang, J. & Shao, Y. 2008. Applications of the strength reduction finite element method to a gravity dam stability analysis. Proceedings of the 10th International Symposium on landslide and engineered slopes. Xi’an. Duan, Q. & Zhang, P.W. 2008. On the treatments for the deformation parameters in the strength reduction finite element method. Proceedings of the 10th International Symposium on landslide and engineered slopes. Xi’an. Duncan, J.M. & Chang, C.Y. 1970. Nonlinear analysis of stress and strain in soils. Journal of Soil Mechanics and Foundation Engineering Division, ASCE 96(5):1629–1653. Duncan, J.M. 1996. State of the art: Limit equilibrium and finite element analysis of slopes. Journal of Geotechnical Engineering 122(7):577–596. Espinoza, R.D., Burdeau, P.L.P.C. & Mohunthan, B. 1994. Unified formulation for analysis of slopes with general slip surface. J. Geotech. Engng, ASCE 120(7):1185–1104. Farzaneh, O. & Askari, F. 2003. Three-Dimensional Analysis of Nonhomogeneous Slopes. Journal of Geotechnical and Geoenvironmental Engineering 129(2). Fellenius, W. 1927. Erdstatisch Berechnungen, Berlin W.Ernst und Sohn revised edition, 1939. Goh, A.T.C. 1999. Genetic algorithm search for critical slip surface in multi-wedgestability analysis. Canadian Geotechnical Journal 36(2):383–391. Griffiths, D.V. & Lane, P.A. 1999. Slope stability analysis by finite elements. Geotechnique 49(3):387–403. Hoek, E. & Bray, J. 1977. Rock slope engineering. The Institute of Mining and Metallurgy. Hoek, E. 1983. Strength of jointed rock masses. Geotechnique 33(3):187–223.

Hoek, E. 1987. General two-dimensional slope stability analysis-Analytical and Computational Methods in Engineering Rock Mechanics. AlIen Unwin, London. Huang, C.C. & Tsai, C.C. 2000. New method for 3D and asymmetric slope stability analysis. ASCE. Journal of Geotechnical and Environmental Engineering 126(9):917–927. Janbu, N. 1954. Application of composite slip surfaces for stability analysis. Proceedings of European Conference on Stability of Earth Slopes. 3:43–49. Sweden. Jiang, J.C. & Yamagami, T. 2004. Three-dimensional slope stability analysis using an extended Spencer method. Soils and Foundations 44(4):127–135. Li, K.S. 1992. A unified solution scheme for slope stability analysis. Proceeding, 6th International symposium on landslides. 481–487. Christchurch. Lowe, J. III. & Karaflath, L. 1960. Stability of earth dams upon drawdown. Proc. 1st Panamer. Conf. Soil Mech, 2:537–552. Mexico City. Matsui, T. & San, K.C. 1992. Finite element slope stability analysis by shear strength reduction technique. Soils and Foundations 32(1):59–70. Michalowski, R.L. 1989. Three-dimensional analysis of locally loaded slopes. Geotechnique 39:27–38. Morgenstern, N.R. & Price, V. 1965. The analysis of the stability of general slip surface. Geotechnique 15(l): 79–93. Morgenstern, 1991. The evaluation of slope stability—a 25 year perspective. Proc. ASCE Conf. on Stability and Performance of Slopes and Embankments, 1:1–26. Berkeley. Morgenstern, 1992. Keynote paper: The role of analysis in the evaluation of slope stability. Proceedings of 6th Internatioanl Symposium of Landslides: 1615–1629. Naylor, D.J. 1981. Finite elements and slope stability. Numer. Meth. In:Geomech., Proceedings of the NATO Advanced Study Institute, Lisbon, Portugal:229–244. Pham, H.T.V. & Fredlund D.G. 2003. The application of dynamic programming to slope stability analysis. Canadian Geotechnical Journal 40:830–847. Sarma, S.K. 1979. Stability analysis of embankments and slopes. Journal of the Geotechnical Engineering Division, ASCE 105(GT12):1511–1524. Sarma, S.K. & Tan, D. 2006. Determination of critical slip surface in slope analysis.Geotechnique 56(8): 539–550. Sokolovski, V.V. 1960. Statics of soil media. (Translated by Jones DH and Scholfield AN). London: Butterworth. Spencer, E. 1967. A method of analysis of embankments assuming parallel inter-slice forces. Geotechnique 17:11–26. Stark, T.D. & Eid, H.T. 1998. Performance of threedimensional slope stability analysis method in practice. Journal of Geotechnical Engineering, ASCE 124: 1049–1060. Tan, C.P. & Donald, I.B. 1980. Finite element calculation of dam stability. Proc. 11th Int. Conf. Soil Mech. and Fnd. Engr. San Francisco. U.S. Army, Corps of Engineers. 1967. Stability of slopes and foundations, Engineering Manual, Visckburg, Miss. Ugai, K. 1985. Three-dimensional stability analysis of vertical cohesive slopes. Soils and Foundations 25(3):41–48.

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Ugai, K. & Leshchinsky, D. 1995. Three-dimensional limit equilibrium and finite element analysis: a comparison of results. Soils and Foundations 35(4):1–7. Wang, Y.J., Yin, J.H. & Chen, Z.Y. 2001. Calculation of bearing capacity of a strip footing using an upper bound method. International Journal for Numerical and Analytical Methods in Geomechanics 25:841–851. Whitman, R.V. & Bailey, W. 1967. Use of computers for slope stability analysis. Journal of Soil Mechanics and Foundation Engineering Division, ASCE 93(SM4). Wright, S.G., Kulhawy, F.H. & Duncan, J.M. 1973. Accuracy of equilibrium slope stability analysis. Journal of Soil Mechanics and Foundation Engineering Division, ASCE 99(SM10):783–791. Wright, S.G. 1978. Slope stability analysis. Proceedings on Analysis and Design in Geotechnical Engineering, Vol. 2. 153. Yamagami, T. & Ueta, Y. 1988. Search for critical slip lines in finite element stress fields by dynamic programming. Proceedings of the 6th International Conference on Numerical Methods in Geomechanics. Innsbruck, Australia: 1335–1339.

Zhu, D.Y., Lee, C.F. & Jiang, H.D. 2003. Generalized framework of limit equilibrium methods for slope stability analysis. Geotechnique 53(4):377–395. Zhu, D.Y., Dun, J.H. & Tai, J.J. 2007. Theoretical verification of rigorous nature of simplified Bishop’s method. Chinese Journal of Rock Mechanics and Engineering 26(3):455–458 (in Chinese). Zhu, D.Y. 2008. Investigation on the accuracy of the simplified Bishop method. Proceedings of the 10th International Symposium on landslide and engineered slopes. Xi’an. Zienkiewicz, O.C. Humpheson, C. & Lewis R.W. 1975. Associated and nonassociated visco-plasticity and plasticity in soil mechanics. Geotechnique 25(4):671–89. Zou, J.Z., Williams, D.J. & Xiong, W.L. 1995. Search for critical slip surfaces based on finite element method. Canadian Geotechnical Journal 32:233–246. Zheng, H. & Li, C. 2002. Solution to the factor of safety by finite element method. Chinese Journal of Geotechnical Engineering 24(5):626–628 (in Chinese).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Improving the interpretation of slope monitoring and early warning data through better understanding of complex deep-seated landslide failure mechanisms E. Eberhardt Geological Engineering, University of British Columbia, Vancouver, Canada

A.D. Watson BC Hydro, Burnaby, British Columbia, Canada

S. Loew Engineering Geology, ETH Zurich, Switzerland

ABSTRACT: The past several years have seen significant advances in landslide monitoring technologies. Remote sensing techniques based on satellite and terrestrial radar can now provide high-resolution full area spatial coverage of a slope as opposed to relying on geodetic point measurements. Automation in the form of wireless data acquisition has enabled the collection of data with increased temporal resolution. These tools provide increased capacity to detect pre-failure indicators and changes in landslide behavior. Yet the interpretation of slope monitoring data, especially that for early warning, still remains largely subjective as geological complexity and uncertainty continue to pose major obstacles. This paper reviews several recent developments in landslide monitoring techniques but questions the phenomenological approach generally taken. Examples are then provided from several recent experimental studies involving ‘‘in situ laboratories’’ in which detailed instrumentation systems and numerical modeling have been used to better understand the mechanisms controlling pre-failure deformations over time and their evolution leading to catastrophic failure. Preliminary results from these studies demonstrate that by better integrating the different data sets collected, geological uncertainty can be minimized and better controlled with respect to the improved interpretation of slope monitoring and early warning data.

1

Comprehensive reviews of these methods are provided by Bhandari (1988), Glastonbury & Fell (2002), Crosta & Agliardi (2003) and Rose & Hungr (2007). Inherently, these approaches are ‘holistic’, disregarding details pertaining to the underlying slope failure mechanism. Whether the displacement measurements are made using an extensometer positioned across a tension crack or a system of geodetic reflectors across a slope, the analysis is often carried out in the same way—surface displacements are recorded over time, which are then extrapolated or analyzed for accelerations in order to predict catastrophic/impending failure. Generally, the kinematics and causes of failure are not well defined, and instead, the surface manifestation of the instability (i.e. surface displacements) is relied upon for predictive analysis. Not surprisingly, only a few cases have been reported where these techniques have been successfully applied as part of a reliable forward prediction (e.g. Rose & Hungr 2007); most involve back analyses.

INTRODUCTION

Monitoring forms a key component of most landslide hazard assessments, providing data that may be used to quantify the nature of the hazard, its extent, kinematics and stability state, sensitivity to triggering mechanisms, response to mitigation works, etc., or to provide early warning of an impending failure especially those where lives or infrastructure may be at risk. In both cases, issues of uncertainty relating to the geological conditions, slope kinematics and failure mode provide major obstacles that contribute to a lack of definition of the problem. Techniques used for forecasting impending failure (i.e. temporal prediction) are largely phenomenological, relying on surface-based point measurements of displacement monitored over time, which are then extrapolated or analyzed for accelerations that exceed set thresholds based on earlier patterns. Fukuzono’s (1985) inverse velocity method is one such example.

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movements can be measured to several millimeters accuracy (Froese et al. 2005). Temporal decorrelation due to vegetation coverage, however, dramatically affects interferometric coherence and limits the detection resolution, although multi-image based approaches like the Permanent Scatterers (Ferretti et al. 2001) and Small Baseline (Berardino et al. 2002) methods work to limit spatial decorrelation effects and topography errors. Several studies have now been published where coherence for a study site could be maintained and DInSAR successfully applied. Most of these are from slides showing coherent movements over larger areas, often involving rock masses or colluvium with strong plastic deformations (e.g. Rott et al. 1999, Froese et al. 2005, Singhroy et al. 2005). Figure 1a shows deformations detected along a major fault and over old mine workings and colluvium towards the bottom

Numerical modeling offers a means to account for complex subsurface processes by breaking problems down into their constituent parts and analyzing the cause/effect relationships (and their evolution), which govern the behavior of the system as a function of changing environmental factors. However, these analyses require tight controls on the representation of geological heterogeneity and structure, soil and/or rock mass behavior, and special boundary conditions. This information may be derived in part from surface and borehole data, but more often it is limited to subsurface projections based on surface observations. Slope monitoring data provides an important means to calibrate and constrain detailed numerical models. At the same time it must be recognized that most in situ measurements are affected by the same issues of rock mass complexity and variability as the numerical analyses they are meant to constrain. In many situations the interpretation of monitoring data is far from straight forward. In turn, it has been demonstrated that numerical modeling can be used to help constrain interpretations of complex field measurements (Eberhardt & Willenberg 2005, Watson et al. 2006). Thus it must be emphasized that a counterbalance and close association should exist between field measurements and analysis to develop a more complete understanding of the slope hazard problem (Sakurai 1991).

2 2.1

CONTINUING DEVELOPMENTS IN LANDSLIDE MONITORING Remote sensing and InSAR

Although technological advancements continue with respect to traditional total station and prism monitoring systems (e.g. combined robotic total station and global navigation satellite systems; Brown et al. 2007), many of the key recent advances in slope monitoring are those related to remote sensing technologies, primarily satellite and ground-based radar but also non-radar variants like airborne and terrestrial laser scanning (e.g. Rosser et al. 2005). At the regional scale, Differential Interferometric Synthetic Aperture Radar (DInSAR) is proving to be a useful means for identifying landslides within large coverage areas (e.g. 100 × 100 km using ERS data; Meisina et al. 2005) to help in the development of landslide inventories. InSAR uses satellite emitted electromagnetic signals to measure the phase difference resulting from the path length change between satellite passes of the same area taken from the same flight path. The difference in phase can be used to determine ground movement in the line of site of the SAR satellite, and with an emitted electromagnetic wavelength of a few centimeters, these ground

Figure 1. a) Vertical surface deformation map based on satellite DInSAR for Turtle Mountain for a 2-yr period (after Singhroy et al. 2005). b) Line of sight displacements derived from terrestrial radar for Randa for a 59-day period.

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long-term in situ monitoring efforts. However, new technologies like fiber optics are also being tested to capitalize on the stability and insensitivity of fiber optic sensors to external perturbations. A review of new developments in fiber optic sensing technologies for geotechnical monitoring is provided in Inaudi & Glisic (2007). Data reliability is equally a key issue. Recent studies involving cases where the displacements being measured are particularly small or where deep inclinometers are involved, have led to improved algorithms and procedures for identifying and correcting systematic errors (e.g. Mikkelsen 2003, Willenberg et al. 2003). Studies involving complex, deep-seated, rock slope instabilities have seen attempts to better integrate multiple geological and geotechnical data sets to improve data interpretation (e.g. Willenberg 2004, Watson et al. 2006, Bonzanigo et al. 2007, Hutchinson et al. 2007). These attempts at data ‘‘fusion’’ are moving towards the adoption of Virtual Reality (VR) technology, where the identification of hidden relationships, discovery and explanation of complex data interdependencies, and means to compare and resolve differing interpretations can be facilitated (Kaiser et al. 2002). Concomitant with data integration is data management. Important new elements include Web GIS services integrated into the operational resources of decision makers. These services are linked to early warning systems through wireless data acquisition and transmission technologies, which enable realtime data from multiple remote monitoring sites to be accessed and viewed off-site by means of the internet. This is proving to be a highly valuable resource where an unstable slope threatens a community, critical facility or, in the case of large open pit mine slopes, worker safety. Furthermore, Hutchinson et al. (2007) propose that spatially and temporally distributed measurements should be combined with a knowledge engine and an evolving rule base to form the hub of a decision support system. One of the more comprehensive systems in place is that installed by BC Hydro for its chain of hydroelectric dams on the Columbia River in British Columbia, Canada (Fig. 3). The system connects dataloggers at six large landslide sites along different dam reservoirs to a central monitoring computer using radio and microwave communication (Fig. 3c). The landslides sites are separated by up to 150 km and range in size from less than one million to over a billion cubic meters. These include: 731 Block, Checkerboard Creek, Downie Slide, Dutchmans Ridge, Little Chief Ridge and Little Chief Slide. The extensive system in place is used for both investigative and predictive monitoring, and involves the continuous monitoring of a large number of piezometers, in-place inclinometers, extensometers, water weirs, load cells and tiltmeters. The basis of the data

of Turtle Mountain in western Canada, location of the 1903 Frank Slide (Singhroy et al. 2005). These drew attention to a possible link between slope movements in the upper slope and de-stressing of the slope’s toe due to the slow collapse of the old workings (Froese & Moreno 2007). In comparison, brittle rock masses showing smaller-scale complex block movements are more difficult to study with satellite-based DInSAR due to the strong spatial variability in block velocities. Instead, ground-based DInSAR (e.g. Tarchi et al. 2003) with a typical pixel size of a few meters may prove more suitable. Figure 1b shows preliminary results from Randa in southern Switzerland, location of the 1993 Randa rockslide, which are being used to provide important information about the area affected by slope movements and to support kinematic analyses. 2.2

Slope Stability Radar (SSR)

Other new developments in the use of radar have involved moving away from synthetic aperture radar and instead using real-aperture from a stationary platform, as in the case of Slope Stability Radar (SSR). With SSR, the system is typically set up 50 to 1000 m from the foot of the slope and the region of interest is continuously scanned, comparing the phase measurement in each image pixel with previous scans to determine the amount of movement. (Fig. 2; Harries & Roberts 2007). The combination of near real-time measurement, sub-millimeter precision and broad area coverage to quickly identify the size and extent of a developing failure is helping to establish SSR as a key tool for managing unstable rock slopes, especially in open pit mining (e.g. Harries et al. 2006, Little 2006, Day & Seery 2007). Further advantages of the system is that it is not adversely affected by rain, fog, dust or haze. Vegetation on the slope, however, may reduce the precision in pixels where there is low phase correlation between scans (Harries & Roberts 2007). Again, the advantage of these remote sensing-type systems is that they provide fast and updatable data acquisition over broad areas as opposed to point measurements, coincident with prism placement, when using traditional geodetic monitoring systems. 2.3

Wireless data acquisition and data management

Conventional slope monitoring (e.g. inclinometers, total stations, tiltmeters, extensometers, crackmeters, etc.), continues to represent the favored means by which to measure and monitor slope deformation directions, magnitudes and rates, both on surface and at depth. Instrument reliability is of paramount importance, for which continuous improvement to the performance of vibrating-wire technology has led to it being widely recognized as the preferred choice for

41

Figure 2. GroundProbe’s Slope Stability Radar (SSR) system showing the continuous monitoring of millimeter-scale movements across the entire face of an unstable open pit mine slope (after Harries et al. 2006).

Figure 3. a) Example of the detailed instrumentation used by BC Hydro to monitor slope displacements, temperature and pore pressures at depth for a rock slope, Checkerboard Creek, above one of their dam reservoirs (after Watson et al. 2007). b) Monitoring and alarm interface used by BC Hydro to remotely monitor several reservoir slopes along their system of hydroelectric dams on the Upper Columbia River. c) Schematics of the wireless data acquisition/transmission system used for the remote monitoring system.

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construction of a deep drainage adit that successfully led to the stabilization of the landslide (Fig. 4, Eberhardt et al. 2007). A second example is that provided by Watson et al. (2006, 2007) for BC Hydro’s Checkerboard Creek (Fig. 3a). Again, detailed monitoring of the slope revealed a similar persistent annual displacement cycle dominated by an active sliding phase in autumn to late winter and inactivity during the spring and summer. Although at first these periods of displacement activity appeared to correspond to periods of increased precipitation as in the case of Campo Vallemaggia, it wasn’t until several years worth of measurements were collected that it was observed that the annual displacement cycle was repeated each year regardless of the amount of precipitation. Instead, the annual displacement cycle was more strongly correlated to seasonal temperature variations in the near surface bedrock and the deformation mechanism explained in terms of thermally induced slip along sub-vertical joints (Watson et al. 2006). In both cases, the assessments and subsequent approaches taken were completed based on the results of investigative monitoring used to develop reliable geological and hydrogeological models later aided by numerical modeling to more fully understand the deformation mechanisms involved. Predictive monitoring systems usually evolve after investigative monitoring and assessment. Without this investigation phase it is unlikely that a predictive system could be designed, or early warning thresholds set, with confidence. Monitoring information must be assessed in the context of the physical setting and the conclusions of the investigation phase. Landslides can change their behavior within a few weeks or a few days. Thus, for a predictive monitoring system to be effective the frequency at which the instruments are monitored must be a fraction of this response time and the system must be in place to react to the instrumentation results. Furthermore, in order for decision makers to be in a position to react correctly to predictive monitoring data, the chance of faulty alarms or misleading instrument readings should be minimized by the use of the most reliable instrumentation possible.

communication is a mix of UHF radio and spreadspectrum radio which allow the dataloggers to communicate with the central monitoring computer. The fundamentals of the communication system include a transmitter, receiver and surge arrestor which allows reliable communication over distances of a few kilometers, up to 100 kilometers with proper antennae. Advantages at these sites over satellite or cell phone communication include cost and ease of use. Disadvantages include the requirement to use repeaters if line of sight on long distances is not possible.

3 3.1

MONITORING OF LANDSLIDE BEHAVIOUR Current state of practice

Landslide monitoring serves two important functions (Moore et al. 1991): i. Investigative Monitoring: To provide an understanding of the slope and thus enable an appropriate action to be implemented. ii. Predictive Monitoring: To provide a warning of a change in behavior and thus enable the possibility of limiting damage or intervening to prevent hazardous sliding. Instrumentation typically includes: piezometers, inplace inclinometers, extensometers/crack monitoring, tiltmeters and surface geodetic monuments. Often for large landslides the most reliable instruments are those installed in deep subsurface exploratory boreholes (e.g. Fig. 3a). Investigative monitoring can be used to obtain a greater understanding of the slope behavior, thus enabling the correct approach to be taken or to confirm that the approach taken was correct. Monitoring for investigative purposes can be as little as an annual visual inspection or as much as continuous measurement of a comprehensive instrumentation network (e.g. Fig. 3c). For the case of slow moving slides, many years of monitoring and annual cycles may be required to identify relationships between water levels, movement and other seasonal effects such as temperature. This was demonstrated by Bonzanigo et al. (2007) for the case of Campo Vallemaggia, an 800 million m3 deep-seated landslide in strongly fractured and weathered crystalline rocks in southern Switzerland that threatened two villages founded on the landslide. Detailed inclinometer and piezometer measurements collected over a five year period were used to cross-correlate the stick-slip behavior of the landslide with pore pressures exceeding a threshold value tied to longer-term precipitation events (Fig. 4). These measurements were subsequently used in the decision making process to go forward with the

3.2 In situ rockslide laboratories Efforts to improve predictive monitoring have seen several recent multi-disciplinary studies focused on improving our understanding of complex landslide deformation mechanisms. These include the Randa In Situ Rockslide Laboratory in Switzerland, the Turtle Mountain Field Laboratory in western Canada and the Åknes/Tafjord Project in Norway. The first of these, the Randa In Situ Rockslide Laboratory (Fig. 5), was a comprehensive experimental investigation into the spatial and temporal evolution of large rock slope failures in fractured crystalline rock.

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Figure 4. Correlation between the downslope velocities of the Campo Vallemaggia landslide and borehole pore pressures measured before and after the opening of a drainage adit to stabilize the slope. Slide velocities were measured using an automated geodetic station; pore pressures are expressed as the hydraulic head in the piezometer (after Eberhardt et al. 2007).

define the rock mass structure (e.g. Fig. 6), aided in the positioning of borehole instruments (e.g. inplace inclinometers) and was essential for reliably interpreting the monitoring data. A similar multi-disciplinary field campaign was carried out for the Åknes/Tafjord Project, where up to 30–40 million m3 of unstable rock moving with a mean rate of 2–4 cm/year has been identified as a potential threat and tsunami generating hazard for people and infrastructure living along the inner Storfjord (Blikra et al. 2005). Geological, geodetic and geophysical studies (including GPS, resistivity, georadar, reflection and refraction seismics, airborne laser scanning, and high-resolution air photography) were carried out to define the geometry and volume of unstable areas. Through these detailed studies it was found that a previously unmonitored section of the slide was moving at 15 cm/year, leading to revisions in the unstable volume and thus the magnitude of the potential hazard (Roth et al. 2006). The fracture network, both existing and newly generated, was a central focus in the designs of the instrumentation networks at Randa, Åknes and Turtle Mountain. In different ways, displacement monitoring targeted resolving the complex displacement field generated by multiple moving blocks. At Turtle Mountain, a combination of crackmeters, wire-line extensometers and tiltmeters, were used to monitor surface tension cracks, enabling seasonal displacement patterns to be resolved (Moreno & Froese 2007). Similar

One of the prime motivating factors was to develop a better understanding of rock mass strength degradation (e.g. through the destruction of intact rock bridges between non-persistent discontinuities) and the progressive development of internal shear zones, and their accommodation of larger slope displacements, leading to increased extensional strains and eventually sudden collapse (e.g. Eberhardt et al. 2004b). Focus was also placed on improving early warning capabilities in the presence of persistent and non-persistent discontinuities, and multiple moving blocks and internal shear surfaces. For this, a high-alpine facility was constructed above the scarp of the 1991 Randa rockslide (Fig. 5a), where ongoing movements of 1–2 cm/year are being recorded in gneissic rock for a volume of up to 10 million m3 . This facility included the installation of a variety of instrumentation systems designed to measure temporal and 3-D spatial relationships between fracture systems, displacements, pore pressures and microseismicity (Fig. 5). The monitoring was complimented by a detailed geophysical field campaign, which included 3-D surface seismic refraction and georadar surveys to resolve subsurface 3-D fracture distributions (Heincke 2005), and crosshole georadar and seismic tomography to identify key geological features (Spillmann 2007). These were then compared to those mapped on surface and in the boreholes to develop a 3-D geological model of the unstable rock mass (Willenberg 2004). This information helped to

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Figure 5. The Randa In Situ Rockslide Laboratory in southern Switzerland. a) Area of investigation (solid white line) and outline of present-day instability (black dashed line) above the scarp of the 1991 rockslide. Photo by H. Willenberg. b) Installation of surface and subsurface monitoring instruments and central data acquisition station housing batteries, power generation sources (solar and wind) and data acquisition and transmission hardware. c) Plan view map showing location of boreholes, geodetic reflectors and geophones relative to the active slide area and open tension cracks (after Willenberg 2004).

at Åknes and Randa. At Randa, the deepest inclinometer was also fitted with an Increx electromagnetic induction sliding extensometer system to enable the profile of 3-D displacement vectors at depth to be determined. Biaxial vibrating wire in-place inclinometers were installed and positioned at depth intervals coinciding with key fractures identified through the borehole televiewer surveys to provide continuous monitoring of subsurface deformations along these structures. Integration of these different displacement data sets with the 3-D geological model showed that the displacements recorded on surface and at depth were localized across active discontinuities (Fig. 7) and that the kinematic behavior of the slope was dominated by complex internal block movements rather than those of a coherently-sliding mass (Willenberg 2004). Willenberg et al. (2003) demonstrated that to resolve these complex displacement patterns, rigorous correction algorithms must be carried out to attain the requisite resolution. The fracture network also plays a controlling role with respect to the distribution of pore pressures at depth and their coupled relationship with unstable rock slope movements. The design of the experimental monitoring networks at Randa, Åknes and Turtle Mountain each included borehole monitoring of pore pressures to correlate with measured displacements. At Randa, piezometers were positioned and packed off along zones indicating potentially higher fracture

Figure 6. Cross-section through the Randa study area showing the network of discontinuities (F-1 to F-3) and faults (highlighted in black) mapped using geological and geophysical methods (after Loew et al. 2007).

monitoring was carried out at the Åknes and Randa project sites. This traditional focus on surface displacement monitoring addresses certain logistic and economic realities in terms of what may be feasible for on-site monitoring, yet it must also be asserted that only so much can be inferred on surface for a problem that develops at depth. For this, deep inclinometer measurements were added to the monitoring networks

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Figure 7. Integrated borehole data set for the 120-m deep borehole at Randa, showing from left to right: fracture frequency log, optical televiewer log (highlighting traces of major fractures), cumulative inclination changes for a two-year period (and corresponding preliminary kinematic interpretation), cumulative axial displacements for the same two-year period, and corresponding 3-D displacement vector magnitudes and orientations (after Willenberg 2004).

ditions at ground surface is observed in landslides showing artesian pressures at depth. The final key monitoring strategy common to the Randa, Åknes and Turtle Mountain projects was the use of microseismic monitoring to detect and study subsurface brittle fracture processes. Spatially clustered microseismic events in numerous fields (e.g. mining, geothermal energy, nuclear waste disposal, etc.) have proven effective in providing critical information with respect to stress-induced tensile fracturing mechanisms and/or shear slip along internal fracture planes. The microseismic network at Randa was the most detailed of the three and included three geophones (28 Hz) mounted in deep boreholes, nine geophones (8 Hz) mounted in shallow boreholes, and

permeability as determined from borehole televiewer data. These data showed several water tables distributed within the rock mass and different types of pore pressure interactions with infiltrating surface water and atmospheric pressure variations (Willenberg 2004). This is a common feature for most deep-seated slides in crystalline rock, where preferential fracture permeability and hydraulic barriers (e.g. from fault gouge) result in isolated compartments of groundwater flow and reaction delays between surface precipitation and pressure responses at depth, making correlations between slope movements and precipitation events extremely difficult (e.g. Moore & Imrie 1992, Bonzanigo et al. 2007). A strong sensitivity of slope movements to changing groundwater recharge con-

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Figure 9. Cross-section showing the relationship between microseismic activity (high PDF values indicate microseismogenic zones), faults (dashed lines) and the approximate limit of slope movements. The cumulative PDF represents the sum of the hypocenter probability density functions for all microseismic events. After Spillmann et al. (2007).

Figure 8. Vertical components of a locatable microseismic event: a) raw; b) 100–500 Hz bandpass filtered signals. Signals are sorted according to the source-receiver distance, with sen-sor A1 being the farthest and B5 the closest. Absolute time scale is arbitrary. After Eberhardt et al. (2004a).

two 24-channel seismographs (Fig. 5c). The spatial distribution of the twelve triaxial geophones was chosen to ensure that the hypocenter parameters generated from the seismic sources could be reliably constrained within the area of interest (Spillmann et al. 2007). One of the initial findings from this system was that the higher frequency content of the recorded microseismic events was strongly attenuated (Fig. 8), pointing to the presence of large open fractures at depth. This was fully compatible with the geological model and borehole televiewer and pore pressure data (Willenberg et al. 2004). Larger low frequency events, such as those generated from natural seismic activity in the region, did not suffer as much from signal quality degradation. Interpretation of the recorded local microseismic activity involved comparing the cumulative hypocenter probability density functions, which incorporates uncertainties in the arrival times and travel times for each event (see Spillmann et al. 2007), with the results from the geological, geophysical and geotechnical investigations. From this, it was observed that the microseismic activity was concentrated in two main zones (Fig. 9): that near the scarp of the 1991 rock slide events and that coinciding with the highest density of faults (Spillmann 2007). These two zones are bound by the geodetically determined limits of the moving mass (e.g. Fig. 5c). A full interpretation of the microseismic activity recorded during the Randa experiment is reported in Spillmann et al. (2007). The system at Turtle Mountain also combines surface- and borehole-mounted geophones, and builds on earlier experiences with seismic monitoring at Turtle Mountain carried out between 1983 and 1992. These identified several different sources including

Figure 10. UDEC modeled thermal-induced horizontal displacements compared to inclinometer measured slope displacements (after Watson et al. 2006).

local earth tremors and rock falls, together with seismicity believed to be related to deformation and stress relief within Turtle Mountain and the ongoing collapse of mine workings at the base of the mountain (Read et al. 2005). Similar attempts will be made at Åknes to classify the microseismic events recorded by their system, based on rock falls, small-scale slides, and those directly related to the main slide body (Roth et al. 2006). As with Randa, signal quality has been noted as being a key limiting factor with respect to constraining source locations and mechanisms. It is expected that the application of emerging monitoring technologies like those used in the Randa, Åknes and Turtle Mountain project studies, will provide an improved means to gain a better understanding of rock slope deformation kinematics and failure.

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Figure 11. Distinct-element modeling of complex rock slope displacements at Randa and comparison between measured and modeled cumulative displacement profiles, assuming slip along discontinuities and elasto-plastic block deformation. Note that model boundaries extend beyond those shown (as indicated by the dashed boundary line).

displacement patterns induced during thermal cycling and those measured in situ by the instrumentation (Fig. 10). Results for the latter revealed that the cyclic nature of the displacements seen in the monitoring data was controlled by the thermal response of the rock mass to seasonal temperature changes in the upper 10 m of ground. This is similar to preliminary findings at Turtle Mountain, which suggest that thermal cycling (i.e. thermal induced stress changes) contribute more towards measured slope deformations than do heavy precipitation events (Moreno & Froese 2007). Thus, not only does monitoring data provide an important means to constrain numerical models, but numerical modeling provides a means to better understand monitoring data (Eberhardt & Willenberg 2005). At both Randa and Åknes, simplified numerical modeling has been carried out together with the mapping and monitoring programs to help identify and constrain possible sliding surface/instability scenarios that would produce displacement patterns similar to those measured in situ. Figure 11 provides an example from a series of distinct element models generated for the Randa study, incorporating the key active geological structures identified through mapping and geotechnical monitoring. The blocks are modeled using a MohrCoulomb elasto-plastic constitutive model where the properties are scaled to those for an equivalent continuum to account for smaller scale discontinuities not explicitly included in the distinct-element model. Results show a correspondence between the measured and modeled block movements with toppling and translational movements in the upper part of the slide and outward rotation of the blocks at depth. The latter suggests that deep-seated yield together with shearing along persistent discontinuities may be an important

At Turtle Mountain and Åknes, these are being integrated into early warning monitoring and response plans that comprise detailed monitoring procedures, threshold and alert level development, notification protocols and emergency response (e.g. Froese et al. 2005). Froese et al. (2005) conclude that the use of multiple systems that provide different spatial and temporal coverage, together with sufficient redundancy, provide a higher level of confidence in interpreting the kinematics of movement and impending failure than would be available from single sensor readings. 3.3 Use of numerical models to interpret slope monitoring responses Numerical analysis can be a useful tool to provide confirmation of the geological model and/or conclusions drawn from investigative monitoring, as well as to explore possible future behavior. An example of this is the numerical analysis completed for the Checkerboard Creek rock slope to explore and confirm the indications from investigative monitoring that a thermal control existed for the deep-seated rock slope movements being measured (Watson et al. 2006). Modeling was initially carried out using the Itasca finite-difference code FLAC, but results suggested that a continuum approach could not capture the pattern of displacements and pore pressures measured in situ. Itasca’s distinct element code UDEC was next used because of its capability to explicitly include joints and shear zones, with groundwater flow restricted to those joints, together with its ability to model thermal and dynamic loads. Models investigated the stability of the slope during a 1 in 10, 000 year earthquake, as well as correlations between modeled

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field mapping and instrumentation data to constrain numerical models but also in the use of numerical modeling to provide a means to better interpret and understand complex monitoring and early warning data. Thus, by better integrating the different data sets collected through all phases of an investigation, from mapping to monitoring to analysis, geological uncertainty can be minimized and controlled with respect to the comprehension of complex rock slope failure mechanisms, thereby improving our ability to effectively assess, monitor, mitigate and predict the potential for catastrophic rock slope failure and provide early warning to those endangered.

contributing factor in the complex block deformation patterns measured. It should be noted that uncertainty does exist for the inclinometer readings over the lower portions of the borehole, and the modeling results incorporate the limitations inherent in a 2-D representation of a strongly 3-D problem. Still, any insights gained into the instability mechanism, either supporting or refuting a current interpretation, provides a means to better plan and design future in situ investigation, instrumentation and monitoring schemes for the site. 4

CONCLUSIONS

Issues related to geological complexity and uncertainty represent a significant obstacle to better predicting the spatial and temporal evolution of catastrophic rock slope failures. Advances in satellite and terrestrial radar technologies (e.g. Slope Stability Radar) and automation through wireless data acquisition are helping to address these issues. However, the findings summarized in this paper emphasize the additional need to not only improve landslide monitoring technologies but to also better integrate the various data sets collected during field-based investigations, monitoring and stability analyses in order to overcome these challenges. Through the integration of these data, a more reliable model of the controlling landslide mechanism can be incorporated into the engineering decision-making process. Several examples are provided from recent studies that aim to improve our mechanistic understanding of deep-seated rock slope behavior. These include experimental efforts involving the construction of high alpine research facilities or ‘‘in situ rockslide laboratories’’, where the integration of state-of-the-art site characterization and instrumentation systems and numerical modeling are being used to better understand the mechanisms controlling pre-failure deformations over time. At depth, passive monitoring of microseismic activity offers a means to detect subsurface tensile fracturing and/or shear slip along internal fracture planes that may provide insights into the evolution of a progressively developing rock slope failure. Data from both surface geodetic and subsurface instrumentation systems can be integrated to obtain a description of the 3-D displacement field. When further integrated with subsurface geological information, derived through geophysical and/or borehole investigations, complex block movements can be resolved relative to major persistent fractures and shears. The results presented also emphasize the importance of numerical modeling to provide support for and/or refute interpretations drawn from investigative monitoring as well as to explore possible future behavior. Focus is placed not only on using

ACKNOWLEDGEMENTS The authors wish to acknowledge and thank the different researchers, whose results were drawn on for the material presented in this paper, including those at BC Hydro, GroundProbe and those connected to the Randa, Åknes and Turtle Mountain rockslide research studies. Special thanks are extended to the Randa In Situ Rockslide Laboratory team, including Dr. Heike Willenberg, Dr. Keith Evans, Dr. Hansruedi Mauer, Dr. Tom Spillmann, Dr. Björn Heincke, Prof. Alan Green and Prof. Doug Stead.

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Journal of Engineering Geology & Hydrogeology 38: 363–375. Roth, M., Dietrich, M., Blikra, L.H. & Lecomte, I. 2006. Seismic monitoring of the unstable rock slope site at Åknes, Norway. In Gamey (ed.), Proc. 19th Annual Symp. on the Application of Geophysics to Engineering and Environmental Problems (SAGEEP), Seattle. CD-ROM. Rott, H., Scheuchel, B. & Siegel, A. 1999. Monitoring very slow slope movements by means of SAR interferometry: A case study from mass waste above a reservoir in the Otztal Alps, Austria. Geophys. Res. Lett. 26 (11): 1629–1632. Sakurai, S. 1991. Field measurements versus analysis in geotechnical engineering problems. In Sorum (ed.), Field Measurements in Geomechanics, Proc. of the 3rd Int. Symp., Oslo. Rotterdam: Balkema, v3, pp. 405–414. Singhroy, V., Couture, R. & Molch, K. 2005. InSAR monitoring of the Frank Slide. In Hungr et al. (eds.), Landslide Risk Management, Proc. of the Int. Conf., Vancouver. Leiden: Balkema, pp. 611–614. Spillmann, T. 2007. Borehole radar experiments and microseismic monitoring on the unstable Randa rockslide (Switzerland). D.Sc. thesis, Applied and Environmental Geophysics, Swiss Federal Institute of Technology (ETH Zurich). 205 pp. Spillmann, T., Maurer, H., Green, A.G., Heincke, B., Willenberg, H. & Husen, S. 2007. Microseismic investigation of an unstable mountain slope in the Swiss Alps. J. Geophys. Res. 112(B07301): doi:10.1029/2006JB004723. Tarchi, D., Casagli, N., Moretti, S., Leva, D. & Sieber, A.J. 2003. Monitoring landslide displacements

by using ground-based synthetic aperture radar interferometry: Application to the Ruinon landslide in the Italian Alps. J. Geophys. Res. 108(B8): 2387, doi:10.1029/2002JB002204. Watson, A.D., Martin, C.D., Moore, D.P., Stewart, T.W.G. & Lorig, L.L. 2006. Integration of geology, monitoring and modeling to assess rockslide risk. Felsbau 24 (3): 50–58. Watson, A.D., Moore, D.P., Stewart, T.W. & Psutka, J.F. 2007. Investigations and monitoring of rock slopes at Checkerboard Creek and Little Chief Slide. In Eberhardt et al. (eds.), Proc. 1st Canada-U.S. Rock Mechanics Symposium, Vancouver. London: Taylor & Francis, v2, pp. 1015–1022. Willenberg, H. 2004. Geologic and kinematic model of a complex landslide in crystalline rock (Randa, Switzerland). D.Sc. thesis, Engineering Geology, Swiss Federal Institute of Technology (ETH Zurich). 187 pp. Willenberg, H., Evans, K.F., Eberhardt, E. & Loew, S. 2003. Monitoring of complex rock slope instabilities—correction and analysis of inclinometer/extensometer surveys and integration with surface displacement data. In Myrvoll (ed.), Proc. 6th Int. Symp. on Field Measurements in Geomechanics, Oslo. Lisse: A.A. Balkema, pp. 393–400. Willenberg, H., Evans, K.F., Eberhardt, E., Loew, S., Spillmann, T. & Maurer, H.R. 2004. Geological, geophysical and geotechnical investigations into the internal structure and kinematics of an unstable, complex sliding mass in crystalline rock. In Lacerda et al. (eds.), Proc. 9th Int. Symp. on Landslides, Rio de Janeiro. Leiden: A.A. Balkema, pp. 489–494.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Effects of earthquakes on slopes Ikuo Towhata University of Tokyo, Tokyo, Japan

Tetsuo Shimomura Ohbayashi Corporation, Tokyo, Japan

Masanori Mizuhashi Public Works Research Institute, Tsukuba, Japan

ABSTRACT: This paper First addresses examples of large and small slope failures that were triggered by earthquakes. Their significance comes from the negative effects to the human community, which consist not only of the number of casualties but also of the difficulties in post-earthquake rescue and restoration. To mitigate the problems, identification of seismic instability and assessment of debris fun-out distance are important. Because the existing practical methods for these have several problems, the direction of their improvement is presented with examples and case studies. 1

the failed slope, and 300 m in depth. The total volume of the failed mass is evaluated to be 24 billion m3 . This event occurred in a prehistoric time, probably 10,000 years ago, according to C14 dating (Watson & Wright 1969). Oberlander (1965) attributed this event to the foot erosion by river, but it is possible that the unstable slope was finally destroyed by seismic shaking. The important feature of this landslide lies in the enormous run-out length of debris, which traveled as long as 20 km, overtopping an anticline hill (Fig. 2). Gigantic slope failures in more recent times often claimed many victims. Figure 3 demonstrates the ruin of Yungay City of Peru that was destroyed by seismic failure of Huascaran Mountain slope upon an earthquake of magnitude = 7.9 in 1970. The number of victims exceeded 17, 000. When a similar slope failure occurred in 1960, the debris flowed along a nearby Rio Santa channel, and the city was protected from it by a hill behind the city (Fig. 3). This experience gave people a wrong idea that debris would never hit the city in future. After the 1970 tragedy, the entire city was relocated to a safer place. One of the consequences of a gigantic landslide is creation of a natural dam. The slope of Tsao-ling in Taiwan failed several times in a large scale in the past: in 1862 and 1941 due to earthquakes and in 1942 due to rainfall (Kawata 1943). Figure 4 shows a lake after the 1999 ChiChi earthquake. Although a lake thus created may become a good tourist spot, problem is the possibility of the collapse of the dam and flooding in the downstream region.

INTRODUCTION

Slope failure is one of the most significant problems during strong earthquakes. It is important in that it might claim thousands of lives and destroy both public and private properties. When local transportation is blocked by slope failures, moreover, rescue and restoration become significantly difficult. It is true, on the other hand, that slope failure is a part of geomorphological processes and may be beyond the human control. Therefore, all what human beings can do is not to worsen the risk of slope failure, but to mitigate negative effects to the human community. The present paper addresses what happened in slopes during past and recent earthquakes, their effects to the public, and various kinds of efforts to mitigate the negative effects. Note that attention is focused mostly on natural slopes, but some events in manmade slopes will be introduced in relation to recent urban developments. 2

EXAMPLES OF ON-SHORE LARGE SLOPE FAILURES DURING EARTHQUAKES

This chapter addresses different kinds of seismicallyinduced slope failures that occurred both on shore and in the sea, in either big or small scale, and their consequences to the human communities. The discussion is initiated with the largest on-shore landslide in the world. Seimareh landslide is located in south-west Iran and measures 16 km in width (Fig. 1), 5 km in length of

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Figure 1. View of Seimareh landslide from Pol-e-Doghtar.

Figure 2. Debris deposit in Seimareh.

Figure 4. Natural dam in Tsao-ling after 1999 ChiChi earthquake.

Figure 3. Ruin of Yungay City in Peru.

Figure 5. Statistics on cause of failure of natural dams (after Tabata et al., 2002).

Tabata et al. (2002) reported that the dam collapse due to overtopping is substantially more frequent than piping failure (Fig. 5). Another negative consequence is the long-term slope instability after the seismic failure. Fig. 6 shows a part of Ohya slide in the upstream area of Abe River, Japan. This slope collapsed upon the 1707 Hoei earthquake (M = 8.4). Since then, the destabilized slope has been producing debris flow at heavy rainfalls frequently (Imaizumi et al. 2005). Consequently, the river floor has raised significantly (Fig. 7) and the chance of flooding and overtopping river dikes

has increased. To mitigate this situation, continuous efforts have been made to improve dikes and stabilize slopes.

3

PROBLEMS OF ON-SHORE MINOR SLOPE FAILURES

When the surface of a slope is subjected to weathering and disintegration, strong earthquake shaking triggers

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Figure 6. Ohya slide in Shizuoka Prefecture. Figure 9. Seismic failure of road embankment during 2004 Niigata-Chuetsu earthquake, Japan.

Instability problems occur in artificial fills as well. Figure 9 illustrates an example in which a road embankment, which was constructed on a slope, failed upon an earthquake. Many local roads are subjected to a similar problem. The cause of the problem is that insufficient construction budget leads to the fill construction without removing the soft surface deposits. Those debris and organic materials at the bottom of the road embankment cannot resist against the seismic force, and the entire embankment collapses. What is important is the choice between earthquake resistance and money saving. In case the seismic failure would not be fatal, it may be reasonable to allow such a failure from the viewpoint of performance-based design and the policy of minimizing the life cycle cost. A similar but more serious problem lies in residential development. Around a sprawling mega city, inexpensive residential development is needed for lowincome citizens, and such a development is conducted in hilly areas. Since cost has to be saved, soft valley deposits may not be removed, and, upon earthquakes, private lands and houses collapse. Such a loss of private properties is never allowed by limited budget of residents.

Figure 7. Valley of Abe River filled with debris.

4 Figure 8. Surface failure of weathered slope (Kashmir of Pakistan in 2005).

PROBLEMS OF SUBAQUEOUS SLOPE FAILURES

It is possible that earthquakes trigger failure of submarine slopes. In addition to the famous examples of Grand Banks in 1929 (Heezen & Ewing 1952) and off Orléansville in 1954 (Heezen & Ewing 1955), the 1964 Alaska earthquake of magnitude = 9.2 destabilized the sea bed deposit in front of Valdez (Fig. 10). A huge volume of soil mass (75 million m3 according to Coulter & Migliaccio (1966)) collapsed upon shaking, and this mass movement caused tsunami in the

falling of the surface soils. Figure 8 illustrates a slope of weathered limestone near Muzaffarabad, Pakistan, in 2005. The surface slide can easily destroy roads and prevent local transportation. Accordingly, emergency rescue and restoration become very difficult.

55

Figure 10. Aerial view of Valdez (photograph taken by Dr. K. Horikoshi). Figure 11. Japan).

Valdez Bay. This tsunami destroyed the Valdez municipality and the town was later relocated in a safer place. The problem was that the submarine deposit of the Valdez Bay was most probably composed of fine nonplastic silt that was produced by the glacial erosion and was transported into the sea by the river in Figure 10. Being fine, silt grains deposit in water very slowly and the resultant density becomes very low. Since there is no cohesion, moreover, its undrained shear strength is very low. The possible excess pore water pressure during undrained shear does not dissipate during shaking, because the grain size is small. These situations made the bay deposit highly vulnerable to undrained failure and liquefaction. Moreover, it should be recalled that the highest tsunami so far recorded was triggered by rock slide impact induced by an earthquake (in Lituya Bay of Alaska: Miller 1960). Stability of submarine slopes is reduced due to rapid rate of sedimentation (or human reclamation). If strong earthquake loading is superimposed on such a situation, slope failure is likely to occur. Note that a submarine landslide is one of the causative mechanisms of tsunami as was the case in North Sea (Long et al. 1989).

5

Reinforcement of cliff behind house (Kamakura,

2. Relocation of human settlement from potentially hazardous areas 3. Emergency evacuation The first choice is costly as stated above. The third choice may not be rapid enough during an earthquake. Both (1) and (2) require identification of potentially unstable slopes with practically reasonable cost. This assessment of failure risk may be made by determining shear strength of soil experimentally or by field investigation, and then running a stability analysis with reasonable seismic load taken into account. This procedure is unfortunately costly and can be conducted only in special situations. Hence, less expensive and time-efficient methodology is required, although the accuracy may be sacrificed to some extent. The identification of potentially hazardous area for (2) further requires assessing the run-out distance of the failed mass. Fig. 12 compares two slope failures that were caused by the 2004 Niigata-Chuetsu earthquake. The one in front was of 110,000 m3 in soil volume and traveled over some distance at the bottom of the valley. Consequently, this soil mass stopped the river flow and formed a natural dam. In contrast, the surface failure in the same figure had approximately 50 cm in thickness, suggesting that only the surface weathered soil fell down. This smaller soil mass did not travel laterally in the bottom of the valley and did not therein affect either the river flow or road traffic. Thus, the flow characteristics of soil mass depends at least partially on its volume as pointed out earlier by Hsü (1975). Furthermore, the effect of water content is illustrated in Fig. 13 in which a dry cliff in Muzaffarabad of Pakistan fell down upon the 2005 Kashmir earthquake. Noteworthy is that the soil did not translate laterally and houses in the bottom of the valley were

MITIGATION OF DAMAGE CAUSED BY EARTHQUAKE-INDUCED SLOPE

The previous sections introduced examples in which human communities were seriously damaged by seismic slope failures. Although it is advisable to stabilize all the potentially hazardous slopes (Fig. 11), financial restrictions do not allow it. Accordingly, there are three alternative options to be taken: 1. Reinforcement of slope by means of retaining walls, ground anchors, and other reinforcements

56

Figure 12. Failure of slope at Naranoki in Yamakoshi Village upon 2004 Niigata-Chuetsu earthquake (Mizuhashi et al., 2006).

Figure 14. Maximum possible epicentral distance of failed slopes during past earthquakes (Yasuda, 1993).

Figure 13.

would occur in a 500 m × 500 m grid during a future earthquake (Table 2). The weight points in Table 1 were determined by regression analyses of past experiences. Therefore, some consistency is expected between evaluation and reality. There are problems, however, in Table 1. First, Table 1 is not intended to evaluate the safety of a particular slope. Second, the weight point for hardness of rock (W4 ) increases as the rock becomes harder, implying more risk. This strange nature comes from the regression data in which failures in hard rock slopes were more numerous than those in soil, probably because slopes of weak soil did not exist or had failed during heavy rainfalls or in other circumstances prior to earthquakes. Consequently, Table 1 does not have a reasonable consideration of soil or rock material properties. To solve those two problems mentioned above, an attempt was made to improve the method by taking into account the characteristics of soil. The weight in Table 1 was counted for individual slopes that failed during the 2004 Niigata-Chuetsu earthquake and their calculated risk was compared with reality (Mizuhashi et al. 2006). The results in Figure 16 are understandable to a certain extent. The good point is that the assessed rank of risk changes with the increasing volume of landslide mass. Since larger events are associated with greater risk assessment, the results are good. However, one of the biggest events at Naranoki is of lower assessed risk, and there is a need for further improvement. The study puts emphasis on the effect of water. In particular, the deterioration of properties due to

Collapse of dry slope in Muzzafarabad, Pakistan.

not affected. This nature is different from the behavior of bigger soil mass in Fig. 12 that was of high water content. 6

ASSESSMENT OF SEISMIC RISK DUE TO SLOPE FAILURE

Any seismic risk depends primarily on the regional earthquake activities. Risk in seismic countries is higher than that in nonseismic countries. This feature is expressed in an empirical diagram in Figure 14 where the maximum distance of slope failure from an earthquake epicenter is plotted against the earthquake magnitude. In case a concerned slope lies within the critical epicentral distance, the failure risk has to be evaluated. Financial reasons, however, eliminate in most cases the use of in-situ tests or laboratory shear tests on shear strength of undisturbed samples collected from a concerned slope. Accordingly, risk is judged on the basis of easily-available information. Table 1 shows an example in which various site factors are evaluated and weight points from them are added to make the final evaluation on the number of slope failures that

57

Table 1. Weighting for factors related to slope instability (See Fig. 15 for W : Kanagawa Prefectural Government, 1986).

Table 2. Assessed number of slope instabilities per 500 m × 500 m grid (Kanagawa Prefectural Government 1986).

Figure 15.

2004. The significant difference in strength was probably caused by deterioration or hydration of soil minerals upon submergence. Furthermore, Figure 18 compares both peak and residual strengths of different soils to confirm the submergence-induced deterioration. With these in mind, it was attempted to add a water-concerned index to risk assessment, which is obtained without running expensive shear tests. The present study employed swelling characteristics and plasticity index. For swelling tests, in-situ soil grains were ground to the size less than 75 microns so that water effects would occur rapidly. Then, specimens for standard oedometer tests were reconstituted in the laboratory. Dry soil was compacted in two layers, with 20 impacts per layer, so that the dry mass density would be 1.061 g/cm3 (60 g for a cylindrical sample). After compression under 20 kPa for 1,000 minutes in a dry state, specimens were submerged in water and the volume change was recorded (Fig. 19 for example). After some contraction occurred during the first 30 minutes, volume expansion (swelling) was observed in water-sensitive soils. Note that the extent of swelling was remarkably variable; some soils exhibited no volume expansion upon submergence. The other index is plasticity index (liquid limit– plastic limit in %). Although this index is not directly related with mechanical properties of soils, it stands for the extent of mineral-water interaction and also is easy to measure in the laboratory. Figure 20 suggests

Vertical cross section of slopes.

Figure 16. Seismic hazard assessment of slopes by using Kanagawa Prefectural method.

water submergence is focused on. Figure 17 illustrates drained shear behaviors of air-dry and watersubmerged specimens of soil collected from the Urase site where rainfall-induced landslide occurred in July,

58

Figure 17. Drained triaxial compression tests on air-dry and submerged specimens.

Figure 19. Water submergence of soil powder collected from Haguro Tunnel site in Niigata-Chuetsu.

Figure 20. ity index.

Correlation between swelling strain and plastic-

Figure 18. Reduction of shear strength of soils collected from rainfall-induced landslide sites.

that there is a positive correlation between swelling strain and plasticity index. Figure 21 demonstrates that the swelling strain is greater for soils collected from sites of greater slope failure. However, the correlation is not clear in this figure because the volume of the soil mass is also affected by the topography and other geological conditions. Thus, swelling strain may be used as one of the indices to better assess the risk of earthquake-induced slope failure. Similarly, Fig. 22 illustrates that there is a positive correlation between the landslide volume and plasticity index, if such exceptional cases as sandy slopes and a reactivated landslide are eliminated from this figure. On the basis of Figures 21, 22, the point, W , in Table 2 was adjusted by employing either W +

Figure 21. Correlation between landslide mass volume and swelling strain.

(swelling strain, %)/10 or W + (plasticity index)/100. The results are shown in Figures 23, 24. The consistency with the damage extent and the determined rank are better than that in Figure 16.

59

Figure 25. Apparent friction angle of landslides obtained from geometry. Figure 22. Correlation between landslide mass volume and plasticity index.

Figure 26.

Figure 23. Correction of Kanagawa Prefectural method by swelling strain.

This diagram was drawn by using data collected by the present study. As is well known, the apparent friction angle, φ, or H /L = tan(φ), decreases as the volume of landslide mass increases. A second approach to assess the travel distance is the use of Newmark sliding block analogy (Newmark 1965). It is more advanced than the foregoing method because it employs the shear strength of soil and the time history of (design) earthquake acceleration. However, there is not yet a clear agreement on what kind of shear strength of soil should be used out of cyclic strength, undrained monotonic strength or else. In the present study, the Newmark method was used to reproduce the gigantic failure of Tsao-ling slope in Taiwan which was triggered by the 1999 ChiChi earthquake. The appearance of a part of the slide is shown in Figure 26, and the geological cross section is presented in Figure 27. It may be found that the slope is made of interbedding sandstone and mudstone (shale). To determine the shear strength of those materials, block samples were collected from the remaining part of the failed slope after the quake. Direct shear tests revealed that the mudstone was of the least shear strength and was likely to be the cause of the failure (Towhata et al. 2002). Figure 28 indicates the stress-displacement relationship of mudstone specimens. It is therein found first that both peak and residual strengths increase

Figure 24. Correction of Kanagawa Prefectural method by plasticity index (Ip ).

7

Tsao-ling slide after 1999 ChiChi earthquake.

ASSESSMENT OF RUN-OUT DISTANCE OF DEBRIS

The extent of risk due to earthquake-induced landslide depends not only on the instability of a slope but also on the travel distance of a failed debris mass. In this regard, attention has to be paid to the dynamic characteristics of debris. Figure 25 illustrates the classical concept of the apparent friction angle by Hsü (1975).

60

Figure 27. Geological cross section of Tsao-ling slope prior to 1999 event (Ishihara, 1985).

Figure 29.

Acceleration time history for Tsao-ling analysis.

Figure 30. kPa.

Calculated sliding displacement with c = 417

by substituting Amax into an empirical formula by Noda (1975),

Kh =

 3

Amax /g/3

(1)

in which g stands for the gravity acceleration. By assuming the slope angle of 14 degrees and the thickness of the sliding rock slab to be 80 m, the pseudo-static factor of safety was obtained to be 0.73. In order to conduct the Newmark-type analyses, the CHY028 acceleration records were converted to a component parallel to the slope (Fig. 29). Then the time history of sliding displacement was calculated. Figure 30, however, shows that the calculated displacement was far less than the reality. Furthermore, the displacement terminated at the end of strong shaking, which is contradictory to the complete failure of the slope in reality. Sassa et al. (2004) used a ring shear device to study in more details the behavior of soil undergoing rapid flow. The problem in underestimation of displacement is that the real failed rock mass was broken into pieces, as will be shown in the next chapter (Fig. 32), and the assumed strength parameters were not relevant. In particular, broken debris hardly had cohesion. Therefore, a second analysis was carried out by using c = 0 kPa, while using the same friction angle. The result in Figure 31 shows that displacement increases towards infinity even after the end of shaking.

Figure 28. Direct shear tests on Tsao-ling mudstone specimen (Towhata et al. 2002).

with the confining pressure. A second finding is that a cycle of unloading and reloading after the post-peak softening does not accomplish the peak strength again. Since the Tsao-ling slope experienced instability several times in the past, the present study decided to use the residual strength for stability analyses. The employed strength parameters were c = 417 kPa and φ = 19.4 degrees. The first analysis was pseudo-static, in which the maximum acceleration (Amax ) recorded at a nearby observatory (CHY028 at 8 km distance) of 748 cm/s2 was converted to a seismic coefficient of Kh = 0.305

61

Figure 31. cohesion.

8

Calculated sliding displacement without Figure 32. slope.

Debris deposit at the bottom of Tsao-ling failed

Figure 33.

Distortion of debris flow with time.

Figure 34.

Layered model of debris flow.

Figure 35.

Internal and base friction angles.

NUMERICAL ANALYSIS ON LAYERED MODEL OF SOIL MASS

The unsatisfactory performance of the Newmark-type analysis in the previous section implies that the calculation of debris over a long distance is beyond its capacity. This is because the Newmark method assumes a rigid block movement, while real debris flow consists of decomposed grains (Fig. 32). As an alternative, therefore, a more advanced analysis such as distinct element analysis is promising. The present paper, however, addresses a simpler choice. To get an idea about the type of analysis, model tests were conducted on the nature of debris flow. Figure 33 shows the change of appearance of a model of flow. At 0.8 second, there are three targets near the front of the flow (in a white frame). Those targets are still visible at 1.0 second. However, they are overlain by following soil at 1.2 second, and have completely disappeared at 1.4 second. This suggests that there is a significant distortion and movement of grains inside the flow mass, which makes Newmarktype analysis totally irrelevant for the present type of problem. Figure 34 illustrates the layered idealization of a debris mass in which the mass is composed of N layers of finite thickness. The internal friction angle between layers is designated by φ, while the base friction by φb . When the bottom friction is greater than the internal friction, the bottom layer stops its movement after some translation, while overlying layers can continue their movement further. This idea seems to be consistent with the observation in Fig. 33. The overall view of the analysis is illustrated in Figure 36 where the initial kinetic energy is supplied by free fall of Hi , and the velocity decreases gradually with the progress of down-slope movement. The apparent friction angle is assessed by the overall falling and lateral displacement (H /L). An example analysis was made by assuming 100 layers (N = 100), the slope angle (θ ) = 10 degrees, the base friction angle (φb ) = 40 degrees, and varying international friction angles (φ). First, Figure 37 indicates the travel distance of the front of the debris that

increases with time. The lower the internal friction is, the longer is the run-out distance. Second, Figure 38 shows the number of layers (n) that arrive at different travel distances. As illustrated in Figure 34, lower

62

Figure 36. analysis.

Initiation and termination of movement in layer

Figure 37. time.

Travel distance of debris front increasing with

Figure 38. distances.

Number of layers that reach different travel

Figure 39. Height of debris deposit changing with horizontal distance.

layers stop their movement at shorter travel distance because of the greater bottom friction, while upper layers can travel longer. This feature makes a significant difference in the ultimate travel distance; the smaller internal friction allows longer travels. Third, the thickness of debris deposit is assessed. When many layers stop within a short distance (case of φ = 30 degrees in Figure 38), the accumulation of debris produces a high debris hill. This feature is somehow assessed by dn/d (distance). As shown in Fig. 39, the greater internal friction creates a high hill of accumulated debris at the front, and, in contrast, the lower internal friction allows more spread deposit over a greater area. The latter corresponds to a flow of liquefied (fluidized) debris with high water content in which the reduced effective stress makes the internal frictional resistance very small. Finally an attempt was made to reproduce the empirical relationship in Figure 25. Because the thickness is represented by the number of layers, N , the volume is approximately represented by N 3 . It is seen

Figure 40. Reproduced apparent friction angle changing with debris volume.

in Figure 40 that H /L (tangent of the apparent friction angle) decreases with increasing volume when the internal friction angle is remarkably smaller than the bottom friction.

63

9

CONSIDERATION ON LAYER THICKNESS IN LAYERED FLOW SIMULATION

The numerical simulation in the previous section did not clearly indicate how to determine the layer thickness and, accordingly, the number of layers, N . At present, it is supposed that the thickness depends upon grain size of the debris. Figure 41 demonstrates a debris deposit in Niigata that was triggered by heavy rainfall. In this valley, the deposits developed a step in the surface and its height was approximately 3 m. Figure 42 shows a deposit of slope failure near Joganji in Nagaoka City that was caused by the 2004 Niigata-Chuetsu earthquake. The height of the step was approximately 50 cm. Finally, Figs. 43 and 44 reveal a fluidized slope failure in Tsukidate during the 2003 Miyagi-ken earthquake. The height of the step near the front was 58 cm.

Figure 43. Earthquake-induced slope failure at Tsukidate during the 2003 Miyagi-ken earthquake.

Figure 41. Rainfall-induced debris flow deposit Chu-ei Tunnel in Niigata-Chuetsu in 2005.

Figure 44. Step of deposit in earthquake-induced slope failure at Tsukidate.

Information was collected from case studies as well as literatures on the above-mentioned layer height and grain size distribution. The results are plotted in Figure 45. There is no good correlation between the layer thickness and the mean grain size, D50 . In contrast, D90 or probably the maximum grain size has some correlation. Tentatively, therefore, an empirical formula of Layer thickness (m) = 30 × D90 (m)

(2)

is proposed. Consequently, the number of layers is determined by Figure 42. Step of deposit in earthquake-induced slope failure near Joganji, 2005 Niigata-Chuetsu earthquake.

N = Total thickness of debris / layer thickness

64

(3)

REFERENCES Coulter, H.W. & Migliaccio, R.R. 1966. Effects of the earthquake of March 27, 1964 at Valdez, Alaska, US Geological Survey Professional Report 542-C. Heezen, B.C. & Ewing, M. 1952. Turbidity currents and submarine slumps, and the 1929 Grand Banks Earthquake, Am. J. Science. 250: 849–873. Heezen, B.C. & Ewing, M. 1955. Orléansville earthquake and turbidity currents, Bull. Am. Assoc. Petroleum Geologist 39(12): 2505–2514. Hsü, K.J. 1975. Catastrophic debris streams (Sturzstroms) generated by rockfalls. Geol. Soc. Am. Bull. 86: 129–140. Imaizumi, F., Tsuchiya, S. & Ohsaka, O. 2005. Behaviour of debris flows located in a mountainous torrent on the Ohya landslide, Japan. Can. Geotech. J. 42: 919–931. Ishihara, K. 1985. Stability of natural deposits during earthquakes, Theme Lecture. 11th ICSMFE, San Francisco, Vol. 1: 321–376. Kanagawa Prefectural Government. 1986. Report on Seismically induced Damage during future earthquake. pp. 13–63 (quoted by Manual for Zonation on Seismic Geotechnical Hazards, 1998, TC4 of ICSMGE). Kawata, S. 1943. Study of new lake created by the earthquake in 1941 in Taiwan. Bull. Earthq. Res. Inst., Univ. Tokyo 21: 317–325 (in Japanese). Long, D.E. & Dawson, A.G. 1989. A Holocene tsunami deposit in eastern Scotland. J. Quaternary Science 4(1): 61–66. Miller, D.J. 1960. Giant Waves in Lituya Bay Alaska, Shorter Contributions to General Geology, US Geological Survey Professional Paper 354-C. Mizuhashi, M., Towhata, I., Sato, J. & Tsujimura, T. 2006. Examination of slope hazard assessment by using case studies of earthquake- and rainfall-induced landslides. Soils and Foundations 46(6): 843–853. Newmark, N.M. 1965. Effects of earthquakes on dams and embankments, Geotech. 5(2): 137–160. Noda, S., Uwabe, T. & Chiba, T. 1975. Relation between seismic coefficient and ground acceleration for gravity quaywall. Report of the Port and Harbor Research Institute 67–111 (in Japanese). Oberlander, T. 1965. The Zagros streams, Syracuse Geographical Series, No. 1. Syracuse University Press. Sassa, K., Fukuoka, H., Wang, G.-H. & Ishikawa, N. 2004. Undrained dynamic-loading ring-shear apparatus and its application to landslide dynamics. Landslides 1(1): 7–19. Tabata, S., Mizuyama, T. & Inoue, K. 2002. Natural dam and disaster Kokin Shoin Publ. pp. 50–53 (in Japanese). Towhata, I., Yamazaki, H., Kanatani, M., Ling, C.-E., & Oyama, T. 2002. Laboratory shear tests of rock specimens collected from site of Tsao-ling earthquake-induced landslide. Tamkang Journal of Science and Engineering 4(3): 209–219. Yasuda, S. 1993. Zoning for slope instability. Manual for zonation of seismic geotechnical hazards, TC 4, Int. Soc. Soil Mech. Found. Eng. 49–49. Watson, R.A. & Wright Jr., H.E. 1969. The Saidmarreh landslide, Iran. Geol. Soc. Am. Special Paper 123: 115–139.

Figure 45. Empirical relationship between thickness of debris layer and grain size.

10

CONCLUSIONS

The present paper discussed knowledge collected from recent earthquakes. In mitigation of negative effects of slope failure to the human community, it is very important to identify potentially hazardous slopes and to assess the run-out distance. In particular, this has to be done with a reasonable cost. With regard to these, the following conclusions are drawn. 1. Problems of large slope failures are well known. Those of minor failures, however, have to be understood more, because they may stop local transportations and make rescue or restoration more difficult. 2. Seismically hazardous slopes should be identified with some consideration on characteristics of soils. 3. For the hazard assessment, a use of either swelling properties or plasticity index, both of which can be measured with minor efforts, is proposed. 4. Hazard assessment further needs to assess the travel distance of debris. 5. Newmark rigid block analogy is not relevant for this purpose. This study proposes a similarly simple layer idealization for the assessment of run-out distance. 6. To facilitate the determination of layer thickness, a correlation between the thickness and particle size was presented. ACKNOWLEDGMENT Direct shear tests on Tsao-ling rock block samples were conducted at the Central Research Institute of Electric Power Industries. Model tests on debris flow were carried out by Mr. Y. Nishimura for his undergraduate thesis. Field studies were made by T. Ito of Fudo-Tetra Corporation. Deep thanks are expressed to those supports and contributions.

65

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Monitoring and modeling of slope response to climate changes H. Rahardjo School of Civil and Environmental Engineering, Nanyang Technological University, Singapore

R.B. Rezaur Department of Civil Engineering, Universiti Teknologi Petronas, Perak, Malaysia

E.C. Leong School of Civil and Environmental Engineering, Nanyang Technological University, Singapore

E.E. Alonso, A. Lloret & A. Gens Department of Geotechnical Engineering and Geosciences, UPC, Barcelona, Spain

ABSTRACT: Shallow slides are often triggered by climate effects. An understanding of the slope failure conditions and effective remedial measures can be achieved by comprehensive field monitoring of climatic and hydrologic changes and the consequent changes in slope responses. Two contributions from two different geographic regions are presented to gain understanding of the complex phenomena involved in slope failure studies. In the first part Alonso et al., contributes theoretical analysis of a stochastic model for the reliability of planar slides in a partially saturated soil, subjected to a rainfall history described as a time series and then presents a case history of shallow mudslides triggered by a Mediterranean climate, analyzed by means of a coupled hydro-mechanical modeling tool. The joint saturated-unsaturated consideration of the slide is necessary to understand field data. In the second part Rahardjo et al., contributes field monitored data from three residual soil slopes in Singapore and demonstrates how field monitored data on climatic, hydrologic, and slope variables were used to evaluate slope responses under subtropical Singapore climate. 1

antecedent rainfall, rainfall intensity and rainfall duration. However, it is impossible to dissociate climate effects from the type of slide. In general, large landslides require specific analysis. Shallow slides, however, react to rainfall conditions in a more predictable manner. The objective of this paper is to present strategies for monitoring and modelling of slope response to climate changes. In this context two contributions from two different geographic regions are presented in two sections (PART-A and B). In the first part (PART-A) Alonso et al., presents a stochastic model for the analysis for the reliability of planar slides in a partially saturated soil, subjected to a rainfall history described as a time series and then presents a case history of shallow mudslides triggered by a Mediterranean climate analyzed by means of a coupled hydro-mechanical modeling tool. The theoretical analysis of the risk of slide associated with a given record of rainfall presented by Alonso et al., considers infiltration conditions through a partially saturated soil and integrates rainfall effects through a random function formulation of the concept of risk of failure. Closed form

INTRODUCTION

Global warming, rising sea level and climatic changes have become important issues of the world in recent decades. Climatic changes have affected rainfall patterns in many parts of the world, causing occurrences of numerous landslides. Many tropical areas are prone to frequent rainfall-induced slope failures (Poh et al. 1985, Pitts & Cy 1987, Tan et al. 1987, Chatterjea 1994, Toll et al. 1999). The problem is escalated by the increasing rate of hillside developments for engineered and fill slopes in many regions. Rainfall of both event-based and antecedent, runoff, infiltration, and their contributions to pore-water pressure changes in a residual soil slope are the variables pertinent to slope responses. The dynamic flux boundary conditions are controlled by the physical properties of the soil, in particular the unsaturated soil in the vadose zone above the water table. Attempts to relate climate conditions to the occurrence or reactivation of landslides are numerous (Wieczorek (1996) and Corominas (2000) provide a detailed account). Most methods combine

67

solutions were found for the probability of failure. The theoretical analysis highlights several aspects of rain-induced risk of failure. The case history for shallow instabilities in a clay formation analyzed by means of a coupled hydro-mechanical analysis presented by Alonso et al., is also valid for saturated and unsaturated slope conditions. In the second part (PART-B) Rahardjo et al., presents field monitored data from three residual soil slopes in Singapore and demonstrates how field monitored data on climatic, hydrologic, and slope variables were used to evaluate slope responses in terms of; (i) pore-water pressure changes; (ii) runoff generation and infiltration amount. Part-I and II together provides understanding of the complex phenomena involved in slope failure studies. 2

Figure 1.

it will be assumed to be constant in the following). The safety margin of an infinite slope against failure along a plane at depth (H − h) (Fig. 1) is given by    H Sr dz + γs (1 − n)(H − h) M = γw n

PART-A. MODELLING RAINFALL EFFECTS ON SHALLOW SLIDES (ALONSO ET AL.)

h

× cos α(cos α tan φ ′ − sin α)

2.1 Rain infiltration and risk of sliding—a simple stochastic model

+ c′ + (pa − pω ) tan φb

Soil moisture variation in time within slopes is recognized as a major source of uncertainty concerning the safety conditions of any slope. Any wetting process modifies several parameters which control the overall stability. In fact, the natural unit weight increases, the available shear strength decreases with a decrease in soil suction and, when the soil becomes saturated, the pore water pressure becomes positive thus modifying effective stresses. In addition, the internal geometry of different soil strata having varying permeabilities, plays a significant role in controlling the hydraulic regime within the slope. An example developed later (Villa Blasi slope) illustrates this comment. The variation of soil moisture conditions in the slope is controlled by the varying rates of infiltration which in turn depend on rainfall intensity, soil conditions and geometry. Except for the case of homogeneous masses of impervious clay soils or extremely pervious formations, experience shows that, in a large class of slopes, failures occur during periods of heavy rainfalls.

(2)

Safety margin is defined as the difference between shear strength and shear stress and it provides a linear measure of safety. In Equation 2, Sr , n and γs are the degree of saturation, the porosity and the solid specific weight respectively. 2.1.2 Infiltration model The classical theory assumes that the net infiltration, q(t), results from a balance between rainfall, corrected by means of a runoff coefficient, and evapotranspiration. On the other hand the stochastic analysis of a time series of data regarding rainfall and evapotranspiration provides the necessary parameters to identify q(t) from a stochastic point of view. The equations governing the flow of water in a rigid partially saturated soil are (Bear & Bachmat 1991) ∂ (Sr n) + div(vω ) = 0 ∂t

pω vω = −Kgrad z + γω

2.1.1 Safety margin of slopes in partially saturated soils Fredlund et al. (1978) proposed the following expression for the strength of a partially saturated soil: τ = c′ + (σ − pa ) tan φ ′ + ( pa − pω ) tan φb

Geometry of the solved problem.

(3) (4)

If a linear expression for the water retention curve is adopted, (5)

Sr = Sr0 + as ( pω − pω0 )

(1)

where Sr0 is the degree of saturation for a reference suction pω0 and as is a constant, Equations 3–5 become

where (σ − pa ) and (pa − pw ) are the net normal stress and suction, (c′ , φ ′ ) are the effective cohesion and friction and tan φb is a suction-related friction coefficient which takes the value of φ ′ for very low suctions and reduces progressively as suction increases. (However,

∂ 2 pω ∂pω = Cs 2 ; ∂t ∂z

68

Cs =

K nas γω

(6)

Then, the safety margin (2) is found as:

This equation may be normalized if, Z=

z , H

T =

Cs t , H2

u = [pω /γω H (q0 /K − 1)] − z/H

M (Z, T ) = ω0 +

q0 being a constant under steady state conditions and it results in, ∂u ∂ 2u = ∂T ∂Z 2

+ (−An 2 + Bn 1 ) sin

Z =0

2πnT T0

(11)

where the different coefficients have been defined in Appendix 1. The mean safety margin, ω0 , is a deterministic function of Z. Figures 3a, b show the evolution of safety margin at different depths of a slope subjected to sinusoidal infiltrations of widely different frequencies. It can

(7)

Boundary conditions for an infiltration q(t) at the surface and a fixed ground water level at depth H are u = 0;

N  2π nT (An 1 + Bn 2 ) cos T0 n=1

(8a)

q(t) − q0 q1 (T ) ∂u = = I (T ) = ∂Z q0 − K q0 − K

(8b)

If q0 is selected as the mean of the process q(t), I (T ) becomes a stochastic process of zero mean, which, on account of the periodic nature of hydrologic events, will be expressed as:  N  du  = I (T ) = An cos λ2n T + Bn sin λ2n T ; dZ Z=1 n=1 λ2n =

2π n T0

(a) (9)

where the coefficients An and Bn are non correlated Gaussian random variables of zero mean and common variance (σn2 ) and T0 is a dimensionless reference period. Figure 2 shows a 6 year record of the process I (t) for a meteorological station in the vicinity of Barcelona. Plotted values represent averages of 10 days. Evaporation data was obtained from actual monthly measurements, in a 1.25 m diameter tank corrected to reduce the free water surface to soil conditions. The solution of equation (7) with (8) and (9) is

u(Z, T ) =

N  n=1

Tn (Z) cos

2πnT 2πnT + Wn (Z) sin T0 T0

(10)

(b) Figure 3. Evolution of safety margin at different depths (p = 0 to 4 m) of a slope (α = 32◦ ) subjected to sinusoidal infiltration: (a) Low frequency; (b) High frequency. K = 0.7 × 10−7 m/s, φ′ = 30◦ ; φb = 25◦ ; as = 10−5 m2 /N; c′ = 1000 zN/m2 .

Figure 2. Infiltration record measured in a meteorological station in Barcelona.

69

νa = (σM˙ /2πσM ) exp(−r 2 /2)

be observed the limited influence of high frequency components at points not very close to the surface. On the other hand, high period oscillations of infiltration have the capability of penetrating deep into the slope. Figure 3a shows also the delaying effect of infiltration into the soil.

where σM˙ is the standard deviation of the derivative in time of the safety margin. Knowing the expression (12) for the autocovariance function, both σM2 and σM2˙ are found as

2.1.3 The safety margin as a random process in time—Correlation structure In order to analyze the slope reliability as a levelcrossing problem associated with the evolution in time of the safety margin, it is first necessary to establish its correlation structure. It may be shown that the autocovariance function of the safety margin is given by CM (T ) =

N  n=1

σn2 (21 + 22 ) cos λ2n τ

σM2 = σM2˙ =

(12)

−νa t 1 − φ(r)



n=1

N  n=1

σn2 (21 + 22 ); σn2 λ2n (21 + 22 )

(15a, b)

2.1.5 Application of the developed formulation to the reliability of slopes in a given hydrologic environment In all the cases presented below the soil properties are maintained constant. They correspond approximately to a partially saturated silty soil and they are φ ′ = 30o ; φ b = 25o ; as = 10−5 m2 /N. The parameterc′ of the strength envelope was assumed to increase slightly with depth (1000p N/m2 , where p is the depth below the surface) in order to represent non homogeneous soil profiles, typical, for instance, in residual deposits. Failure probabilities correspond always to a yearly period. Figure 4 shows the variation of probability of failure with the slope angle for different positions of the failure plane. It is interesting to note the transition from deep-seated surfaces, more likely to occur in steep slopes, to shallow failures in gentle slopes. In fact the influence of rainfall induced moisture changes is determinant to provoke shallow instabilities. On the other

2.1.4 The level crossing problem For Gaussian processes, assuming that the rate of crossing a certain barrier follows a Poisson distribution and that the time intervals spent in the ‘‘safe’’ regions have a common exponential distribution whose mean is the inverse of the mean rate of crossings, Vanmarcke (1975) derived the following expression for the probability of no crossings in the interval 0 to t: 

N 

All of these expressions are used subsequently to predict the reliability of an earth slope in a given hydrologic environmental condition (Barcelona area).

The individual variances associated to each of these periodic terms reflect both the variability of the ‘‘external’’ infiltration terms (through σn2 ) and the process of percolation, the modification of strength parameters of the soil and its weight and the (mechanical) definition of stability—(through the term (21 + 22 ), which is defined in the Appendix). Once the covariance is determined, standard procedures to analyze stochastic time records (see for instance Tretter (1976)) were used to determine the power spectrum density function of the safety margin. The analysis of a slope subjected to the Barcelona infiltration record shows how the influence of high frequencies of hydrologic variation is increasingly damped at depth.

LB (t) = [1 − φ(r)] exp

(14)

(13)

where r is the normalized crossing level, φ(r)√is related to the error function (φ(r) = 1/2[1 − erf (r 2)]) and νa is the mean rate of crossings above the level r. In our case, if one works with the zero-mean safety margin process, M ∗ = M − ω0 , the normalized barrier indicating slope failure is r = −ω0 /σM , where σM is the standard deviation of safety margin. The mean rate of crossings above level r is given by

Figure 4. Probability of failure of an infinite slope in partially saturated soils for different slope angles and depths of failure plane.

70

important result is, however, that the slope is especially prone to shallow failure surfaces for a wide range of slope angles. In other words, the critical slope angle which marks the boundary between shallow or deep failure planes (approximately at 37◦ if H = 5 m in the example solved, see Figures 5 to 7) increases when H increases. It is finally noted that the above results were all obtained for rainfall records averaged over 10 days. No significant influence was found when the average period changed from 5 to 20 days. 2.2 A case history: ancona slides 2.2.1 Background In December 1982, a large flow slide in overconsolidated clays destroyed the suburbs of the city of Ancona (Italy). Several years later, a research project was launched to investigate the behavior of shallow slides in clays in a Mediterranean climate. Villa Blasi slope was selected for the study and a field investigation was set out. Inclinometers, as well as piezometers, were installed and specimens were taken for specialized testing. Rainfall was recorded. A representative profile of the slope is shown in Figure 7. Three layers (α, β, γ) could be distinguished. The surface layer α (3–4 m thick) is made of remoulded brown Ancona clay with a significant proportion of organic matter. It overlies a 4 to 10 m thick layer (β) of stiff brown silty clay with sandy inclusions (brown Ancona clay). At depth, the Pliocene substratum consists of stiff blue silty clay with sandy inclusions (blue Ancona clay).

Figure 5. Influence of soil permeability. H = 5 m; φ′ = 30◦ ; φb = 25◦ ; as = 10−5 m2 /N; c′ = 1000 p N/m2 .

2.2.2 Soil properties Dominant clay minerals are montmorillonite, illite and chlorite. Calcite amounts to 22% of the total mineral composition. Brown and blue Ancona clays are Figure 6. Influence of depth of ground water level in failure risk for different slope angles and position of failure plane.

hand steep slopes have small (deterministic) safety margins, decreasing with depth according to the stability model adopted, and their influence overrides the strong moisture induced effects close to the surface. The marked influence of permeability is shown in Figure 5. Other conditions being equal the increase in permeability reduces fast the risk of failure. For a given infiltration history the developed pore pressure gradients are smaller for larger K values and therefore changes in soil strength are small. The depth of groundwater level (Fig. 6) controls the mean ‘‘dryness’’ of the soil in the sense that both Sr and pw are reduced. Larger strengths (implied by higher suctions) result in increased reliabilities. The

Figure 7. Longitudinal profile of slope and position of the phreatic level. The position of piezometers (A, B, C, D, E and P) is also shown.

71

Figure 8. Water retention curve of brown Ancona clay.

Figure 9. Variation of apparent cohesion with suction for Brown Ancona clay.

classified as CH or MH (WL = 52–64%; PI: 25–34%; clay fraction: 40–55). Water content is low in these formations: 21%–32%. Oedometer tests indicated a significant overconsolidation stress: 2 to 3 MPa for layer β and 3 to 4 MPa for layer γ. In drained triaxial tests, the brown Ancona clay exhibited a brittle behaviour (brittleness index, IB = 0.40). Suction controlled oedometer tests were performed to determine the effect of suction on volumetric deformations and on permeability. Experimental data on the variation of permeability and degree of saturation with suction is shown in Figure 8. A coupled finite element flow-deformation code (NOSAT) for saturated-unsaturated analysis, developed at the Department of geotechnical Engineering and Geosciences of the UPC, was used in calculations. NOSAT solves the balance equations for water and air, and the mechanical equilibrium.

phases. Equilibrium conditions were first defined. An idealized geological sequence was simulated. The soil was initially deposited under normally consolidated conditions. Then, erosion (unloading) took the slope to its actual geometry. For this phase on an elastoplastic hyperbolic model, having parameters c′ = 95 kPa, ϕ ′ = 27◦ , was used. This sequence tried to reproduce the actual preconsolidation stresses measured in oedometer tests. The phreatic surface of the deep hydrogeological regime was also defined at this stage (see Fig. 7). In the second phase of the analysis, rainfall infiltration is simulated. Average uniform monthly flow rates were imposed as flow boundary conditions at the slope surface. If a positive pore water pressure is calculated at the surface (which is an indication that the rainfall rate was larger than the infiltration capacity of the soil), the boundary condition is changed to zero suction, which is a flooding condition. The third stage in the analysis concerns deformations and safety factors. Local safety factors were determined by comparing at some depths within the slope available shear strength and existing shear stress. The rainfall record actually used in computations is shown in Figure 10. There was some uncertainty in the actual field permeabilities of layers α, β and γ. This is key information to understand stability conditions. Time records of pore pressure changes measured in piezometers could be used, however, to define a permeability profile. The idea was to perform a sensitivity analysis varying the permeabilities of layers α, β and γ. A total of 10 different permeability profiles, shown in Table 1, were subjected to the same rainfall record in the manner outlined above. Then, the most probable permeability layering was identified by comparing piezometer measurements and calculated values.

2.2.3 Modelling Villa Blasi slope The soil, in this case, was characterised by a non-linear elastic behaviour. Volumetric deformations induced by suction changes were introduced by means of state surfaces relating void ratio, suction and mean net stress. Figure 9 shows the variation of apparent cohesion with suction for the intermediate layer (brown clay). A first step in modelling was to approximate the actual hydrogeologic conditions. Two flow regimes were identified. A deep flow regime was characterised by a phreatic level at nearly constant elevation in both wet and dry seasons (depth varies between 4 and 14 m). Figure 7 shows some piezometer readings and the calculated position of the phreatic level. In addition, a surface flow regime was directly controlled by rainfall. Positive and negative pore water pressures were measured during the year in the upper few meters. Calculations were performed in three

72

Figure 10.

Rainfall records used in calculations. Figure 11. Comparison of measured and calculated piezometer readings (piezometer C). Depth of measuring chamber 9.80 m (open tube piezometer in substratum).

Table 1. Combination of soil layer permeability for sensitivity analyses. Saturated water permeability kws (m/s) Case

Layer α

Layer β

Layer γ

a b c d e f g h j k

10−9 10−9 10−8 10−7 10−6 10−8 10−7 10−6 10−7 10−6

10−9 10−9 10−9 10−9 10−9 10−8 10−8 10−8 10−7 10−7

10−9 10−9 10−9 10−9 10−9 10−9 10−9 10−9 10−9 10−9

2.2.4 Computational results Figures 11 and 12 show two examples of the comparison made between field data and calculated pore pressures. Figure 11 indicates the readings in an electrical piezometer located in α layer. Some combinations of layer permeability lead to a consistent agreement with field data. (Case h, for instance). Suctions were often recorded. There is another interesting consideration about the relationship between failure risk and soil permeability. Examining the results of all computations, it became clear that some combinations of layer permeability and layering sequence led to the ‘‘strongest’’ response of the slope, in the sense of producing the highest pore pressures for a given sequence of rainfall. The general idea is that, given a climatic record, there are particular combinations of soil permeability which are critical in terms of pore water pressure development. This idea is illustrated in Figure 12 which is a plot of

Figure 12. Computed depth of water level at piezometer P5699 for time t = 350 days after the beginning of the prediction exercise (October 1st, 1992). Depth of sensor: 2.80 m.

the computed response of piezometer P5699, 350 days after the initial date for the modelling. The combination Kα = 10−7 m/s, Kβ = 10−8 m/s, leads to the maximum value of the (positive) pressure at the point considered. There are no easy rules to predict a critical permeability profile because the computed water pressure integrates many

73

Figure 14. Profiles of local safety factors (based on FE computations) at a borehole C (Case h, homogeneous strength). Figure 13. Computed and measured profiles of horizontal displacements at borehole B.

phenomena: surface flow, infiltration, storage capacity of the soil, previous wetting history of the soil etc. The calculated and measured profile of horizontal displacements is shown in Figure 13. It is clear that a distinct shear surface has developed at the interface between α and β layers. Most probably, a previously existing shear surface was reactivated. 2.2.5 Stability analysis—Local safety factors Field observations and the results of inclinometer readings indicate that a simple planar model could represent the shallow slides. Since stresses were determined also in a theoretically more accurate analysis, an expression for the safety factor may be found. It uses stresses and water pressures calculated in the numerical model: c′ + F(FEM ) =



Figure 15. Profiles of FE-based local safety factors at borehole C (Case h, heterogeneous strength).

(σx + σy ) (σx − σy ) − 2 2  cos 2α − τxy sin 2α − pw tan ϕ ′

(σx − σy ) sin 2α − τxy cos 2α 2

(17)

Calculated variations of local safety factors with time and depth are shown in Figure 14. Absolute values of F are high, however. If the soil strength is assumed to be heterogeneous and remoulded parameters (c′ = 0; φ ′ = 24o ) are assigned to the upper α layer, the calculated profiles of the local safety factor are shown in Figure 15. Minimum safety factors are now found at the α−β interface, where shear slips were detected by the inclinometer. However, the minimum calculated value (F = 1.5), is not able to explain the failure.

Figure 16. Profiles of water pressure for a permanent flooding of the slope. (Case h, heterogeneous soil strength).

The analysis performed so far only includes the rainfall recorded in 13 months. Heavier rains certainly had occurred in the immediate past of the slope. An extreme rainfall event is to assume that the slope

74

Figure 17. Profiles of FE-based local safety factor. Heterogeneous strength distribution. Case h. Figure 18. Location of the study slopes, generalized geological map of Singapore, schematic diagram of relative position and arrangement of field instruments.

surface is flooded during a certain time period. This condition is illustrated in Figure 16, which shows the calculated profiles of water pressure. Three days of permanent surface saturation leads to almost hydrostatic pore water pressure conditions in the upper remoulded α layer. This situation is compatible with suction values maintained in the lower β layer. This situation brings a different picture of the safety conditions of the slide (Fig. 17). Now surface instabilities are predicted after 12 hours of surface saturation of the slope. The entire upper α clay layer becomes soon unstable under extreme rainfall events.

3

3.1

Table 2.

PART-B: SLOPE HYDROLOGY AND RESPONSES TO CLIMATE CHANGES (RAHARDJO ET AL.)

Description of the slopes.

Variable

Yishun

Mandai

NTU-CSE

Area (m2 ) Vegetation cover (%) Slope angle (◦ ) Slope height (m) Slope length (m) Aspect Slope form

165 100 23 7 16.5 West Plain

180 100 31 11 18.0 West Plain

140 100 27 7 14.0 West Plain

Service Singapore 1997). The three recorded highest rainfalls in Singapore are; 512 mm in 1978, 467 mm in 1969 and 366 mm in December 2006. The rainfall is usually greatest in the months of November to January (the north-easterly monsoon) but rain falls in all months of the year, with an average of 179 rainy days in a year. Rainstorms are short, intense and generally have a limited spatial extent, with intensities typically ranging between 20 and 50 mm/h, although short duration (5 min) rainfall intensities can exceed 100 mm/h (Sherlock et al. 2000). The potential evaporation rate in Singapore was calculated to vary between 5.16–7.53 mm/day (Gasmo 1997). All four slopes had grass as the vegetative cover. The geology of the study sites consists of residual soils from two major geological formations (Public Works Department 1976); (a) the Bukit Timah Granite (Yishun and Mandai) which occupy the north and central-north region of Singapore; and (b) sedimentary rocks of the Jurong Formation (NTU-CSE), occupying the west and southwest region of Singapore (see Figure 18). These two residual soils comprise two-thirds of Singapore’s land area.

Study area

Three slopes were selected for instrumentation to provide data for rainfall-induced slope failure studies. Out of these three slopes (see Figure 18) one was located in Yishun, one in Mandai and two in Nanyang Technological University (NTU) campus (hereafter called the NTU-CSE slope). The characteristics of the four slopes are shown in Table 2. These slopes were selected because they were located in two major geological formations in Singapore, subjected to frequent shallow landslides. The climates at the study sites are hot and humid equatorial, with no marked dry season. The temperatures vary little throughout the year with an annual average temperature of 26.6◦ C and a mean relative humidity of 84% (Meteorological Service Singapore 1997). The average annual rainfall in Singapore varies between 2000 mm around the fringes of the island to about 2300 mm in the central region (Meteorological

75

of the slopes as obtained from pressure plate tests are shown in Figure 20. Elaborate discussion on the soil properties of these slopes are presented in Rahardjo et al. (2004b).

3.2 Field instrumentation and data collection The slope responses were characterised through measurement of the following variables: (i) rainfall inputs to the slopes (ii) runoff generation from the slopes (particularly NTU-CSE slope) and (iii) changes in pore-water pressures in response to rainfall. Rainfall was recorded at each slope with a tippingbucket rain gauge. Runoff was measured using a perspex flume (Rahardjo et al. 2004a). Corrugated zinc sheets, 300 mm high and driven about 100 mm into the ground bordered each instrumented slope (Figure 18). The boundaries guided the runoff into the perspex flume at the lower end of the plot where the surface runoff was measured using a capacitance water-depth probe installed in the flume. The waterdepth probe was connected to a data logger that stored runoff data every 10 s during rainfall. Runoff measurements for both simulated and natural rainfalls were conducted particularly in NTU-CSE slope. Pore-water pressure changes in response to climatic changes were recorded using jet-fill tensiometers. In addition to the tensiometers, piezometers and temperature sensors were installed at various locations in the slopes. Figure 18 shows the details of field installation of the instruments. All sensors within each instrumented slope were monitored continuously and automatically by a field data acquisition system (DAS). The DAS was programmed to scan the sensors at 4 h intervals during periods of no rain and at 10 min intervals during rainfall and continuing at the same rate until 30 min after cessation of rainfall. A rainfall event would trigger the rain gauge, and data for surface runoff were collected during rainfall events only. More details on field instrumentation can be found in Rahardjo et al. (2007).

3.3.2 Evaluation of slope responses in terms of pore-water pressure changes due to rainfall Slope responses to rainfall were assessed through the pore-water pressure changes in the slope. The tensiometer readings from the slope crest (row A) were plotted with the corresponding rainfall readings to indicate the seasonal pattern of pore-water pressures at Yishun slope (Figure 21). The tensiometers recorded increases in pore-water pressures at all levels on wet days. The deeper soil layers (3 m depth) at the slopes’ crest and toe maintained positive pore-water pressures for all the 420 monitored days, while frequent changes in pore-water pressure from negative to positive were recorded for shallow depths (50 cm) in response to rainfall (Figure 21). The soils at shallower depths are in close proximity to the atmosphere and slopes’ vegetation. As a result, the soils at shallow depths are easily and frequently influenced by rainfall and evapotranspiration compared to deeper soil layers. During the monitoring period, negative pore-water pressure development as low as −57 kPa was observed at shallow depths (Yishun slope, Figure 21). Positive pore-water pressures were also observed at all soil depths after a significant rainfall and appear to be a common phenomenon in the monitored slope (Figure 21). Infiltration into the slope does not lead to a constantly wet soil condition, as is seen in the rapid matric suction recovery during dry periods (Figure 21). The average daily matric suction recovery rates were 5 kPa/day for shallow depths and 1 to 3 kPa/day for greater depths. The instant response of pore-water pressure to the infiltrating rainwater indicates considerable infiltration on the grass covered slopes. To illustrate the sensitivity of pore-water pressures to rainfall and to reflect the hydrologic response of the slopes during a wet and a dry period, the record of pore-water pressures and rainfall for a period of 6 weeks was selected from the time series of the Yishun slope as shown in Figure 22, which illustrates the decreasing pore-water pressures during the dry period from February 12 to March 8, 1999. A small storm on March 8 only affected the pore-water pressures at a shallow depth (0.5 m) due to the small amount of infiltrated water. After the storm the pore-water pressures at 0.5 m depth started to decrease again due to the evapotranspiration processes. On the other hand, several rainfall events during the wet period from March 9 to March 16, 1999, caused the pore-water pressures at all depths (0.5, 1.7, and 2.9 m) to increase.

3.3 Results and discussion The results of the field tests, laboratory tests and field monitoring are presented in two forms. First, field test and laboratory test results used to characterise the engineering properties of the slopes are presented in the next section. Second, field monitoring and data analyses results are presented in the subsequent sections to show how slope responses were evaluated in terms of (i) pore-water pressure changes and (ii) runoff generation and infiltration amount. 3.3.1 Site observation The geotechnical properties of the residual soils at the four slopes derived from laboratory tests and field tests are shown in Table 3. The information obtained from site investigation and laboratory tests was used to produce simplified soil profiles for the slopes as shown in Figure 19. The soil water characteristic curve (SWCC)

76

77

2 3

1

4

1 2 3

Moderate to high plasticity silt and clay Low plasticity clay Silty sand

Silty sand Clayey sand Moderate to low plasticity silt Silty sand

Moderate to low plasticity silt Moderate plasticity silt Silty sand

Description

30 19 12

36 23

MH–ML SM–MM

CH–MH CL to SC–CL SM-ML

25 26

35 52 19

w (%)

SM SC

ML MH SM

USCS

Index

65 36 29

52 45

55 88

45 57 42

LL (%)

Average* properties

35 23 23

34 27

31 34

31 40 26

PL (%)

94 74 27

60 46

41 40

58 78 34

Fines (%)

– 2.10 2.32

1.89 1.95

2.02 1.94

1.88 1.76 2.05

ρ (Mg/m3 )

– 2.725 –

2.680 2.680

2.684 2.686

2.688 2.714 2.667

Gs

139 107 167

10 0

12 14

12 13 35

c′ (kPa)

34 36 37

29 28

30 31

33 29 31

φ ′ (◦ )

Shear strength

0.5 – –

0.24 0.20

0.19 –

0.30 0.64 –

Cc

1.9 × 10−7 8.0 × 10−7 2.8 × 10−9

– –

9.5 × 10−6 –

7.2 × 10−7 – –

Hydraulic ks (m/s)

USCS = Unified Soil Classification System (ASTM, 1997); w = Water content; LL = Liquid Limit, PL = Plastic Limit; ρ = Total density; c′ = Effective cohesion; φ ′ = Angle of friction; Gs = Specific gravity; Cc = Compression index; and ks = saturated permeability. * Average of 2 to 9 samples, on some occasions there were only one sample. – Results were not available due to sample damage or technical difficulty.

NTU-CSE

Mandai

1

Yishun

2 3

Layer

Slope

Table 3. Soil properties of the slopes.

Figure 21. Time series of rainfall and pore-water pressure at Yishun slope.

Figure 19. Generalized soil profile of the four slopes (a) Yishun slope, (b) Mandai slope, (c) NTU-CSE slope.

Figure 22. A trace of pore water pressure and rainfall record in Yishun slope showing tensiometer response to a dry period and wet periods.

Tensiometer readings at all depths and all locations (6 rows × 5 depths) in the Yishun slope for two selected periods (March 8 and March 16, 1999) were used to produce the pore-water pressure contours shown in Figures 25 and 26. These figures illustrate the magnitude and distribution of pore-water pressure within the entire slope profile and identify the movement of water in the subsurface regime. These two specific events were chosen to observe the magnitude and distribution of pore-water pressures across the entire slope (that is slope response) at the end of a dry period (March 8) and a wet period (March 16) period (Figure 22). Figure 25 shows widespread negative pore-water pressure prevailing across the entire slope during the dry period. The groundwater table from the piezometric readings (near slope crest, midslope, and toe) for this period is also shown in Figure 25. However, the slope experiences more matric suction development at the crest than

Figure 20. Soil-water characteristic curves of Yishun, Mandai and NTU-CSE slope.

Pore-water pressure records in row A at all depths of the Yishun slope are plotted for four different times during a dry period (Figure 23) and four different events during a wet period (Figure 24). These figures show the pore-water pressure distribution across the soil depth and the progressive vertical movement of water (infiltration and evapotranspiration) during these two periods. Figure 23 indicates that the soil profile dries from the surface downwards, owing to evaporation and transpiration during dry periods. Figure 24, in conjunction with the rainfall events shown in Figure 22, shows the successive increase in pore-water pressure along the soil depths in response to cumulative rainfall amounts during the wet period.

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Figure 25. Magnitude and distribution of pore-water pressure along the soil profile in Yishun slope on 8 March 1999 at the end of a dry period. Figure 23. Pore water pressure profile in row A of Yishun slope showing the advancement of a drying event at different depths.

Figure 26. Magnitude and distribution of pore-water pressure along the soil profile in Yishun slope on 16 March 1999, after a wet period with 98 mm of cumulative rainfall.

saturation to near saturation conditions throughout the soil profile at higher rainfall rates. The downslope movement of water becomes more obvious from the contours plots of the distribution of hydraulic heads shown in Figures 27 and 28. As water flow is caused by the hydraulic head gradient, a comparison of total hydraulic heads (composed of the pressure head and elevation head, assuming velocity head to be negligible) at the slope crest and toe (both during dry condition, Figure 27, and wet condition, Figure 28) shows the existence of a hydraulic gradient during both dry and wet periods. The hydraulic gradient during the wet period is greater than during the dry period, which suggests there is more downslope flow during a wet condition than during a dry condition (assuming hydraulic isotropy). Decreases in matric suction were observed during the analyses of pore-water pressure response to rainfall at almost all magnitudes of rainfall events. Therefore, the correlation between the increase in pore-water pressure uw , (The difference in pore-water pressure before and after a rainstorm) and rainfall amount, R, was tested. This was done by splitting the rainfall data into daily totals and taking the algebraic difference in pore-water pressure before and after daily

Figure 24. Pore-water pressure profile in row A of Yishun slope showing the advancement of the wetting at different depths during a wet period.

at the toe. This reflects the variability in hydrologic response within the same slope and could be because, in addition to evaporation, there is preferential lateral movement of water downslope. Figure 26 shows the widespread development of positive pore-water pressure across the entire slope after the wet period, with about 98 mm of cumulative rainfall (from Figure 22). The groundwater table location from the piezometric readings for this period was not conclusive as they did not show any significant rise in the water table (instrument malfunction), and therefore the water table is not shown in Figure 26. The pore-water pressure distribution during this period is fairly uniform along the slope face, indicating that significant rainfall brings uniformity of pore-water pressures in the soil profile. The widespread development of positive pore-water pressures is a consequence of the increasing degree of

79

Figure 27. Magnitude and distribution of hydraulic head along the soil profile in Yishun slope on 8 March 1999 at the end of a dry period. Figure 29. Relationship of increase in pore-water pressure at 50 cm depth to daily rainfall amount for all instrumented slopes.

to assess the infiltration effects on the slope due to natural rainfall events because natural rainfall intensity and duration vary within a rainfall event and from event to event. To understand infiltration effects on the residual soil slope under controlled rainfall conditions 10 simulated rainfall experiments (5 experiments during February of 1998 and another 5 experiments between 17 November 1998 and 5 January 1999) were conducted on NTU-CSE slope. The runoff data were analysed for the 27 natural and 10 simulated rainfalls to determine total runoff volume, total infiltration, peak intensity of each rainfall event. The total runoff amount was calculated from integration of the area under the runoff hydrograph. From the time series of rainfall and pore-water pressures shown in Figures 21 and 22 and similar plots for other slopes, rainfall amounts of less than 0.5 mm per day appear to have no significant effect on the pore-water pressures, even at shallow depths. Therefore, interception by the slope vegetation (grass) was estimated at 0.5 mm from the time series of pore-water pressures. Total infiltration was estimated for all runoff measurements by subtracting total runoff and an interception loss of 0.5 mm from total rainfall. Figure 30 plots runoff amounts against rainfall amounts for the 27 monitored natural rainstorms. A linear regression fitted to the data points shows an intercept on the x-axis, indicating that a threshold rainfall amount of about 10 mm must be exceeded to produce a significant runoff. Hydrologic responses of slopes are locally and geographically variable, causing difficulty in comparing results. The limited references from Southeast Asia further add to this difficulty. The threshold rainfall observed in this study is comparable to that reported by Tani (1997) for hilly area in Okayama, Japan, but is about half of that reported by Premchitt et al. (1992) for Hong Kong in a sub-humid tropical climate.

Figure 28. Magnitude and distribution of hydraulic head along the soil profile in Yishun slope on 16 March 1999, after a wet period with 98 mm of cumulative rainfall.

rainfall, in row A at the depth of 50 cm. A regression analyses was performed to fit the data sets of the increase in pore-water pressure, uw , and daily rainfall amount, R. A semi-logarithmic equation of the form, uw = a2 + b2 log R, where a2 and b2 are coefficients, best described the data set. Figure 29 shows the relationships between the increase in pore-water pressure at 50 cm depth of row A, and the daily rainfall amount for the three slopes. The equations in Figure 29 indicate that both coefficients a2 and b2 increase with increasing particle sizes (see Table 2). Mandai, being rich in coarse particles, shows higher values of uw than other slopes. This suggests that the increase in pore-water pressure is not only dependent on the rainfall amount, but it is also affected by soil properties and antecedent soil moisture. Although uw values are different for Yishun, Mandai, and the NTU-CSE, they appear to increase with the rainfall amount, and the rate of increase tends to decline at a daily rainfall amount greater than about 10 mm. 3.3.3 Evaluation of slope responses in terms of runoff generation and infiltration Rainfall and runoff data were collected for 27 natural rainstorms in the NTU-CSE slope. It is difficult

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Figure 32. Rainfall and runoff hydrograph from a simulated rainfall event on 9 February 1998 in the NTU-CSE slope.

Figure 30. Relationship between storm rainfall (natural rainfall) and runoff amount at NTU-CSE slope.

Figure 33. amount. Figure 31. Rainfall and runoff hydrograph of a natural storm event on 12 December 1998 recorded at NTU-CSE slope.

Percent infiltration as a function of rainfall

Rainfall and runoff hydrograph from a simulated rainfall event during February 1998 on the NTU-CSE slope is shown in Figure 32. The infiltration rates shown in Figure 32 are derived by subtracting runoff rates from rainfall rates. The hydrograph (Figure 32) shows that there was no runoff during the early part of the simulated rainfall event. During this period all the rainfall water is lost as infiltration (a fraction may have also been retained by the slope vegetation as interception). It is noted that during the rainfall event of 9 February 1998 the runoff rate did not exceed the infiltration rate and the rise in runoff rate is slow. This is due to the relatively dry soil conditions that prevailed before the rainfall event. In Figure 33, infiltration amounts (as a percentage of total rainfall) are plotted against rainfall amounts from 27 natural and 10 simulated rainfalls monitored in the NTU-CSE slope. It appears from Figure 16 that rainfall events producing small total amounts of rainfall may contribute fully to infiltration. This again

Figure 31 shows the rainfall and runoff hydrographs of a natural composite storm recorded at the NTU-CSE slope on December 12, 1998, with a total rainfall of 45 mm, and two peak rainfall intensities of 240 and 120 mm/h, resulting 18 mm of total runoff. This composite storm is selected because it reflects the slope response to rainfall during dry and wet soil conditions. The first runoff hydrograph is characterized by a flash flood, which decreases quickly and represents overland flow under dry soil conditions. The first rainfall with 28 mm of rainfall produced 10 mm of runoff, which is about 35% of the rainfall. In contrast, the second rainfall event with similar rainfall characteristics and a total amount of 17 mm, but under wet soil conditions, produced about 8 mm of runoff, a consequence of increasing saturation condition of the slope. This runoff is about 47% of the rainfall.

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• In relatively gentle slopes shallow failures induced by moisture changes are more likely than deep seated failures. • However, if the slope angle increases deep failures become more frequent • For the cases analyzed here, increasing the permeability reduces the risk of failure • In general, the deeper the water tables the lower the risk of sliding. Shallow failures tend to develop, with more probability, if the water table is deep.

suggests the existence of a threshold rainfall amount. Any rainfall below this amount will not produce any runoff, and the whole rainfall may end up as infiltration. With reference to Figure 33 (broken line), this threshold appears to be about 10 mm of total rainfall. Beyond the threshold rainfall, the percentage of rainfall contributing to infiltration decreases with an increase in total rainfalls. The infiltration amount could decrease to about 40% of the rainfall (Figure 33) for rainfall events that produce a high total amount of rainfall. This, however, does not mean that the total infiltration amount is less during rainfall events with a higher total rainfall than the total infiltration during rainfall events with a smaller total rainfall. For example, with 40% of the rainfall contributing to infiltration, a 100 mm rainfall (higher total amount) would result in a total infiltration of 40 mm. With 100% of the rainfall contributing to infiltration, a 10 mm rainfall would produce only 10 mm of total infiltration. The data suggest that in residual soil slopes total infiltration could range between 40% and about 100% of the total rainfall depending on the rainfall amount. The relationship (Figure 33) derived from the rainfall records in the residual soil slope has practical significance. If the rainfall amount is known, Figure 33 could indicate the fraction of the rainfall that could become infiltration. This may be useful for seepage analysis that requires this information as flux boundary conditions. More rigorous interpretation on the hydrological responses of these slopes can be found in Rezaur et al. (2003); Rahardjo et al. (2005).

4

4.2

PART-A: The case history

Two deformation mechanisms were identified in the slope: a creep-type displacement, which was detected along the full depth of the soil investigated (around 12 m) and a surface planar slide. The first mechanism is interpreted as a deformation associated with volume changes of the over-consolidated clays as the water pressures change in time as a reaction to rainfall events. Water pressures at the upper two weathered layers are controlled by the atmospheric weather. A fairly good agreement between computed and measured water pressures is achieved when the three identified layers are characterized by three different permeabilities. For a given climatic record, the critical situation of a given slope (in the sense of reacting with the maximum development of water pressures) is obtained for a particular combination of layer permeabilities. In other words, given a soil profile and geometry and its associated permeabilities and additional water flow parameters, there exist rainfall records which lead to a maximum ‘‘reaction’’ of the slope in terms of pore pressure development. Permeability and water retention are therefore fundamental properties in slope stability analysis. As deformation accumulates, peak strengths are attained in surface layers and eventually remoulded and even residual strength conditions develop. Weathering mechanisms result also in a change in permeability, which is stronger the closer to the surface. Water pressures recorded are consistent with a decrease of permeability with depth. (10−6 m/s, 10−8 m/s, 10−9 m/s). Permeability transitions lead to peak pressures computed at the interfaces. Strength degradation (accumulated straining) and peak water pressures (positive or negative) result in the development of a sliding surface at the α−β interface, where minimum safety factors are consistently found. Once a planar sliding surface has developed the conceptual model of slope motion is simple: an upper reworked layer slides on top of a critical surface at residual or near-residual conditions. Periods of activity are dictated by suction changes in the upper few meters which depend critically on

CONCLUSIONS

4.1 PART-A: The stochastic model A method to analyze the risk of failure of slopes in partially saturated soils induced by climatic changes has been presented. Spectral and correlation functions for the safety margin of the slope were explicitly derived in terms of the spectral representation of the infiltration record. Some results of the theory of random processes were then used to find expressions for the failure risk in a given period (1 year). The method has been applied to the hydrologic conditions prevailing in the mediterranean coast close to Barcelona and to slopes in a partially saturated silty soil whose mean strength, permeability and suction characteristics were estimated from previous work and past experience. It was found that, • Cyclic infiltration/evaporation records exhibiting a large time period of occurrence can penetrate deep into the soil. High frequency components only affect shallow depths and dampen fast at depth.

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the slope geometry, flow boundary conditions, rainfall record, flow parameters (permeability, water retention properties) and its spatial variation. 4.3

Corominas, J. 2000. Landslides and climate. Proc. 8th Int. Conf. Landslides. Cardiff. pp. 1–31. Fredlund, D.G., Morgenstern, N.R., & Widger, R.A. 1978. The shear strength of unsaturated soils. Canadian Geotechnical Journal 15(3): 313–321. Gasmo, J.M. 1997. Stability of Unsaturated Residual Soil Slopes as Affected by Rainfall. Master of Engineering Thesis. School of Civil and Structural Engineering, Nanyang Technological University, Singapore. Meteorological Service Singapore. 1997. Summary of Observations (annual publication), Singapore. Pitts, J., & Cy, S. 1987. In situ soil suction measurements in relation to slope-stability investigations in Singapore. E.T. Hanrahan, T.L.L. Orr, and T.F. Widdis, eds., Proc., 9th European Conf. on Soil Mechanics and Foundation Engineering. Vol. 1, Balkema, Rotterdam, The Netherlands, 79–82. Poh, K.B., Chuah, H.L. & Tan, S.B. 1985. Residual granite soils of Singapore. Proceedings of the 8th Southeast Asian Geotechnical Conference, Kuala Lumpur, Malaysia. 11–15 March, 1985. 1(3):1–9. Premchitt, J., Lam, T.S.K., Shen, J.M. & Lam, H.F. 1992. Rainstorm runoff on slopes. GEO Rep. 12, Geotechnical Engineering Office, Hong Kong. Public Works Department. 1976. The geology of the Republic of Singapore, Singapore. Rahardjo, H., Aung, K.K., Leong, E.C. & Rezaur, R.B. 2004b. Characteristics of residual soils in Singapore as formed by weathering. Engineering Geology 73: 157–169. Rahardjo, H., Lee, T.T., Leong, E.C. & Rezaur, R.B. 2005. Response of a residual soil slope to rainfall. Canadian Geotechnical Journal 42: 340–351. Rahardjo, H., Lee, T.T., Leong, E.C. & Rezaur, R.B. 2004a. A flume for assessing flux boundary characteristics in rainfall-induced slope failure studies. Geotechnical Testing Journal 27(2): 145–153. Rahardjo, H., Leong, E.C. & Rezaur, R.B. 2007. Effect of antecedent rainfall on pore-water pressure distribution characteristics in residual soil slopes under tropical rainfall, Hydrological Process. 21: (in press). Rezaur, R.B., Rahardjo, H., Leong, E.C. & Lee, T.T. 2003. Hydrologic behavior of residual soil slopes in Singapore. Journal of Hydrologic Engineering, ASCE 8(3): 133–144, Sherlock, M.D., Chappell, N.A. & McDonnell, J.J. 2000. Effects of experimental uncertainty on the calculation of hillslope flow paths, Hydrological Processes. 14(14): 2457–2471. Tan, S.B., Tan, S.L., Lim, T.L. & Yang, K.S. 1987. Landslide problems and their control in Singapore. Proc., 9th Southeast Asian Geotechnical Conf., Southeast Asian Geotechnical Society, Bangkok, Thailand. Vol. 1, 25–36. Tani, M. 1997. Runoff generation processes estimated from hydrological observations on a steep forested hillslope with a thin soil layer. Journal of Hydrology 200(1–4): 84–109. Toll, D.G., Rahardjo, H. & Leong, E.C. 1999. Landslides in Singapore. 2nd Int. Conf. on Landslides, Slope Stability and the Safety of Infra-structures, Singapore. 269–276. Tretter, S.A. 1976. Introduction to Discrete Time Signal Processing. Wiley. NY.

PART-B: Slope hydrology and responses to climate changes

The pore-water pressure profile of the slopes presented in part-II shows distinct differences in slope hydrologic response during dry and wet conditions. The relationships between the increase in pore-water pressure and daily rainfall allows for an estimate of the rise in pore-water pressure due to rainfall. However, use of these relationships should be restricted to the field conditions under which measurements have taken place. The results of natural and simulated rainfall— runoff experiments conducted on the test slope suggest that a large proportion of the rainfall contributes to infiltration in the residual soil slope. A rainfall may contribute from 40% to about 100% of its total rainfall as infiltration (assuming negligible interception losses) depending on the rainfall amount. This information is useful for seepage analyses that require the total infiltration amount as an input parameter. There appears to be a threshold rainfall of about 10 mm to generate runoff. The characteristics of infiltration processes, runoff generation, and pore-water pressure changes identified in this study have relevance for the assessment of rainfall-induced slope instability in residual soil slopes under similar climatic conditions in different geographic regimes. ACKNOWLEDGEMENT This work presented in Part-B was funded by a research grant from National Science and Technology Board, Singapore (Grant: NSTB 17/6/16). The authors gratefully acknowledge the field assistance of the Geotechnics Laboratory staff, School of Civil and Environmental Engineering, Nanyang Technological University, Singapore, during the field instrumentation, trouble shooting and data collection for this study. REFERENCES ASTM. 1997. Annual book of ASTM standards, Philadelphia, 04.08–04.09. Bear, J. & Bachmat, Y. 1991. Introduction to Modeling of Transport Phenomena in Porous Media. Kluwer. Dordrecht. Chatterjea, K. 1994. Dynamics of fluvial and slope processes in the changing geomorphic environment of Singapore. Earth Surface Processes and Landforms 19: 585–607.

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Vanmarcke, E. 1975. On the Distribution of the First Passage Time for Normal Stationary Random Processes. Journal of Applied Mechanics Division. A.S.M.E. March: 215–220. Wieczorek, G.F. 1996. Landslide triggering mechanisms, In A.K. Turner & R.L. Schuster (eds.) Landslides: investigation and mitigation. TRB Special Report, 247. National Academy Press, Washington, 1996, 76–90.

ω0 = Sr0 − as pω0 − γω (q0 K −1 )

5

1 = 11 β11 − 12 β12



H +h 2



γω n + γs (1 − n)

× (tan φ ′ cos α − sin α) cos α (H − h) 

q  0 −1 + c′ + tan(φb ) pa − hγω K

APPENDIX: COEFFICIENTS FOR EQUATIONS (10) AND (11)

2 = 12 β11 + 11 β12

Tn = ω21 Cn + ω22 Dn Wn = −ω22 Cn + ω21 Dn

f1 (Bn − An ) + f2 (Bn + An ) Cn = 2αn (f12 + f22 )

11 =

ω1 H (−ω11 + ω12 ) − ω2 ω21 2αn

12 =

ω1 H (ω11 + ω12 ) − ω2 ω22 2αn

  ω11 = cos αn (e−αn + eαn ) − cos αn Z e−αn Z + eαn Z

f2 (Bn − An ) − f1 (Bn + An ) 2αn (f12 + f22 )  α n = 1 2λ n   f1 = cos αn e−αn + eαn   f2 = sin αn −e−αn + eαn   ω21 = cos αn Z e−αn Z − eαn Z   ω22 = sin αn Z −e−αn Z − eαn Z

ω12 = sin αn (e−αn − eαn ) − sin αn Z(e−αn Z − eαn Z )

Dn =

β11 =

f2 − f1 2αn (f12 + f22 )

f2 + f1 2αn (f12 + f22 ) 

q   0 − 1 cos α cos α tan φ ′ − sin α ω1 = γω2 nas H K

q  0 −1 ω2 = tan φb γω H K β12 =

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Soil nailing and subsurface drainage for slope stabilization W.K. Pun Geotechnical Engineering Office, Civil Engineering Department, Government of the Hong Kong Special Administrative Region, China

G. Urciuoli Department of Geotechnical Engineering, Via Claudio, Napoli, Italy

ABSTRACT: A wide range of slope stabilization and protective measures are available. Soil nailing and subsurface drainage are amongst the very commonly used techniques for slope stabilization. The concept of soil nailing involves creating a stable block of composite material by strengthening the insitu ground with soil nails. The interaction between the ground and the soil nail is complex, and the mobilization of forces in the soil nail is dependent on many factors. The mechanism of subsurface drains in slopes involves a decrease in pore water pressures in the subsoil, and consequently an increase in effective stresses and soil shear strength in the whole drained domain. This paper gives an overview of the mechanism of soil-nailed system and subsurface drainage measures and presents some geotechnological developments related to their applications. 1

Zaruba & Mencl 1982, Leventhal & Mostyn 1987, Schuster 1992, Hausmann 1992, Fell 1994, Holtz & Schuster 1996, Perry et al, 2003, Ho 2004). This paper focuses on the developments in geotechnology in the two of the common stabilization measures, viz. soil nailing and subsurface drainage.

INTRODUCTION

Landslides have resulted in the loss of human lives and properties in many parts of the world. To combat landslide risk, a wide range of risk mitigation measures are available. These range from hard engineering measures of slope stabilization and landslide protective works to soft community means of public education. Stabilization works aim at reducing the likelihood of failure of a slope whereas the other measures reduces the risk by minimising the consequences of slope failures. The range of slope stabilization works may be categorized as follows (Ho 2004): a. b. c. d. e. f. g. h. i.

2

SOIL NAILING TECHNOLOGY

2.1 Introduction The soil nailing technique was developed in the early 1960s, partly from the techniques for rock bolting and multi-anchorage systems, and partly from reinforced fill technique (Clouterre 1991, FHWA 1998). The New Austrian Tunnelling Method introduced in the early 1960s was the premier prototype to use steel bars and shotcrete to reinforce the ground. With the increasing use of the technique, semi-empirical designs for soil nailing began to evolve in the early 1970s. The first systematic research on soil nailing, involving both model tests and full-scale field tests, was carried out in Germany in the mid-1970s. Subsequent development work was initiated in France and the United States in the early 1990s. The result of this research and development work formed the basis for the formulation of the design and construction approach for the soil nailing technique in the subsequent decades. The concept of soil nailing involves creating a stable block of composite material by strengthening the

surface protection and drainage, subsurface drainage, slope regrading, retaining structures, structural reinforcement, strengthening of slope-forming material, vegetation and bioengineering, removal of hazards, and special materials and techniques.

Further detail breakdown of the various categories of slope stabilization works and landslide protective measures are depicted in Figure 1. The development, characteristics, application and performance of various types of slope stabilization works have been reported by many researchers and practitioners (e.g. Hutchinson 1977, Veder 1981,

85

Surface protection and drainage

Subsurface drainage

Slope regrading Retaining structures

Structural reinforcement

'impermeable' cover vegetated cover with geotextile protection toe erosion protection measures infilling of tension cracks surface drainage channels inverted granular filter over discharge points short relief drains trench (counterfort) drains cut-off drains raking drains vertical drainage wells drainage galleries/adits siphon drains geosynthetic drains vertical sand/gravel drains pressure relief wells electro-osmosis drainage vacuum dewatering blasting

pumped gravity drainage

cut back toe weighting embedded walls (cantilever or tie-back) gravity walls composite walls materials: metallic, polymeric, reinforced fill woven/non-woven, fibre glass, electro-kinetic geo-synthetics construction: reinforced earth/anchored earth (usually horizontal reinforcement, but may have vertical reinforcement), hybrid system (e.g. segmental block and gabion) insitu reinforcement

soil nails (usually unstressed but may be prestressed) prestressed anchors (large presumes) soil dowels

chemical admixtures

recompaction

Strengthening of slope-forming material (exclude structural reinforcement and drainage)

shear trenches grouting

others

Vegetation and bioengineering

Defensive measures Removal of h azards Special materials and techniques

reticulated micropiles rock bolts and rock dowels lime nails/piles other techniques (e.g. electro-osmotic tie-back) aerated cement suspension and mortar lime flyash/lime slurry Calcite In-situ Preciptation System (CIPS) replacement of failed material with compacted fill surface recompaction of loose fill slopes dynamic compaction compaction induced by displacement piles trenches backfilled with compacted gravel or no-fines concrete fracture grouting compacting grouting permation grouting void infilling dentition, buttressing of loose rock blocks, various ground improvement techniques

tree root reinforcement bioengineering - live staking, live fascines, brushlayering, willow poles, tree root reinforcement branchpacking, etc. ('living approach') reinforced grass system, Biobund system, etc. ('combined living and non-living' approach) netting and meshing (including concrete grillage) rock ditches debris barriers scaling of loose rock blocks removal of boulders trimming of local over-steepened slope excavation of local loose fill lightweight fill, EPS, Pneusol, Rubber Soil TM, reuse of waste materials underdrainage to landslide trail, thermal treatment, short-circuit conductors, ion-exchange

Figure 1. Classification of slope stabilization and protective measures (after Ho 2004).

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materials: metallic, polymeric installation: drill & grout, jacked, driven, ballasted large diameter bored piles driven piles hand-dug caissons

• Long soil nails are difficult to install, and thus it renders the use of the soil nailing technique more difficult in dealing with deep-seated landslides and sizeable slopes. • As soil nails are not prestressed, mobilisation of soil-nail forces will be accompanied by ground deformation. The effects on nearby structures, or services may have to be considered, particularly in the case of soil-nailed excavations. • Soil nails are not effective in stabilising localised steep slope profiles, back scarps, overhangs or in areas of high erosion potential. Suitable measures, e.g., local trimming, should be considered prior to soil nail installation.

insitu ground with soil nails. This requires that the soil nails are installed at close spacing, both horizontally and vertically. The following are typical merits of adopting the soil nailing technique in respect of ease of construction, cost and performance (GEO 2008): • It is suitable for sites with difficult access because the construction plant required for soil nail installation is small and mobile. • It can more easily cope with site constraints and variations in ground conditions encountered during construction, e.g., by adjusting the location and length of the soil nails to suit the site conditions. • During construction, it causes less environmental impact than cutting back and retaining wall construction as no major earthworks and tree felling are needed. • There could be time and cost savings compared to conventional techniques of cutting back and retaining wall construction which usually involve substantial earthworks and temporary works. • It is less sensitive to undetected adverse geological features, and thus more robust and reliable than unsupported cuts. In addition, it renders higher system redundancy than unsupported cuts or anchored slopes due to the presence of a large number of soil nails. • Its failure mode is likely to be ductile, thus providing warning signs before failure.

2.2 Mechanisms of soil-nailed system 2.2.1 Load transfer mechanism The soil nailing technique improves the stability of slopes, retaining walls and excavations principally through the mobilisation of tension in the soil nails. The tensile forces are developed in the soil nails primarily through the frictional interaction between the soil nails and the ground as well as the reactions provided by soil-nail heads and the facing (Fig. 2). The tensile forces in the soil nails reinforce the ground by directly supporting some of the applied shear loadings and by increasing the normal stresses in the soil on the potential failure surface, thereby allowing higher shearing resistance to be mobilised. Soil-nail heads and the facing also provide a confinement effect by limiting the ground deformation close to normal to the slope surface. As a result, the mean effective stress and the shearing resistance of the soil behind the soilnail heads will increase (Fig. 3). Soil-nail heads and the facing also help preventing local failures near the surface of a slope and promote an integral action of

Like every other stabilization technique, soil nailing has its limitations: • The presence of utilities, underground structures, or other buried obstructions poses restrictions to the length and layout of soil nails. • The zone occupied by soil nails is sterilised and this site poses constraint to future development. • Permission has to be obtained from the owners of the adjacent land for the installation of soil nails beyond the lot boundary. This places restrictions on the layout of soil nails. • The presence of high groundwater levels may lead to construction difficulties in hole drilling and grouting, and instability problem of slope surface in the case of soil-nailed excavations. • The effectiveness of soil nails may be compromised at sites with past large landslides involving deepseated failure due to disturbance of the ground. • The presence of permeable ground, such as ground with many cobbles, boulders, highly fractured rocks, open joints, or voids, presents construction difficulties due to potential grout leakage problems. • The presence of ground with a high content of fines may lead to problems of creeping to soil nails.

Figure 2.

87

Load transfer mechanism of soil nailed structure.

ground, the nail inclination, and the tensile strength, shear strength and bending capacity of the soil nail. Generally speaking, the axial strain will mobilize tensile or compressive forces, and the lateral strain will mobilize shear force and bending moment in the soil nail. Due to relatively slender dimensions of soil nails, the reinforcing actions from shear and bending are limited by the small flexural strength, and they are usually negligible (FHWA 1998). The effect of inclination and bending stiffness of soil nail are discussed further in the subsequent sections of this paper. Compressive and shear strains are developed in the soil behind a soil-nail head in response to the ground deformation in the active zone (Fig. 2). If the resultant strain is close to the direction perpendicular to the base of soil-nail head, the head-ground interaction will be predominantly in the form of a bearing mechanism. However, if the resultant strain is in a direction that deviates significantly from the normal to the base of the soil-nail head, the head-ground interaction will be a combination of bearing and sliding mechanisms. These interaction, particularly the bearing mechanism, gives rise to tensile loads at the heads of soil nails. The tensile loads at the soil-nail heads are taken up by the soil-nail reinforcement. The interaction increases as the size of the soil-nail heads or the coverage of facing increases, resulting in larger tensile loads. Further discussion on the effect of soil-nail heads is given in Section 3.5 below. The mobilization of pullout resistance along a soil nail in the passive zone depends on many factors. Theoretically, the bond strength between the soil nail and the ground depends on the contact stress and the interface coefficient of friction. Where a soil nail is installed by the drill-and-grout method, the process of drilling reduces significantly the radial stress at the circumference of the drillhole. The drillhole remains stable by soil arching. Subsequent grouting will restore a certain level of the radial stress in the soil around the drillhole. The contact pressure at the drillhole face is generally small compared to the overburden pressure except where pressure grouting is adopted. This seems to imply small bond strength at the ground/grout interface. In reality, as the drillhole face may be irregular and rough, the mechanical interlocking between the cement grout and the ground also contributes a significant portion of the bond strength. Upon pulling of the soil nail, shearing may occur within the ground mass in a finite zone surrounding the soil nail. If the soil is dilative, the effect of restrained soil dilatancy will come into play. The effect of this can be significant and can lead to high soil-nail friction (Pun & Shiu 2007). Soil nails are considered to tie the active zone to the passive zone. It should be noted that the two-zone concept is only an idealisation for design purpose. In reality there is a complex shearing zone subject to shear distortion, unless the failure is dictated by joint

10kPa 30kPa

nail head position (400mm)

50kPa 70kPa 90kPa 110kPa 130kPa 150kPa 170kPa

(a) Contours of mean effective stress in soil nailed slope 10kPa

30kPa 50kPa 70kPa 90kPa 110kPa 130kPa 150kPa 170kPa

(b) Contours of mean effective stress in unreinforced slope

Figure 3. Contour of mean effective stress in (a) soil-nailed slope, and (b) unreinforced slope.

the reinforced ground mass through redistribution of forces among soil nails. The resistance against pullout failure of the soil nails is provided by the part of soil nails that is embedded into the passive zone. When there is a tendency of ground movement in the active zone, the soil nail will experience both axial and lateral strains through two fundamental mechanisms of nail-ground interaction. They are: (i) the nail-ground friction that leads to the development of axial strains in the soil nails, and (ii) the soil bearing on the soil nails and the nail-ground friction on the sides of soil nails that lead to the development of lateral stains in the soil nails. In these two mechanisms, the interactions between the ground and the soil nails are complex and the forces developed in the soil nails are influenced by many factors such as the bearing capacity of the ground to resist reaction force from the soil nail, the relative stiffness of the soil nail and

88

settings where the failure surface is distinct. The effect of the shear zone on the mobilization of forces in soil nails is discussed in Section 3.4 below. 2.2.2 Effect of nail inclination Unlike the reinforcements in reinforced fill structures, which are placed in horizontal direction, soil nails can be installed in the ground at various inclinations. In cramped sites, soil nails are sometimes installed at large inclinations. Different nail inclinations may produce different effects on the behaviour of soil-nailed (a) Principal Strains in Unreinforced Sand after Peak structures. In this paper, nail inclination, α, is the angle of a soil nail made with the horizontal; and nail orientation, θ, is the angle between a soil nail and the normal to the shearing surface. The typical relationship between α and θ is presented in Figure 4. Jewell (1980) investigated the fundamental behaviour of reinforced soil by carrying out a series of direct shear box tests on sand samples reinforced with bars and grid reinforcements. One of the significant findings of his work was that the shear strength of the reinforced soil is dependent on the orientations of the reinforcements. Jewell’s investigation shows that rein(b) Principal Strains in Sand Reinforced by a Grid at an forcement significantly modifies the state of stress and Orientation = +30º (legend: double arrow represents principal strain in soil, and that by varying the orientation of the tensile strain) reinforcement, the reinforcement can either increase or decrease the shear strength of the soil. Figure 5 Figure 5. Incremental strains at peak shearing resistance in compares the pattern of strain in soil between the unreinforced and reinforced sand (after Jewell 1980). unreinforced and reinforced tests. The presence of the reinforcement causes a significant reorientation of the principal axes of strain increment of the soil. The soil strains close to the reinforcement are small because the reinforcement inhibits the formation of the failure plane. When the reinforcement is orientated in the same direction of the tensile strain increment of the soil, tensile forces are induced in the reinforcement through the friction between the soil and the reinforcement. Likewise, compressive forces are induced in the reinforcement if the reinforcement is placed close to the compressive strain increment of the soil. Figure 6 shows the orientations of the reinforcement in which compressive or tensile strain increments Figure 6. Increase in shear resistance for reinforcement placed at different orientation (after Jewell & Wroth 1987).

are experienced. The shear strength of the soil starts to increase when the reinforcement is placed in the direction of tensile strain increment, and it reaches a maximum when the orientation of the reinforcement is close to the direction of the principal tensile strain increment. When the reinforcing elements are oriented in a direction of a compressive strain increment, there is a decrease in shear strength of the reinforced soil.

Figure 4. Relationship between nail inclination and orientation.

89

Figure 7. Geometry and material parameters of model slope.

This shows that in order to optimise strength improvement of soil, the reinforcement should be placed in the directions of principal tensile strain in the soil. When the reinforcement deviates from its optimum orientation, strength improvements decrease. Results of laboratory investigation of the behaviour of soil reinforced with bars were also reported by Marchal (1986), Hayashi et al (1988), and Palmeira & Milligan (1989). They showed that the orientation of reinforcements played a significant role in the improvement of the shear strength of the reinforced soil mass. A negative orientation can lead to the development of compressive force in the reinforcement and consequently loss of shear strength of the soil. There exists an optimum reinforcement orientation in terms of strengthening of the soil. These findings are consistent with that of Jewell (1980). The effects of nail inclination on the safety margin of a slope was studied by Shiu & Chang (2005) in Hong Kong by means of numerical analysis using the two-dimensional finite difference code, Fast Lagrangian Analysis of Continua (FLAC) (Itasca 1996). A simulated slope of 20 m in height, standing at an angle of 55◦ , and with an up-slope of 10◦ in gradient was adopted for the analysis. Figure 7 shows the geometry of the slope and the material parameters used in the numerical analysis. Each soil nail was 20 m long with a 40 mm diameter steel bar in a 100 mm grouted hole. A Mohr Coulomb model was assumed for the soil. A cable element was used to represent the soil nail as the bending stiffness of the soil nail was not considered. Developments of the tensile forces in the soil nails were governed either by the tensile strength of the nail or the peak shear strength at the soil-grout interface. Slope stability analysis was first carried out on the unreinforced slope. From the results of the analysis, the unreinforced slope has a minimum factor of safety (FoS) close to 1.0 for the initial soil strength parameters of c′ = 10 kPa, and φ ′ = 43◦ . In slope engineering, the FoS is conventionally defined as the

ratio of the actual soil shear strength to the minimum shear strength required for equilibrium. As pointed out by Duncan (1996), FoS can also be defined as ‘‘the factor by which the shear strength of the soil would have to be divided to bring the slope into a state of barely equilibrium’’. FoS can therefore be determined simply by reducing the soil shear strength until failure occurs. This strength reduction approach is often used to compute FoS using finite element or finite difference programs (Dawson et al 1999, Krahn 2003). Figure 8 shows that for soil nails with a small inclination of 20◦ , tensile forces are developed in all the soil nails. On the other hand, when the soil nails are inclined steeply at an inclination of 55◦ , compressive forces are developed in the top four rows of soil nails whereas tensile forces are mobilized only in the bottom three rows of soil nails. Tensile forces in the soil nails can improve the slope stability whereas compressive forces can have opposite effect. Increases in FoS ( FoS) due to the soil nails were calculated for different nail inclinations. Figure 9 shows the relationship between the calculated FoS and nail inclinations (α) for the model slope. The FoS is close to 1 with little variations for the range of α between 0◦ and 20◦ . The FoS decreases quickly as α increases beyond 20◦ , reflecting that the reinforcing effects of the soil nails reduce rapidly with increasing nail inclinations. At α = 55◦ , the value of FoS is almost zero. These studies show that the nail inclination can significantly affect the reinforcing action of the soil nails. Increase in nail inclination would decrease the efficiency of the reinforcing action of the soil nails. For steeply inclined soil nails, axial compressive forces may be mobilized in the soil nails. The compressive forces would reduce the stability of the soil-nailed structure. 2.2.3 Effect of bending stiffness of soil nail Steel reinforcements can sustain shear forces and bending moments, and thus this ability of steel soil

90

Figure 9. Variation of increase in factor of safety against inclination of soil nails.

Figure 8. slope.

laboratory and theoretical study reported by Pedley (1990) and Jewell & Pedley (1990, 1992). In the study of Pedley (1990), a series of direct shear tests were carried out in a large-scale direct shear apparatus (1 m × 1 m × 1 m). Three different types of circular elements were tested. They were solid steel bars (16 to 25.4 mm in diameter), metal tubes (15.88 to 25.4 mm in external diameter and 13.19 to 22.36 mm in internal diameter) and grouted bar (50.8 mm in diameter with steel bar diameters 6.71 to 16 mm). Figure 10 shows the distributions of: (a) the reinforcement bending moment (M) normalised by the plastic moment capacity (Mp ); (b) the reinforcement shear force (Ps ) normalised by the full plastic axial capacity (Tp ); and (c) the lateral stress on the reinforcement (σ ′l ) normalised by the limiting soil bearing stress (σ ′b ). These distributions were the stress conditions of the reinforcement at a shear displacement of soil of 60 mm. The test results show that even when the measured bending moments were close to the fully plastic moment (M/Mp = 1) in all the tests, the maximum shear force (Ps ) in the reinforcement was less than 6% of the plastic axial capacity (Pp ). Pedley (1990) confirmed the laboratory test results by theoretical analysis. He derived elastic and plastic models for determining the maximum shear force mobilised in the reinforcement bar. These models were also reported in Jewell & Pedley (1990, 1992). The elastic analysis simply defines the stress condition before the reinforcement reaches plasticity; it does not represent the failure condition. As such, only the plastic analysis is discussed here. The plastic analysis will always give a larger reinforcement shear force than the elastic analysis. The limiting plastic envelope for a bar of rectangular cross-section is given by (Calladine 2000):

Axial force distribution in soil nails in model

nails may enhance the shear strength of soil. The development of shear force in soil nails involves a mechanism which is dependent on the relative stiffness of the soil nail and the ground, the soil bearing strength, the orientation and shear deformation of reinforcement, and the thickness of shear zone. The effect of the bending stiffness of a soil nail on nail forces and displacements has been investigated by many researchers, e.g. Schlosser (1982), Marchal (1986), Gigan & Delmas (1987), Pluemelle et al (1990), Pedley (1990), Jewell & Pedley (1990, 1992), Bridle & Davies (1997), Davies & Le Masurier (1997), Smith & Su (1997) and Tan et al (2000). The most notable and comprehensive investigation was the

M + Mp

91



T TP

2

=1

(1)

Figure 11. Plastic analysis of soil nail interaction (after Jewell & Pedley 1992).

bar, σy is the yield stress of the bar and σb′ is the limiting bearing stress between soil and reinforcement. The limiting bearing pressure (σb ’) between the soil and the reinforcement required to achieve the plastic equilibrium is:

Figure 10. Profiles of bending moment, shear force and lateral stress in reinforcement (after Pedley 1990).

(σb′ )max =

Equation 1 is slightly conservative for circular bar. Since no simple relationship can be derived for circular bar, this equation has been adopted by Jewell & Pedley (1990). The relation between maximum shear force Ps and the maximum moment Mmax depends on the magnitude and distribution of the lateral loading on the reinforcement bar. For plastic analysis for a soil nail under the lateral loading shown in Figure 11, the equations for this distribution of lateral loading are: 4Mmax (P s ) = ls 

4σy ls = D 3σb′

8Mmax ls2 D

(4)

The theoretical plastic limiting maximum shear force Ps that can be generated in an ungrouted round bar that also supports axial force, P, is:   2  T 8 Ps = 1− Tp 3π(ls /D) Tp

(5)

Details of explanations and derivation of Equations (1) to (5) can be found in Pedley (1990). Jewell & Pedley (1992) computed and presented envelopes of limiting combinations of shear force Ps and axial force T for a grouted reinforcement bar of 25 mm diameter with typical soil parameters and showed that the magnitude of the limiting shear force in the reinforcement was only a small proportion of the axial force capacity. This was the case even when the reinforcement was oriented so as to mobilize the maximum shear force.

(2)

(3)

where ls is the distance between the points of maximum moment on either side of the potential shear surface, (Fig. 11), D is the diameter of reinforcement

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Pedley (1990) also back-analysed an instrumented 6 m high soil-nailed wall, which was loaded to failure. He found that the highest contribution of reinforcement shear force to soil strength improvement was less than 3% to that due to reinforcement axial force. This is in agreement with the theoretical study result that only very small amount of shear force can be mobilised in a soil nail. The effect of bending stiffness of soil nails on nail forces and displacements and the safety margin of soilnailed slopes has also been studied by Shiu & Chang (2005) in Hong Kong by means of numerical simulations using the finite element code PLAXIS. The slope model is the same as that shown in Figure 7. The FoS of the soil-nailed slope, the tensile forces, shear stresses and bending moments developed in the soil nails at different inclinations were computed. The maximum axial force developed in a soil nail is Tmax . Figure 12 shows the sum of the maximum tensile forces mobilised in all the soil nails (Tmax ) at limit equilibrium condition of the slope model. The maximum shear force in a soil nail at the location where the shear plane intersects the soil nail is Psmax . The sum of the maximum shear forces (Psmax ) mobilized in the soil nails at limit equilibrium condition of the model are also plotted in Figure 12. The value of Psmax rises steadily with increasing nail inclination. The rise is small, from 31 kN/m at α = 10◦ , to 76 kN/m at α = 55◦ . In contrast, the value of Tmax decreases rapidly with increasing nail inclination. For small nail inclinations, Tmax is much larger than Psmax . Comparing Figures 9 and 12, it can be seen that both FoS and Tmax decrease with increasing nail inclinations. This similarity illustrates that FoS is strongly influenced by the nail axial force. The FoS is not sensitive to the mobilized shear resistances in the soil nails. The modeling results show that small shear forces are mobilized in soil nails and they have little effect on the factor of safety of the slope, except perhaps at very steep nail inclination where dowel action starts

to play a role. However, soil nails are not effective in providing dowel action. For that purpose, other types of structural element should be considered, e.g. large diameter piles. The above studies show that the contribution of shear force and bending stress of soil nails on enhancing the shear strength of the soil mass is very small. Large soil displacements are required to mobilise shear and bending forces in the soil nails. When failure conditions are approached, the contribution of shear and bending action may be more significant but is still small. For these reasons, the soil nail design practice in most countries, such as the USA (FHWA 1998), the UK (Department of Transport 1994), Japan (JHPC 1998) and Germany (Gässler 1997), ignores any beneficial effects from the mobilisation of shear force or bending stress in the soil nails. An exception to this is the French design approach (Clouterre 1991) in which the contributions of shear and bending of the nails are considered. Clouterre (1991) emphases that shear forces are mobilised in soil nails only when the structures are near failure. Although the bending stiffness of soil nail has little contribution to the shear resistance of a soilnailed system, the beneficial effect of shear ductility of soil nail reinforcement should not be ignored. For example, steel has large shear ductility. As a result of the mobilization of shear and bending ductility at large deformations, a soil-nailed system comprising steel reinforcement tends to exhibit ductile rather than brittle failure. 2.2.4 Effect of thickness of shear zone The reported laboratory investigation and numerical studies generally considered sliding along a shear plane. To investigate the influence of the thickness of shear zone on the mobilisaion of shear force and bending moment in soil nails, a study involving numerical simulations of large shear box tests was conducted by the Geotechnical Engineering Office (GEO) in Hong Kong. The two-dimensional finite element code PLAXIS was used. Three cases were considered, ranging from a well-defined shear plane to a wide shear zone (Fig. 13). The model shear box was 3 m deep and 6 m long. A reinforcement was placed in the middle of the box across the vertical slip surface. The box was assumed filled with homogeneous sand. The model and the assumed parameters are shown in Figure 14. A Mohr-Coulomb model was assumed for the soil. In the numerical simulation, the shear box was initially restrained to move. An overburden pressure of 80 kPa was applied on the top of the box to model a 5 m high earth pressure. In the case of sliding along a shear plane, an imposed downward uniform displacement δ of 5 mm, 10 mm, 25 mm and 50 mm was applied in sequence at the right-half part of the shear box.

Figure 12. Variation of total maximum tensile force (Tmax ) and total maximum shear force (Ps max ) with nail inclination (α).

93

Figure 13. Cases considered in numeral study of effect of shear zone.

Figure 15.

Shear force along steel bar in shear box model.

Figure 16. model.

Bending moment along steel bar in shear box

25 mm and 50 mm was again applied in sequence in the downward direction beyond the shear zone along the top and bottom of the right-half part of the box. Two values of z, 100 mm and 200 mm respectively, were considered. In the case of wide shearing zone, the shear zone was extended to cover the entire righthand part of the shear box. The results of the numerical simulations are shown in Figures 15 and 16. The mobilisation of shear force and bending moment in soil nails is affected by the thickness of the failure shear zone. The narrower the shear zone, the higher is the shear force and bending moment in the soil nails.

Figure 14.

2.2.5 Effect of soil nail head and facing Our understanding of the magnitude and distribution of loadings developed at soil-nail head is not as good as our knowledge of the development of tensile forces in soil nails (FHWA 1998). This is because of the lack of good quality field monitoring data. The available data from instrumented soil nails are generally difficult to interpret in the vicinity of soil-nail heads, where bending effects of soil nail tend to be more significant arising from the weight of the soil-nail head (Thompson & Miller 1990). There has been little field

Model for numeral study of effect of shear zone.

In the case of a narrow shearing zone, a linearly varying downward displacement was applied across the shear zone of width z during the loading phases. An imposed uniform displacement of 5 mm, 10 mm,

94

monitoring data obtained using load cells at soil-nail heads probably because of the difficulties in placing load cells between soil-nail heads and soil (Stocker & Reidinger 1990). Despite the lack of good quality field monitoring data, a number of studies including model tests (e.g. Muramatsu et al 1992, Tei et al 1998), fullscale field tests (e.g. Gässler & Gudehus 1981, Plumelle & Schlosser 1990, Gutierrez & Tatsuoka 1988, Muramatsu et al 1992) and numerical simulations (e.g. Ehrlich et al 1996, Babu et al 2002) have been carried out. The results of these studies provide useful insight into the role and behaviour of soil-nail heads. Many of the studies are related to soil-nailed retaining walls where soil-nail heads are integrated into a concrete facing. Gutierrez & Tatsuoka (1988) reported loading tests performed on three model sand slopes: (i) unreinforced slope, (ii) slope reinforced with metal strips but without a facing, and (iii) slope reinforced with metal strips and with a facing (Fig. 17). The slopes were loaded at the crest by a footing with a smooth

base. Result of the tests is shown in Figure 18. It indicates that the reinforced slope with facing can sustain a higher load than the reinforced slope with no facing, and a much higher load than the unreinforced

Figure 19. Shear stain contour in model tests by Gutierrez & Tatsuoka (1988).

Figure 17. Model slope reinforced with metal strips and with a facing tested by Gutierrez & Tatsuoka (1988).

Figure 18.

Figure 20. Relationship between factor of safety and soilnail head size.

Results of tests of Gutierrez & Tatsuoka (1988).

95

Figure 21.

Variation of axial nail forces for (a) 800 mm soil head and (b) no nail head.

in nails with heads of 800 mm wide. For the soil nails without soil-nail heads, no tensile force is developed at the front end of the soil-nail whereas for the soil nails with soil-nail heads, large tensile forces are mobilised in the soil nails at the connections to the soil-nail heads. The maximum tensile forces mobilized along the soil nails are much larger in the latter case. A series of centrifuge tests has also been conducted in the Geotechnical Centrifuge Facility of the Hong Kong University of Science and Technology to investigate the reinforcing effect of soil nails and soil-nail heads (Ng et al 2007). Figure 22 shows an instrumented model used in one of the nailed slope centrifuge tests. The test results support the results of the numerical simulations that soil-nail heads can substantially improve the stability of a soil-nailed slope. Results of the above model tests, field measurements and numerical simulations highlight the importance of soil-nail heads in the soil nailing applications. They show that soil-nail heads, whether in the form of individual concrete pads or as part of concrete facing, greatly enhance the stability of a soil-nailed slope.

slope. Figure 19 shows the failure planes of the slopes when loaded at the crest. Deep and shallow failure planes were observed in the unreinforced slope. For the reinforced slope with no facing, failure took place close to the slope face. This was because the reinforcement alone was not effective at retaining the active zone. For the reinforced slope with facing, the failure was observed at a greater depth. The maximum tensile force generated in the reinforcement was larger than that in the reinforced slope with no facing. Furthermore, substantial tensile force was induced in the reinforcement at the connection to the facing. The fullscale field tests by Plumelle & Schlosser (1990) and Muramatsu et al (1992) also showed that the presence of a facing could enhance the stability of a reinforced slope and helped prevent shallow failures. Soil-nail heads used in slope stabilization works in Hong Kong are usually in the form of isolated reinforced concrete pads. To investigate the effect of soil-nail heads on stability of nailed slopes, a series of numerical simulations using the two dimensional finite element code FLAC was conducted (Shiu & Chang 2004). Figure 7 shows the slope model. Strength reduction technique (Dawson et al 1999) was employed to compute the factor of safety (FoS) of the model slope. In the simulations, soil-nail heads of different sizes were considered. The slope without any soil nails (i.e. unreinforced) had a minimum FoS close to 1. Based on the FLAC analysis, Figure 20 shows the relationship between the calculated FoS of the model slope and soil-nail head sizes. The FoS increases from 1 for the unreinforced slope to 1.2 for the soil-nailed slope with no soil-nail heads. Substantial increases in the FoS are observed when soil-nail heads of sizes ranging from 400 mm wide to 800 mm wide are provided. The trend of increase in FoS levels off for soil-nail head sizes larger than 800 mm wide. It shows that soil-nail heads can significant enhance the stability of a soil-nailed slope. Figure 21 compares the axial tensile forces developed in soil nails without soil-nail heads with those

2.3

Modelling and design

2.3.1 Design approach and standard A soil-nailed structure is required to fulfil fundamental requirements of stability, serviceability and durability during construction and throughout its design life. Other issues such as cost and environmental impact are also important design considerations. The design for stability generally entails the setting up of ground and design model, consideration of potential failure mechanisms, stability analyses, determination of soil-nail design capacity, soil-nail head and facing design, and detailing. The failure mechanisms of nailed structures can broadly be classified as external failure and internal failure. External failure refers to the development of

96

50 0

Upstream drainage board

Upstream temporary reservior

Model box 280

100

20 ¡ã PPT13 PPT14 Nail F PPT12 Nail E PPT11 PPT10 CoarseCDG Nail D block PPT9

PPT7

Inlet hole

PPT5 PPT1

codes. Driven by the Eurocode, all the European countries under the European Union use the partial factors approach in soil nail design. In the USA, it appears that it is at a transition stage of changing over from global safety factor approach to partial safety factors approach. The American design code permits the use of either load and resistant factor design approach (which is similar to partial factors approach) or service load design approach. In Hong Kong, the soil nail design approach is essentially a combination of global safety factor approach (permissible stress design) and partial safety factor approach. Table 1 summarises the design approaches recommended by the different design codes. Calculation methods involving trial wedges (singlewedge or double-wedge) and limit equilibrium methods (LEM) of slices on circular, spirial, or other non-circular slip surfaces are commonly used. While these methods are good enough for design purpose, none of them can account for the actual behaviour of a soil-nailed structure, which is a strain compatibility problem. It is possible to define a wide variety of nail length patterns that satisfy stability requirements but that may not satisfy serviceability requirements. It is essential to have a good understanding of the principles behind the calculation methods so that the appropriate method is used and the results are interpreted correctly. For instance, the factor of safety of a soil-nailed slope computed using the simplified Janbu method is in sensitive to the location of the applied soil nail force. This is an inherent limitation of the method, and it may give rise to an over- or under-estimation of the true safety margin. In light of this, GEO (2008) recommends that only stability analysis methods that satisfies both moment and force equilibrium should be used in soil nail design. In Hong Kong, more than 3,000 slopes and retaining walls have been stabilized using soil nails. The vast amount of soil nail designs had allowed the development of a prescriptive design approach for the design of soil-nailed soil cut slopes and retaining walls. Prescriptive measures are pre-determined, experience-based and suitably conservative modules of works prescribed to a slope or retaining wall to improve its stability or reduce the risk of failure, without detailed ground investigation and design analyses. Using prescriptive measures has the technical benefits of enhancing safety and reducing the risk of failure, by incorporating simple, standardised and suitably conservative items of works to deal with uncertainties in design that are difficult to quantify. There would also be savings on time and human resources, by eliminating detailed ground investigation and design analyses. The prescriptive soil nail design guidelines in Hong Kong comprise standard soil nail layouts and a set of qualifying criteria for the application of the prescriptive measures to ensure that the prescriptive design

PPT Unit:mm

LV DT1 LV DT2

Downstream drainage board 50 0

Nail C PPT8 50 PPT6 65¡ã Nail B PPT2 40 0 PPT4 PPT3

Downstream temporary reservior Outlet hole 100

1130

(a) Set-up of a nailed-slope model in centrifuge test.

(b) Front view of the slope model in centrifuge test

Figure 22.

Centrifuge test to study soil-nailed slope.

potential failure surfaces essentially outside the soilnailed ground mass. The failure can be in the form of sliding, rotation, bearing, or other forms of loss of overall stability (see Fig. 23a). Internal failure refers to failures within the soil-nailed ground mass. Internal failures can occur in the active zone, passive zone, or in both of the two zones of a soil-nailed system. In the active zone, internal failure modes could be: • failure of the ground mass, i.e., the ground disintegrates and ‘flows’ around the soil nails and soil-nail heads • bearing failure underneath soil-nail heads • structural failure of the soil nail under combined actions of tension, shear and bending • structural failure of the soil-nail head or facing, i.e., bending or punching shear failure, or failure at headreinforcement or facing-reinforcement connection • surface failure between soil-nail heads, i.e., washout, erosion, or local sliding failure In the passive zone, the failure mode is mainly pullout failure of soil nail along soil-grout interface or reinforcement-grout interface. The various internal failure modes are illustrated in Figure 23b. The approach of limit state design incorporating partial factors is adopted in many soil nail design

97

Figure 23.

Principal modes of failure of soil-nailed system.

reviews, laboratory investigations (e.g. Law et al 1998, Ng & Chiu 2003) and numerical modelling (e.g. Cheuk 2001), specific guidelines on the use of soil nails in fill were developed by HKIE (2003). The following are salient points of the design recommendations:

approach was applied safely within the bound of past experience. The guidelines were developed based on the findings of review of several hundreds of soil nail designs (Pang & Wong 1997, Pun et al 2000, Lui & Shiu 2004). Soil nail designs using the prescriptive approach have been successfully applied to many soil cut slopes and retaining walls since the promulgation of the design guidelines in Wong et al (1999) and Lui & Shiu (2005). Soil nailing is also a feasible option for the stabilization of loose fill slopes. However, as loose granular fill material exhibits contractive behaviour upon shearing, there is concern that the loose fill may lose strength at such a rate that the forces mobilised in the soil nails will not be able to compensate for the loss of shear strength in the fill. Following comprehensive literature

• Large-strain steady-state undrained shear strength should be assumed for loose fill in the design. • Global stability should be provided for by bonding soil nails into a competent stratum. • Local surface stability should be enhanced by the provision of a concrete grid structure covering not less than 50% of the slope surface and connecting soil nail heads. • Soil nails should be closely spaced horizontally and vertically.

98

Table 1. Summary of approach and method of soil-nailed slopes design. Place

Document

Design approach

Common method of analysis

China

Geotechnical Engineering Handbook 1994 Eurocode 7 : Geotechnical Design, 1995 Soil Nailing Recommendations Clouterre 1991

Partial factors approach

Limit equilibrium method using single wedge mechanism No specific method defined for soil nail design Limit equilibrium method based on Bishop and two-part wedge mechanism Limit equilibrium method of slices on non-circular slip

Europe France Hong Kong

Guide to Soil Nail Design and Construction (Geoguide 7)

Japan

Japanese Design Guide: Design and Construction Guidelines for Reinforced Cut Slopes

South Africa

Lateral Support in Surface Excavations: Code of Practice 1989 Design Methods for the Reinforcement of Highway Slopes by Reinforced Soil and Soil (HA 68/94) BS 8006: 1995 Code of Practice for Strengthened/Reinforced Soils and Other Fills

UK

UK

USA

Manual for Design & Construction Monitoring of Soil Nail Walls

Limit state approach incorporating partial safety factors Limit state approach incorporating partial safety factors Global safety factor approach combined with separate factors of safety on tensile strength of steel and pull-out resistance Global safety factor approach combined with safety factors for tensile strength of steel and pull-out resistance Global factor of safety combined with separate factors of safety on strength of steel and pull-out resistance Limit state approach incorporating partial safety factors Limit state approach incorporating partial safety factors Limit state approach incorporating partial safety factors or Service Load Design

Calculation models based on conventional circular or linear slip surface analyses Limit equilibrium method using trial sliding wedges Limit equilibrium method using two-part wedge mechanism Limit equilibrium method with reference two-part wedge, circular slip and log-spiral methods Design models based on Load and Resistance Factor Design and Service Load Design Limit equilibrium based on two-part wedge and slip circle method is adopted

zone is to be checked. The UK practice also requires the checking of the bearing capacity failure in soil. The French, Japanese and German methods use empirical earth pressures which are related either to the maximum tension developed in the soil nail (Tmax ) or Coulomb earth pressure. The U.S. and Japanese practice consider directly the strength of soil-nail head when determining the magnitude and distribution of nail forces along the length of the soil nails. If the beneficial effect of the soil-nail head is not considered, the pull-out resistance of the soil nail at the active zone would be significantly reduced. This may lead to more number of soil nails being required. In all the design methods, the size, thickness and reinforcement details of soil-nail heads are determined on the basis of the earth pressure acting on the soil-nail heads. Two main design aspects are considered: the bearing capacity of the soil beneath the soil-nail head and the structural strength of the soil-nail head itself. Many of the design methods (such as those used in France and U.S.) were developed mainly for soil-nailed

• The grid structure should be designed to withstand bending moments and shear forces generated by the loose fill it is retaining; it should be adequately founded on a competent stratum. • The potential of leakage from water-carrying services should be duly considered. 2.3.2 Soil-nail head design Guidelines on the design of soil-nail head are available in the design codes of the UK (Department of Transport 1994), France (Clouterre 1991), USA (FHWA 1998), Japan (JHPC 1998) and Germany (Stocker & Riedinger 1990). All these documents recognize the soil-nail head or facing as a significant component of the overall soil nail system, and they provide specific recommendations for design pressures. They also recognize that the magnitudes of pressures induced in the soil-nail heads are controlled by many factors such as the density and length of the nails and the stiffness of the soil-nail head. Both the UK and the Japanese practice require that the pull-out failure at the active

99

walls, where the soil-nail heads form part of the concrete facing. In these cases, bearing failure of the soil beneath soil-nail heads or facing is unlikely to occur and as such little guidance has been provided in respect of soil bearing failure. A method on the design against bearing failure of the soil behind isolated nail heads is given in the UK guidance document HA68/94. For the development of soil-nail head design guidelines in Hong Kong, a series of numerical analysis was carried out. In the study, FLAC analysis was performed to examine the bearing failure beneath square soil-nail heads. A small slope model of 5 m in height was used and various slope angles were considered (Fig. 24). In the analysis, the soil-nail head was pushed into the ground by a nail force to simulate the situation of soil moving out from a slope and pressing against the soil-nail head. The nail forces used are determined from the allowable tensile strength of steel bars. Figures 25 and 26 show the shear strains and the displacement vectors respectively at the point of bearing failure for a 600 mm x 600 mm soil-nail head on a 45◦ slope. Typical results of the analyses in terms of c’-φ’ envelope for limit equilibrium (i.e. when bearing failure occurs) are plotted in Figure 27. In this plot, the soil-nail head loads are expressed as diameters of steel bars. A number of the plots have been developed for different slope angles and soil-nail head sizes.

Figure 24.

Knowing the shear strength parameters of the soil, the steel bar diameter and the slope angle, a designer can determine the size of soil-nail head from these plots. A design table has been derived from the plots for different combinations of slope angle and angle of shearing resistance of soil (Table 2). 2.3.3 Pullout resistance Pull-out capacity is a key parameter for the design of soil nails. At present, methods for estimating pullout capacity are not unified as reflected by the many approaches used in different technical standards and codes of practice, such as effective stress method (GEO 2008), empirical correlation with SPT N values (JHPC 1998), correlation with pressuremeter tests (Clouterre 1991), and correlation with soil types (FHWA 2003). The merits and limitations of the various methods are summarized in Table 3. The effective stress method is adopted in Hong Kong. The allowable pullout resistance provided by the soil-grout bond strength in the passive zone, TP , is given by (Schlosser & Guilloux 1981): TP =

c′ Pc Lc + 2Dσv′ μ∗ Lc FP

(6)

Slope model for bearing capacity analysis. Figure 26.

Typical displacement vector plot.

14 40 mm diameter bar

12 10 32 mm diameter bar 8 6 25 mm diameter bar

4 2 0 31

Figure 25.

Typical shear strain plot.

32

33

34

35

36

37

38

39

40

41

42

43

44

Figure 27. Shear strength required for 600 mm × 600 mm nail heads on a 45◦ slope to mobilise allowable tensile strength of nails of specified diameters.

100

Table 2. Recommended sizes of isolated soil-nail heads in Geoguide 7 (GEO, 2008).

Soil shear strength parameter near the slope surface ϕ’ 34◦

34◦

38◦

40◦

c’ (kPa) 2 4 6 8 10 2 4 6 8 10 2 4 6 8 10 2 4 6 8 10

45◦ ≤ Slope angle < 55◦

55◦ ≤ Slope angle < 65◦

Slope angle ≥ 65◦

Diameter of soil-nail reinforcement (mm)

Diameter of soil-nail reinforcement (mm)

Diameter of soil-nail reinforcement (mm)

25 800 600 600 600 400 600 600 600 400 400 600 600 400 400 400 600 400 400 400 400

32 800 800 800 600 600 800 800 600 600 600 800 600 600 600 600 600 600 600 600 600

40 800 800 800 800 800 800 800 800 800 800 800 800 800 800 800 800 800 800 600 600

25 600 600 400 400 400 600 400 400 400 400 400 400 400 400 400 400 400 400 400 400

32 600 600 600 600 600 600 600 600 600 600 600 600 600 600 400 600 600 600 400 400

40 800 800 800 800 600 800 800 800 600 600 800 800 600 600 600 800 600 600 600 600

25 600 600 400 400 400 600 400 400 400 400 600 400 400 400 400 600 400 400 400 400

32 600 600 600 600 600 600 600 600 600 400 600 600 600 400 400 600 400 400 400 400

40 800 800 600 600 600 800 800 600 600 600 600 600 600 600 600 600 600 600 600 600

Notes: (1) Dimensions are in millimetres unless stated otherwise. (2) Only the width of the square soil-nail head is shown in the Table.

Table 3. Merits and limitations of the methods for determining ultimate pull-out resistance. Method

Merits

Limitations

Empirical correlation

Related to field performance data; can better account for influencing factors.

Need a large number of field data and take a long time to establish a reasonable correlation; a general correlation may not be applicable to all sites.

Pull-out test

Related to site-specific performance data.

Need to carry out a considerable number of field pull-out tests during the design stage; not feasible for small-scale project; time consuming.

Undrained shear strength

Based on soil mechanics principles; easy to apply.

Generally not suitable for Hong Kong; many factors that affect the pull-out resistance are not accounted for.

Effective stress

Based on soil mechanics principles; easy to apply.

Many factors that affect the pull-out resistance are not accounted for.

Pressuremeter

Related to field performance data; can better account for influencing factors.

Need a large number of field data to establish a reasonable correlation; a general correlation may not be applicable to all sites; pressuremeter test is not common in Hong Kong.

101

where c′ is effective cohesion of the soil, Pc is outer perimeter of the cement grout sleeve, Lc is bond length of the cement grout sleeve in the passive zone, D is outer diameter of the cement grout sleeve, σν′ is vertical effective stress in the soil, μ∗ is coefficient of apparent friction of soil (μ∗ may be taken to be equal to tan φ ′ , where φ ′ is the effective angle of shearing resistance of the soil), FP′ is factor of safety against pullout failure at soil-grout interface. For design purpose, the vertical effective stress in the soil is calculated from the overburden pressure, which implies that the contact pressure at the soil-grout interface is governed by the overburden pressure. This assumption is not necessarily true because the normal stress at the face of the drillhole is reduced to zero after drilling due to arching effect and the grouting pressure is generally so low that only a small contact pressure can be restored. The contact pressure is likely much less than the overburden pressure. The effects of hole drilling process, overburden pressure and grouting pressure on pullout resistance was investigated by means of laboratory pullout tests by Yin & Su (2006). The test set up is shown in Figure 28. Compacted fill of completely decomposed granite was used in the tests. The study showed that (a) the drilling process during soil nail installation led to stress reduction in the soil around the drillhole and the pullout resistances of the nails were not dependent on the amount of vertical surcharge applied if gravity grouting was adopted (Fig. 29); and (b) pullout resistances of the soil nails increased with an increase of grouting pressure (Fig. 30). Pullout tests are routinely carried out in sacrificial test nails in Hong Kong for the verification of design assumptions. In order to examine the significance of the potential stress reduction due to the arching effect, the results of about 900 pullout tests were reviewed. The pullout resistance measured in the field was compared with the theoretical values estimated by the effective stress method. About 84% of the tests were conducted in granite or volcanic saprolite. The rest were conducted in other types of material such as fill, colluvium and moderately decomposed rock. Many of the test nails were not loaded to bond failure because the ultimate pullout resistance (Tult ) of the bonded section was higher than the yield strength of steel. The pull-out tests were stopped when the test load reached 90% of the yield strength of steel to avoid tensile failure of the steel reinforcement. Figure 31 shows the plot of the ratio of the field pullout resistance to that estimated using the effective stress method against the overburden pressure. The field pull-out resistances were generally several times higher than those estimated, but the safety margin (i.e. Tult(field) /Tult(estimate) ) gradually decreases when overburden pressure increases.

Elevation

End View Figure 28.

Schematic set-up of the laboratory pullout test.

Figure 29. pressure.

Plot of pullout resistance against overburden

The difference between the measured and the estimated pullout resistance is due to many factors including soil arching, restrained soil dilatancy, soil suction, roughness of drillhole surface, and over-break, which are hard to quantify in design. All these factors except soil arching tend to result in higher pullout resistance than the design value. The finding of the review gives assurance on the adequacy of the effective stress method.

102

Figure 30. Plot of pullout resistance against grouting pressure.

20

Tp (Fill) Dry Tult (Colluvium) Dry Tp (Colluvium) Dry Tult (C/HDG) Dry Tult (C/HDG) Wet Tp (C/HDG) Dry Tp (C/HDG) Wet Tult (C/HDV) Dry Tult (C/HDV) Wet Tp (C/HDV) Dry Tp (C/HDV) Wet

estimate

18 16 14 12

resistance

10 8 6 4

Mean field/ estimate =

2

1

0 0

100

200

300

400

500

600

Figure 31. Plot of field (Tp + Tult ) to estimated pull-out resistance against overburden pressure.

2.3.4 Potential effect of blockage of subsurface drainage by soil nailing works Soil nails installed in the ground may impede groundwater flow and as a result dam up the water level. To study the significance of this effect, a number of numerical models were set up in both 2-D and 3-D for various geological settings, subjected to infiltration (Halcrow China Limited 2007). Typical nail spacings of 1 m to 2 m were adopted in the models. Figure 32 illustrates an example of computed flow nets and water table distributions for a slope under three conditions: (a) without soil nails; (b) soil nails with excessive grout loss, and (c) soil nails with no grout loss. Results of the numerical modeling show that under typical conditions where there is little grout loss during the grouting operation, there should be no significant blockage of the drainage paths. It is also found that the influence of soil nails on groundwater flow can be significant if excessive grout escapes laterally to affect large volumes of the country rock. Therefore, measures should be taken to avoid excessive grout loss. Where excessive grout loss occurs during installation of soil nails, the cause should be investigated and, if necessary, measures taken to monitor rises in hydraulic

Figure 32. Flow Patterns in a slope (a) without soil nails, (b) soil nails with excessive grout loss, (c) soil nails with no grout loss (after Halcrow, 2005).

head and to take action to drain the ground upstream of the nails. 2.3.5 Long-term durability of soil nails Durability is an important aspect of soil nailing system. The long-term performance of soil nails depends on their ability to withstand corrosion attack from the surrounding ground. Soil nails of different ages exhumed from the ground in Hong Kong revealed that localized corrosion could occur even if hot dip galvanization was provided, particularly in areas where voids existed

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Figure 33.

Localized corrosion in exhumed soil nail.

in the cement grout (Shiu & Chang 2003). Figure 33 shows the corrosion on an exhumed soil nail. The design for durability of a steel soil nail entails the assessment of the corrosivity of soil at the site and the provision of corrosion protection measures. Different soil corrosivity assessment schemes are available in the world, and different countries may adopt different corrosion protection measures. Eyre & Lewis (1987) developed two soil corrosivity assessment schemes, one for general assessment purposes and another for reinforced earth and culvert application. Two types of assessment scheme were subsequently developed by others in the UK. The first was that of Murray (1993), who incorporated the scheme for general assessment purposes into the specifications for soil corrosivity assessment in respect of soil nailing in the UK. Tests are conducted on soil samples only. The soil aggresivity classification scheme in Hong Kong (GEO 2008) was developed based on that of Murray (1993). The UK Department of Transport made reference to the scheme for reinforced earth and culvert application and the recommendations of Brady & McMahon (1993) and developed a corrosivity assessment scheme for corrugated steel buried structures under roadways (Department of Transport 2001). In this scheme, both soil and water samples are collected and tested. CIRIA (2005) has recommended a corrosivity assessment scheme based on that developed by the Department of Transport (2001). Depending on the soil aggresivity of a site, the required design life and the intended degree of protection, different measures may be adopted to protect steel bars against corrosion. The common corrosion protection measures are cement grout, sacrificial thickness to the steel, sacrificial metallic coating to the steel (e.g. hot-dip galvanizing with zinc coating), sacrificial nonmetallic coating to the steel (e.g. epoxy coating), and corrugated plastic sheathing. Cement grout can prevent corrosion by forming a physical and a chemical barrier. The cement grout physically separates the steel from the surrounding

soil. Due to the alkalinity of cement grout, a tight oxide film is also formed on the surface of the steel bar. This further protects the steel from corrosion. However, micro-cracks will occur in the cement grout when the soil nail is subject to tensile stress. Shrinkage cracks may also be formed during the setting of the cement grout. Once cracked, the function of the cement grout in corrosion protection is not reliable. The provision of sacrificial steel thickness is a simple and widely used method. It allows for corrosion of the steel by over-sizing the cross-section of the steel bar. Products of corrosion also form a protective coating between the steel and its surrounding. Whilst this coating offers no physical protection to the steel, it may slow down the rate of corrosion by changing the kinetics of the chemical reactions. Zinc is the most common type of metal used to provide corrosion protection to steel bars. The galvanizing zinc coating is strongly resistant to most corrosive environments. It provides a barrier protection and a cathodic protection function to the steel. Non-metallic coatings in the form of fusion-bonded epoxy have been used in the USA to protect steel bars from corrosion. The epoxy coatings do not conduct electricity and they isolate the steel bars from the surrounding environment. To be effective, the coating has to be impermeable to gases and moisture and free from cracks. The interface between the steel and the coating has to be tight. When a high level of corrosion protection is needed, corrugated plastic sheaths can be used in conjunction with cement grout. The sheath prevents ingress of water and corrosive substances even if the cement grout is cracked. To overcome the problem of corrosion of metallic reinforcement, non-metallic soil nails may be used. An alternative to steel reinforcement is composite material made of fibres embedded in a polymeric resin. It is generally known as fibre-reinforced polymers (FRP). FRP is highly corrosion resistant. The common types of fibre used in composites for civil engineering works are carbon FRP (CFRP), glass FRP (GFRP) and aramid FRP (AFRP). Shanmuganathan (2003) gave a state-of-the-art review of the development and application of FRP composites in civil and building structures. There are reported cases of using CFRP reinforcements to slope stabilization works in the USA, the UK, Spain, Greece, Japan and Korea (e.g. Unwin 2001, Ground Engineering 2004). Carbon fibres are the primary load-carrying component in CFRP reinforcement, which are characterized by low weight, high strength and high stiffness. The primary function of the resin is to provide a continuous protection medium to the fibres and to transfer stresses among fibres. CFRP reinforcement is anisotropic in nature and is characterized by high tensile strength in the direction of the fibres. It is

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non-corrosive and has a much better strength-toweight ratio than steel reinforcement. However, CFRP reinforcement does not exhibit yield behaviour. The lack of ductility necessitates special consideration in its application as a soil nail. Figure 34 illustrates schematically the stress-strain behaviour of CFRP reinforcement in comparison to steel. According to ACI (2001), the tensile strength of CFRP reinforcement ranges from 600 MPa to 3,690 MPa (c.f. 460 MPa for high yield steel bar). Specific test results on tensile testing of CFRP reinforcement strips (Fig. 35) which were used in a field trial by the GEO in Hong Kong are summarized in Table 4. The tests gave tensile strength ranging from 1,990 MPa to 2,550 MPa, with an average value of

2,280 MPa. This average tensile strength is about five times that of high yield steel. The shear strength of CFRP reinforcement is generally much lower than its tensile strength. Benmokrane et al (1997) reported that the shear strength of some CFRP reinforcements is only about 11% of its tensile strength. While there are a number of national design and construction guides on the use of CFRP reinforcement in concrete structures (e.g. Japanese Ministry of Construction 1997, JSCE 1997, IStructE 1999, ACI 2001), international standard on the use of CFRP reinforcement as soil nails is lacking. An interim design and construction guideline for CFRP soil nails has been developed by the GEO for use in its slope upgrading programme. The following are salient points of the GEO guideline: • A suitably conservative estimate of the design tensile strength is made using a partial material safety factor of 3.3 to cater for the uncertainty in material properties and to compensate for the lack of ductility of CFRP. • A partial safety factor of 1.4 on bond strength between CFRP reinforcement and grout is adopted following the recommendation of IStructE (1999). • The inclination of the CFRP soil nails to limited to within 15◦ from the horizontal so as to optimize the reinforcing efficiency of the soil nails and to limit the slope movement for the mobilization o the tensile force in the soil nails. • Only CFRP reinforcements with a shear strength equal to or greater than that of steel are used. • CFRP reinforcements with polyester resins should not be used because of their relatively ease of degradation in highly alkaline environment.

Figure 34. Stress/strain curves of typical high yield steel bar and CFRP bar.

Loading Figure 35.

Failure

Tensile testing of CFRP strips.

2.3.6 Aesthetic considerations A good soil nail design should give due attention to the aesthetic aspects in addition to safety and serviceability considerations. GEO Publication No. 1/2000 (GEO 2000) provides comprehensive guidance on the landscape treatment and bio-engineering for slopes and retaining walls. The publication contains general principles and good practice for enhancing the appearance of engineered slopes and illustrates these by a large number of case examples. While the principles of aesthetics and landscape treatment are given for unreinforced slopes, they are generally applicable to soil-nailed systems. The following design principles are worth considering for enhancing the appearance of soil-nailed slopes and retaining walls: • Make the appearance of soil-nailed systems compatible with and minimize visual impact to the existing environment. • Identify and preserve, wherever practical, mature trees on slopes and near their crests and toes.

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Table 4. Results of tensile test on CFRP reinforcement bars. Specimen no.

Width (mm)

Thickness (mm)

Rupture load (kN)

Tensile stress at rupture (MPa)

Modulus of elasticity (GPa)

Rupture strain (%)

1 2 3 4 5

29.0 29.0 28.9 29.0 28.7

4.3 4.5 4.5 4.4 4.6

318 330 262 253 314

2,548 2,522 1,996 1,985 2,363

120.2 119.4 111.6 114.9 103.5

2.12 2.11 1.79 1.73 2.28

Average

2,283

113.9

2.01

granitic or volcanic saprolite of a gradient up to 60◦ have been successfully vegetated using this technique. Where the provision of vegetated surface cover on a slope is practically not feasible, hard landscape treatment can be provided to improve its appearance. Possible methods are masonry block facing, ribbed or other patterned concrete finishes, toe planters, colouring and planter holes. More fancy techniques such as decorative artwork and artificial rock may also be used. 2.4 Construction technology

Figure 36. Fixing details of steel wire mesh and erosion control mat on slope face with soil nail heads.

• Locate soil nails and other engineering features away from tree trunks and roots. • Pay attention to the design and location of manmade features such as surface drainage channels, stairways and catchpits in order to minimise their visual impact, e.g. concrete aprons on either side of drainage channels can be designed using geotextiles or other bioengineering techniques. • Route maintenance stairways with care to minimise visual impact and paint railings in sympathetic unobtrusive colours. • Place soil nails in a regular, rather than a random pattern. • Recess isolated soil-nail heads and treated them with a matt paint of a suitable colour to give a less intrusive visual appearance. A large variety of greening techniques are available in the market. The Hong Kong experience shows that a simple method which involves the use of an erosion control mat in conjunction with a steel wire mesh allows the provision of vegetation covers to steep slopes (Fig. 36). Many steep soil-nailed cut slopes in

2.4.1 Construction method The choice of installation method depends on a number of factors such as cost, site access, working space, and ground and groundwater conditions. A brief description of the commonly available soil-nail installation methods is summarised below. a. Drill-and-Grout. This is the most common installation method. In this method, a soil-nail reinforcement is inserted into a pre-drilled hole, which is then cement-grouted under gravity or low pressure. Various drilling techniques, e.g., rotary, rotary percussive and down-the-hole hammer, are available to suit different ground conditions. The advantage of this method is that it can overcome underground obstructions, e.g., corestones, and the drilling spoil can provide information about the ground. The size and alignment of the drillholes can be checked before the insertion of reinforcement, if needed. Potential construction difficulties are hole collapse and excess grout loss. The drilling and grouting process may also cause disturbance and settlement to the adjacent ground. b. Self-Drilling. This is a relatively new method when compared with the drill-and-grout method. The soil-nail reinforcement is directly drilled into the ground using a sacrificial drill bit. The reinforcement, which is hollow, serves as both the drill rod and grout pipe. The installation process is rapid as the drilling and grouting are carried out simultaneously. Instead of using air or water, cement

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grout is used as the flushing medium, which has the benefit of maintaining hole stability. No centralisers nor grout pipes are needed and casing is usually not required. However, self-drilling soil nails may not be suitable for the ground containing corestones as they cannot penetrate through rock efficiently. It may be hard to ensure the alignment of long soil nails due to the flexibility of reinforcement. Durability may also be a concern if it relies on the provision of grout cover and corrosion protective coatings to steel reinforcement as corrosion protection measures. c. Driven. In this method, soil-nail reinforcement is directly driven into the ground by the ballistic method using a compressed air launcher, by the percussive method using a hammering equipment, or by the vibratory method using a vibrator. During the driving process, the ground around the reinforcement will be displaced and compressed. The installation process is rapid and it causes minimal ground disruption. However, due to the limited power of the equipment, this method can only be used to install soil nails of relatively short length. Moreover, the soil-nail reinforcement may be damaged by the excessive buckling stress induced during the installation process, and hence it is not suitable for sites that contain stiff soil or corestones. Since the soil-nail reinforcement is in direct contact with the ground, it is susceptible to corrosion unless non-corrodible reinforcement is used. The experience in Hong Kong shows that soil nails can generally be constructed by means of the drill-and-grout method without many difficulties. However, under some unfavourable ground conditions, construction problems may be encountered. The following geological conditions are susceptible to excessive grout leak during soil nail installation: f • fill, containing a significant proportion of coarse materials, i.e., boulders, cobbles, gravel, and sand; • colluvium and fluvial deposits with a high proportion of coarser material; • erosion pipes which may be partly infilled by porous and permeable material; • material boundaries within colluvium, and between colluvium and in-situ material, and within corestone-bearing saprolite, especially at the margins of corestones, open joints, faults and shear zones, and other discontinuities (e.g., zones of hydrothermal alternation, etc.) that are weathered and eroded, and so are open; • landslide scars, tension cracks, and other features related to slope deformation, as these may include voids within transported and in-situ materials; and

• drainage lines intersecting slopes, within which colluvium may be present, erosion pipes may be developed, and preferred groundwater throughflow indicated by seepage locations/horizons may also occur. 2.4.2 Use of non-destructive testing method for quality control Like other buried works, it is difficult to verify the quality of an installed soil nail. In the context of this paper, the quality of an installed soil nail refers to the as-built length and the integrity of cement grout. In order to enhance the quality control of soil nailing works, non-destructive testing (NDT) methods could be carried out on installed soil nails. With the help of NDT, the overall picture of the quality of installed soil nails can be built up, which facilitates the identification of the areas for follow-up actions. A number of NDT methods including sonic echo method, Mise-a-la-Masse method, magnetometry, electromagnetic induction method and time domain reflectometry (TDR) have been examined in Hong Kong. Amongst these, the TDR method was found to be reliable, simple and not expensive (Cheung 2003, Lee & OAP 2007). The principle of TDR technique was derived in 1950s from that of radar. Instead of transmitting a 3-D wave front in radar, the electromagnetic wave in the TDR technique is confined in a waveguide (O’Connor & Dowding 1999). TDR is commonly used in the telecommunications industry for identification of discontinuities in transmission lines. In the 1980s, the application of the technique was extended to many other areas such as geotechnology, hydrology, material testing, etc (Dowding & Huang 1994, Siddiqui et al 2000, Liu et al 2002, Lin & Tang 2005). TDR is based on transmitting electromagnetic pulses through a transmission line, which is in the form of coaxial or twin-conductor configuration, and receiving reflections at the locations of discontinuities. By measuring the time for the pulses to travel from the pulse generator to the point of discontinuity, one can determine its location using Equation (7). L = vp t

(7)

where L is the distance between the pulse generator and the point of discontinuity, and t is the respective pulse travel time. The pulse propagation velocity, vp , is related to the electrical properties of the material in the close proximity to the pair of conductors by the following expression (Topp et al 1980): vc vp = √ ε

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(8)

Figure 37. Analogy of a soil nail with pre-installed wire as a twin-conductor transmission line.

where vc is the speed of light in vacuum (3×108 m/s) and ε is the dielectric constant which measures how a material reacts under a steady-state electric field (for air ε ≈ 1, for cement grout ε ≈ 10, for water ε ≈ 80). If a wire is pre-installed alongside a soil-nail reinforcement, which is generally a steel bar, as shown in Figure 37, the configuration becomes analogous to a twin-conductor transmission line and the end of the reinforcement-wire pair becomes a discontinuity. This suggests that TDR can be used to determine the length of installed steel soil nails. As indicated in Equation (1), the two key parameters that have to be known for the estimation of soil-nail length are (i) the time for a pulse to travel from the reinforcement head to its end, t, and (ii) the pulse propagation velocity, vp . Equation (8) further suggests that the pulse propagation velocity, vp , along a reinforcement-wire pair in air will be much greater (2 to 3 times) than that in cement grout. Hence, the pulse travel time along a soil nail with voids in grout will be less than that in a fully grouted soil nail of the same length. Apart from the effect on pulse propagation velocities, a reflection will be induced whenever an electrical pulse reaches the location of discontinuity in the grouted reinforcement-wire pair (e.g. the end of a soil nail or a void). The magnitude and polarity of the reflection depend on the amount of changes in electrical impedance at the location of discontinuity, which can be expressed in terms of the reflection coefficient, Ŵ(Hewlett Packard 1998): Ŵ=

Vr Z − Zo = Vi Z + Zo

(9)

where Vr is the peak voltage of the reflected pulse, Vi is the peak voltage of the incident pulse, Z is the electrical impedance at the point of reflection and Zo is the characteristic electrical impedance of the grouted reinforcement-wire pair. Figure 38 shows a theoretical TDR waveform of a cement grouted reinforcement-wire pair with void section in the middle. There will be reflections at the location of the void as well as the end of the pair.

According to Equation (9), a positive reflection will be returned at the discontinuity when there is an increase in electrical impedance (e.g. reflection 1 at the interface of grout/void and reflection 3 at the end of the pair), whereas a negative reflection will be returned otherwise (e.g. reflection 2 at the void/grout interface). Moreover, the pulse travel time is less than that in the fully grouted pair. In other words, one can in-principle determine the quality of an installed soil based on a TDR waveform. In order to investigate the feasibility of applying TDR technique in the estimation of soil-nail length, TDR tests were conducted on prefabricated soil nails of various known lengths. Figure 39 shows the TDR test results where reflections are returned from the respective soil-nail ends and the time of pulse propagation is found to be proportional to the length of the soil nail. Based on the contrast in pulse propagation velocity in air and grout and the occurrence of reflections where these is a change in impedance along the reinforcement-wire pair, TDR results in-principle can be interpreted to infer the grout integrity of a soil nail. To examine this, TDR tests were conducted on prefabricated soil nails with built-in grout defects of varying void sizes at different locations along the soil nails as shown in Figures 40 and 41. These model test results indicate that soil nails with significant grout defects will result in shorter TDR-deduced length with some characteristic patterns in the TDR waveform. These patterns depend on the location as well as size of the defects. Up to the end of 2007, over 10,000 soil nails at about 850 sites have been tested using TDR in Hong Kong (Fig. 42). In general, the percentage difference between the TDR-deduced length and design length of the soil nails is small and lies within the uncertainty limit of the test. There were a small number of soil nails (less than 1%) with such difference exceeding an alert limit and displaying anomalous TDR wave forms. Further investigation was conducted where needed. The anomalies encountered so far were found to be due to grout defect in the soil nails. The experience shows that the TDR technique can be an effective tool to supplement site supervision in the quality control of soil nailing works, which cannot be checked easily after construction. While TDR, like any other NDTs, does not give definitive answer to the cause of anomalous test results, it flags up soil nails that warrant further examination and coupled with appropriate NDTs, the conditions of the soil nails can be ascertained (Pun et al 2007). 2.5

Performance of soil-nailed system

Since the introduction of the soil nailing technology to Hong Kong in late-1980’s (Watkins & Powells 1992),

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Figure 38.

Theoretical TDR waveform of a soil nail with defect in grout sleeve.

Figure 39. length.

TDR results on soil nails of various known Figure 40. TDR waveforms for soil nails with a void section between grouted sections.

more than 3,700 soil cut slopes have been upgraded by means of soil nailing in Hong Kong. The performance of these soil-nailed slopes gives an indication of the reliability of the soil nailing technique. In the period between 1993 and 2006, a total of 25 landslide incidents on permanent soil-nailed slopes were reported to the GEO. The landslides were generally of small scale, involving local shallow sliding failure or washout, with failure volumes ranging from less than 1 m3 to a maximum of about 35 m3 . The average annual failure rate of such relatively minor landslides on soil-nailed slopes between 1997 and 2006 is 0.078%. This average annual failure rate is

of a similar order of magnitude as that of engineered, unsupported soil cut slopes in Hong Kong. Amongst the 25 reported landslide incidents, 15 cases were reviewed in detail. All the 15 landslide incidents involved active zone failures. There was no report of external failure or passive zone failure. The slopes were steep and had a vegetation cover before failure, four of which had a gradient equal to or exceeding 45◦ and the other 11 exceeding 50◦ . Twelve of the 15 cases were associated with surface

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Figure 43.

Figure 41. TDR waveforms for soil nails with two void sections between grouted sections.

Figure 42.

Shallow failure at a soil-nailed slope.

Conducting TDR test at a soil nail.

erosion and detachments from the near-surface materials between the nail heads (see example in Fig. 43). In two cases, the soil-nail heads were partially exposed but the nail reinforcements and grout sleeves remained intact. In the remaining case, the soil-nail heads were undermined and the soil nail reinforcement bars were bent (see Fig. 44). The common factors contributing to the landslides were inadequate slope protection, inadequate drainage provisions or presence of adverse geological or hydrogeological conditions. The review shows that steep soil-nailed slopes with vegetated covers are fairly vulnerable to minor failures, as the potential for shallow small detachments between soil-nail heads within the active zone of the soil nail system cannot be prevented effectively by means of the soil nails. The efficiency of the soil nails, which are typically shallowly inclined for the upgrading of substandard soil cuts in Hong Kong, is not high in so far as prevention of shallow detachment is concerned. This is because the nails are not orientated at an optimal

Figure 44. Failure of a soil-nailed slope undermining soilnail heads.

inclination in relation to the steep slip surfaces with regard to the mobilization of tension forces in the soil nails. In case of shallow failures on a steep vegetated soil-nailed slope, there is little horizontal displacement to mobilize the tensile forces in the nail bars and hence their effectiveness in stabilizing potential vertical/subvertical failure surfaces may be limited. No failure with a volume larger than 50 m3 on soilnailed slopes has been reported in Hong Kong so far.

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As a comparison, the average failure rate of landslides with a volume larger than 50 m3 for engineered, unsupported soil cuts in Hong Kong is 0.018%. Soil nails appear to be effective in preventing large-scale failures. It is also worth noting that no landslides have been reported on soil-nailed slopes with a hard surface cover in Hong Kong. A hard surface cover is effective in minimizing surface infiltration and provides a better protection against surface erosion than a vegetated cover. However, a hard surface cover may not be acceptable from the environmental and aesthetic points of view. A variety of flexible structural facing such as tensioned wire mesh may be used to enhance the stability of vegetated slopes. This would however increase the construction cost. A proper design has to balance between the risk of possible minor failure, cost and environmental considerations. In this regard, risk mitigation measures, such as debris traps, toe barriers or buffer zones, may be considered as an integral part of the slope design to cater for possible minor detachments from vegetated slopes.

3

SUBSURFACE DRAINAGE

3.1 General aspects In saturated soils, subsurface drains are widely used as control measures against slope instability, as they are less costly than other types of stabilization works and suitable for a large number of cases, even when the landslide is very deep and structural measures are not effective. The mechanism of drains inside slopes involves a decrease in pore pressures in the subsoil and consequently an increase in effective stresses and soil shear strength in the whole drained domain. In particular, the increase in soil shear strength along the potential sliding surface of the landslide body, due to the function of drains, is responsible for the slope stability improvement. Therefore the first step in the design of a drainage system is the determination of the pore pressure change that is required to increase the factor of safety of the slope to the design value (Fig. 45). The next step is to design the geometric configuration of drains that will result in the required pore pressure change. The effect of the drainage system is usually analyzed for the steady-state condition, which is attained some time after drainage construction (i.e. in the long term). After drain installation, a transient phenomenon of equalization of pore pressures occurs, provoking subsidence of the ground surface. The magnitude of subsidence depends on (i) the compressibility of the soils concerned, (ii) the thickness of the drained

Figure 45.

Approach to design drains to stabilize a slope.

domain, and (iii) the amount of lowering of the water table. Problems related to excessive ground settlements are expected when the drained soil is very thick, as in the case of deep drains. As regards the transient phase, two aspects have to be evaluated in the design: a. whether the delay until the drains are completely effective is affordable, b. whether settlements associated with de-watering will damage buildings and infrastructures at the ground surface. The steady-state condition is usually analyzed by assuming continuous infiltration of water at the ground surface to recharge the water table. In the literature, results of steady-state analysis are often presented in non-dimensional design charts, that practitioners generally use to design drainage systems. The water flow captured and discharged by drains depends largely on the permeability of the drained soils. In steady-state condition the permeability of the ground does not affect the amount of lowering of the pore pressures in the subsoil, which depends on the hydraulic conditions at the boundaries of the examined domain and the geometry of the drainage system. Thus the quantity of discharge is not an indicator of the performance of drains, which has to be investigated by means of piezometers to measure the change in the level of water table by drains. Indeed, pore pressure changes are the most direct and useful indicators of drains being in good working condition. Measurements of surface and deep displacements are good indicators of overall slope stability. These measurements complete the instrumentation framework (Fig. 46). 3.2 Drain types Among the measures for slope stabilization, drains are probably the most commonly used. Because of their widespread application in very different ground conditions and geomorphology, the technology in this field

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Figure 46. Measurements to evaluate that drains are in good working condition.

Figure 48. Superficial and deep trenches, with main and secondary branches: a) map, b) longitudinal section.

Figure 47. Superficial and deep trenches, with only main branches: a) map, b) cross section.

Figure 49. Horizontal drains inserted from the ground surface: a) map, b) longitudinal section, c) cross section.

is advancing continuously. Nowadays the most common types of drains used in geotechnical engineering applications are as follows: • Superficial and deep trenches, with main (Fig. 47) and possibly secondary (Fig. 48) branches, • Horizontal drains (Fig. 49), executed from the ground surface, • Wells, with or without horizontal drains (Fig. 50), • Tunnels, with or without horizontal drains. Drain trenches should be excavated deep enough to intercept the regions of positive pore pressures. Superficial trenches can be excavated by means of an excavator up to a depth of approximately 5 m from the ground surface. The width of the trench is dependent on the type of excavator being used and may vary from 0.5 to 1.0 m. In open areas, trenches can have sloping sides, the gradient of which is based on stability consideration. Where there is not enough space, trench sides have to be formed to vertical and should be properly supported. Guidelines on the design of lateral support to excavation are given in many publications, e.g. BS 6031:1981 (BSI 1981). Problems of trench instability can be reduced by opening up trenches in short

Figure 50. Drain wells with and without horizontal drains: a) map, b) longitudinal section.

112

Figure 51.

Scheme of a superficial trench.

Figure 52.

lengths and backfilling the trench within a short time after excavation. Deep trenches can be excavated by means of grab shells. The sides of the trenches, being vertical, should be supported by slurry, e.g. polymeric mud. Trenches need to have a high discharge capacity to avoid the saturation of the backfilling material or of the lower portion of it. This can be achieved providing a drainage layer of gravel materials or installing at the bottom of the trench a perforated pipe (with slots on the upper part). The perforated pipe should be wrapped with a geotextile to prevent the clogging of the slots by fine soil particles (Fig. 51). A compacted clay cover should be placed on the top of the trench to prevent ingress of surface water, which should be drained by means of a system of surface drainage network. The impermeable cover should have a minimum thickness of 0.5 m and should be compacted in layers. Trenches should be constructed starting from the lowest point in the area to be drained, so that they can drain water during construction. Inspection wells that intercept the trenches should be installed to allow: • monitoring of the working condition of the drainage system, possibly by measuring the flow; • maintenance, possibly flushing of the perforated pipe. Horizontal drains involve the drilling of holes in the ground. The diameter of the hole is usually 100120 mm, and it is drilled with a tricone or drag bit. A PVC slotted pipe, protected by a geo-textile, is inserted in the hole (Fig. 52). The maximum length of horizontal pipes is around 100 m, but in some cases it has been possible to reach 300 m. Deposits of calcium salts and iron oxide can block horizontal drains; regular maintenance by flushing the pipes with a high pressure water jet, should therefore be programmed. In the absence of maintenance, drain pipes cannot remain functional for a long time.

Scheme of a horizontal drain.

To reduce precipitation of calcite it is good practice to drill the hole at an inclination slightly above horizontal, such that the pipe is not continuously submerged. Conversely, there are other chemical phenomena, favored by bacterial activity, that are due to aeration (Walker & Mohen 1987). At the portion of a horizontal drain near to the slope surface, it is recommended to use a 3–6 m long un-perforated pipe, grouted all around with cement, to prevent the penetration of tree roots into the pipe, which could block the water flow. Wells and tunnels are costly and complex to construct; for this reason they tend to be used in special circumstances, e.g. in deep landslides where other types of drain are unable to reach the sliding surface. Drain trenches and horizontal drains are the most commonly used drain systems in slope stabilization. Therefore these are treated in detail in the next part of this paper. 3.3

Influence of groundwater on landslides in saturated soils

Investigation carried out in many parts of the world (Kenney & Lau 1984, Urciuoli 1998) showed that pore pressures at shallow depths are strongly influenced by seasonal atmospheric conditions. Urciuoli (1998) showed, on the basis of piezometer measurements, that a critical line can be drawn inside clay formations, separating the zone in which groundwater regime is transient, due to variation in atmospheric conditions, from the deep zone, in which pore pressures remain essentially constant throughout the year (Fig. 53). The division of the saturated soil domain into two zones is useful to interpret the mechanism of landslides and its relation with pore pressures, as is described below. Active shallow landslides are usually characterized by the presence of a sliding surface in the zone affected

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analyzed by considering the groundwater regime as a steady-state phenomenon. 3.4

Groundwater regime

3.4.1 Equation If the subsoil is assumed to be a saturated porous medium, characterized by isotropic permeability K, the equation governing transient flow can be obtained by imposing the conservation of the liquid mass. Darcy’s Law is inserted therein. In this way the subsequent equation is obtained: Figure 53. Pore pressure in clayey formation: a) measurements at Casagrande piezometers installed at different depths (Basento Valley, Italy), b) scheme of pore pressure regime in the subsoil.

−K · (hxx + hyy + hzz ) = ε˙ v where h is the piezometric head: h=ξ+

by a transient groundwater regime. In this case, the safety factor of the landslide body varies with time along with pore pressure fluctuations. Pore pressures attain their maximum values during wet seasons: shallow landslides can reactivate, as a direct consequence of reduction in shear strength along the sliding surface. The rate of slope movement is characterized by a seasonal trend (Pellegrino et al. 2004b). In active deep landslides, slope movement may well be due to intense plastic strains, occurring in some regions of the subsoil where shear stresses are close to the failure envelope of the soil. In other words, displacement points of the landslide body may not be localized on a sliding surface. The rate of movement is generally very small and constant (i.e. not characterized by a seasonal trend), according to the steady groundwater regime. In both cases drains can play an important role in contributing to slope stability. In potential or active shallow landslides drains might prevent the rise of the water table, which is a consequence of atmospheric changes, up to the critical level that endangers slope stability. Nonetheless, methods of analyzing the stabilization effect of drains commonly available in the literature (e.g., Hutchinson 1977, Desideri et al. 1997) model the groundwater regime as a steady-state phenomenon (seepage) and assume the presence of a film of water on the ground surface. In areas where the weather is not very rainy, such as in southern Europe, this assumption underestimates the effects of drains on slope stability. In potential or active deep landslides drains reduce the steady-state pore pressures, which in turn increase the difference between mobilized shear stresses and the soil shear strength. In this way the plastic strain rate in the overstressed zone is drastically reduced. The effectiveness of drains is correctly

(10)

u γw

(11)

ξ is the geometric head, u is the pore pressure and γw is the water unit weight; ε˙ v is the volumetric strain rate of the soil. Assuming the solid skeleton is an isotropic linearly elastic medium, with Young’s modulus E and Poisson index ν, volumetric strain is expressed as: εv = dp′ ·

3(1 − 2ν) E

(12)

where dp′ is the variation in the mean effective stress during the transient phenomenon. Under the assumption that mean total stresses p( p = p′ + u) remain constant in all the volume subjected to the analysis we obtain: dp′ = −du = −γw dh

(13)

Substituting eq. (13) in eq. (12) and eq. (12) in eq. (10), we obtain:   ht − cv hxx + hyy + hzz = 0 cv3D =

KE (x, y, z) ∈  0 < t ≤ T 3(1 − 2ν)γw

(14)

where cv3D is the coefficient of consolidation in 3D condition;  and T are respectively the spatial and time integration domains. The first of equations (14) can be simply specialized to 2D and 1D cases; contextually the expression of the coefficient of consolidation has to be modified, according to table d5. Once the hydraulic boundary conditions are fixed, the transient solution tends to a steady distribution

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Table 5. Consolidation equation and coefficient of consolidation for 2D and 1D conditions. Conditions

2D

Equation

ht − cv2D (hxx + hzz ) = 0

p′ (mean effective stress)

(σx′

1D

+ σz′ )(1 + ν) 3

εv (volume strain)

(1 + ν)(1 − 2ν) d(σx′ + σz′ ) E

assumption

d(σx + σz ) = 0

cv (coefficient of consolidation)

cv2D =

KE 2 · γw (1 + ν)(1 − 2ν)

ht − cv1D hzz = 0 σz′ (1 + ν) 3(1 − ν)

1 − ν − 2ν 2 ′ dσz (1 − ν)E

dσz = 0 cv1D =

KE · (1 − ν) γw (1 − ν − 2ν 2 )

of pore pressures u(∞, x,y,z), that can be obtained directly by integrating equation (15): (hxx + hyy + hzz ) = 0

(x, y, z) ∈  0 < t ≤ T

(15)

From eq. (15), it is clear that the steady solution does not depend on the properties of the soil. At equations (10) or (15), hydraulic conditions at boundaries must be added. Among them, water flow through the ground surface affects pore pressures in subsoil more than others. To take this aspect into account, two different hydraulic conditions at ground surface are considered in this paper: • a film of water continuously present, • a flux of water varying with a seasonal trend. When, in practice, drainage works is analysed by means of numerical codes (DEM or FEM), the problem may be solved by taking soil stratigraphy and heterogeneity into account. Pore pressures can be calculated all along the critical sliding surface; then they can be used in slope stability analysis. Practitioners very often prefer to estimate pore pressure, lowered by drains, by means of non-dimensional charts obtained for homogeneous soil and very simple geometric schemes. Design charts are a general tool: they cannot consider hydraulic conditions at ground surface with a seasonal trend, which necessarily depends on typical climatic features of the region being considered. Hence the design charts presented in this paper are obtained under the most generic assumption of a water film at the ground surface. 3.4.2 First solution (film of water at ground surface) In this section steady-state solutions are presented for drains operating in 3D conditions, assuming a film of

Figure 54. Usual geometry of drain systems: a) drain trenches, b) horizontal drains. The cells analysed in this paper are delimited with bold lines.

water fixed at ground surface. From this configuration 2D solutions are derived as particular cases. Cells are considered, each containing a single drain and confined on lateral boundaries with impermeable planes (Fig. 54). The scheme represents a case in which an infinite distribution of drains is working, so that it is possible to isolate the soils drained by a single drain, between two consecutive planes of symmetry, that behave as impermeable surfaces. Hydraulic conditions along cell boundaries are indicated in Figure 55. At ground surface the following condition, which represents the film of water at the ground surface and is able to recharge the water table, is imposed: h=z

(16)

Under this assumption all the examined domain is submerged.

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Groundwater regime in natural conditions, before the construction of drains, can be very often schematized through the model of infinite slope (1D), which is a very simple case to analyse, the pore pressure regime being described by the relationship below: u=

Figure 55. Analysed cells, with indication of hydraulic boundary conditions.

Condition (16) is imposed along the drain boundaries1 ; it physically means that inside the drains the pore air pressure is atmospheric. The presence of drains inside soils modifies the water level in a complex way. In the analysed cell represented in Figure 56 this situation is shown for the case of drain trenches, by means of pressure head envelopes on three planes parallel to the ground surface at three different depths. The lowering of the water table caused by drains is not homogeneous in : it depends upon the distances of the examined point from the drain boundaries and from the ground surface. The drainage effect is weaker in the deepest zone of the slope. Drains cause a non-hydrostatic pore pressure distribution, involving a flux downwards. The vertical component of head gradients are larger near the drain boundaries where pore pressures can be much lower than those of hydrostatic ones. The effect of drain surface on pore pressure distribution is also evident in Figure 57, where the case of three trenches all around the cell is compared to the case with two trenches positioned in front of one another. The action of horizontal drains is represented in Figure 58, through the pore pressure distribution and its evolution in time, along three vertical axes and three horizontal ones. To use design charts it is useful to refer the problem to a simple framework that describes pore pressure regime in slopes, as indicated in the following.

1 For horizontal drains, Marino (2007) has shown that the hole can be schematized as a segment coincident with its axis. With this simplification she obtained results very close to the more rigorous ones in which the hole is correctly schematized as a cylinder.

γw · (D − Hw ) 1 + tan α · tan β

(17)

where D and Hw are the depths of a generic point and the water table surface respectively, both measured along the vertical from the ground surface (Fig. 59); α and β are the slopes of the ground surface and of the water flux respectively. Supposing the water flux is parallel to the ground level and using the quantity zw , which expresses the depth below the water level measured along the normal to the slope, eq. (17) is changed into eq. (18): u = γw zw cos α

(18)

As has been shown above, distribution of pore pressure caused by drains is complex. Therefore simplification is required to handle the problem more manageably. Accordingly, 3D pore pressure distribution resulting from the action of drains can be schematized as a 1D distribution, equivalent to 3D distribution as regards its influence on slope stability. 1D distribution can be obtained by replacing the pressure head envelope on each plane parallel to the ground surface with a uniform one. In Figure 60, on the longitudinal section of the slope, is represented the equivalent pore pressure regime that reconnects the problem to that of an infinite slope. Assuming 1D equivalent condition in groundwater regime, pore pressure is easily calculable. On a planar sliding surface parallel to the ground level, pore pressure is uniform and is calculable by means of eq. (17). Along a circular sliding surface Ŵ, pore pressure uŴ (t, x) is a function of the depth of the point of Ŵ. To simplify the analysis mean pore pressure on Ŵ, u¯ (t, Ŵ), can be defined as follows: u¯ (t, Ŵ) =

Ŵ



uŴ (t, x)dx

l

(19)

where l is the length of the unstable mass on the ground level, between the two ends of the sliding surface. Because a film of water is assumed at ground surface, sliding surface is completely submerged, and the average pore pressure u¯ (t, Ŵ) along it can be evaluated at the depth 2/3zŴ , where zŴ is the maximum depth below the water level in direction z, normal to the slope. The effect of drainage is often evaluated by means of the efficiency E(t, x, y, z), which in each point of  expresses the difference between the initial and current

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Figure 56. Pressure head envelope between trenches on three planes at different depths: a) D/H0 = 0.5, b) D/H0 = 1, c) D/H0 = 1.5.

Figure 57. Pressure head envelope on the plane at the base of trenches for the case of 3 drain surfaces (a) and 2 drain surfaces (b), adjacent to the analysed cell.

value of pore pressure (at a generic time t), normalized to the initial value: E(t, x, y, z) =

u(0, x, y, z) − u(t, x, y, z) u(0, x, y, z)

(20)

When design charts are used, the function E, defined in each point of , is not particularly useful for evaluating the overall stability of slopes. We need a function that expresses the global evolution of pore pressure along the critical sliding surface Ŵ. By means of u¯ (t, Ŵ) we can define the average efficiency E¯ along the sliding surface: ¯ Ŵ) = u¯ (0, Ŵ) − u¯ (t, Ŵ) E(t, u¯ (0, Ŵ)

(21)

Finally, with a view to representing the steady-state solution, the function E¯ ∞ can be used: u¯ (0, Ŵ) − u¯ (∞, Ŵ) E¯ ∞ (Ŵ) = u¯ (0, Ŵ)

Figure 58. Pore pressure distribution determined by horizontal drains, along 3 vertical axes (a, b, c) and 3 horizontal axes (e, f, g).

The function E¯ ∞ plays a key role in designing slope stabilization by drains, because it represents the final distribution of pore pressure [¯u(∞, Ŵ)], used in the calculation to obtain the desired improvement in slope stability. In practice, E¯ ∞ (Ŵ) is calculated, after determining the function u¯ (∞, Ŵ) from slope stability analysis, as the pore pressure distribution that guarantees the safety factor chosen by the designer. From E¯ ∞ (Ŵ), by means of non-dimensional charts, the designer can determine the geometric characteristics of the drain system. On the basis of the analysis presented in this paper, the value of E¯ ∞ (D) has been calculated and represented in design charts (see Appendix A) as a function of the following parameters:

(22)

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H = depth of analysed volume , H0 = depth of drain,

Figure 59.

Infinite slope with a generic water flux. Figure 61. Water flux recharging water table during the year, obtained from monthly rainfall, from which runoff and evapotranspiration are subtracted (from D’Acunto and Urciuoli 2006, modified).

Figure 60. Transformation of the real pressure regime (represented in the cross section) in a equivalent one (in the longitudinal section).

D = depth of the plane on which efficiency is evaluated (correspondent to sliding surface), Ly = longitudinal length of the analysed volume  (in the case of trenches it is the spacing between principal branches of drain trenches), S = spacing between secondary branches of drain trenches, i = spacing between horizontal drains, l2 = length of secondary branches of drain trenches, l1 = Ly − l2 . 3.4.3 Second solution (variable water flux at ground surface) The solution presented in this section regards the effect of an unsteady hydraulic condition at the ground surface, where a function of time, representing water flux which recharges the water table, is applied (Fig. 61). Water recharging the water table is calculated by subtracting runoff and evapotranspiration from rainfall, using a study on groundwater regimes in relation to climatic features in southern Italy (D’Acunto & Urciuoli 2006). The adopted function expresses the concept that when rainfall is abundant a film of water forms at the

ground surface and an amount of rain infiltrates. If the climate is cool, evapotranspiration is low and most of the infiltrated water recharges the water table. If the climate is warm, rain evaporates and the quantity that infiltrates is removed by evapotranspiration. This case is solved only for a system of drain trenches analysed in 2D conditions so as to quantify the influence of hydraulic conditions at the ground surface on drainage efficiency. Having fixed a flux at the ground surface, instead of a head, the position of the water table is unknown and is obtainable from analysis. In the cases studied, the ground water level is always located a few metres from the ground level. In this case, the sliding surface is partially submerged. In order to calculate efficiency (E), in all points above the water table it is assumed u = 0 (neglecting suction); thus efficiency varies from 0 to 1 during the phenomenon of pore pressure equalization. As regards the depth at which efficiency has to be calculated if the sliding surface is circular, mean pore pressure is first calculated along the submerged part of the sliding surface. It corresponds to the pore pressure at an equivalent depth zw = 2/3zŴ . Hence this value has to be adjusted to take into account the part of the sliding surface above the water level, along which pore pressure is zero. To this aim, in Figure 62 a coefficient [η] is proposed to make this adjustment, based on simple geometric considerations on the examined problem that are not illustrated here to save space. By means of η the equivalent depth must be calculated as 2/3·η ·zŴ . While at work, drain trenches are partially submerged. This situation influences hydraulic conditions on the contacts between drains and soils. Given that backfilling material has a degree of saturation more or less equal to zero, water can flow only from the subsoil to the trenches and not vice versa. In this case, a double boundary condition is necessary on the trench:

118

Figure 62. Corrective coefficient to evaluate the equivalent depth at which efficiency must be calculated in the case of partially submerged circular sliding surface.

hn = 0, if u < 0 in  (the trench boundary must be considered impermeable), h = z, if u > 0 in , where n is normal to the trench boundary. In Figure 63 the solution of a case with a flux applied at the ground surface, taken from D’Acunto et al. (2007), is reported. The piezometric surface is compared to the pressure head envelope on the plane through the bases of trenches at different times: at the initial condition, one year after, eight years after the beginning of trench work (data on soil and geometry are reported in the figure). In each diagram two axes are reported: on the left there is the height above the z = 0 plane (positioned at 10 m from the ground surface), which is useful to locate the water table; on the right there is the pore pressure scale, to represent the envelope of pore pressure on the plane through the base of the trenches. The position of the water table allows us to determine the height of the trench which is working at the time considered (it is the height below the water table). The envelope of pore pressure on a deep plane allows the efficiency of the trench to be assessed on that plane. It can be seen that the water table (whose position is essential to fix the boundary condition on the trench wall) is well above the pore pressure envelope. This means that much of the height of the trench works long after the beginning of drainage, and trenches are always ready to discharge the water flow of heavy rains. This explains the ability of drain trenches to avoid pore pressure peaks (as is shown in Figure 53 for natural regimes) during wet seasons. Measurements carried out on instrumented sites where drains were constructed show that the water table is not subject to seasonal fluctuations where drains are in good working condition. This is clearly shown in the next section.

Figure 63. Evolution of water table (WT) and pore pressure (PP) on the plane at the base of trenches in the case of a water flux applied at the ground surface: a) initial condition, b) 1 year after the beginning of the drainage process, c) 8 years after (from D’Acunto et al, 2007). In figure a) the water table and pore pressure are coincident; the difference is due to different scales used for the representation of WT and PP.

3.5 Case history 3.5.1 The site In 1990, in southern Italy, during highway works along the Sele river some ancient landslides were reactivated as a consequence of trench excavations. After these events further works was needed to assure the

119

safety of the highway. In particular, in the urban district of Contursi, the highway crosses an extensive unstable zone affected by several landslides. The unstable area was stabilized by drains, consisting of trenches in some zones and wells with horizontal drainage pipes in others (Fig. 64). The groundwater regime was investigated before and after drain construction, using Casagrande piezometers installed inside boreholes in the whole area. The investigation program was fairly extensive, aiming to measure pore pressure fluctuations inside and around the landslide body and in the substratum, in order to determine the hydraulic conditions at the boundary of the unstable area. Piezometer measurements were carried out continuously from 1991 onwards (Fig. 65). Data are grouped according to zones. As regards zones far away from drains (Fig. 65a), where the groundwater regime is affected by natural conditions, pore pressure undergoes seasonal fluctuations of the order of some metres, more significant in the upper cells but noticeable also in the lower cells. As regards the drained zone, measurements at piezometers installed between trenches show that pore pressures are well below those measured far from the drained area. Moreover, seasonal fluctuations more or less vanish. This is more evident in zone B, which is the closest to the drain system, while in zone A the behaviour is intermediate.

Figure 65. Pore pressure measurements at Contursi in zones: a) far from drainage, b) intermediate, c) very close to drainage (from Pellegrino et al. 2004a).

These observations show the greater stabilizing effect of drains, since the maximum pore pressure, attained during winter and early spring, is the cause of reactivation of seasonal landslides. This aspect of drain action consisting in the ‘‘lamination’’ of the seasonal trend has not been stressed in the literature despite its great importance in slope stability. At the Contursi site, soils were investigated extensively both in the laboratory and in the field: the most reliable value of permeability k was obtained from fallen head tests carried out inside piezometers. Results were very scattered, with an average close to 10−8 m/s. As regards the coefficient of consolidation cv , a value of 3.8 · 10−5 m2 /s was calculated from the permeability obtained by in situ tests and oedometer ed modulus Eoed , by laboratory tests (cv = k·E γw , γ w = unit weight of water).

Figure 64. The slope uphill of the highway along the Sele river in Contursi; stabilized with drain trenches and wells: a) map, b) stratigraphic section (from Pellegrino et al. 2004a).

3.5.2 Analysis of groundwater regime in the site of Contursi Referring to the described site the slope was modelled as a trapezium (Fig. 66), constituted by homogeneous

120

Figure 66. Analysed domain with hydraulic conditions adopted at the boundaries (D’Acunto and Urciuoli 2006).

soil, whose bases are 105 m (H ) and 70 m, and length 280 m(L). The slope α is 7◦ 1′ . Numerous parametric analyses were performed in order to study the response of the water table to different conditions at the boundaries. It emerged that water flow through the ground surface was the hydraulic condition that most influenced pore pressures in subsoil around the slip surface. This result coincides with the fact that the slip surface is relatively shallow (its depth is always less than 9 m). Hence major efforts were made to examine the hydraulic condition at the ground surface, which allowed the function in Figure 61 to be obtained. The vertical boundaries on the uphill side were considered a section of an almost infinite slope, since there was no perturbation either of the geometry or soil stratigraphy nearby. It was then supposed that around the boundary the pore pressure was constant along the direction, parallel to the ground surface. All along the base of the trapezium, the piezometric head was assigned according to a linear function, starting at the downhill side where the water level was at 65 m (in the Sele River). As regards the hydraulic condition on the downhill vertical side of the examined volume it should be noted that this boundary is very close to the alluvial soils of the Sele river. Hence it is considered to have a greater capability of draining water from the analysed subsoil. The analysis was developed in transient conditions, regulated by the function in Figure 61. Thus results were obtained that represent the pore pressure fluctuations without and with drains as a consequence of the variable hydraulic conditions at ground surface. When drains are working, fluctuations are drastically reduced: Figure 67a shows the seasonal trend of pressure head in a point at a depth of 6.6 m from the ground surface. Numerical results are compared (Fig. 67c) with measurements at piezometers inside two secondary branches of trenches (piezometers TC3, Sc and SF in Fig. 67b). It can be seen that the solution obtained for a permanent film of water at the ground surface overestimates pore pressure in the subsoil. Measurements are well fitted by the analytical solution with flux imposed at the ground surface.

Figure 67. Groundwater regime at Contursi. Comparison between measurements and numerical results (from D’Acunto and Urciuoli, 2006): a) calculated pore pressure without and with drains, b) position of piezometers with respect to drain trenches, c) comparison between measurements and numerical results for different boundary conditions.

3.5.3 Influence on efficiency of the hydraulic condition at the ground surface By considering the imposed flux at the ground surface, instead of the water film, the following results were obtained: • the water table is strongly depressed compared with the conditions without drains, • seasonal fluctuations are lowered by drains.

121

To evaluate efficiency, reference should be made to maximum values assumed by functions uo and u∝ during the year. It is thus possible to calculate that efficiency is up to 40% greater than that corresponding to a film of water fixed at the ground surface (D’Acunto & Urciuoli 2006).

Geotechnical Engineering in Naples. The related studies was mainly led by the second author with contributions made by Prof. Berardino D’Acunto, Ms. Nunzia D’Esposito (Eng.) and Ms. Roberta Marino (Eng.), that are gratefully acknowledged.

REFERENCES 4

CONCLUSION

Soil nailing is a robust slope stabilization measures and is generally applied to stabilize man-made slopes, retaining walls or excavations. The interaction between the ground and the soil nails is complex. Nevertheless, the findings of numerous researches have provided some understanding of the mechanism and behaviour of soil-nailed systems, particularly on the effect of nail inclination, soil-nail heads and bending stiffness of reinforcement bars. Design practice is well established. In contrast, there is room for advancement of the construction technology and the application of innovative materials as soil nail reinforcement. The use of non-destructive technique can greatly enhance the confidence of quality control of soil nail construction. Subsurface drainage stabilizes the ground by preventing significant rise in porewater pressure. Compared with other types of engineering works, it is a very cost-effective stabilization measures, especially at large site. Different types of drains have been developed and they serve different purposes. Trench drains and horizontal drains are more commonly used than other types of drains, and simple design charts have been developed to facilitate designers to assess the effects of the drains on pore pressures.

ACKNOWLEDGEMENT This paper is published with the permission of the Head of the Geotechnical Engineering Office and the Director of Civil Engineering and Development, Government of the Hong Kong Special Administrative Region. The part of this paper on soil nailing technology was based on the review and research work carried out in the Geotechnical Engineering Office in Hong Kong. The soil-nail related studies was mainly led by the first author with contributions made by many colleagues, particularly Mr Y.K. Shiu, Dr W.M. Cheung, Dr G.W.K. Chang and Dr D.O.K. Lo. All contributions are gratefully acknowledged. The part of this paper on drainage was based on the research work carried out in the Department of

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Appendix A. Example of design charts.

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Special lectures

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Loess in China and landslides in loess slopes Z.G. Lin & Z.J. Xu Northwest Research Institute of Engineering Investigations and Design, Xian, China

M.S. Zhang Xian Geological and Mineral Resources Research Institute, Xian, China

ABSTRACT: Loess in china is outstanding for its stratigraphical intactness, huge thickness, vast expanse of distribution and metastable geotechnical properties. During the Quaternary Period the arid, semi-arid climatic environment provided the stage on which material sources came into being, grain particles were wind-transported and deposited and loess strata were thus formed. The paper puts emphasis on the granulometry and engineering properties peculiar to loessial deposits as well as the role they might play in the initiation and evolution of loess landslides. From this, the paper proceeds to such problems as geomorphological zonation, landslide distribution, morphological and structural features of loess landslides as well as landslide classifications, giving due attention to seismic landslides. This is followed by a summing-up of the methods of landslide prevention and remediation. Finally, the issues that need more attention and further in-depth research are raised.

1

INTRODUCTION

Loess covers large and wide parts of Northwest, North and Northeast China, its distribution being concentrated on the so-called Central Loess Plateau which comprises Shaanxi, Gansu and Shanxi provinces as well as most of Ningxia Autonomous Region. Here the loess deposit formed in the unique Quaternary climatic environment is outstanding for its stratigraphical intactness, sustained distribution and huge total thickness. The granulometric composition of loess is remarkable for stableness, maintaining in the meanwhile a well-oriented gradual change from Northwest to Southeast. In engineering properties the soil is characterized by its state of consolidation (underconsolidated in the upper part and slightly overconsolidated in its lower part). These properties are likewise remarkable for their regular and oriented change in the general direction from Northwest to Southeast all across the Loess Plateau. In such climatic, sedimentation and stratigraphical environment conditions, geomorphological evolution has followed correspondingly unique paths resulting in the formation of loess landslides peculiar to such conditions. In the push for the development of the West, it is important to have a correct understanding of loess landslides and use proper methods of investigation, prevention and remediation. This paper tries to give a brief summing-up of the research results achieved and

experiences accumulated up to now so as to contribute to the development of new research programs and new prevention and remediation methods. 2

DISTRIBUTION OF LOESS IN CHINA AND ITS ENGINEERING PROPERTIES

2.1 Distribution of loess in China Loess in China is mainly distributed in Gansu, Shaanxi, Shanxi, Henan and Qinghai provinces, the Ningxia Autonomous Region and the Inner Mongolian Autonomous Region, blanketing the Central Loess Plateau and its neighboring areas. Next in distribution are Hebei, Shandong, Liaoning and Heilongjiang provinces, and the Xinjiang Autonomous Region .The land surface covered by loess in China is estimated to be 6.3 × 105 km2 which is equal to about 6.6% of the total land area of the country. See Figure 1. 2.2

Stratigraphical features

Stratigraphical intactness and huge total thickness are prominent features of loess in China, owing to which almost complete records of the geographical, climatic, depositional and biological changes and evolution over the entire Quaternary Period have been preserved. What is shown in Figure 2 is a typical stratigraphical profile of the loess in China.

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Figure 1.

Distribution of loess in China.

Figure 2. A typical sketch profile of loess in China.

Figure 3.

Granulometric zoning of the loess plateau.

Figure 4. China.

Engineering geological zonation of the loess of

2.3 Granulometry Throughout the Quaternary Period the granulometric composition kept stable. However against this background a gradual change from coarse to fine in the general direction from NW to SE is discernible (Fig. 3). This stableness in composition is reflective of the geographical, climatic and depositional environment conditions of that period and evidences the eolian origin of loessial deposits. Of course the existence of the numerous reddish brown paleo-soil layers bears evidence of the climatic fluctuations in the generally stable environment. 2.4 Engineering properties Of the engineering properties of loess four deserve special attention. They either play important roles in the initiation and evolution of landslides or are contributive to the behavior of man-made slopes. 2.4.1 Collapsibility Loessial terrains (or construction sites) fall into two types, i.e. the type which collapses under the pressure

from its overburden weight when saturated and that which collapses under the combined pressure from its own overburden and additional loads (fills, foundations, embankments etc) when saturated. In Figure 4 is shown the engineering geological zonation of the loess in China and in Table 1 are given the soils laboratory test results (mean values of physico-mechanical indices) for Zone I, Zone II and Zones I, II, III, IV, V and VII as a whole. Figure 5 is a generalization of the afore-mentioned regularity of oriented gradual change (amelioration) in engineering properties of loess in the direction from NW to SE all across the Central Loess Plateau and its neighboring regions. And Figure 6 is a photo which tells of the grave consequences that could be brought about by flooding in overburden-collapsible loessial terrains (Qian et al. 1988, Lin 1994, Wang et al. 1990). In the meanwhile, it should be noted that in step with increase of water content the soil sample invariably tends to increase in compressibility with corresponding decrease in collapsibility, as shown in Figure 7.

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Age

Q3 Q22 Q12

Q3 Q22 Q12 Q1

Q3 Q22 Q12 Q1

Zone

Gansu Prov. (Zone I)

North. Shaanxi and Longdong (Zone II))

Six Zones (Zone I-V) and Zone VII)

13.21 12.85 13.09 20.10

13.38 15.30 18.10 20.10

8.98 6.79 4.90

Nat.wat. cont. w(%)

15.26 16.26 17.23 18.55

14.70 16.40 18.30 18.50

14.58 15.37 16.20

Unit wt. γ (kN·m−3 )

1.016 0.884 0.783 0.773

1.105 0.924 0.760 0.773

1.032 0.868 0.742

Void rat. e

28.62 29.11 29.01 31.05

29.9 30.9 30.6 31.1

27.60 26.87 26.40

Liquid lim. wL (%)

10.86 10.79 11.25 11.65

11.2 12.2 12.3 11.7

9.37 7.34 8.90

Plast. index Ip /(%)

Init. pressure Psh (MPa) 0.135 0.270 1.000 0.092 0.416 0.684 0.137 0.419 0.981

Coef.of overburd. collaps. δzs 0.052 0.054 0.051 0.023 0.014 0.008 0.003 0.015 0.018 0.026 0.003

Table 1. Soils laboratory test results (mean values of physico-mechanical indices) for different zones.

Figure 5. Characteristic generalization of the initial collapse pressure-overburden pressure relationships.

Figure 6. General view and close-up of ground subsidence in and around test pit caused by flooding (Q4 and Q3 loess).

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0.580 1.380 2.148 1.855

0.835 1.224 1.680 1.856

0.508 2.047 3.800

Precons. press.(nat.) Pc (MPa)

0.158 0.453 1.036 1.615

1.615

0.097 0.473

0.137 0.287 1.040

Precons. press.(sat.) Pc (MPa)

10.96 15.19 12.25 13.00

13.20 15.00 12.00 13.00

11.20 10.00 9.00

Sand grains (%)

70.70 68.62 69.87 70.00

65.80 68.10 64.00 70.00

78.75 76.00 77.00

Slit cont. (%)

18.34 16.19 17.88 17.00

21.00 16.90 24.00 17.00

10.05 14.00 14.00

Clay cont. (%)

Figure 7. Interrelationship between the coefficient of collapsibility, δs , and the coefficient of compressibility, a, as a function of the degree of saturation, Sr .

Figure 9. 75 kPa.

Normalization is feasible under condition σ3 ≥

Figure 10. Stress-strain curves of soil samples from Xian (depth ≤ 10 m).

Figure 8. K0 becomes constant when consolidation pressure σ3 ≥ σ3cr .

2.4.2 Structurality and structural strength Loess is outstanding for its structurality which is due to its environment of deposition and the diagenetic process it underwent, its micro-structural features, the existence of soluble salts and the soil matrix suction. As shown in Figure 8 the coefficient of lateral pressure K0 turns constant only after the consolidation pressure reaches and exceeds σ3cr . Figure 9 shows that normalization by P(=(σ1 + 2σ3 )/2) of the (σ1 − σ3 )/P vs axial strain ε1 curves of the soil sample is feasible only after the consolidation pressure exceeds 75 kPa. Figure 10 and Figure 11 respectively depict the q(=(σ1 − σ3 )/2) vs ε1 relationships of loess samples from two different depths and consolidated under different pressures σ3 or, in other words, the soil

Figure 11. Stress-strain curves of soil samples from Xian (depth: 13.2–17.4 m).

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suction-dependent strength becomes very low or even vanishes when the soil water content is over 25% or degree of saturation exceeds 65%. Figure 12 (Xing 2001) gives an example depicting this soil-water characteristic.

Figure 12. Soil-water characteristic curve of a loess (Q3 ) sample from Yangling, Xi’an, Shaanxi Province.

2.4.4 Shear strength of sliding bed soil types In addition to its suction-dependency and like many other soils, the shear strength of loess is also remarkable for the large drop from peak to residual parameter values. Table 2 gives the classification of sliding bed soil types in loess slides and in Table 3 are given the geotechnical properties of those soil types (Wang 2005).

Table 2. Classification of sliding bed soil types in loess slides.

3

Sliding bed soil type and age Malan Loess (Q3 ) Loess

Physical properties of bed soil

3.1 Loess landslides as a peculiar geomorphological element of the Loess Plateau

Silty bed soil

Figure 13 (Derbyshire et al 1994) is a schematic representation of the geomorphological divisions (zoning) of the Central Loess Plateau. Each division has its characteristic and representative landform element or elements such as Yuan (platforms or residual platforms) as shown in Figure 14, Liang (ridges) as shown in Figure 15, Mao (mounds, hills) also shown in Figure 15, fluvial plains and fluvial terraces as shown in Figure 16. Loess landslides as peculiar landform elements are scattered among them.

Lishi Loess (Q2 ) Wucheng Loess (Q1 )

Clayey bed soil

Mixtures of loess and underlying bedrock material (Q4 ) Weak intercalations in bedrock

DISTRIBUTION AND TYPES OF LOESS LANDSLIDES

Weak rock-turned bed soil

3.2 Distribution of loess landslides structurality vs soil age relationships (Qian et al. 1988, Lin 1994, Wang et al. 1990). Structural strength of loess invariably weakens as water content increases until the state of saturation is reached. In this state the structural strength of loess can be termed the ‘‘residual structural strength’’ or, usually in subsoil treatment and foundation design in China, the ‘‘initial (collapse) pressure’’. It is to be pointed out that for loess the terms ‘‘residual structural strength’’, initial (collapse) pressure’’ and ‘‘preconsolidation pressure’’ are technical synonyms, having the same numerical value. 2.4.3 Suction or matrix suction As a kind of unsaturated clayey soils and like swelling clays, loess distinguishes itself by particularly high suction when it is in a state of low moisture content. This suction in loess manifests itself as a kind of negative pore pressure effect and enables the soil to manifest high strength which however decreases in step with increase of water content. Usually, this

Figure 17 (Jin et al. 1996) is a schematic map of landslide distribution in the Loess Plateau and its contiguous areas, on which the clusterly nature of landslide distribution is discernible. Each large cluster has its combination of natural conditions and slidecausative factors. For examples, the NS trending lengthy cluster near the right side margin of the map is just where the Yellow River gorge and the west slope of the Lüliang mountain range are; the linear assemblage at the lower middle of the map is representative of the 180 or more old landslides which are closely spaced one after the other and extend eastward from Baoji along the north shore the Wei River; and the clusters of landslides west of the Lüliang Range on the map came into being during the 1920 M = 8.5 Haiyuan Earthquake. Landslides and their distribution can be located and delineated with precision by means of satellite images and air photos. The successful use of infrared color imagery for Xi-Yu expressway is an example, as shown in Figure 18.

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Table 3. Geotechnical properties of different sliding bed soil. Geotechnical properties Strength parameters

Sliding bed soil type

Permeability k(cm · s−1 ) (3–7) × 10−4

Silty bed soil Clayey bed soil (5–6) × 10−6 Weak rock-turned bed soil 0.4, suction is lower than the values obtained by conventional laboratory tests (curve 1 in fig. 9b). This effect is probably related to the different rate of suction decrease imposed in the two procedures accounting for the pyroclastic nature of the particles. The size of the pores between the particles (intergranular pores) is significantly higher than that of internal pores (intragranular pores) (see fig. 4). Consequently the amount of retained water required to cause a given suction decrease is a function of the exposure time. Indeed only long exposure times (as in the conventional tests) enable the intragranular pores to absorb water. 2.2.2 Results of infiltration tests on layered artificial deposits In this section we discuss the results of phase IVa of the layered FL20 test with a pumiceous layer of soil A interbedded between two ashy layers of soil B. In this test the slope angle was of 40◦ and a pervious boundary was reproduced at the base (side 1); rainfall intensity was kept constant at 18 mm/h for 4 hours and 50 minutes. The results in terms of suction against time are shown in figure 10. As in the case of homogeneous deposits a marked suction decrease appears throughout the soil mass, but in this case an instability condition was not reached. The comparison between the measurements in the two ash layers across contact with

0 5 10 15 20 section A-A'

25

section B-B' section C-C'

30 0

40

80

120

160

200

240

280

Figure 10. Suction measurements during FL20 at three different depths.

208

t = 00' 00 t = 92' 44 t = 293' 38

0.2

section 2.1. The unsaturated permeability functions are described by the Brooks and Corey (1964) expression. The adopted retention curves are described by the van Genuchten (1980) expressions: for both topsoil and ash (B) the retention curves of figure 9b were adopted; for pumices (A and C) the retention curve of figure 11b; for altered ash (D) data from the literature (Fredlund and Rahardjo, 1993).

0.8

t = 44' 03 t = 175' 33

flume infiltration test van Genuchten, 1980

0.6

θw

ashes 0.15

vG: m=0.3 n=7 α=3.3 θr=0.25 θs=0.55

0.4

pumices 0.1 ashes

0.2

z [cm]

0.05

θw

0 0.0

0.2

0.4

0.6

0.0 0.8 0

ua-uw [kPa] 20

40

60

Figure 11. Volumetric water content profiles retrieved by TDR during experiment FL20 (a) and Water Retention Curve of pumices derived from TDR.

the pumiceous layer shows an immediate response in the top layer with a different trend and a marked delay in the bottom layer. Furthermore, the comparison between the two suction measurement performed in the same bottom ash layer highlights a negligible difference despite the different depth. Figure 11a gives the θw profiles at different times obtained with the TDR inverse profiling methods using a single probe installed through the three layers. Since the suction assumes the same value at the ashpumice interface, coupling the measurements of the suction in the ash with the θw evaluated in the pumices, it is possible to extract the water retention curves of pumices experienced by the soil during the infiltration process (Figure 11b). The experimental points were fitted by the expression proposed by van Genuchten (1980) (θr = 0.25; θs = 0.55; m = 0.3; n = 7; α = 3.3). 2.3 Analysis of infiltration The complexity of infiltration process on unsaturated soils requires the use of a numerical model. The reliability of the model can be improved through a preliminary calibration (given the variability in soil parameters) and a validation through a back-analysis of in situ monitoring. In this paper the former one was obtained using data from laboratory tests on natural samples and simulating infiltration tests on the slope model; the latter through back-analysis of in situ suction measurements using initial and boundary conditions derived from monitoring. Numerical analysis was carried out using an I-MOD3D (volume finite method) program developed in VBA application for ARCOBJECTTM /ARCGIS 9.2TM to automate the mesh-generation starting from a Digital Terrain Model. The analysis was performed under isothermal conditions for a undeformable unsaturated porous medium neglecting the flux of the gas phase. The saturated permeability of volcanic ashes (B and D) and pumices (A and D) was assumed equal to the maximum values obtained in laboratory tests and reported in

2.3.1 Calibration of numerical model trough a back-analysis of infiltration tests Calibration of numerical model was carried out on the basis of results of infiltration tests on homogeneous (FL10) and layered slopes (FL20). A 3D analysis was performed schematizing the slope model with a mesh characterized by dx = dy = dz = 1 cm, considering the boundary conditions indicated in tables 2 and 3 and assuming for the initial condition a constant value of suction equal to the mean value recorded at the beginning of tests. As an example of this calibration stage, in figure 12 the experimental data of FL10 test are compared with the results of two numerical simulations, considering for the ashes B the van Genuchten parameters obtained either in conventional laboratory tests (simulation 1 with curve 1 in Fig. 9b) or in the flume infiltration tests (simulation 2 with curve 2 in Fig. 9b). As expected, the best fitting was obtained by using retention curve 2 that allows both the trend and the final values of suction at different depths to be captured. Conversely, simulation 1 cannot reproduce the trend of the deepest tensiometers and the value of suction at the end of the test. Retention curve 2 probably leads to higher suction than measured since it was obtained for equalization times longer than the test duration, which was insufficient to involve the water inside the intragranular pores.

0

2 20

40 4 measured data simulation 1 simulation 2 60 6 0

10

20

30

Figure 12. Test FL10: comparison between suction measurements and numerical simulation at section A-A′ in fig. 7.

209

80

0

70

40

60

80 nest 1 nest 2 nest 3 nest 4 nest 5

40 30

Mar-03

Jan-03

Feb-03

Dec-02

Oct-02

Nov-02

Sep-02

Jul-02

Aug-02

Jun-02

320 Apr-02

280

0 May-02

240

0 40

intermediate devices depth 0.90-1.00m

40

nest 3 nest 3 nest 4 nest 4

30 20

120 160 200

Mar-03

Feb-03

Jan-03

Dec-02

Nov-02

Oct-02

Sep-02

Aug-02

Jul-02

280 Jun-02

240

0 Apr-02

10

May-02

h [mm]

80

50

2.3.2 Validation of model: comparison between field data and numerical analysis At 2002 a complex monitoring system was installed on the natural grassed area surrounding the catastrophic flowslide of Cervinara (on the right side of the landslide 3a in fig. 1). The equipment consists of a pluviometer and five tensiometer stations. The tensiometers (jet-fill type manufactured by SoilMoisture) were installed at different dephts within the top soil and the volcanic ashes (B and D). Figure 14 shows the suction measured at different depth from April 2002 to April 2003. During the winter season, suction reached very low values, ranging between 2 kPa (in the top soil) and 15 kPa (in layer B). The dry season started in June as indicated by the lack of rainfall during the period June-July. In this dry period the suction increased up to values of some tens of kPa with a peak in middle July, when the shallow device of nest 1, installed at depth of 60 cm in the evapotranspiration zone, indicated a maximum value near to 50 kPa. Subsequently the summer had an abnormal rainfall regime for the Mediterranean area. On August 7th the daily rainfall was of 165 mm only slightly lower than

200

10

60

In the figure 13 the results of the layered FL20 test are compared with the numerical simulation considering, for both the soils, pumices and ashes, the retention curves obtained in flume tests. Numerical analysis correctly simulated the response of both points located across the pumice layer and points located within the ash layer below the pumices, reproducing the different trend recorded above and below the pumice. The results show the importance of model parameter calibration based on flume infiltration test results, since infiltration rates and boundary conditions applied to the model slope resemble real slope conditions.

160

20

70

Figure 13. Test FL20: comparison between suction measurements and numerical simulation at three depths at section A-A′ in fig. 7.

120

h [mm]

superficial devices depth 0.60m

50

Figure 14. Suction measured at the five tensiometer stations as a function of daily rainfall.

the previous 30 years maximum (180 mm in December 19th 1968). Furthermore the total rainfall during the summer (550 mm) was almost half the annual precipitation. Due to the summer rains, suction fell from the end of July with a faster decrease at the end of September. In the months of October and November, it fluctuated around values of a few kPa. Monitoring also shows that during summer a significant reduction in suction occurred as a consequence of prolonged rainfall, as after July, 19th and not during the most intense and short events, as on August 7th. This is a consequence of the low hydraulic conductivity of the shallow unsaturated soils, which prevents water infiltration during short precipitations. In contrast, infiltration is facilitated by the increase of degree of saturation occurring for long lasting events. The period between 13/7/2002 to 8/8/2002, characterized by abnormal rainfall after a prolonged dry period, was selected to validate the numerical model. The slope has been schematized with a 3D mesh derived by DEM (with dx=dy=0.5 m) and by stratigraphy of fig. 1 using a dz=0.12 m. At the ground surface two conditions have been used: average daily rainfall intensity from monitoring or evaporation flux during dry period considering the minimum value suggested by Wilson (1990) for unsaturated soil (ranging between 0.33 mm/h (first day) and 0.012 mm/h (after three days)). For the lateral and base surfaces have been considered a condition of free flow. The initial condition in terms of suction has been established from field measurements.

210

80

0

70 nest 1 0.6m

80

nest 2 0.6m

120

simul. 0.24m

40

160

simul. 0.36m

30

200

simul. 0.6m

20

240

10

280

0 Aug-02

Aug-02

nest 3 0.9m

60

nest 3 1m

50

nest 3 1.9m

40 80

simul 1.08m

40

0

simul 2.28m

30

120 160

m 500

h [mm]

ua - uw [kPa]

70

Aug-02

Jul-02

Jul-02

Jul-02

320 Jul-02

80

h [mm]

ua - uw [kPa]

50

Figure 16.

200

20

30-Aug-02

20-Aug-02

10-Aug-02

31-Jul-02

280 21-Jul-02

0 11-Jul-02

240

01-Jul-02

10

400 300

Figure 15. Comparison between in situ measurements and numerical analysis.

200 100

A comparison between the monitoring results and numeric simulations at three depths is shown in figure 15. During this wet period a adequate agreement has been obtained in both top soil and volcanic ash layer B, although the measured values are discontinuos and a higher decrease of suction is predicted during rainfall. As observed by field measurements and confirmed by numerical analysis prolonged rainfall (13/07–19/07) have an effect on suction higher than a shorter and heavier rainstorm (7/08) in the shallowest layers. In layer D the suction variation does not seem to be strictly influenced by rainfall events and the inadequate prediction of numerical analysis is probably related to the assumption on the initial condition. During dry period the predict trend is qualitatively in agreement with measurements even if characterized by lower value. This is probably due to the effect of transpiration at the upper boundary due to the presence of a grassed surface that has not been considered in the analysis. 3

Forgia Vecchia

1988 landslide

40

60

LANDSLIDES AT ‘‘LA FOSSA’’ CONE

Instabilities occur along the flanks of La Fossa pyroclastic cone that has a base diameter of 1200 m at sea level and is 400 m high (Fig. 16). Major instabilities affected the NE sector (where investigations concentrated) that is formed by a sequence (Fig. 17) consisting from bottom to top of (Dellino et al. 1990): – laminated sands from dry surge deposition, up to 150 m thick (unit A);

0 -100

Shadow relief of La Fossa edifice.

SE

wet surge deposit (C) fumarole

(C)

NW

Punte Nere fall products with dry surge intercalations (B) P.te Nere basal dry surge (A)

(B) Punte Nere lava flow

(A) Punta Roia lavas

Figure 17. Schematic SW–NE profile of La Fossa cone based on data from Dellino et al. (1990). The dashed line indicates the approximated location in the sequence of the 1988 slide.

– layers of coarse grained (up to 200 mm in diameter) fall products with thin intercalations of sands from dry surge deposition (unit B); – a lava flow closing the eruptive cycle. This sequence is overlaid by a wet-surge deposit (varicoloured tuffs) up to 20 m thick (unit C). At present volcanic activity is essentially hydrothermal and concentrates at La Fossa edifice. Degassing is both concentrated at fumaroles (i.e. particular points where structural, stratigraphic and stress conditions created preferential paths) or is diffused throughout the whole cone, favoured by the high permeability of the pyroclastic formations. Around fumaroles pyroclastic materials are altered, often up to the complete argillification; volumes and geometry of the argillified material (material D) are highly variable, depending on flow paths and physicochemical characteristics of the hydrothermal fluids. Small sheet slides occur in the Varicoloured tuffs but the largest documented instability was the translational slide occurred along a partly altered dry surge horizon within the Punte Nere fall deposit (unit B).

211

Table 4. Physical properties of pyroclastic materials. Mat.

γ kN/m3

wn %

ρs Mg/m3

Sr %

e0

CF %

wL , IP %

C A B D1 D2

14.70 13.84 14.62 15.70 14.91

13.52 13.00 20.8 74.5 68.7

2.52 2.67 2.66 2.74 2.61

36–45 28.2 47.5 100 95.0

0.9 1.13 1.16 2.04 1.89

10 – – 57 63

75, 23 85, 41

D1: 1988 slip surface; D2: outer cone rim.

400 dry surge normal to bedding dry surge parallel to bedding wet surge argillified pyroclastite

300

200

100

0

0

Figure 18.

100

200

300

400

500

Figure 19. View of the 1988 slide. At the top on the right the main scarp white altered materials is apparent.

Direct shear tests on pyroclastites.

3.1 Geotechnical characterization Undisturbed samples of dry surge materials (from units A and B), varicoloured tuffs (unit C) and argillified pyroclastites (material D) were taken from cuts. The average values of the physical properties are reported in Table 4. The material forming the varicoloured tuffs is a silt with sand and traces of clay. It has low saturation, apart from local increases around fumaroles. The deposit is homogeneous and has the mechanical behaviour of a non-cohesive soils. Direct shear tests on saturated specimens performed at normal stresses lower than the fragile-ductile transition provide, similarly to triaxial CIU tests, a friction angle ϕ ′ of 36◦ with negligible cohesion (Fig. 18). Dry surge deposits at the base of the Punte Nere sequence and those intercalated in the overlying fall products are both uniform medium sands formed by extremely irregular, clean particles at times bonded together. Strength data from direct shear tests, reported in Figure 18, show slight non-linearity and anisotropy (ϕ ′ from linear regressions forced through the origin varies between 35◦ and 38◦ depending on shearing direction). Otherwise if a cohesion intercept c’ is considered regression yields values of c′ = 30 kPa and ϕ ′ = 30◦ .

Hydrothermal alteration around fumaroles has completely changed mineralogy, texture and hence mechanical behaviour of the parent pyroclastites. Argillified materials are fully saturated and have medium-to-high plasticity deriving from a high smectite content (D1 and D2 in Tab. 4). Materials sampled at different sites show unexpected similarity, from both mineralogical and geotechnical point of view, which indicates that alteration has the same effect on pyroclastites with different texture. Argillification drastically reduces shear strength: friction angle decreases to 26 or 21◦ depending on whether null intercept is imposed or not (in the latter case a cohesion of 21 kPa is obtained). Alteration also affects hydraulic conductivity which reduces to 3.5 × 10−10 m/s. 3.2

Instability phenomena

Small slips occur around the cone within the Varicoloured tuffs but the largest instability recorded at Vulcano is the 1988 slide (Fig. 19). On April 20th about 2 × 105 m3 of pyroclastic material detached on the eastern side of the cone along a dry surge horizon of the fall deposit (B) (Tinti et al., 1999). The main scarp developed at an active fumarole and the uppermost part of the sliding surface was altered up to argillification. Seepage of water condensed from fumarole vapours

212

3.3 Monitoring

Cumulative rainfall (mm)

1000

April 20th landslide

1972-73 1961-62

800

1987-88 1969-70 600

400

200

0

J

A

S O N

1987

D

J

F M

A M J

1988

Figure 21. Average cumulative rainfall between 1965 and 1995 (bar chart) and actual cumulative rainfall in particularly wet years (line plots).

was observed along fractures forming the scarp even some years after the slide. The slide occurred during a major period of unrest, characterized by intensification of seismic activity and degassing (Fig. 20). However failure does not seem to have been triggered by dynamic actions because the strongest seismic event (M = 4, Neri et al. 1991) occurred 20 days before the slide with epicentre at 20 km from the island. Rasà & Villari (1991) invoke also reiterate inflation-deflation cycles and localized action of high-temperature hydrothermal fluids as general causes of strength reduction in the volcano flank. The analysis of hydraulic conditions (groundwater circulation and suction), degassing and rainfall regime at La Fossa cone suggest that failure could be initiated by a sharp change in pore fluid pressures due to the volcano unrest which was possibly favoured by a preceding rainy wet season. The sole rainwater infiltration seems instead to be not sufficient to induce instability. Analysis of hydrologic data between 1958 and 1988 indicate that other three histories of cumulative rainfall were equal or higher than that recorded during the months preceding the slide (Fig. 21).

Sudden increase in volcanic activity

Progressive normalization of volcanic activity

78

so d e i l tem ga s s p. c ing lo po s e t o int (VGa O

74 70

)

66

40

62 30

air tem peratu re

58

20

54

10 soil temp. far from fumarole (VCD)

0 1

rainfall

1.5

16 12

2 2.5

8

3

suction at VGO

3.5 4

0

20

0.5

s at ucti VC on D

Figure 20. Steam output at fumaroles located at La Fossa crater (after Bukumirovic et al. 1997, re-drawn).

In order to collect experimental evidence of the link between soil suction and degassing activity, in 2007 continuous monitoring of soil temperature and suction was started by INGV-Palermo and 2nd University of Naples at two stations (figure 16). Increase in soil temperature revealed, in fact, to be associated to the rise of steam output. Probes were installed in altered pyroclastites, close to a degassing point (VGO station), and far from fumaroles, in fresh pyroclastites (VCD station). During the preceding years, suction was also measured manually by means of portable tensiometres at other locations. Soil temperature and suction are plotted versus time in Figure 22. In order to separate the effect of degassing from those of evaporation and rainwater infiltration, air temperature and rainfall plots were also included. Data allow to draw some preliminary considerations. Soil temperature markedly increases in middle July 2007 and maintains at the same level over all summer. During this period of more intense volcanic activity suction at VGO station progressively decreases in spite of the absence of rainfall and of high air temperature (i.e. higher evaporation). In this respect short suction increments only occur at air temperature peaks. The effect of the increased degassing lasts until the end of November as the insensibility of suction to the increased rainfall demonstrates. Successively suction slightly increases as an effect of scarce rainfall and normalization of degassing. Finally, far from degassing points (station VDC), where soil and air temperature are similar, suction is virtually influenced only by rainfall.

4 J

A

S

O

N

D

J

Figure 22. Soil/air temperature, suction and daily rainfall versus time.

213

3.4 Hydraulic conditions and slope stability Monitoring data seem to support the hypothesis that condensed vapour increases water content of pyroclastites. Furthermore, observations around fumaroles and in wells drilled at the cone foot (Madonia, pers. com.) suggest that condensed vapour could change seepage in the cone and hence shallow groundwater circulation. In fact continuous seepage of warm water from fractures forming the 1988 slide scarp was noticed even during the driest season of the last fifty years (Sept. 2003). Furthermore in fboreholes at the cone foot (i.e. BL1 borehole in Figure 16) high pore water temperature was measured in the dry season within the first 15 m, accompanied by intense alteration. These two elements seem to indicate that warm water from the cone permanently circulates in the shallower layers. The complete displacement of the colder sea water (the borehole top is at sea level) also suggests that flow is continuous and seepage velocity is significant. Evaluating amount and distribution of saturation changes and seepage processes in the pyroclastic deposits is extremely difficult. In order to make preliminary considerations on flow paths and distribution of pore pressures, steady-state and transient 2D seepage analyses were carried out under simplified hypotheses using water retention curves determined in the laboratory. Analyses wee conducted through the SEEP/W code (Geo-Slope International, 2004). Stratigraphy and boundary conditions of the model are reported in Figure 23: the seepage domain was delimited by the ground surface at the top, the slip surface at the bottom, and the fumarole fracture at the upper side where inflow of condensed vapour is imposed (Fig. 23). Along the slip surface the argillified portion is fully saturated whilst downslope suction is equal to average values measured in situ. Inflow is evaluated from average flow measurements in the shallow groundwater (Madonia pers. com). Seepage parallel to layering establishes throughout the partially saturated varicoloured tuffs and in most of the underlying B unit, being more defined at the interface with the argillified horizon. The influence of the fumarole inflow is greatly increased accounting for material anisotropy (kh /kn = 10); under this assumption a maximum matric suction of 70 kPa at the slope top is calculated. If during the unrest the inflow is increased (four times according to the steam output reported by Bukumirovic, 1997), suction decreases by some 20 kPa in the upper third of the slope and a continuous zone of positive pore pressures forms within the varicoloured tuffs, which extends downslope for few tens of metres. Limit equilibrium analyses run using strength parameters reported in section 3.2. indicate that the

in t

e rv

al

of

su

c ti

on

normal inflow of condensed vapour

co

n to

urs

20

kP

a

a)

saturated zone

in t

e rv

inflow of condensed vapour during unrest al

of

su

c ti

on

co

nto

u rs

10

kP

a

ANISOTROPIC

b) 100 m

Figure 23. Increase in concentrated degassing on seepage and suction distribution in the 1988 slide area (a) considering or (b) otherwise anisotropy of the pyroclastic materials; suction contours start at zero value on left boundary.

1988 slide mobilized a shear strength which is intermediate between that of the fresh and of the argillified dry surge material. In the case of dry slope the percentage of argillified slip surface that causes failure does not match in situ observations. Therefore positive pore pressures and/or partial filling of the tension crack at the main scarp are to be considered. Finally the drop in suction evaluated from seepage analyses in the partially saturated Varicoloured tuff blanket could be sufficient to trigger the small slips and sheet slides observed around fumaroles. 4

CONCLUSIVE REMARKS

Comparison between the results of model tests, numerical simulations and in situ measurements allowed us to set up a reliable tool for predicting slope response to rainfall infiltration for shallow layered pyroclastic deposits. The analysis of tests on Cervinara ash showed that analysis is correct if the highest saturated permeability obtained from laboratory tests is adopted and

214

the retention curve is extracted from the results of infiltration test characterized by boundary conditions and infiltration rates similar to real slope conditions. Comparison between homogeneous and layered slope models indicates the strong influence of the pumice layer interbedded between ash layers that substantially modify the flow regime in the lowermost ash layer, preventing soil wetting. This result is confirmed by field monitoring. On the flanks of pyroclastic cones of volcanoes characterized by active hydrothermalism, intensification of degassing can result in a decrease in suction due to localized or diffuse release of condensed vapour. At La Fossa crater the phenomenon was evidenced by in situ monitoring of soil suction and soil temperature. In situ observations and measurements indicate that seepage of condensed vapour is appreciable. Simple models based on the geotechnical characterization of pyroclastic materials suggest the hypothesis that variations in suction can be significant to stability of volcano slopes when these are very close to limit conditions and if material hydraulic anisotropy is considered. The validation of this hypothesis requires further monitoring data during periods of intense unrest and more comprehensive models that account for non-isothermal multiphase pore fluid pressure and groundwater circulation, influencing the state of stress and hence stability. ACKNOWLEDGEMENTS The research was supported by the INGV-DPCI contract V5/12 (2005–2007) and PRIN 2006 Project. Special thanks are due to P. Madonia (INGVPalermo) for monitoring data and suggestions on volcanic activity, to E. Damiano (2nd Univ. of Naples) for model slope tests and to V. Savastano (AMRA) and V. Grana (CNR-IGAG) for numerical simulations. REFERENCES Brooks, R.H. & Corey, A.T. 1964. Hydraulic properties of porous media. Hydrology Paper No. 3, Colorado State Univ., Fort Collins, Colorado Bukumirovic, T., Italiano, F. & Nuccio, P.M. 1997. The evolution of a dynamic geological system: the support of a GIS for geochemical measurements at the fumarole field of Vulcano, Italy. J. Volcanol. Geotherm. Res., 79: 253–262. Calcaterra, D. & Santo, A. 2004. The January, 10, 1997 Pozzano landslide, Sorrento Peninsula, Italy. Engineering Geology, 75: 181–200. Dellino, P., Frazzetta, G. & La Volpe, L. 1990. Wet surge deposits at La Fossa di Vulcano: depositional and eruptive mechanisms. J. Volcanol. Geotherm. Res., 43: 215–233.

Elsworth, D., Voight, B. & Taron, J. 2007. Contemporary views of slope instability on active Volcanoes. Volcanic Rocks, Proc. Workshop W2—11thISRM Congress, Malheiro A.M. & Nunes J.C. (Eds.), Azores, 3–9. Taylor & Francis. Fredlund, D.G. & Rahardjo, H. 1993. Soil Mechanics for Unsaturated Soils. In Wiley-Interscience Pubblication, John Wiley & sons, inc. Gardner, W.R. 1958. Some steady state solutions of the unsaturated moisture flow equation with application to evaporation from water table. Soil Sci., 85(4): 228–232. Geo-slope International 2004. SEEP/W Finite-element code for groundwater seepage analyses. Greco, R. 2006. Soil water content inverse profiling from single TDR waveforms. Journal Hydrol., 317: 325–339. Kunze, R.J., Uehara, G. & Graham, K. 1968. Factors important in the calculation of hydraulic conductivity. Proc. Soil Sci. Soc. Amer., 32: 760–765. Neri, G., Montalto, A., Patanè, D. & Privitera, E. 1991. Earthquake space-time-magnitude patterns at Aeolian Islands (Southern Italy) and implications for the volcanic surveillance of Vulcano. Acta Vulcanologica, 1: 163–169. Olivares, L. & Picarelli, L. 2001. Susceptibility of loose pyroclastic soils to static liquefaction—Some preliminary data. Proc. int. conf. Landslides—Causes, countermeasures and impacts. Davos Olivares, L. & Picarelli, L. 2003. Shallow flowslides triggered by intense rainfalls on natural slopes covered by loose unsaturated pyroclastic soils. Géotechnique, 53(2): 283–288. Olivares, L. & Damiano, E. 2007. Post-failure mechanics of landslides: laboratory investigation of flowslides in pyroclastic soils. Journal of Geotechnical and Geoenvironmental Engineering ASCE, 133(1): 51–62. Olivares, L., Damiano, E., Greco, R., Zeni, L., Picarelli, L., Minardo, A., Guida, A. & Bernini, R. 2008. An instrumented flume for investigation of the mechanics of rainfall-induced landslides in unsaturated granular soils. Submitted to ASTM Geotechnical Testing Journal. Picarelli, L., Evangelista, A., Rolandi, G., Paone, A., Nicotera, M.V., Olivares, L., Scotto di Santolo, A., Lampitiello, S. & Rolandi, M. 2006. Mechanical properties of pyroclastic soils in Campania Region. Proc. 2nd int. work. on Characterisation and Engineering Properties of Natural Soils, Singapore, 3: 2331–2383. Rasà, R. & Villari, R. 1991. Geomorphological and morphostructural investigations on the Fossa cone (Vulcano, Aeolian Islands):a first outline. Acta Vulcanologica, 1: 27–133. Tinti, S., Bortolucci, E. & Armigliato, A. 1999. Numerical Simulation of the landslide-induced tsunami of 1988 on Vulcano Island, Italy. Bulletin of Volcanology, 61: 127–137. Topp, G.C., Davis, J.L. & Annan, A.P. 1980. Electromagnetic determination of soil water content: measurement in coaxial transmission lines. Water Resour. Res., 16: 574–582. van Genuchten, M. Th. 1980. A closed-form equation for predicting the hydraulic conductivity of unsaturated soil. Soil Sci. Soc. Am. J., 44: 615–628.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Prediction of landslide movements caused by climate change: Modelling the behaviour of a mean elevation large slide in the Alps and assessing its uncertainties Ch. Bonnard Formerly Soil Mechanics Laboratory, Swiss Federal Institute of Technology, Lausanne, Switzerland

L. Tacher Engineering and Environmental Geology Laboratory, Swiss Federal Institute of Technology, Lausanne, Switzerland

M. Beniston University of Geneva, Caroug, Switzerland

ABSTRACT: The consideration of predicted climate change conditions in the hydrogeological and geomechanical modelling of a large landslide allows the assessment of its future behaviour in case of crisis. This application shows that the predictions are not necessarily pessimistic, despite of the uncertainties of the needed assumptions.

1

INTRODUCTION AND OBJECTIVES

For many decades there has been a clear consensus within the scientific community to express various quantitative or semi-quantitative relations between climatic conditions and general landslide movements of different kinds, as well as to use these experimental or empirical relations to try to predict future movements (Terzaghi 1950, Wieczorek 1996). However such relations have often proved to be deceiving as many short-term or long-term complex factors influence the crises of landslides. One of the possible predictive approaches with neural networks has tried to combine observed past movement data and climatic information to predict future movements (Vulliet et al. 2000) Such a prediction is nevertheless reliable only in a short-term perspective and when no critical situation is likely to occur (Bonnard 2006), which is indeed the case when a reliable forecast is wished! On the other hand, because of the numerous impacts that a changing climate can have on many elements of the planetary environment, it is of essence to predict the future course of climate forced by enhanced concentrations of greenhouse gases in the atmosphere. Predicting the speed and amplitude of climatic change can thus provide a measure of guidance to decision makers and climate-impact specialists.

In terms of land-surface processes, and particularly slope instability events, climatic factors are often assessed as a key factor in the triggering and/or the amplification of various forms of landslides, rockfalls, debris flows, etc. Precipitation is certainly today the dominant driving mechanism for many forms of slope instabilities, through water loading in soils beyond a critical threshold, or through excessive runoff that will lead to rapid surface erosion and debris flows. Both heavy but short-lived precipitation or moderate but continuous rainfall can thus provoke various forms of slope response, either in a natural or a man-made surrounding. Extremes of temperature can also contribute to slope instability, notably through repetitive freeze-thaw mechanisms at high elevations that tend to weaken rocks by progressively enlarging fractures. In addition, permafrost degradation in high mountains resulting from milder atmospheric temperatures can also contribute to slope instabilities by reducing the cohesion of slope material currently embedded within subsurface ice. In a changing global climate, it is thus of interest to know how temperature and precipitation patterns may change, both in space and time, and also in terms of mean and extreme conditions. Such changing conditions may result in increased or even new forms of slope hazards compared to those encountered under current climate.

217

It is also essential to model the complex infiltration process of rainfall and snowmelt in large slides, in order to be able to establish a transient distribution of groundwater pressure at any point of the landslide mass and within the slip surface, so as to model the movements induced by these pressure changes in a FEM. The objective of this paper is thus to present the global trend of climate change and then to illustrate its possible long-term effect in the case of a large slide. The obtained results will show that in some cases, the so-called evidences are not granted for sure. It is also important to trace and quantify all kinds of possible uncertainties in this multiple process in order to assess the reliability of long-term predictions. 2

TREND OF GLOBAL CLIMATE CHANGE

Perhaps the most exhaustive source of information concerning future climatic change is provided by the Intergovernmental Panel on Climate Change (IPCC). In the succession of reports published in 1996, 2001 and 2007 (IPCC 2007), a number of global climate simulation models have been applied to assess the response of the climate system to anthropogenic forcing in the 21st century, based on a number of greenhouse-gas emission scenarios developed by Nakicenovic et al. (2000). According to the scenario, itself a function of assumptions on population growth, economic growth, technological choice, and policy decisions, the global mean temperature change over present ranges from 1.5–5.8◦ C, as illustrated in Figure 1. This represents an amplitude of change that is probably one order of magnitude larger than changes reconstructed for the past 20, 000 years (i.e., since the last glacial maximum), and a speed of change that is up to two orders of magnitude greater than typical natural fluctuations of climate. Climate model solutions suggest that the change in temperature will be stronger in the high latitudes compared to the equatorial region. This is because of the strong positive feedback that can be expected as a result of smaller areas covered by snow in the northern continents, a shorter winter period, and reduced sea-ice cover in the Arctic Ocean. Reduced snow and ice cover will substantially modify the surface energy balance, particularly through increased absorption of solar energy. Temperature change will also be greater over the continents than over the oceans, because of the larger heat capacity of the oceans. While a warmer climate will enhance the hydrologic cycle, precipitation will not necessarily increase everywhere. The latest climate models published by the IPCC (2007) suggest that the northern latitudes may experience greater precipitation than currently, but that rainfall in Mediterranean and arid climates

Figure 1. Range of climate futures according to greenhouse-gas emission scenarios; the dark gray zone represents the more likely range of global temperature change (IPCC 2007). For the emission scenarios, see Nakicenovic et al. (2000).

may decrease (i.e., many semi-arid and arid regions could become even drier in the future). Precipitation totals will probably increase by 2100 in the Monsoon climates of India and China, and in the inter-tropical convergence zone around the equatorial belt. Temperature in Europe will increase on average by 4–6◦ C, with strong regional and seasonal differences; for example, summertime warming in southern Europe is expected to be greater than during the winter, because of the positive feedback effect of dry soils during this season (Seneviratne et al. 2006). In the Alps, wintertime temperatures will rise by 3–4◦ C by 2100 compared to current climate, according to the level of greenhouse gases. Summer temperatures may rise by more than 6◦ C during the same period, as a result of positive feedback effects from dry soils and reduced snow and ice cover in the Alps (Beniston et al. 2007). Figure 2 shows the difference between summer temperatures for current (1961–1990) and future climates in Basel, Switzerland, not only for mean conditions, but also in terms of the upper quantiles that essentially represent heat-wave conditions. Simulated results for the low emissions B2 scenario and the high emissions A2 scenario are shown; interestingly, the difference in temperature between the high and low emissions scenarios is less than between the B2 scenario and current climate. This implies that even with rather stringent policies to abate greenhouse gas emissions, the increase in temperatures as seen for the B2 scenario will result in summer heat waves that are as intense, or even stronger, than the 2003 heat wave, with an even greater potential for strong heat waves in the A2 scenario. Indeed, statisticallyspeaking, the 2003 heat wave could occur one summer out of two in a future climate (Schär et al., 2004).

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Figure 2. Observed summer temperatures for 1961–1990 (means and 90% quantile values), and simulations for 2071–2100 for the low emissions B2 and the high A2 emissions scenarios established by the IPCC (2007). See text for further details.

In a future climate, whatever the emissions scenario considered, the freezing level will thus rise by about 500–600 m in winter and close to 1, 000 m in summer, thereby accelerating glacier retreat and exacerbating the natural hazards associated with deglacierized landscapes. Positive temperatures at increasingly high elevations will penetrate into permafrost layers, leading to its progressive melting and thus reducing the cohesion of soils. Christensen & Christensen (2003) have shown that northern Europe will experience more precipitation on average, while in a large band stretching from France to the Black Sea and beyond, summer precipitation is projected to diminish by as much as 40%. Simultaneously, many regional climate models point to a strong increase of short-lived but very intense precipitation events in certain regions that are already prone to such hazards, such as parts of central Europe, the Alps, southern France, and northern Spain. In Switzerland, simulations of climate forced by high greenhouse-gas emissions for the period 2071–2100 compared to the reference 1961–1990 period shows a marked shift of the seasonality of mean precipitation (e.g., Beniston 2006), with strong increases in winter and spring, and substantial reductions in summer and fall, as seen in Figure 3 for four different regional climate model simulations (the Danish HIRHAM; the Swiss CHRM; the Italian ICTP; and the Swedish RCAO models); while the absolute value of change differs from one model to another, all models agree on the seasonal sign of change. The principal cause of these changing patterns is related to the strong summer warming and drying in the Mediterranean zone that would also affect the Alps and regions to the north, and the enhanced winter precipitation that a milder climate may bring to the region, rather under the form of rain than of snow. As a result of the change in mean precipitation, the frequency of extreme rainfall

Figure 3. Percentage change in average seasonal precipitation for the control and scenario climates as simulated by 4 different regional climate models; see text for further details.

events also changes in seasonality compared to current climate. Model simulations suggest that springtime extremes will increase the most, while summer events may decline by as much as 50%. However, when heavy precipitation occurs in a warmer climate, it may be even more intense because of the additional energy provided by a warmer atmosphere than today. In terms of the potential for floods, natural hazards and damage that the changes in means and extremes of precipitation may trigger in the Alps, it should be emphasized that heavy precipitation is a necessary but not sufficient condition for strong impacts. For example, Stoffel and Beniston (2007) have shown that while debris-flows of the past have occurred mostly during wet summers, it is conceivable that in a greenhouse climate the frequency of such events could decrease because of the shifts in the occurrence of extreme precipitation from summer to spring or fall by 2100. The response of slopes and watersheds to high precipitation levels varies from one event to another for a number of reasons, in particular the prior history of precipitation, evaporation, permeability of soils and the buffering effect of snow during an event. That may lead to decrease at mean altitude. The more elevated the freezing level, the greater the potential for strong runoff and high intensity of erosion since there is a larger surface area upon which water can flow off the slopes. Under current climate, the most intense events are observed to occur during the summer months, where the freezing level is higher than 3,500 m above sea level. In a future climate, on the other hand, the freezing level associated with heavy rainfall will be on average 500 m lower because many events will take place either in spring or in autumn, at a time of the year where conditions will be cooler than for current summers. However, even if their frequency is likely to decrease, the magnitude and impacts of future summertime debris flows, mudslides or rock-falls could be greater than currently because of warmer temperatures and higher precipitation intensities.

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3

TYPES OF LANDSLIDES AND SLOPES LIKELY TO BE AFFECTED BY CLIMATE CHANGE

It is evident that all kinds of shallow slides and improperly drained engineered slopes, as well as potential debris flow creeks are likely to present a higher hazard level in the future, as one of the major characteristics of climatic conditions in the 21st century is to display more intense storms occurring probably with an increased frequency. Another reason for this increased hazard is the always extending area of impervious zones in the concerned watersheds, due to the development of urbanization and roadways. This situation causes higher peak floods in streams or excessive discharges in inappropriate sewage or drainage ducts that are likely to divert sudden flows at the surface of slopes through manholes or pipe failure and thus generate destructive mud flows, if specific retention works are not foreseen. In the case of high mountain slopes, the increasing melting of the permafrost zone due to higher summer temperatures can also be a cause of unexpected debris flows, even outside of a rainfall event. The loosened fan material at the toe of mountain cliffs provides more sediments for the debris flows that can generate more extensive impacts in the valley floor; this critical situation is especially due to the intense tourist development of chalets in zones providing space with a view and easy access. In all these cases there is a nearly simultaneous occurrence between the storm triggering the landslide and the development of the landslide process leading to severe visible impacts on buildings or agricultural land, so that there is no doubt about the relation between the climatic conditions and the consecutive damage. However in the cases of larger slides such a correlation is not evident to demonstrate and a nearly similar rainfall pattern can cause a slight increase of the movements of a slide during one winter and a severe crisis during the next winter, as it was observed and monitored at La Chenaula landslide in Switzerland in 1982 and 1983 (Noverraz & Bonnard 1992). In the case of large to very large slides, extending on an area of one to several km2 , with differences of elevations between scarp and toe that can reach several hundreds of meters, the situation is even more complex, as several factors may influence the reaction of the landslide mass, namely: – variation of rainfall amount with elevation – offset of snowmelt episodes along the slope – capacity of snow cover to absorb a large amount of rainfall before infiltration occurs – variation in vegetation implying different interception and evapotranspiration patterns

Figure 4. Relationships between horizontal displacements and computed flux entering the La Frasse landslide in a cumulated representation. Thin line: infiltration issued from COUP model (considering temperature, snowmelt, vegetation, soil, sun exposure . . . ). Thick line: weighted and truncated infiltration over several years in the past. The best fit of the thick curve with the displacements (grey area) is concluded from the shape of the curves rather than from the value of the corresponding points.

– distribution of permeability fields at the surface of the slide and in the landslide mass – possibilities of differentiated infiltration along streams flowing on the slide as well as in cracks – occurrence of floods in the stream flowing at the toe of the slide and likely to cause erosion – development of urbanization and collecting water pipes and ducts – differences of reaction of the landslide mass along the slope (swelling and depletion zones). In such cases it is also necessary to consider the effect of climatic impact not only for one specific event, but for a crisis period that can be caused by antecedent conditions extending to several years before the observed crisis. In the case of La Frasse Landslide, in Switzerland, it has thus been shown (Tacher et al. 2005) that several factors like interception of modest rainfall and evapotranspiration had to be considered in order to express a good correlation between infiltration flow and movements (Fig. 4). It is then necessary to analyse the changing climatic situation through the formulation of scenarios implying different episodes along the yearly cycle. It is of course essential to gather enough long-term data on the movement of the slide as well as continuous displacement information on so as to understand its global behaviour and to confront the observed and modelled displacement vectors. 4

LONG-TERM MONITORING OF THE IMPACT OF CLIMATE CHANGE ON LANDSLIDE BEHAVIOUR

In order to determine the possible effect of climate change conditions on the behaviour of large slides,

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it is first necessary to gather available data providing information on their long-term movements. Such an investigation has been carried out in Switzerland within a national research project (PNR31), considering a dozen of very large slides, extending over areas from one to some 40 km2 , and for which ancient geodetic survey data had been collected (Noverraz et al. 1998). In the specific case of Lumnez Landslide, in the Canton of Graubünden in Eastern Switzerland, the position of the spires of 7 village churches has been regularly monitored for more than a century (up to 17 monitoring campaigns beginning in 1887). The results showed in general a very constant average velocity, varying from 3 to 20 cm/year (Fig. 5). Only one point located near the toe of the slide (village of Peiden) displayed a clear reduction of velocity after the years 1940, which can be partially explained by a series of dry years and then by the construction of a dam on the river Glenner flowing at its toe, upstream of the slide. As far as the annual rainfall is concerned, the long-term trend is not so marked in this region (average value of 950 mm/year—Figure 5) as in western Switzerland, in which a clear increase of annual rainfall by some 10% has been observed since the years 1980 (this fact has induced the Swiss hydrological service to change the long-term reference rainfall value from the period 1901–1960 to that extending from 1961–1990 for all rain gauge stations). It can thus be observed that most of the very large slides monitored display a fairly constant velocity even if the are affected by long periods of higher precipitation. This fact can of course be due to their size and depth, inducing a certain mechanical inertia, as well as to the large storativity of their hydrogeological conditions. But this observation is not necessarily valid for all slides.

Figure 5. Cumulated displacements of several topographic points at Lumnez landslide over a period of more than one hundred years.

In a few cases, indeed, some monitored points of these large slides have displayed an acceleration phase that may last from several months to a year, like in the case of La Frasse landslide, for which crisis periods of a few months duration have been recorded in 1966, 1982–83 and 1993–94 (Tacher et al. 2005). In most of the duly monitored slides, this major acceleration phase does not imply the whole landslide mass, but a part of it, generally located at its toe or eventually in an area in which the depth of the slide is reduced. Such a situation was clearly put forward in the case of Chlöwena Landslide in Switzerland, in 1994: the crisis lasted for 4 to 5 months, with a peak velocity of 6 m/day at the end of July that was reduced to a few cm/year at the end of September (Vulliet & Bonnard 1996). A comprehensive approach of such complex phenomena therefore requires first a long-term monitoring and then a detailed modelling in order to understand the hydrogeological and geomechanical conditions that explain the crisis episodes. In a second step it is possible to determine the probable effects of climate change in a quantitative way, considering several crisis scenarios. Such an approach has been applied to various slides and in particular to the Triesenberg landslide.

5

MODELLING THE CLIMATE CHANGE CONDITIONS : THE TRIESENBERG LANDSLIDE

The Triesenberg landslide extends over a significant part (i.e. 5 km2 ) of the Principality of Liechtenstein (160 km2 ), located to the East of Switzerland (Figure 6). It also includes two villages, Triesen at its toe and Triesenberg at mid-slope, the infrastructures of which incur occasional damage, in particular during crisis episodes. The movements of this landslide are quite ancient and date back to the end of the Wurmian period; presently they are generally slow (i.e. some mm/year to cm/year) in normal conditions and locally may reach velocities of a few dozens of cm/year during severe crisis periods. As the slide displays a relatively slow movement, many buildings have spread on the slope in particular during these last decades, due to the real estate development. The objectives of the research were the following: – to determine the critical hydrogeological conditions that cause an acceleration of the slide and that may justify the triggering of an alarm system; – to foresee the behaviour of the slide under the possible effect of climate change, so as to establish bases for the sustainable development of the slope.

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Figure 7. Geological model of the Triesenberg landslide. The draped topographic map is lifted to display the geological units. Figure 6. Location of the Principality of Liechtenstein and of the Triesenberg landslide.

The first aim of the models developed does not consist in determining the possibilities of stabilizing the overall slope, as it can be expected that such works would by far pass the planned investments by the authorities of the Principality. What is aimed at is to live with the slide, and not to slow it down. The specific difficulties presented by the Triesenberg landslide mainly refer to its large area, to its essentially unsaturated hydrogeological conditions and to the slow movement velocities. After calibrating the model parameters with respect to the crisis of 2000, the impact of climate change has been analysed by modifying the boundary conditions of the hydrogeological model, on the basis of the relevant climatic scenarios, as set up by the Swiss Commission for the assessment of climate change (OcCC, 2004). Then the respective computed groundwater pressures have been introduced in the geomechanical model, as it was done for the year 2000. 5.1

Main features of the landslide

5.1.1 Morphology and geology The slope is oriented from North-East (up) to SouthWest (down). It presents some small undulations but is generally fairly regular. Based on a digital terrain model, the mean slope is 24◦ (Figures 7 and 8). Three parts are distinguished: – In the upper part, deep-seated slope movements occurred, probably at the end of the Wurmian glacial retreat (14,000 years); they are now underlined by a terrace in the topography at the top of the slope (Figure 8) and were triggered by a deep landslide,

Figure 8.

Geological vertical cross-section.

the so-called Prehistoric landslide. This upper zone, largely inactive, is not considered to cause a driving force on the slope. Approximately, it covers 1.7 km2 with a volume of 74 million m3 . – The prehistoric landslide is known by some boreholes. It is more than 80 m deep and is made of flysch (clayey shales). This zone is today stabilised; moreover, no movement has been observed at the toe of the landslide, where it lies under the Rhine river alluvia (gravels). – The active landslide (Table 1) covers the prehistoric one. It is also composed of flysch and takes place on a slip surface located at an average depth of 10–20 m. According to the inclinometer data, available from 1995 to 2002, the slip surface is approximately one meter thick. The analysis of the observed intensities and directions of the movements (Figure 9) showed that the area is indeed composed of three instability zones that can be considered as independent. This is confirmed by the reduced depth of the slip surface close to the assessed boundaries of the three areas and by the spatial distribution of damage to infrastructures (Frommelt AG 1997). The practical consequence of the decomposition of the landslide into three distinct systems is to allow defining three different modelling areas for the 3D

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The yearly observation of displacements shows a close dependence of the movements on the seasons. A reactivation is generally perceived in the spring, which corresponds to the snowmelt period. This indicates that the main driving force of the movements is the variation of pore water pressure in the slope. However, a reactivation may also occur following a storm event. The tectonic Arosa zone (Fig. 8) is a very important feature of the hydrogeological system due to its low permeability: a part of the Valüna Valley groundwater (Fig. 7) flows on the Arosa zone and feeds the basal surface of the landslide, causing the Triesenberg groundwater basin to be much larger than its topographic watershed. This mechanism is proven by several observations (Tacher & Bonnard 2007). Such a double feeding is also effective outside intensive infiltration periods. Both a hydraulic balance of the Triesenberg slope (Bernasconi 2002) and a numerical model calibration suggest that about one half of the inflow in the landslide is supplied by a base flow from the Valüna Valley through the sandstones covering the Arosa zone (ca. 9 mio m3 /year). Groundwater discharge occurs through some one hundred springs distributed over the landslide, as well as at its toe, in the River Rhine alluvia. The water table is located about 20 to 30 m below the soil surface at the top of the landslide, whereas at the bottom, it almost reaches the ground surface.

Table 1. Main features of the Triesenberg active landslide. Aspect

Characteristics

3.1 km2 min. 460 m, max. 1500 m a.s.l. 2300 m 1500–3200 m 10–20 m 37 millions m3 24◦ 0 to 3 cm/year Flysch (clayey shales) including elements of limestone and sandstone Vegetation Pasture land and some wooded zones Investigations Hydrogeology, boreholes with inclinometers, GPS, RMT geophysical methods, laboratory tests, modelling Possible damage Infrastructures of two villages Area Altitudes Length Width Mean depth Volume Mean slope Mean velocity Soil

5.2

Figure 9. Total displacements between 1976/1981 and 1996/1997 (i.e. during some 20 years) in the whole instability area, and boundaries of the three so called independent landslides.

mechanical modelling, supposing negligible kinematic interactions between them (the hydrogeological model involves a single regional area). 5.1.2 Hydrogeology Besides supplying the geomechanical models with hydraulic pressures, the hydrogeological analysis and modelling aims at a better understanding of the particular hydrogeological behaviour of the slope.

Hydrogeological modelling of the year 2000

The year 2000 was chosen to perform the modelling; during this year, a critical phase with a reactivation of the movements was observed, showing a good correlation with the snowmelt phase in April. A violent thunderstorm also occurred on August 6th. This year has a return period of the annual rainfall of 42 years, according to the Gumbel law. Another reason of this choice is the availability of calibration data for both hydrogeological and geomechanical models. As the slide is very thin, the unsaturated zone is of relatively high importance, which justifies computing groundwater flows in unsaturated regime, i.e. the flows are governed by Richards’ equation (Hillel 1980). From the model results, in terms of volumes, the direct infiltration reached 7.52 mio m3 in 2000, while the inflow through the Arosa zone was about 9.86 mio m3 (Figure 10). The cumulated rate of the springs reached 1.06 mio m3 , which represents only a few percent of the total outflow; the balance flow seeps in the Rhine river alluvia. The outflow curve is smoother than the inflow events, due to the capacitive function of the landslide. Typically, the August 6th storm response was absorbed and delayed. In May, the snowmelt episode did not lead to spectacular changes in the hydraulic

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balance because, due to the slope topography, the melting occurred progressively from the bottom to the top. For example, when the snowmelt occurred in the Valüna Valley, it had already finished on the landslide several days to weeks earlier. More relevant from the geomechanical point of view is the piezometric behaviour. It is illustrated by piezometer B8 (Fig.11). The respective calibration was carried out by comparing the water table data with the hydraulic head computed at the bottom node at this site, i.e. at the slip surface. Both main events of the year 2000 led to a peak more than 2 m high. Just after these peaks, the head decrease was slower in the model than in the reality. This can be explained by the relative smoothing of the parameter field, mainly over depth. Heterogeneities are also responsible for another observation: during the snowmelt event, the model reacted with a delay of some days with respect to

Figure 10. Hydraulic balance of the model for the year 2000. Thin line: Feeding by the Arosa zone. Dot line: Direct infiltration on the slope. Thick line: Outflow rates through springs and the River Rhine valley.

Figure 11. Measured and computed hydraulic heads in piezometers B4 and B8. Thin line: B4 measured. Thick line: B4 computed. Thin dotted line: B8 measured. Thick dotted line: B8 computed.

the monitoring data. Local pervious heterogeneities that were not considered in the model accelerated the piezometer response to inflows in the Valüna Valley. Such a delay did not occur at the beginning of August since inflows concern both the Triesenberg and Valüna basins. The numerical results suggest that the model globally fits with reality, despite a simplification of the parameter fields, a rough estimation of the unsaturated parameters and a minimal knowledge of the real hydraulic balance. Computed hydraulic pressures are thus suitable as an input in the hydro-mechanical models in order to describe the direct causes of the movements during crises.

5.3 Geomechanical modelling of the year 2000 The effect of the hydraulic head variation with time, as determined by the hydrogeological modelling, on the mechanical behaviour of the whole slide, has been modelled by a FE code, Z-SOIL (2-D and 3-D), using a Biot-type formulation, implying the conservation of mass and momentum of both fluid and solid phases (François et al. 2007). In the 3-D model, the maximum displacement values are in general slightly lower, but they appear within zones where damage has been reported (Fig. 12). Parametric studies have also been carried out to evaluate the effect of the selected friction angles (between 30 and 21◦ ) and of the range of water pressure variation (the computed data through the hydrogeological model were multiplied by 1.25 and 1.5 respectively). Both simulations display nearly linear variations and prove that, even in extreme conditions, it is not expected that the movements will lead to a catastrophic behaviour of the whole slope.

Figure 12. Distribution of the obtained displacements after 291 days of simulation, from January 1 to October 18, 2000 (water pressure data multiplied by 1.25) and location of the more active zone in the central slide (3 cm).

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5.4

Modelling of Climate change impact and related uncertainties

According to (IPCC 2007), the air temperature should increase in the medium term, especially in summer, and the rainfall should increase in winter, but decrease in summer. The climatic scenario for 2050 used in this study is issued from the Swiss ‘‘Organe consultatif sur les Changements Climatiques’’ (OcCC 2004), more specific to the North of the Swiss Alp context (Fig. 13). According to this scenario, it can thus be expected that: – In winter, the total infiltration would increase and rain would partly replace snow accumulation. On the other hand, snowmelt at the beginning of the spring would be less important. – In summer, the storm events would remain similar, if not slightly worse, but the total infiltration would be smaller than today because of higher evapotranspiration. In this study, the rainfall in 2050 is considered to increase by 2 mm/day in winter and to decrease by 2 mm/day in summer. Those values are added as a one year sinusoidal transformation to the records for the year 2000. Similarly, the temperature curve for 2050 is obtained by adding a one year sinusoidal function to the records of the year 2000, considering a warming of 1.5◦ C in winter and 3.5◦ C in summer (Figure 14). Considering the impact on landslides, such a scenario is not obviously more severe, mainly for the landslide zones in altitude. Indeed, besides the total infiltration, the groundwater pressure fluctuations have a major effect on the movements. By diminishing the rather massive infiltration period of snowmelt, the 2050 scenario smoothes out the groundwater head curve at spring time. In particular in the Valüna valley, the fast snowmelt at the beginning of May 2000 might be replaced by a succession of less important episodes of rain, falling on a thin accumulation of snow.

Figure 13. Climatic scenario for 2050 for the North of the Alps, after OcCC. Horizontal axis: months of the year. Black bars: Rainfall change in % with respect to present average seasonal values. Grey bars: Temperature change in ◦ C.

The target of the models is here to consider the most unfavourable scenario as far as the landslide movements are concerned. Thus these worst case infiltration conditions for 2050 are as follows, even if they are not the most plausible: – No consideration of the decreasing of gross rainfall in summer. The infiltration curve is left intact from May 1st, – Keeping the snowmelt event of the end of April, – In winter, adding infiltration periods without decreasing the accumulated snow height. In practice, the 2050 infiltration scenario implies to add infiltration days between January 1 and April 20 to the year 2000 conditions. For all altitude classes of infiltration, a 5 mm/day event is introduced each ten days (Fig. 15). This represents an additional infiltration of 55 mm/year.

Figure 14. Climatic scenario for 2050 in the Valüna valley. Upper curves: Temperatures in 2000 (solid line) and in 2050 (dotted line). Lower curves: Gross rainfall in 2000 (solid line) and in 2050 (dotted line).

Figure 15. Scenario 2050. Infiltration conditions for altitudes below 625 m.

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The results of both hydrogeological and geomechanical models with such modified boundary conditions are very similar to those obtained for the year 2000. Typically, the hydraulic heads in piezometer B8 (Fig. 11) are changed by some centimetres only. Considering uncertainties on climatic changes, the modelled scenario appears to be probably the worst case. In such conditions, the computed velocity field for 2050, if the parameter calibration on year 2000 is considered as reliable, is a rather pessimistic global assessment of the slope. However, the transition to a stormier climatic regime may have local consequences (hectometric slides, mudflows) not considered in this regional modelling.

6

CONCLUSIONS

A detailed hydrogeological and geomechanical modelling as it was recently applied at the Triesenberg and La Frasse landslides allows a significant modelling of large landslide movements during crises, provided sufficient information is available. The application of predicted climatological conditions in the future then supplies quantitative values of possible movements, considering appropriate scenarios. However, extremely rare conditions with a very remote probability cannot be modelled reliably, as the boundary conditions may significantly differ from the ones considered in the original model. The analysis of several large landslides in other contexts (Bonnard et al. 2004) also shows that the effect of climate change on landslides within the next 50 years or so must not be overemphasized. Indeed, as shown here, the progressive snowmelt that will begin earlier than before tends to reduce the occurrence of critical situations in the spring or summer. On the contrary, it is clear that the expected increase of storm intensity, as foreseen by some climatologists, may produce more violent and frequent small slides and debris flows; but this specific prediction is not relevant for large landslides and cannot justify a development of more severe disasters related to this type of phenomena. Indeed, due to the heterogeneity of the material at a large scale, to the increased range of altitude where infiltration occurs, to the capacitive function of the landslide mass and to the more complex hydraulic relationships with the bedrock, the response to climatic events may be significantly smoothed and delayed, which explains this relatively optimistic vision.

ACKNOWLEDGMENTS The authors wish to thank the authorities of the Principality of Liechtenstein for supporting this research,

Dr Riccardo Bernasconi geological office for supplying data and advice, as well as all the colleagues who participated to the modelling.

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Seneviratne, S.I., et al. 2006. Land-atmosphere coupling and climate change in Europe. Nature 443:205–209. Stoffel, M. & Beniston, M. 2006. On the incidence of debris flows in the Swiss Alps since the early Little Ice Age and in a future climate. Geophysical Research Letters, 33, L16404. Tacher, L., Bonnard, C., Laloui, L. & Parriaux, A. 2005. Modelling the behaviour of a large landslide with respect to hydrogeological and geomechanical parameter heterogeneity. Landslides Journal 2 (1):3–14.

Tacher, L. & Bonnard, Ch. 2007. Hydromechanical modelling of a large landslide considering climate change conditions. Lecture at International conference on ‘Landslides and Climate Change—Challenges and Solutions’. Ventnor, Isle of Wight, UK. 21–24 May 2007. Vulliet, L. & Bonnard, Ch. 1996. The Chlöwena landslide: Prediction with a viscous model. Proc. VIIth Int. Symp. on Landslides, Trondheim Vol. 1:397–402.

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Geology, geotechnical properties and site characterization

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Geotechnical appraisal of the Sonapur landslide area, Jainita hills, Meghalya, India R.C. Bhandari, P. Srinivasa Gopalan & V.V.R.S. Krishna Murty Intercontinental Consultants and Technocrats Private Limited, New Delhi, India

ABSTRACT: The occurrence of landslides particularly on cut slopes along the roads with in Jainita hills in Northern Eastern part of Himalayas of Meghalaya state in India are common features. These slope failures causes considerable loss of life and property along with many inconveniences such as disruption of traffic along highways. The paper present deals with geotechnical synthesis of slides as per recommendations of RHRS system and stabilization measures suggested. The landslide is located on North South trending ridge on eastern bank of river Sonapur. The rocks involved in sliding are highly jointed sandstone shale of Oligocene series. The total inclined length of the affected slopes is around 800 meter. The jointing in rock is attributed to the nearness of the area to a major thrust. These studies indicate the failure along the slopes is ‘‘Rock fall-cum debris flow’’. 1 1.1

INTRODUCTION Occurrence of land slide

October. The month of June and July experiences maximum precipitation. The annual precipitation varies from 4000 mm to 8000 mm during the year maximum precipitation of 1200 mm is recorded in month of June and July. Maximum temperature reaches to 24◦ C in month of September while minimum temperature recorded is 10◦ C in month of February.

A land slide occurs when due to gravity forces the rock/soil mass moves down wards due to heavy precipitation, run off or ground saturation. The flow occurs generally during period of intense rainfall, on steep hill slopes where the rocks are tectonically disturbed. The flow/fall from many different sources can combine in channels, and their destructive power is greatly increased. They continue flowing down hills and through channels, growing in volume in addition of water sand, mud, boulders, up rooted trees and other material. When the flows reach flatter ground, the debris spread over a broad area thus affecting the considerable length of road.

2

1.2

3

Location of area

The occurrence of land slides, particularly along the road in the Jaintia hills of Meghalaya state is common feature during rainy season. This cause considerable loss of life and property along with many inconveniences such as disruption of traffic between Shillong and Silcher every year. Sonapur land slide occurring near vicinity of township and is located between KM 141.100 to KM 141.350 on Shillong—Silcher highway. 1.3

Climate

The climate of the area is moderate. The rainfall occur during monsoon period which extends from May to

REGIONAL GEOLOGY OF THE AREA

2.1 Rock types The rock types in the region available, varies in age from Archeans to Tertiary the general sequence of the rocks available in the area is as follows (Figure 1).

3.1

LOCAL GEOLOGY AND GEOMORPHOLOGY Geology

The Sonapur landslide is located on North–South trending ridge, on east bank of river Sonapur/Lubah. The river Lubah flows at elevation 33.0 m from mean sea level, and the present road elevation 44.0 meter. The highest elevation at top of the crown of slide is 481.25 meters. Thus the total height affected by slide along the slopes are also observed slopes is around 440 meters. The rock exposed in the area are sand stones and shales of Oligocene age. The rocks are highly jointed and local folding along the slopes are also observed.

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4

Table 1. Regional geology. Rock types Feldspathic sandstones, pebbles and conglomerates Shales Sandstone and marls (Kopiliformation) Limestone and Sandstone (Sylhet formation) Gneissses and granites

Group

Age

Brail

Oligocene

Garo

Miocene

Jaintia Gniessic complex

Eocene Archeans

HISTORY OF SLIDE

For information see table on next page. 5 5.1

STRUCTURAL ANALYSIS Geometry of slide

The geometry of the slide is as under. (For plan and section of land slide area please refer to Figures 2 and 3). The area is being drained by two prominent drainages. The drainage, which is located on southern side is being continuously recharged through minor

Figure 1. Regional geological map of the area. Figure 2.

Geological plan of slide area.

Figure 3.

Section along the central line of slide area.

3.2 Geomorphology The area is occupied by undulating topography characterized by presence on hills and valleys. The area is dissected by number of streams and network of their tributaries. The drainage pattern in the entire district represent a most spectacular feature revealing extra ordinary straight course of rivers and streams, evidently along master joints and faults which are impressions of major geological activity in the area. The magnificent gorges scooped out by the river in the southern part on Jainitia district are result of massive headword erosion by anticedent streams along joints of sedimentary rocks, exposed in the area.

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233

Rolling down of boulders from top 200 m above road level along slopes

Loss of cohesion in soil along slopes due to heavy rains

Rolling down of boulders from top 200 m above road level along slopes

2000 Aug–Sep

2001 01 to 15 June

2004 July

1999 June–July

Rolling down of boulders from top 200 m above road level along slopes

i) Loss of cohesion in soil along slopes due to high precipitation ii) Earth quake of 1987, possibly resulted loosening of shear strength between jointed blocks Flow of debris material from higher slopes

1998 July

1989 September

Possible cause

History of slide.

Date of occurrence

Table 2.

Construction of culvert Retaining wall, Wire crated wall with jute mesh

Total number of incidence occurred. Four road was closed for 5–6 days. Size of boulders 2 × 2 × 3 and 7 × 10 × 3

High-magnitude slide, Boulders of 3.75 m × 2.5 m × 8.8 m rolled down along slopes Road blocked for 10 days one oil tanker, truck with loaded goods, one passenger bus buried down. Number of casualties not known

Land slide frequently occurred. The road was blocked for total period of 22 days

Road blocked 2–3 days Land slide of less magnitude but size of boulders was 5.5 m × 1.8 m × 4.5 m

Clearance of debris Diversion of water at road level ii) Repairs of gabion structures at road level Clearance of debris Diversion of water at road level ii) Repairs of gabion structures at road level Construction of Retaining wall, Check wall water chute Concrete pavement and geotextiles

Intensity was less; road was blocked daily for 1–2 hours for 15 days

Slide of high intensity Road closed for more than one month

Remarks

Clearance of debris and maintenance of remedial measures taken were same as in 1998

Clearance of debris Construction of gabion walls and road level diversion the drain

Remedial steps taken

6

Table 3. Geometrical parameters of slide. Base—Width at road level

300–350 m

Width at center Crown—Width Inclined Length (from River Bed to Crown of slide) Slope from river bed level to point A Slope from Point A to B Slope from point B to C (Crown)

180 m 70 m

The geological structures of the study area are characterized by strongly jointed rocks of variable strength mainly sand stone, and shales of Oligocene age. The synthesis of geological date indicates that the slope stability problems in the area are associated with surcharging of slopes during heavy precipitation and with inflow of ground water from fissures. A great loss of strength of the rock mass results, particularly in zone of weathering and causes its subsequent displacement along slopes.

800 m (approx) 35◦ 25◦ >60◦

Table 4. Wedge analysis. Amount and Relationship with Wedge Intersection of direction of slope stability (on No. Joint set plunge southern face) 1

1 and 2

2

2 and 3

3

1 and 3

SLOPE STABILITY ANALYSIS AND ROCK FALL MECHANISM

1. The dislodged blocks from higher reaches rolls down along the slopes are mixed with debris and moves down along slopes, thus forming the slides as ‘‘Complex Slide’’. The schematic geological plan and sections are shown as Figures 2 and 3 respectively. 2. The slope analysis is divided in two parts

50◦ S 50◦ W Forms slope most unstable as the axis of wedge plunge towards open face 44◦ : S 33◦ W Forms unstable slopes as axis of wedge plunge towards face by angle of 4′ 12 cubicyard

81 27

High precipitation period or continual water on slope Many falls

27

9

9 9

27 369

less than 300 are assigned, a very low priority where as the slopes with rating excess 500 are identified as areas requiring urgent remedial measures. Since in the present case the total rating is more than 300 but less then 500, the area falls under category requiring long term planning for slope stabilization after detailed geological investigation. 8

REMEDIAL MEASURES ADOPTED

The following remedial measures have been adopted so sorto control the damage at road level time to time.

STUDY OF AREA AS PER ROCK FALL HAZARD SYSTEM (RHRS)

The Highway and railway construction in mountainous regions present a special challenge to engineers and geologists. The rock fall hazard rating system ‘‘RHRS’’ was developed by Pierson et al 1990. The following analysis of Sonapur slide as has been made as per Rock Fall Hazard Rating System. The RHRS system does not include recommendation on actions to be taken for different rating this is because decisions on remedial action for specific slope depends upon many factors such as budget allocation for highway work. However slopes with rating

1. 2. 3. 4.

Construction of retaining wall at road level Construction of culvert at road level Diversion of drain at road level Construction of check dams across perennial drains along slope. 5. Concrete pavement 6. Construction of chute along the slope To keep the road free for vehicular movement through-out the year, as a part long term planning the road maintenance Engineers are proposing construction of twin landslide gallery, Alternatively construction of approx. 500 m long tunnel by passing

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slide—zone has also been recommended as a measure to the traffic smoothly by passing the slide zone. 9

CONCLUSION AND RECOMMENDATIONS

The rock falls, are due to falling of blocks of sand stone/shale from top i.e. near crown area of slide. Which are generated due to intersection of joints, with in sand stone and shale. The pre-existing old landslide debris material lying from river bed level to a height of around 200 meters along slopes is mixed with clay and boulders. The loose blocks resting on saturated slopes increases the load on soil. Under these circumstances when the soil gets saturated with water, looses its cohesion and angle of internal friction added with increase of overlying weight, starts flowing with velocity along slopes. When such material with great momentum hits out at remedial structures such as retaining wall, breast wall, water chute concrete pavement and Geo-textile causes complete damage of these structures. The studies carried out earlier has suggested construction of ‘‘Twin Slide Shed’’ at the road level, allow to pass the debris material over it, alternatively construction of approximately 500 meter long tunnel has been also recommended. By studying the survey data and detailed section of land slide area. The protection measures which is appears to be more appropriate are

i. Construction catch pit at RL ± 200 m of (bigger dimension 25 × 10 × 5) to arrest the movement of water, boulders coming from up slopes ii. Construction of cascading chute from catch pit to road level along perennial drainage. iii. Construction of gabion wall at road level. iv. Construction of culvert at road level. v. Construction of chute from Invert level of culvert at road level to river bed for an inclined length of 50 m. ACKNOWLEDGEMENT Authors are thankful to Chairman and Managing director ICT Shri. K.K. Kapila for encouraging the authors to write the manuscript of this paper. REFERENCES Hoek & Bray J.W. 1981. Rock slope engineering. Revised third edition. The institution of Mining and Metallurgy, London. Pierson L.A., Davis S.A. & Van Vickle R. 1990. Rock fall hazard Rating system—implementation manual. Federal highway administration (FHWA) Report FHWA OREG 90-91. FHWA Department of transportation.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

The viscous component in slow moving landslides: A practical case D.A. González, A. Ledesma & J. Corominas Department of Geotechnical Engineering & Geosciences, Tech. University of Catalonia (UPC), Barcelona, Spain

ABSTRACT: The availability of continuous records of both velocities and groundwater table in monitored landslides has increased the interest of the scientific community for the dynamics of the slow moving landslides. In this paper we analyse the role of the viscous component in large landslides by using ten years record of monitoring data from the Vallcebre translational landslide, located north of Barcelona (Spain). Previous research showed that a viscosity term should be considered in order to reproduce the measured displacements in this landslide. We discuss here the hysteretical behaviour of the landslide velocity records observed during some acceleration events produced by the rise and withdraw of the groundwater table. The conceptual model we have used shows that a constant viscous component of the movement is not able to explain the hysteretic behaviour and that consequently, other mechanisms should be searched to explain such a behaviour.

1

INTRODUCTION

In many slow moving landslides a close relationship between landslide velocity and position of the groundwater level has been observed. Acceleration changes in these landslides are usually controlled by water table. Despite this evidence, research works describing in detail the dynamics of such kind of relations are still scarce. This is partly due to the difficulty in the past to measure, in a continuous manner, both the velocities and the position of the groundwater table. Nakamura (1984) showed for a particular landslide in Japan that the amount of landslide movement in the rising limb of groundwater level is larger than that observed in the lowering limb for the same groundwater level. Figure 1 shows that points indicating the groundwater level at the time of measurement line up in a circle counterclockwise with time. A number at the side of each point indicates the month, and ‘‘e’’, ‘‘m’’ and ‘‘l’’ indicate early part, middle part and the last part of the month, respectively. The author compares the velocity of the landslide (amount of displacement in 10 days) in different events for one year. The landslide response is influenced by the different soil conditions and particular characteristics of each event. Bertini et al (1984, 1986) and Picarelli (2004) performed a similar analysis, based on measurements in the Fosso San Martino landslide (Italy) suggesting different velocities for the rising and lowering limb of the piezometric levels (Figure 2). Van Asch et al (2007) coincides with the latter authors saying that for a given groundwater level,

Figure 1. Groundwater level in relation to amount of landslide movement for a landslide in Japan. After Nakamura (1984).

velocities are higher when water table is increasing than when it is decreasing. In this paper we analyse the relationship between groundwater level changes and the landslide displacements and velocities using data from the Vallcebre

237

been carried out every 20 minutes. Piezometric readings have indicated that changes in groundwater levels occur quickly. In-hole wire extensometers have recorded sudden changes in displacement rates that can be directly related to the fluctuations of the water table which is governed by rainfall (Corominas et al 1999). The wire extensometer measurements show that the landslide has never stopped completely. It has been moving since we started the continuous monitoring in November 1996, although velocities slow down significantly during dry periods (Corominas et al 2000). On the other hand, the history of displacement of the extensometers reflects that different parts of the landslide mass move synchronically but with a different rate of displacement.

Figure 2. Displacement rate versus piezometric level for Fosso San Martino landslide. After Bertini et al (1986).

2.3

landslide. The analysis is restricted to acceleration events occurring in the period between November 1996 and August 1998. To check whether these results are consistent, finally a basic conceptual model is presented as well. 2

THE VALLCEBRE LANDSLIDE

2.1 General setting The Vallcebre landslide is a large, active slope failure located in the upper Llobregat river basin, in the Eastern Pyrenees, 140 km north of Barcelona, Spain. The landslide is situated on the western slope of the Serra de la Llacuna. The mobilised material consists of a set of shale, gypsum and claystone layers of continental origin gliding over a thick limestone bed, all of which are of Upper Cretaceous—Lower Palaeocene age. The dimensions of the slide mass are 1200 m long and 600 m wide. The entire landslide involves an area of 0.8 km2 that shows superficial cracking and distinct ground displacements. The toe of the landslide extends to the Vallcebre torrent bed, and is pushing it towards the opposite bank. As a result of this, the Vallcebre torrent has been shifted to the west more than ten meters and the foot of the landslide has overridden the opposite slope to form a back tilted surface. The torrent undermines the landslide toe during floods, causing erosion and local rotational failures which decrease the overall stability. A comprehensive description of the landslide is found in Corominas et al (2005). 2.2 Monitoring of the landslide Since 1996, systematic recording of rainfall, groundwater level changes, and landslide displacements has

Hydrological changes and landslide response

The data show that groundwater reacts almost immediately to rainfall inputs, suggesting that water infiltration is controlled by fissures and pipes rather than by soil porosity. The role of the karstic network in the gypsum lenses is unclear but all the observed field features are very shallow (up to 3 m depth), which is well above of the normal groundwater level fluctuation. Because of this, we have assumed that karstic network (piping) play only a secondary role. A close relationship between the groundwater level changes and landslide activity was observed at borehole S2 (Figure 3). There exists a strong level of synchronism between the two records. Figure 4 shows an interesting relationship between observed velocities and the depth of water table at borehole S2 for the period considered. A cubic curve may be fitted to the data. These data are going to be analyzed thoroughly by considering the landslide acceleration events separately. Each event is defined by the rising and lowering groundwater level limb. Note in Figure 4, that when depth of groundwater table is close to 6 m, velocities tend to be nil. That is, there is a level of water table below which landslide stops. In a previous analysis (Corominas et al 2005), we have considered that beside frictional resisting forces, additional resisting forces (i.e. viscous forces) were necessary to explain the rate of displacement of the landslide. 3 3.1

ANALYSIS OF THE DATA Field data observations

The data shown in Figure 4 correspond to different rainfall events occurred in the period considered without discrimination whether the groundwater level is in a rising or lowering limb.

238

Figure 3. Velocities and water table depths for January, February and March 1997.

Figure 4. Velocities versus water table depths for November 1996 to August 1998 period. Data correspond to mean daily values.

Figure 5. 1996.

Figure 5 shows the relationship between groundwater level changes and landslide velocities for one of the biggest rainfall events of the study period, that of December 1996. This figure shows that velocities in both limbs (rising and lowering) are very close. The difference between two points with the same groundwater level is negligible. However, other events of the same period of data show a completely different behaviour. Figures 6 and 7 show the changes in velocities and groundwater table for the events of January–March 1997 and December

1997–January 1998. Figure 6 shows that the landslide velocities of the rising limb and those of the lowering limb do not coincide. Velocities of the rising limb are slower than the velocities in the lowering limb which is opposite to what has been found by authors mentioned previously (in particular, Nakamura, 1984, Bertini et al., 1986). This particular behaviour was also observed in other events (Figure 8) mainly for small rainfall events where the difference between the values of velocity for the same groundwater level is very small.

239

Velocities versus water table depths for December

At the beginning of the event, velocities increase gradually. Close to the maximum level of water table, velocities increase very quickly. Velocities are still high when water table starts to withdraw but eventually they decrease as well. Once the rainfall has stopped the pore water pressures start to decrease and also the velocity with a decreasing rate which is different to the one during the rising limb. Therefore, a preliminary conclusion for this landslide is that the landslide-velocity response to the rise and withdraws of water table depends on the initial hydrological conditions of the ground and on the magnitude of the event. 3.2

Figure 6. Velocities versus water table depths for January, February and March 1997.

Theoretical Analysis

An attempt to simulate the landslide hydromechanical behaviour was considered, and for that purpose, a basic conceptual model based on the classical equilibrium differential equation was developed: 

F=m·

d2 u 2

dt

+C·

du +k·u dt

(1)

where F = Equilibrium forces; m = mass; u = displacement; t = time; C = damping coefficient related to viscosity; and k = stiffness. The analysis is based on the equilibrium of a infinite slope, using the basic equations described in Corominas et al (2005). A representative soil block with unit length and width and the initial parameters presented in Table 1 were used for the analyses. The stiffness was assumed nil and the soil parameters were obtained from laboratory experiments simulating residual conditions on ring shear equipments allowing for large displacements. Two different situations were considered: i) Model including a viscous component; and ii) Model without a viscous component (i.e., C = 0). In Figure 9 the relationship between velocities and depth of groundwater level taking into account a viscous component is shown. Note that with this model,

Figure 7. Groundwater table depths versus time. Event for January, February and March 1997.

Table 1. Initial model parameters from the Vallcebre landslide.

Figure 8. Velocities versus water table depths for December 1997 and January 1998.

Setting

Value

Damping Coefficient (C) Stiffness (k) Slope angle (α) Friction angle (φ) Cohesion (c′ )

2.46 × 1011 Ns/m 0.0 10◦ 14◦ 0.0

240

the response of velocities to the rising and lowering limb of groundwater level is the same. The results corresponding to the other case analyzed is shown in Figure 10. Now the viscosity term has been neglected. Note that in the rising limb the velocity is increasing and in the lowering limb continues increasing without deceleration of the mass movement when the groundwater level is low. That is, the viscosity provides with a mechanism for dissipating energy in the system, and eventually helps reducing velocity and stabilising the landslide. This is the result of a simple analysis and a simple model, but it may be useful to understand the dynamics of the landslide and the effect of the viscosity term in the movement. According to this conceptual model, a constant viscosity will give a ‘‘reversible’’ response in terms

Figure 9. Results from the model: velocities, u′ versus water table depths, H-d, including a viscous component.

of velocity versus groundwater table. Therefore, a hysteretic behaviour seems to be related either with a non-constant viscosity or with another effect not considered in the conceptual model. For instance, in Vallcebre the toe of the landslide is eroded by the Vallcebre torrent in the case of heavy rains and this may affect the equilibrium conditions. In such particular situations, one may expect a different behaviour of the landslide in the rising limb or in the lowering limb of the velocity curve. 4

CONCLUSIONS

The recorded data available for Vallcebre landslide has been used in this work to analyse the dynamics of the movement during rising and lowering of the ground water table. The velocity of the movement has been considered as main variable to be analysed as dependent of the ground water table level. In some rainfall events the velocities in the rising limb are very close or the same that in the lowering one and comparing this case with the theoretical analysis implies that the mechanism of slope movement has to include a viscous component. However, other events have shown a different behaviour with velocities in the rising limb lesser than velocities in the lowering limb. A simple conceptual model based on the infinite slope analysis was considered to understand the relationship between velocities and groundwater table. It was found that for a single event one should expect a ‘‘reversible’’ behaviour of velocities, and that has been measured in many situations in Vallcebre. However, sometimes a hysteretic effect has been observed in single events, a situation that has been reported by other authors as well. That may be due to a non constant value of the viscosity, or to changes in the conditions acting on the landslide, i.e. the toe of the moving mass has been eroded by the torrent and therefore the conditions when rising or lowering the water table can not be compared directly. ACKNOWLEDGEMENTS

Figure 10. Results from the model: velocities, u′ versus water table depths, H-d, without a viscous component.

First author is on leave from University Centreoccidental ‘‘Lisandro Alvarado’’ from Barquisimeto, Venezuela and express his thanks for the financial support. Moreover, grant provided by Gran Mariscal de Ayacucho Foundation (FUNDAYACUCHO) is also recognized. This research work has been supported by the Spanish Science and Education Ministry (MEC), project number CGL2005-05282 (MODEVALL) and by the Institute of Geomodels (UPC-UB-CSIC).

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REFERENCES Bertini, T., Cugusi, F., D’Elia, B. & Rossi-Doria, M. 1984. Climatic conditions and slow movements of colluvial covers in central Italy. Proceeding of the 4th International Symposium on Landslides. Toronto, Canada. 1: 367–376. Bertini, T., Cugusi, F., D’Elia, B. & Rossi-Doria, M. 1986. Lenti movimenti di versante nell’Abruzzo Adriatico: Caratteri e criteri di stabilizzazione. Proc. XVI Convegno Nazionale di Geotecnica, Bologna maggio 1986: 91–100. Corominas, J., Moya, J., Ledesma, A., Rius, J., Gili, J.A. & Lloret, A.1999. Monitoring of the Vallcebre landslide, Eastern Pyrenees, Spain. Proceedings Intern. Symp. on Slope Stability Engineering: IS-Shikoku’99. Matsuyama. Japan, 2: 1239–1244.

Corominas, J., Moya, J., Lloret, A., Gili, J.A., Angeli, M.G. & Pasuto, A. 2000. Measurement of landslide displacements using a wire extensometer. Engineering Geology, 55: 149–166. Corominas, J., Moya, J., Ledesma, A., Lloret, A. & Gili, J.A. 2005. Prediction of ground displacements and velocities from groundwater level changes at the Vallcebre landslide (Eastern Pyrenees, Spain). Landslides 2: p. 83–96. Nakamura, H. 1984. Landslides in silts and sands mainly in Japan. Proc. IV Int. Symp. On Landslides, Toronto 1984, 1: 155–185. Picarelli, L., Urciuoli, G., & Russo, C. 2004. Effect of groundwater regime on the behaviour of clayey slopes. Canadian Geotechnical Journal, 41: 467–484. Van Asch, Th.J.W., Van Beek, L.P.H. & Bogaard, T.A. 2007. Problems in predicting the mobility of slow-moving landslides. Engineering Geology 91: 46–55.

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The systematic landslide investigation programme in Hong Kong K.K.S. Ho & T.M.F. Lau Geotechnical Engineering Office, Civil Engineering and Development Department, Hong Kong SAR, China

ABSTRACT: The Geotechnical Engineering Office (GEO) has been collecting data and conducting annual reviews of rainfall and landslides since the 1980s. During this time, significant landslide incidents have been selected for detailed study for the purposes of advancing the understanding of landslides. Under the GEO’s systematic landslide investigation programme which was implemented in 1997, all reported landslides are examined and significant landslide cases selected for study to document the failure, establish the probable causes and identify the lessons to be learnt and the necessary follow-up actions. This paper provides an overview of the systematic landslide investigation programme in Hong Kong. 1

THE GEOTECHNICAL ENGINEERING OFFICE

Hong Kong is vulnerable to landslides due to its hilly terrain with dense urban development, the presence of a large number of substandard man-made slopes mostly formed before the 1970s without adequate geotechnical input and control, deep weathering profiles and high seasonal rainfall. In the aftermath of several serious landslides with multiple fatalities, the Geotechnical Control Office (renamed Geotechnical Engineering Office (GEO) in 1991) was established by the Hong Kong Government in 1977 to regulate the planning, investigation, design, construction, monitoring and maintenance of slopes in Hong Kong. Much of the enhanced slope engineering practice in recent years has originated from an improved understanding of landslides in Hong Kong. In particular, the systematic landslide investigation programme of the GEO, which was implemented in 1997, has played a key role in advancing the state of knowledge on slope performance and facilitated a better understanding of the causes and mechanisms of slope failures (Wong & Ho, 2000a).

2

Kong (e.g. Hencher et al. 1984; Wong & Ho, 1995; Wong et al. 1998a, b). As a result of the Kwun Lung Lau landslide on 23 July 1994, a new systematic landslide investigation (LI) programme was implemented by the GEO. This fatal landslide occurred on a 100-year old masonry wall located within a public housing estate, resulting in 5 fatalities and 3 serious injuries, and temporary evacuation of more than 3,900 residents (Figure 1). The GEO carried out a comprehensive investigation into the causes and mechanism of the landslide (GEO, 1994) and an international geotechnical expert, Professor N R Morgenstern, was engaged by the Government to conduct an independent review of the technical investigation. The investigation established that thin masonry walls are liable to fail in a brittle manner without appreciable prior warning. The landslide also highlighted the adverse effects of leakage from buried water-carrying services on slope stability.

THE SYSTEMATIC LANDSLIDE INVESTIGATION INITIATIVE

Between the 1980s and the early 1990s, significant landslides were selected for detailed studies by GEO’s in-house professionals as research and development projects to enhance the understanding of causes and mechanisms of landslides. The technical findings from these selected landslide studies provided insights for improvement to slope engineering practice in Hong

Figure 1.

243

1994 Kwun Lung Lau landslide.

These findings led to the issue of guidelines to rationalize the assessment of the stability of masonry walls and a Code of Practice on Inspection and Maintenance of Water-carrying Services Affecting Slopes (ETWB, 2006). In his independent review (Morgenstern, 1994), Professor Morgenstern concluded that ‘‘Practice in Hong Kong with respect to evaluation of slope stability is excessively influenced in a restricted manner by the slope catalogue and is not sufficiently responsive to indications of potential problems on a project or development scale’’. One of his recommendations was for Government to introduce a more integrated approach into the slope stability assessment process through review of landslides. In response to this, a systematic LI programme was launched by the GEO in 1997. Following a 3-year trial implementation to develop a new LI methodology for long-term use, the systematic LI work has been integrated with the Landslip Preventive Measures (LPM) Programme since 2000. The average annual cost of the systematic LI work is about HK$25 million (about US$3 million).

3

OBJECTIVES OF SYSTEMATIC LANDSLIDE INVESTIGATIONS

The main goals of the LI programme are illustrated in Figure 2 and described as follows: a. identification of slopes in need of early attention before the situation deteriorates to result in a serious problem; b. improvement in knowledge on the causes and mechanisms of landslides so as to formulate new ideas for reducing landslide risk and enhancing the

reliability of landslide preventive or slope remedial works; c. provision of data for reviewing the performance of the Government’s slope safety system and identifying areas for improvement; d. provision of evidence in forensic studies of serious landslides that may involve coroner’s inquest, legal action or financial dispute.

4

In undertaking the investigation of significant landslides, it is important to attend to the sites as soon as practicable in order to collect crucial field evidence that could otherwise be destroyed or removed as part of the debris clearance operation or emergency repair works. Since a large number of landslides may occur within a short period of time during severe rainstorms (e.g. over 250 landslides were reported to the GEO during and immediately following the 19–22 August 2005 rainstorm), an adequate supply of standby resources is essential for the prompt mobilization of a sufficient number of investigation teams. The GEO has been outsourcing the LI work to consultants under a standby arrangement, which has worked well in meeting the operational needs. The advantage of engaging consultants to review the performance of Government’s slope safety system through study of landslides is the impartiality of an independent party. This is especially important for forensic investigations of fatal landslides from a public accountability point of view. Also, overseas landslide experts can be mobilized as members of the landslide investigation teams on a need basis for serious landslides.

5

Figure 2. Main goals of the landslide investigation programme.

IMPLEMENTATION OF THE SYSTEMATIC LANDSLIDE INVESTIGATION PROGRAMME

METHODOLOGY OF SYSTEMATIC LANDSLIDE INVESTIGATIONS

Under the LI programme, all reported landslide incidents are examined to collate data for analysis. The landslides are screened by a panel of experienced geotechnical professionals to identify cases that warrant follow-up inspections and detailed investigations. On average, about 300 landslides are reported to the GEO every year. About 20% of the cases would be selected for inspection by the LI consultants (the vast majority of the reported cases would be inspected by the GEO under the emergency system in providing advice to government departments). Typically, about 10% of the landslides are found to deserve follow-up studies.

244

The following are some of the relevant considerations in screening the landslide incidents for inspections and follow-up studies:

6 6.1

– large-scale failures; – failures with serious consequences, e.g. casualties, major evacuation and significant social disruption; – failures with technical interest, e.g. sites with special geological or hydrogeological features; – failures involving slopes which were previously designed and checked to the required safety standards; – failures of special engineered slopes, e.g. soil-nailed slopes; – slopes with major signs of distress; – slopes with landslide clustering or a history of repeated failures. The following types of landslide studies are carried out under the LI programme: a. Landslide Examination—all the available information on landslide incidents are examined shortly after they are reported to collate data for analysis and the identify cases which deserve further studies. b. Landslide Review—these cover salient aspects of selected landslide and focus on the most important elements of the incident. This type of study is particularly relevant where the incident in itself does not warrant a detailed landslide investigation, and will enhance cost-effectiveness and ensure more effective use of resources. c. Landslide Study—these comprise in-depth studies of selected landslides examination of the history of the failed slope and identification of the causes and mechanisms of failure. d. Forensic investigation—these comprise detailed investigations of fatal or serious landslides to the highest possible rigour of proof in order to prepare a report that can be presented as evidence in legal proceedings. In order to retain in-house expertise in landslide investigations, a small number of landslides continue to be studied by the GEO. In addition to the studies of individual landslides, a diagnostic review of all the landslide data and findings from landslide studies is carried out by the GEO every year to consolidate experience and make recommendations to enhance slope engineering practice and landslide risk management. Integrated thematic studies (e.g. review of slope surface drainage with reference to landslide studies, review of landslides at active construction sites, review of landslides involving slopes affected by water-carrying services, review of soil-nailed slope failures, etc) are also conducted.

KEY FINDINGS FROM SYSTEMATIC LANDSLIDE INVESTIGATIONS Performance of engineered slopes

There was a perception in the past that the slope safety problem in Hong Kong was dominated by failure of old substandard slopes formed before 1977 and that the stability of engineered slopes built after 1977 to a high safety standard should be of little concern. An important development in slope engineering practice in Hong Kong over the more recent years has been the realization of the fact that even engineered slopes have quite a high failure rate and that there is a need to further improve the practice in order to reduce the rate of failure. From the systematic review of the landslide data between 1997 and 2006, the annual average failure rate for engineered slopes is about 0.015% for ‘major’ landslides (defined as landslide with a failure volume ≥50 m3 ), and about 0.068% for minor landslides (viz. 1, the inertia terms are negligible, the last equation of eq. (5) becomes

(4)

u¯ =

1−ε ε

in which U (t) is the total mass of fluid and grains at a transect. The momentum equations are:



=



1−ε ε

εu + (1 − ε)us = U (t)

  ⎧ ∂u ∂u ⎪ +u [(ε − q)ρ + qρs ] ⎪ ⎪ ⎪ ∂t ∂x ⎪ ⎪ ⎪ ⎪ 2 ⎪ ⎪ ⎪ = −ε ∂p − ε (u − us ) − [(ε − q)ρ + qρs ]g ⎪ ⎪ ⎪ ∂x k(ε, q) ⎪ ⎪ ⎪   ⎪ ⎪ ∂u ∂u ⎪ ⎪ ⎪ ×[(ε − q)ρ + qρ ] + u s ⎨ ∂t ∂x   (5) ⎪ ∂u ∂u ⎪ s s ⎪ ⎪ +(1 − ε)ρs + us ⎪ ⎪ ∂t ∂x ⎪ ⎪ ⎪ ⎪ ⎪ ∂p ∂σ e ⎪ ⎪ =− − − [(ε − q)ρ + qρs ]g ⎪ ⎪ ⎪ ∂x ∂x ⎪ ⎪ ⎪   ⎪ ⎪ ∂Q ∂Q ⎪ ⎩ −(1 − ε)ρs g − + us (u − us ) ∂t ∂x in which p is the pore pressure. Here k is assumed as following

k(ε, q) = k0 f (q, ε) = k0 (−αq + βε)

(6)

in which α, β are parameters and 1 < β 90%), the highly humified Irish blanket peat having the lowest LLs among this group (c.f. Fig. 1). Hobbs also suggested that bog peats can be readily distinguished from fen and transitional peats according to whether the natural water content is greater or less than the liquid limit, with LL = 700% being

100 Cranberry Moss Welsh bog

Loss on Ignition (%)

80

Shropshire & Sizewell fens Avonmouth (buried peat) Kings Lynn (buried peat)

60

Skempton & Petley (1970) MacFarlane (1969)

40

Miyakawa (1960) Shropshire & Sizewell fens

20

Irish blanket bogs

0 0

200

400

600

800

1000

1200

1400

1600

Liquid Limit (%)

Figure 4. Variations in liquid limit with organic content as indicated by ignition loss for different UK peatlands, including summary relationships obtained from Japan (Miyakawa) and Canada (MacFarlane) (after Hobbs, 1986, Figs 16 & 17b).

identified as the fen-bog boundary (Fig. 3, also applicable to Fig. 4). Hobbs explicitly excluded blanket peat from this analysis, although the data from Irish blanket bogs (Yang & Dykes 2006) do plot within the ‘amorphous bog peat’ zone. It can therefore be seen that engineering works involving blanket peat deposits cannot rely on published general relationships between index peat properties because of their characteristically high degree of humification, negligible mineral content and botanical composition associated with acidic plant species (c.f. generally alkaline fen peats: Hobbs, 1986). Analysis and interpretation of natural failures in blanket bogs will also therefore need to be primarily based on data obtained from site-specific peat samples. 4. Blanket bog is typically fully saturated all year, thus in its natural state it exerts a negligible (or even zero) normal stress on its basal layer. If the bog is slightly unsaturated, for example if the water table falls up to 0.5 m below the surface, this representing an exceptional degree of drying in western Ireland, the basal peat may experience an effective normal stress of up to around 5 kPa. The low to zero effective normal stresses that apply within natural blanket bogs raise serious doubts about the validity of any shear strength data obtained using standard procedures, in which the peat consolidates significantly under the applied normal loads. Preliminary experiments were therefore undertaken using a direct shear apparatus to provide results for comparison with data obtained previously from the same blanket peats using standard procedures. Small block samples of undisturbed basal peat extracted from near the margin of an Irish bog slide were fully

saturated then sheared in a 100 mm square shearbox. These samples were not consolidated prior to shearing, and the normal loads of between 1 and 10 kPa were applied to each respective sample as shearing commenced, possibly generating some matrix pore water pressures broadly similar to the in-situ field condition. The samples were sheared rapidly, at 0.2 or 0.5 mm min−1 , to reflect the field evidence of extremely rapid failure with the associated likelihood of some undrained shearing effects, again better representing field conditions. The problems of this approach are recognised, including the absence of pore water pressure monitoring or control and the use of normal loads far below the generally accepted minimum for this type of equipment. However, the results from two replicated sets of experimental tests were sufficiently similar to allow preliminary interpretation and to guide subsequent research. What the results appear to show is: (i) the natural basal peat has very little in-situ strength in its undisturbed state with minimal effective stresses, consistent with backanalyses of stability and with field experience of extracting samples for testing; and (ii) further lowstress direct shear and triaxial strength tests are warranted to investigate the validity and potential utility of this approach for use with existing or improved stability assessments and analyses.

4

CONCLUSIONS

The possibility of more frequent natural peat failures as a consequence of climatic changes, and the

343

30

S1 SE5 SE5 low pre-consolidation

25

E6 drained (Kirk, 2001) Peak shear stress (kPa)

E6 no pre-consolidation 20

E6 no pre-consol. (replicated) Landva & La Rochelle (1983) S1

15

SE5 SE5 low pre-consolidation E6 drained (Kirk, 2001)

10

E6 no pre-consolidation 5

0 0

5

10

15

20

25

30

35

40

Normal stress (kPa)

Figure 5. Shear strength results obtained from direct shear tests of Irish upland blanket bog peat (S1, SE5—Dykes & Warburton, 2008; E6—Dykes & Kirk, 2006), also showing the range of ring shear test results (upper and lower envelopes) obtained from Canadian H3–4 Sphagnum peat by Landva & La Rochelle (1983). All samples were normally consolidated prior to rapid first-time shearing except those identified as ‘no pre-consolidation’ (in which the normal loads were applied as shearing commenced, with consolidation occurring during the initial shearing), ‘low pre-consolidation’ (consolidated under 5 kPa normal loads prior to shearing, with the full normal loads applied as shearing commenced) and ‘drained’ (consolidated-drained tests using very low shear rates).

increasing numbers of windfarms and other activities proposed for upland environments that require engineering such as construction of access roads, suggest that the need for improved methods for analysing and modelling the stability of upland blanket bogs is urgent. Conventional stability analyses of blanket bog covered slopes may be appropriate, as has thus far been assumed, but there are as yet insufficient data to verify this or to permit the development of more reliable failure models for peat deposits. However, conventional geotechnical analyses of this material are clearly inappropriate. The research priorities are therefore to investigate the botanical controls on the geotechnical properties of peat, to establish a reliable method for determining the shear strength of peat, and to identify or develop a reliable method for analysing the stability of blanket bog covered slopes.

ACKNOWLEDGEMENTS The Irish blanket peat data were obtained during projects supported financially or otherwise by the UK Natural Environment Research Council (Grant Refs. NER/A/S/2003/00888–9), the Limestone Research Group, Fermanagh District Council & Huddersfield

University. The contributions of many colleagues to this work over several years are gratefully acknowledged.

REFERENCES Adams, J.I. 1965. The engineering behaviour of a Canadian muskeg. Proceedings, 6th International Conference on Soil Mechanics and Foundation Engineering 1: 3–7. AGEC. 2004. Reports on the Derrybrien Windfarm—Final Report on Landslide of October 2003. Unpublished report, Applied Ground Engineering Consultants Ltd., Ireland. Delap, A.D., Farrington, A., Praeger, R.L. & Smyth, L.B. 1932. Report on the recent bog-flow at Glencullin, Co. Mayo. Scientific Proceedings of the Royal Dublin Society 20: 181–192. Dykes, A.P. & Kirk, K.J. 2001. Initiation of a multiple peat slide on Cuilcagh Mountain, Northern Ireland. Earth Surface Processes and Landforms 26: 395–408. Dykes, A.P. & Kirk, K.J. 2006. Slope instability and mass movements in peat deposits. In I.P. Martini, A. Martínez Cortizas & W. Chesworth (eds), Peatlands: Evolution and Records of Environmental and Climate Changes: 377–406. Amsterdam: Elsevier. Dykes, A.P. & Warburton, J. 2007a. Mass movements in peat: A formal classification scheme. Geomorphology 86: 73–93.

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Dykes, A.P. & Warburton, J. 2007b. Geomorphological controls on failures of peat-covered hillslopes triggered by extreme rainfall. Earth Surface Processes and Landforms 32: 1841–1862. Dykes, A.P. & Warburton, J. 2008. Failure of peat-covered hillslopes at Pollatomish, Co. Mayo, Ireland: analysis of topographic and geotechnical influences. Catena 72: 129–145. Dykes, A.P. & Warburton, J. In press. Characteristics of the Shetland Islands (UK) peat slides of 19 September 2003. Landslides. Dykes, A.P., Gunn, J. & Convery (née Kirk), K.J. In review. Landslides in blanket peat on Cuilcagh Mountain, Northern Ireland. Geomorphology. Farrell, E.R. & Hebib, S. 1998. The determination of the geotechnical parameters of organic soils. Proceedings of the International Symposium on Problematic Soils, IS-TOHOKU 98, Sendai, Japan: 33–36. Hanrahan, E.T. 1964. A road failure on peat. Géotechnique 14: 185–202. Hanrahan, E.T., Dunne, J.M. & Sodha, V.G. 1967. Shear strength of peat. Proceedings of the Geotechnical Conference, Oslo, Vol. 1: 193–198. Helenelund, K.V. 1967. Vane tests and tension tests on fibrous peat. Proceedings of the Geotechnical Conference, Oslo, Vol. 1: 199–203. Hobbs, N.B. 1986. Mire morphology and the properties and behaviour of some British and foreign peats. Quarterly Journal of Engineering Geology 19: 7–80. Holden, J. & Burt, T.P. 2003. Runoff production in blanket peat covered catchments. Water Resources Research 39: 1191, doi:10.1029/2002 WR001956. Hollingshead, G.W. & Raymond, G. 1972. Field loading tests on Muskeg. Canadian Geotechnical Journal 9: 278–289.

Kirk, K.J. 2001. Instability of blanket bog slopes on Cuilcagh Mountain, N.W. Ireland. Unpublished PhD thesis, University of Huddersfield, UK. Landva, A.O. 1980. Vane testing in peat. Canadian Geotechnical Journal 17: 1–19. Landva, A.O. & La Rochelle, P. 1983. Compressibility and shear characteristics of Radforth peats. In P.M. Jarrett (ed.), Testing of Peats and Organic Soils: 157–191. Philadelphia: ASTM Special Technical Publication 820. Long, M. 2005. Review of peat strength, peat characterisation and constitutive modelling of peat with reference to landslides. Studia Geotechnica et Mechanica XXVII: 67–90. Long, M. & Jennings, P. 2006. Analysis of the peat slide at Pollatomish, County Mayo, Ireland. Landslides 3: 51–61. MacFarlane, I.C. (ed.) 1969. Muskeg Engineering Handbook. Toronto: University of Toronto Press. Mills, A.J. 2002. Peat slides: morphology, mechanisms and recovery. Unpublished PhD thesis, Durham University, UK. Miyakawa, I. 1960. Some aspects of road construction in peaty or marshy areas in Hokkaido. Sapporo, Japan: Civil Engineering Research Institute, Hokkaido Development Bureau. Skempton, A.W. & Petley, D.J. 1970. Ignition loss and other properties of peats and clays from Avonmouth, King’s Lynn and Cranberry Moss. Géotechnique 20: 343–356. Warburton, J., Holden, J. & Mills, A.J. 2004. Hydrological controls of surficial mass movements in peat. Earth Science Reviews 67: 139–156. Yang, J. & Dykes, A.P. 2006. The liquid limit of peat and its application to the understanding of Irish blanket bog failures. Landslides 3: 205–216.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Stability problems in slopes of Arenós reservoir (Castellón, Spain) J. Estaire & J.A. Díez Laboratorio de Geotecnia, CEDEX, Mo de Fomento, Madrid, Spain

C. Olalla ETSI Caminos, Canales y Puertos, Universidad Politécnica de Madrid, Madrid, Spain

ABSTRACT: This paper presents the analysis of the stability conditions of two slopes situated in Arenós reservoir, taking too into account that a town, named Puebla de Arenoso, is located on the top of one of the two slopes. With the information provided by the geotechnical investigation carried out, some soil profiles were elaborated that made it possible to identify the types of soils existing in the slopes: sandy coluvial sediments on a 10 m thick layer of altered marls. The interpretation of all the data available made it possible to deduce that the slope was in a quite strict equilibrium situation that produced movements in the coluvial sediment layer amplified by the successive increases and decreases of water level in the reservoir. The slope situated in front of Puebla de Arenoso was also studied. This slope suffered a great dimension slide at the beginning of Quaternary, whose stability should be verified. In this case, the main problem that can be originated by a reactivation of that huge slide was the formation of a wave that could affect some buildings situated in low parts of the town or even the dam. To face these problems, some soil treatments were carried out.

1

INTRODUCTION

Arenós dam and reservoir are situated in Castellón, at the east of Spain, on Mijares river. The slopes of the reservoir have presented, since dam construction, instability problems and slides. Among those slides, there are some of great dimensions that limit the maximum volume of water that can be stored in the reservoir. The slopes studied in this paper, as

shown in Figure 1, are slopes 1 and 2, that suffered a sliding at the beginning of Quaternary and a small reactivation in 1989, and slope 4 in whose top Puebla de Arenoso is situated. In the paper those slopes are referred as ‘‘paleoslide’’ (slopes 1 and 2 and ‘‘Puebla de Arenoso slope’’ (slope 4). Slope 3 was also studied but it does not have any important problem of stability.

2

DESCRIPTION OF PROBLEMS

The potential instability of those slopes make necessary to analyze the following aspects:

Figure 1. Diagram showing potential slides in the studied area.

a. Determination of the possibility of a reactivation of the great slide occurred in Quaternary that could affect almost all the surface of slopes 1 and 2. b. Determination of the possible maximum volume of the slides with highest probability of occurrence, taking into account the present geomorphological and hydrological conditions of the slopes.

347

c. Determination of the velocity of the material flow in its way to the reservoir, in case of a slide occurrence. That velocity is one of the factors with more influence in the height of the waves produced in the water stored in the reservoir. d. Determination of the possibility that the material fallen from the slopes could form an artificial dam, whose failure could produce very big waves with a great destructive potential, affecting mainly the low parts of the town or the dam. e. Determination of the stability condition of slope 4 and the definition of the remedial treatments, taking into account its importance as there are some houses inhabitant in its top. 3

Figure 2. Arenoso.

GEOTECHNICAL INVESTIGATION

Position of boreholes drilled in Puebla de

The works performed to analyze the stability of the different slopes are described below.

4

3.1 Paleoslide (slopes 1 and 2)

Based on the data collected in the geological and geotechnical investigation, some geological soil profiles were made.

a. Geological study: it consisted in the interpretation of stereoscopic pairs of photos and ‘‘in situ’’ ground investigation, whose data were used to elaborate geomorphological maps at 1:10.000 and 1:5.000 scales. b. Campaign of geotechnical investigation: five boreholes were drilled, of more than 100 m length each. Inclinometric tubes were installed in all the boreholes. Before installation, narrow and lengthwise incisions were made in the tubes to be used also as piezometric tubes to analyze the variations in water level. The surrounding hole was filled with fine gravel to allow water go into the tubes. c. Control of water levels in the boreholes. d. Search of the eleven boreholes drilled in 1989, after the reactivation of a slide in the toe of slope 1 and analysis of their present state. e. Campaign of laboratory tests made mainly with samples obtained in the altered rock substratum. 3.2 Puebla de Arenoso slope a. Control of the present state of the 21 boreholes drilled in 1972, in the town, before the construction of the dam. The position of those boreholes can be seen in Figure 2. b. Drill of other six boreholes equipped with inclinometric tube to be used also as piezometric tube. Its position can also be seen in Figure 2. c. Inclinometric measurements were made, one per month, during six months. d. A map of cracks existing in the buildings and streets of the town was made to determine the main directions of the ground movements.

SOIL PROFILES

4.1 Paleoslide (slopes 1 and 2) In slopes 1 and 2, situated in front of Puebla de Arenoso and in the left margin of Mijares river, there is a great slide formed by a principal, old and of great dimension slide and other minor slides produced in its toe. The principal slide possibly occurred at the beginning of Quaternary epoch, during Pleistocene, that means it is between 1, 8 and 0, 07 million of years old. Afterwards, and during recent Holocene, other minor slides were generated in the slope toe due to the erosive action of Mijares river. Nowadays that zone is covered by water stored in Arenós reservoir. In 1989 a rotational slide occurred in the slope toe after a period of heavy rains. Those slides can be seen in Figure 3. The upper part of the great paleoslide is situated between 750 and 850 m, above sea level, and its toe is in the bottom of Mijares river valley, at about 540 m, above sea level, as can be seen in Figure 4. It is interesting to remark that, according to different studies, the flow of this slide closed the valley and went up in the other margin. The rests of this flow form the area in which Puebla de Arenoso is situated, about 50 m above the bottom of the valley. The slope is formed by different slide masses that suffered several readjustments. The slide material is formed by an up to 90 m thick disorganized mass with metric and decimetric loose limestone blocks, embedded in sandy-clay matrix. The base of the paleoslide is formed by Aptiense marls and, in the toe area, by Weald clays, that they are in contact with the marls by a fault. In some boreholes, metric and decimetric thick breccia zones were

348

Figure 3. General view of the paleoslide.

Figure 4. Soil profile of the paleoslide.

Figure 5. Aspect of the breccia material found in boreholes.

detected, near the contact between the slide material and the substratum. The slide movement very probably occurred through those breccia zones which are formed by angled fragments of marly limestone, of milimetric and centimetric size, embedded in a marly-clay, blackish grey matrix, as it can be seen in Figure 5. The piezometric level in the upper part of the slope was in the contact zone between slide material and substratum. In the boreholes situated in the low part

Figure 6.

General view of the town near the top of the slope.

of the slope, the piezometric level coincided with the reservoir level. 4.2 Puebla de Arenoso slope As it was said previously, Puebla de Arenoso is located on slide materials belonging to the toe of the paleoslide occurred in the opposite margin of Mijares river, as it can be seen in Figures 3 and 6.

349

Table 1. Material geotechnical parameters deduced by stability back analysis. Material

Cohesion (kPa)

Friction angle (◦ )

Slide materials Substratum altered layer Substratum Weald clays

15 0 0

28.0 12.5 25

Table 2. Figure 7. Soil profile of Puebla de Arenoso slope.

Figure 7 shows the geological soil profile of Puebla de Arenoso slope. The average thickness of slide materials is about 50 m, with a detected maximum value of 56 m. Great blocks and rocky fragments in a chaotic disposition, embedded in a marly-clay matrix, form the slide materials. The substratum, formed by marls that are altered in its upper part, appears below. The points where the boreholes made in 1972 are cut are also drawn. Those points form a possible cinematically sliding surface, whose top coincides with some of the cracks observed in that slope area. This sliding surface goes through the mass of slope materials and it has its toe some meters above the river level.

5 5.1

DATA ANALYSIS Paleoslide (slopes 1 and 2)

The first step was to perform 2D stability back analysis of the different slides occurred in the slope: the great paleoslide occurred in Quaternary, the middle sized slide produced in Holocene and the last and smallest one, occurred in 1989. In these calculations, using Morgenstern-Price method, the sliding surface of the great paleoslide was supposed to develop through the altered rock substratum layer. In the other slides, the sliding surface can be seen in Figure 4. The results of those calculations were the geotechnical parameters of the materials existing in the slope whose values are given in Table 1. The main conclusion of these first analyses is that the reactivation of the paleoslide is not probable as the strength values of the altered substratum layer deduced from stability calculations are clearly lower than the ones determined in laboratory (c = 10 kPa; φ′ = 18◦ ). The second step was to analyse the present slope stability situation, in the calculation hypothesis shown in Table 2. The first results of this second set of calculations is that the most probable future sliding surface is similar

Different hypothesis made in the stability analysis.

Reservoir level (m)

Phreatic level

Safety factor

575 575 600 600 600

Flat High Flat High Fast drawdown

1.06 1.08 1.05 1.08 1.01

to one corresponding to a partial reactivation of the slide occurred in Holocene. The slight difference between the results obtained is due to the fact that the volume of materials affected by an increase in the reservoir water level is relatively small. The main conclusion of this second set of calculations is that present stability situation of the toe of the slope is quite precarious, as the maximum safety factor is below 1.1. 5.2 Puebla de Arenoso slope The potential sliding surface drawn in Figure 7 seems to indicate that the slope is in a very precarious equilibrium situation that produces readjustment movements in the slide mass, increased by the successive oscillations of the reservoir water level. However, it is important to remark that, since the drilling of the boreholes in 1972, no important slide has occurred in this slope. The stability back analysis performed with that surface show that the slide material has the following strength parameters: cohesion between 0 and 10 kPa, combined with friction angles ranging between 20 and 22◦ . Those values are in correspondence with the characteristics of the matrix of the slide mass formed by sandy-clay coluvial with a slight cohesive component. Those values could not be contrasted with laboratory tests due to the great size of the limestone blocks embedded in the sandy-clay matrix. On the other hand, in the inclinometric measurements some movements were detected in three of the boreholes, at depths corresponding with zones near the contact between the slide materials of the paleoslide

350

and the grey marls of the substratum. These measurements were performed during a period of draught and with a slow and progressive decrease in the reservoir water level. 6 6.1

CONCLUSIONS OF THE STUDY Paleoslide (slopes 1 and 2)

a. It is not probable that the great paleoslide has a sharp and sudden reactivation, without some slow movements appear previously in the slope that make it possible to take some remedial treatments. b. However small movements at the toe of the great paleoslide can be produced, as it occurred in 1989. Taking into account the precarious stability equilibrium of the toe of the slope, a period of heavy rains can be an important activation factor for those small movements to be produced. c. The estimated volume of that future potential slide at the toe is about 2,1 millions of m3 . Based on geological considerations, it is estimated that the volume of material that could reach the reservoir would be 1,55 millions of m3 . d. Taking into account the characteristics of the materials existing in the slope, it is not foreseeable that its flow into the reservoir was going to be quick enough not to be detected by a intensive system of auscultation, before a sharp and sudden movement was produced.

7.1

Installation of auscultation

14 GPS stations were installed to monitor the slope movements, as it can be seen in Figure 8, (Solanes, 2007). Besides, the inclinometers, installed during the geotechnical investigation campaign, are read periodically. By now, the movements detected do not exceed 5 mm and are quite slow. 7.2

Analysis of wave height

Some calculations were performed to determine the wave height and the transient elevation of the reservoir water level due to a more or less sudden entrance of material, coming from some of the potential unstable slopes of the reservoir, (Segura, 2007). As the velocity of the potential slopes depends on some physical parameters that are difficult to quantify, calculations were made with a wide range of times of material flow (3, 6, 10 and 30 minutes). Before that, a data recompilation from reference literature was made. The values of velocity of material flow in landslides similar to the one studied here are represented in Figure 9 (Segura, 2007). It can be seen that the velocities used in this study are clearly greater to the ones measured in most of the real cases referenced.

6.2 Puebla de Arenoso slope a. The analysis performed with the boreholes drilled in 1972 makes it possible to draw a possible cinematically sliding surface, whose top coincides with some of the cracks detected in that zone of the slope. This sliding surface goes through the mass of slide material existing in the slope and it has its toe some meters above the reservoir water level. b. The interpretation of the inclinometric measurements performed in the boreholes of the town, during a period of draught and with a slow and progressive decrease in the reservoir water level, made it possible to determine that the slope was in strict equilibrium. Those measurements indicated that movements are mainly produced in the contact between the slide material and the marly substratum. 7

DESCRIPTION OF SOLUTIONS

Taking into account the previous conclusions, some remedial measurements were taken to try to solve the stability problems of the slopes.

Figure 8. slope.

351

View of one of the GPS station installed in the

This 190 m long structure is formed by a slab of a variable thickness between 1.6 and 2.0 m, and vertical walls of height between 8 and 10 m and thickness between 1.2 and 2.0 m. Their main function is to avoid water erosion of the stabilization fills placed in both margins. To design this structure some numerical hydraulic simulations were performed and a scaled physical model was tested. Figures 12 and 13 show a general view of the works performed in the Puebla de Arenoso slope.

Figure 9. Reference values of velocity of material flow in landslides (Segura, 2007).

In spite of it, the obtained results indicate that the maximum reservoir water level is not greater, in any case, to 3 m, in Puebla de Arenoso slope, and to 1 m, in the dam. These values must be considered as acceptable as the existing clearance are greater in both zones. 7.3

Figure 11. Section of the works performed in Puebla de Arenoso slope.

Remedial works

The solution for the stabilization of Puebla de Arenoso slope consisted in the construction of fills made with granular soils, protected by rockfills, in both margins of the river and of a concrete structure to channel the river (Solanes, 2007). A sketch of the solution can be seen in Figures 10 and 11. These fills, with stabilization effects in the toes of both margins, are between 30 and 40 m high. Their inclination is 2H:1V. They have draining layers to dissipate water pressures in case of a fast drawdown. In the riverbed, a layer of 5.5 m of granular material was placed as a foundation layer of the concrete structure.

Figure 10. Plan of the works performed in Puebla de Arenoso slope.

Figure 12. General view of the works performed in Puebla de Arenoso slope.

Figure 13. works.

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General view of the slope after the end of the

ACKNOWLEDGEMENTS The authors of this paper want to acknowledge Mr. Julián Cuesta, geologist of Eptisa, due to his effort during the geological study, and Mr. Fernando Solanes and Miss María Irles of Spanish Ministerio de Medio Ambiente (Ministry of Environment) due to their support and help during the execution of the works.

de Arenós (Castellón). Jornadas Técnicas sobre Estabilidad de Laderas en Embalses. pp 295–332. Zaragoza. Conf. Hidrográfica del Ebro. Segura N. & Fernández L. 2007. Estudio de la variación del nivel de embalse producida por la entrada de material de laderas en el embalse de Arenós. Jornadas Técnicas sobre Estabilidad de Laderas en Embalses. pp 455–476. Zaragoza. Conf. Hidrográfica del Ebro.

REFERENCES Solanes F., Calderón P., Escuder I. & Martínez J. 2007. Estabilización de la ladera de Puebla de Arenoso en el embalse

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The 22 August, 2006, anomalous rock fall along the Gran Sasso NE wall (Central Apennines, Italy) G. Bianchi Fasani, C. Esposito, G. Scarascia Mugnozza & L. Stedile ‘‘Sapienza’’ University of Rome, Department of Earth Sciences and ‘‘CERI’’ Research Center on Geological Risks, Rome, Italy

M. Pecci IMONT, National Mountain Institute of Italy, Rome, Italy

ABSTRACT: It is described the rock fall event occurred along the Gran Sasso massif (Central Apennines, Italy) on 22 August, 2006, when a limestone block, with an estimated volume of about 30,000 m3 , fell from the subvertical NE wall nearby the Corno Grande peak, the highest peak of the Italian Apennines. Despite the small rock volume involved in the landslide, the rock fall deposits covered an area of about 35,000 m2 , a giant and abrasive dust cloud was generated by the atmospheric pressure waves (air blasts) induced by the rockfall impact and determined destructive effects over an area of about 110,000 m2 at the base of the slope. Moreover the dust cloud covered a distance of about 3 km, thus reaching the village of Casale San Nicola and the A24 motorway that was temporarily closed for security reasons. The seismic noise generated by the rock fall was recorded by the National Institute of Nuclear Physics seismometric devices located in the Gran Sasso underground laboratories (LNGS).

1

INTRODUCTION

On August 22nd, 2006 a limestone block of about 30,000 m3 detached from the NE slope of the Gran Sasso d’Italia (the Apennines’ highest peak, 2912 m a.s.l.) at an elevation of 2,800 m asl and was involved in a rock fall along the 1,500 m high cliff, also known as the ‘‘Paretone’’ (big wall). Because of the huge dust cloud generated by the rock fall, the motorway A24, located few kilometres downhill, was closed for security reasons. The seismic noise generated by the rock fall until the final impact over the cliff base was recorded by the seismometric devices located in the underground laboratories (LNGS) run by the National Institute of Nuclear Physics. The rock block involved in the fall consists of Triassic dolomitic-limestone which thrust marly-calcareous and marly-arenaceous formations of Cretaceous-Tertiary age within a very complex structural setting. Prompt site investigations allowed to estimate the volume of the felt rock block by means of laser telemetry measurements from the nearest peak; the effects induced by the impact of the rock fall were also observed and the run out area at the steep cliff toe measured as well. In this paper we analyse the possible causes which led to the slope failure, the dynamics of the fall and the mechanisms which generated the giant dust cloud. We also compare the present case

history with previous unusual rock falls (Morrissey et al. 1999; Wieczorek et al. 2000) in order to predict hazard scenarios potentially induced by future rockfall events along the Gran Sasso NE wall.

2

GEOLOGY AND CLIMATE

The unique landscape of this zone within the Apennines is dominated by the Gran Sasso Massif that lies just like a ‘‘huge boulder’’ (from which the name itself) (Fig. 1) made of Triassic and Jurassic massive limestone and dolomite over relatively smooth valley slopes incised in bedded marly-limestone and sandy-marlstone. The geomorphological setting strictly reflect the structural setting of the area: it is featured by numerous thrust sheets, superimposed one each other during the Pliocene compressive tectonic phases (Ghisetti & Vezzani 1990, Calamita et al. 2002). The more evident sheet crossing the sub-triangular wall of the Corno Grande (Fig. 1) is also the easier mountaineering route to reach the top. The so called ‘‘Jannetta route’’ overlaps the main overthrust, bringing into contact the stratigrafical-structural Corno Grande sub unit (sector above the Jannetta route, containing the rock failure) with the same formation, but

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system elevated the Corno Grande structural high during the extensional phase at the beginning of medium Lias and have been re-activated during the Messinian by the Apennine compressive tectonic phase. Glacial and periglacial processes typify the summit area of the massif which hosts the southernmost glacier (Calderone) in Europe, in a fast reduction phase (Pecci 2007). Extensive historic and prehistoric rockfall and debrisflow deposits have accumulated at the base of the wall (D’Alessandro et al. 2003), where some villages and important lifelines are located. In fact, earthquakes, snowmelt, freezing-melting cycles and frost-wedging effects have caused rock falls in the area, like the historically recorded 1897 event which made a scar known as ‘‘Farfalla’’ (Butterfly) due to its shape (see the circle in Figure 2). The NE wall is a very popular and very challenging climbing cliff, thus imposing further risk conditions (Amanti et al. 1994)

Figure 1. Virtual 3D view of the Gran Sasso NE wall (from the website http://maps.live.com).

3

Figure 2. Frontal view of the landslide paths along the rock slope.

belonging to the stratigrafical-structural Corno Piccolo sub unit in the sector below (Adamoli 1992). This structural element cuts all the eastern face and corresponds to an original normal faults system. Such

DESCRIPTION OF THE EVENT

The event has been observed just after the release until the impact (Fig. 3), while time histories (Fig. 4) recorded by the seismic antenna at LNGS (Laboratori Nazionali del Gran Sasso—National Laboratories of the Gran Sasso) and field investigations helped in better frame the dynamics of the rock fall. It occurred at 7.30 am GMT, when a 30,000 m3 rock block detached from the ‘‘Guglia Bambù’’ (Bamboo Pinnacle) along a slickensided fault surface. After the failure, the already jointed rock mass underwent a crumbling process while experiencing a fall as high as 1,500 m along the steep cliff which features this Gran Sasso slope. Along this section the debris motion took place mainly by free fall and bouncing. In addition, shortly after the detachment the falling mass split into two parts: a first one fell down along an almost straight, ballistic trajectory, while a second one was channelled within a chute (Vallone Jannetta; see the white arrows in Figure 2) following a more winding path and entrapping rock fragments and debris along the path. In both cases, the fallen rock fragments reached the base of the steep slope where they spread over an area of about 35,000 m2 ; a part of the debris was then channelled within the upper section of a deeply cut valley. In addition, immediately after the rock impact at the wall base, a huge and dense dust cloud developed and rapidly moved downhill covering a distance of about 3 km until the A24 motorway and the village of Casale San Nicola (Fig. 3); because of this dense and rapidly moving dust cloud the motorway was temporarily closed for security reasons. The rock fall occurred on a sunny day without any apparent trigger event such as an earthquake or

356

Figure 3. Photographics sequence of the rock fall event, showing the spreading of the dust cloud.

Figure 4. Seismogram recorded by the LNGS in concomitance with the rock fall event. Figure 5. Rock free face of the detached block from different points of view.

heavy rainstorm. During the preceding days some rainfalls were recorded, due to some afternoon thunderstorms, a typical phenomenon during this hot period of the year. Dark streaks observable on the detachment free face indicated paths where water had seeped through the jointed rock mass (Fig. 5). As

regards the geologic controls on the failure mechanism, the shape and volume of the detached block is clearly guided by the presence of tectonic slickensided surfaces (Fig. 5); the jointing conditions within the

357

rock mass are on their turn connected with the local structural setting, featured by tectonic lines of regional importance. Laser-GPS telemetry measurements from the nearest peak also allowed to determine the actual detached rock volume (30,000 m3 ), once compared with images showing the shape of the Bamboo Pinnacle before the fall. The remainder of the pinnacle still in place is an overhanging rock dihedral estimated to be 20,000 m3 in volume. Site investigations along the steep cliff pointed out the numerous joints within the rock mass and the presence of unstable rock blocks.

4

RECORDED EFFECTS

The most significant feature of the 22 August rock fall is definitely represented by the high energy dissipation derived from the impact and rock fragmentation at the slope toe and the so generated dust cloud (Fig. 3). Actually, according to the above reconstructed kinematics of the event and as inferred from the time histories recorded at LNGS (Fig. 4), we hypothesize two distinct impacts due to a gap in the arrival times by the rock fragments which followed the two different paths. As regards the mechanisms involved in such highly dissipative process, it can be assumed that the energy of the fallen rock volume was transferred to both ground and atmosphere, thus creating air pressure waves (airblast effect) and a consequent dense dust cloud. The latter has been driven by the airblast with high velocity and has been able to preserve an abrasive effect over a quite large area (110,000 m2 ) (Fig. 6), and reached the village of Casale San Nicola located 3 km downhill the impact area. Such an abrasive effect is shown by the uprooting and snapping of hundreds of trees and bushes as well as by the debarking (Fig. 7). Airblasting is usually generated by landslides involving rock volumes in the order of 106 m3 . Only few cases have been previously reported (e.g. Yosemite, 10 July 1996 event by Morrissey et al. 1999 and Wieczorek et al. 2000) regarding relatively small magnitude landslides able to generate air pressure waves. In both cases the airblast generation is due to the high impact velocity and the two separated impacts, as inferred from the seismograph time histories. During the time interval between the impacts, the initial atmospheric conditions were changed after the formation of a first dust cloud. Such high density fluid could have favoured the sudden variation in air pressure (Morrissey et al. 1999). The rock debris within the landslide deposit shows grain size within fine sands and boulder (1 m3 ), and

Figure 6. Aerial view showing the extension of the rock fall deposit and the zone affected by air blast.

Figure 7.

Abrasive air blast effect on the vegetation.

a reduction in rock element dimensions versus distance from impact point are recorded. The larger landlside debris channeled within a stream and came to halt 250 m far from the impact point, while the finer material blanketed areas within 3 km distance. 4.1 Considerations about further risk scenarios In addition, during the survey phases that were carried out to better reconstruct the discussed rock fall, it was possible to preliminarily define the overall geomorphic, geomechanical and, finally, slope stability conditions along the whole NE wall of the Gran Sasso. For this purpose, data derived from direct surveys in the accessible areas were coupled with data inferred

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by remote techniques (such as laser telemetry), performable through devices easy to transport on steep and quite unstable tracks. As a result, other blocks within the rock mass prone to detachment have been recognized. Even if most of the so identified blocks have dimensions quite comparable with those of the hereby presented rock fall, it was possible to observe also some huge rock ‘‘pillars’’ with volumes ranging between 105 m3 and 106 m3 (Fig. 8). Furthermore, field surveys pointed out the presence of even largesized (tens of cubic meters) limestone blocks also in some areas downslope of Casale San Nicola village and the motorway, thus testifying the occurrence of past massive and catastrophic rock slope failure. Based on this evidence, it is possible to hypothesize risk scenarios for both the motorway and the village. Even if the gravity-induced landscape evolution under the present boundary conditions seems

Figure 8. DEM adopted for the volume calculations of the unstable blocks.

Figure 9. Possible morphologic constrain able to produce the channelling of granular flow deriving from larger sized landslide event.

to be characterized by frequent, small-sized rock fall events, the potential occurrence of massive rock slope failures involving the ‘‘pillars’’ under specific conditions, could evolve in a dry granular flow. As a matter of fact, the presence of deeply incised channels at the base of the wall can represent the geomorphic ‘‘constraint’’ able to convey the highly fragmented debris, allowing for a long run-out (Fig. 9). 5

CONCLUDING REMARKS

As before the event no earthquake was recorded, no seismic trigger can be invoked for the failure. The only possible causes can be ascribed to the intense jointing within the rock mass and to the frost (thermal) wedging and to the possible permafrost degradation phenomena. The latter is justified by the progressive retreat of the Calderone glacier nearby the source area at the same elevation (Pecci 2007). According to recent studies in the Alps (Davies et al. 2001, Fischer et al. 2006), rockfall events can be referred to climate changes in high mountain environment, thus determining a significant permafrost or seasonal ice pattern modification. Annual mean temperature can influence the stability of rock masses with ice bearing joints. This can be related to the water seepage evidences observed on the free face of the 22 August event. As a consequence, the presented case history could be considered as a further example of rock slope failure induced by climatic change in a very sensitive environment such as the high peaks of the Gran Sasso Massif, the southernmost glacial and periglacial environment in Europe. REFERENCES Adamoli, L. 1992. Evidenze di tettonica d’inversione nell’area del Corno Grande-Corno piccolo (Gran Sasso d’Italia). Bollettino della Società Geologica Italiana III: 53–66. Amanti, M., Pecci, M., Scarascia Mugnozza, G. & Vallesi, R. 1994. Comparison and critical review of quick field data collection methods on rock slopes: a contribution from climbing techniques and experiences. Atti del Convegno ‘‘Man and mountain ’94’’, 20–24 giugno 1994, Ponte di Legno (bs), pp. 189–198. Calamita, F., Scisciani, L., Adiamoli, M., Ben M’Barek, M. & Pelorosso, M. 2002. Il sistema a thrust del Gran Sasso d’Italia (Appennino Centrale). Studi Geologici Camerti 1/2002: 19–32. D’Alessandro, L., De Sisti, G., D’orefice, M., Pecci, M. & Ventura, R. 2003. Geomorphology of the Summit Area of The Gran Sasso d’Italia (Abruzzo Region, Italy). Geografia Fisica e Dinamica Quaternaria 26: 125–141. Davies, M.C.R., Hamzal, O. & Harris, C. 2001. The effect of rise in mean annual temperature on the stability of rock slopes containing ice-filled discontinuities. Periglac. Process 12: 137–144.

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Fischer, L., Kaab, A., Huggel, C. & Noetzli, J. 2006.Geology, glacier retreat and permafrost degradation as controlling factors of slope instabilities in a high-mountain rock wall: the Monte Rosa east face. Nat. Hazards Earth Syst. Sci 6: 761–772. Ghisetti, F. & Vezzani, L. 1990. Stili strutturali nei sistemi di sovrascorrimento della Catena del Gran Sasso (Appennino Centrale). Studi Geologici Camerti vol. spec. 1990, 37–50. Morrissey, M.M., Savane, W.Z. & Wieczorek, G.F. 1999. Air blast generated by rockfall impacts: Analysis of the 1996 Happy Isles event in Yosemite National Park. Journal of Geophysical Research 104, n◦ B10: 23189–23198.

Pecci, M. 2007. The shrinkage of the central Mediterranean cryosphere in a changing mountain environment. Mountain Forum Bullettin VII Issue 2, ISSN 1029–3760 (http://www.mtnforum.org/rs/bulletins/mf-bulletin-200707.pdf). Wieczorek, G.F., Snyder, J.B., Waitt, R.B., Morrissey, M.M., Uhrhammer, R.A., Harp, E.L., Norris, R.D., Bursik, M.I. & Finewood, L.G. 2000. Unusual July 10, 1996, rock fall at Happy Isles, Yosemite National Park, California. GSA Bullettin 112: 75–85.

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New formulae to assess soil permeability through laboratory identification and flow coming out of vertical drains J.C. Gress Ecole Nationale des Travaux Publics de l’Etat, Lyon, France

ABSTRACT: Mastering a lot of landslides through deep dewatering drainage by one or more lines of drains, being pumped either by siphoning pipes or by electropneumatic pumps® , it has appeared to us that water tests were not reliable for different reasons. We suggest here a procedure through laboratory identification tests and new formulae to assess the permeability, we can wait for, and a new formula to assess the flow which will come out of a line of vertical drains in a slope. 1

INTRODUCTION

For past twenty years, the mastering of landslides, in France, has taken great benefit of the experience of a great number of works of deep drainage through siphon drains® or electropneumatic drains® . But it has clearly demonstrate that it was difficult to have a good estimation of the flow, we could wait for. After having analysed why, we propose hereunder new formulae, in order to have a better approach of deep drainage efficiency through lines of vertical drains, formulae to assess soil permeability through identification tests and a formula to assess flow coming out of vertical drains in a slope.

2

– reach the aquifers to be drained, – be bored with the good tools, in order to avoid the decrease of permeability, – must be equipped like wells (proper slotted pipes, filter). The efficiency between adjacent drains must be designed; this will lead to a better assessment of the spacing between drains. If the wanted drawdown, under the soil surface, is less than eleven meters, then the drain will be pumped through siphoning pipes, a hydraulic accumulator,

BASIC PRINCIPLES OF VERTICAL DRAINS

Lines of vertical drains are placed, on the site, in order to dewater the landslide and stop the instability. These drains, with an average spacement of 5 meters must: Manhole Ø 800

Siphoning pipe

Water table before treatment

Outlet manhole Automatically regulated flushing system

Drain

Water tab le

Sliding surface

after tre atment Horizontal reference plane

Tank permanently full of water

Scale:

1 meter

Figure 1. Cross section through a siphon drain® .

Figure 2. Electropneumatic pump® and view of the compressor chamber.

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equipped with a flushing system, regulating the flow above the critical one. If it is greater, then we use electropneumatic pumps® , air under pressure being fed by a compressor placed in a chamber, the pump being equipped with a sensor, analysing if the pump is empty or full and then regulating the feeding with air. Then, we can pump up to depth of 100 meters. 3

CLASSICAL MEASUREMENTS OF THE PERMEABILITY

In order to study the feasibility of such dewatering scheme, there is always a preliminary geotechnical and hydrogeological study with borings, samplings, laboratory tests and water tests on site and in the laboratory. It appears very often that water tests didn’t give a good order of the real permeability, because of: – bit to do the boring being not the good one, – too small diameter, – water test run in an injection way and not in a pumping way, – had we to run a Lefranc or Nasberg test? These difficulties can be mastered. But it stays the major difficulty, that is to say, that the permeability is not homogeneous and can vary vertically through the same geological layer, due to the variation of the clayey fraction or to the superposition of layers of sand and clay, the thickness of these layers being maybe less than decimetric. To put this in evidence, the best thing is to do intact continuous samplings and to analyse the variation of density, granulometry and qualify the activity of clay through Atterberg limits or methylene blue tests. We have then tried to obtain formulae giving a rough assessment of permeability through the different parameters measured. 4

The oedometer formula: log k = Cke

with Ck ≈ 0.5eo (Tavenas, F. et al. 1983) is giving only a variation of k with the void index e for clayey soils. We have worked on correlations, we had through water tests and identification tests of different soils and on the works of Nagaraj et al. 1986 and Sivapullaiah P.V., et al. 2000. It appears that the formulae proposed hereunder fits relatively well, correlating log10 k with WL = liquidity limit, when the particle of the soil have a maximum size of 400 µm; otherwise, VBS = methylene blue value of the total soil; e = void index, %2μ = percent finer than 2 μm. The methylene blue value is very frequently measured in France. The methylene blue value of the 0/400 μm fraction is well correlated to the plasticity index Ip and to the liquidity limit WL : Ip ≈ 0.045 VB0400μ

(3)

WL ≈ 0.14 + 0.063 VB0400μ

(4)

the methylene blue value of a granulometric portion o/d being linked to the percent finer than d through the formulae: VBod x % od = VB0400μ x % 400μ = VB02μ x % 2μ

(5)

When the maximum size of the soil particle is 400 μm, we propose the formulae hereunder, in order to have a rough estimation of soil permeability: if WL < 0.25 log k = −(1.41 + 25.55 WL ) + (4.46 − 3.5 WL )e

(6)

if 0.25 ≤ WL < 0.80 log k = −(5.23 + 9.2 WL ) + (4.6 − 4.11 WL )e

ASSESSMENT OF THE PERMEABILITY THROUGH SOIL IDENTIFICATION PARAMETERS

(2)

(7)

When the maximum size D of the particles is greater than 400 μm, then we propose:

There are not much formulae allowing us to have a good assessment of the permeability, these formulae working for gravely soils to clay. The Hazen formula is well known:

if VBS < 1.5

k = K(D10 )2

if 1.5 ≤ VBS < 10

(1)

k in cm/s and D10 being the diameter in cm of the screen allowing ten per cent in weight of the soil to go through. But it works only for sands and sandy gravels.

log k = −(4.99 + 1.61 VBS) + (3.97 − 0.22 VBS)e (8)

log k = −(6.52 + 0.58 VBS) + (4.03 − 0.259 VBS)e (9) where VBS is the methylene blue value of the total soil.

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e 0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

Water feeding line

0

L

Qw

-1

Horizontal initial piezometric surface

-2

2rw

VBS = 0.1

-3

VBS = 0.5 -4

H

VBS = 1

-5

hw

VBS = 1.5

-6

VBS = 3

-7

VBS = 6

-8

Figure 4. Cross section after dewatering (horizontal initial piezometric surface).

-9

VBS = 8 -10

VBS = 10.5

-11

log(k) -12

Figure 3. Values of log10 k with e and VBS. water feeding line Spring line

We can then with these formulae, through simple laboratory tests, check the in-situ water tests and have a relatively better knowledge of the different levels of permeability.

a

L

Well (drain)

5

NEW FORMULA TO ASSESS THE FLOW COMING OUT OF THE DRAINS

For a line of drains, dewatering an horizontal piezometric surface, for an unconfined aquifer, the flow of each drain is given by:   a 2Qw L Qw H 2 − h2w = + Ln ka πk 2π rw

Ho +hw a po k

Ho2 − h2w

1 + πk Ln 2πarw

Overview of a set of drains.

Initial Parabola q1=Qw/a

(10)

where Qw = flow coming out of each well; rw = radius of each well; k = permeability of the aquifer; a = distance between each drain. Each drain penetrating totally the aquifer, and a feeding line being located at a distance L: In case of an inclined piezometric surface having an initial po slope, we propose this formula: qw =

Figure 5.

po

H0

hw

q0=poHok

Initial upstream flow

q2

X

0

x

Surface of drawdown 0

Figure 6. down).

Cross section after dewatering (surface of draw-

(11)

where Ho = total initial height of the sheet of water; hw = thickness of the dewatered sheet; po = initial slope of piezometric surface; a = distance between each drain; k = permeability; rw = radius of the drain. The initial upstream flow qo is equal to po Ho k. The flow coming out of the vertical drains per meter of line is equal to q1 = Qw a .The downstream flow q2 is equal to q0 − q1 .

The substratum is supposed to be horizontal. Example if: Ho = 20 m hw = 10 m po = 8% a=5m k = 5 × 10−6 m/s rw = 0.08 m

then Qw = 71 liters per hour.

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6

CONCLUSIONS

Vertical drains, either drained by siphoning pipes® or electropneumatic pumps® , have allowed the stabilization of more than too hundred lanslides, these last twenty years. The hereabove proposed formulae of the permeability and of the flow coming out of the drains should lead to an improvement in the design of the scheme of drains, these formulea having to be adjusted to the further experience of new works. REFERENCES Bomont, S. 2002. Drainage with electropneumatic drains® . Conference JNGG 2002. Nancy, France. Bomont, S. 2004. Back experience from four landslides stabilized through lines of siphon drains® in Normandy, France. 9th International Symposium on Landslides 2004. ISL Rio. Bomont, S. et al. 2005. Two applications for deep drainage using siphon and electropneumatic drains® . Slope works for Castlehaven Coast Protection Scheme, Isle of Wight (UK) and slope stabilisation for the Railways Agency, France. In, Proceedings of the International Conference on Landslide Risk Management. 18th Annual Vancouver Geotechnical Society Symposium.

Clark, A.R. et al. 2002. The planning and development of a coast protection scheme in an environmentally sensitive area at Castlehaven, Isle of Wight. Proc. Int Conf on Instability, Planning & Management, Thomas Telford, 2002. Clark, A.R. et al. 2007. Allowing for climate change; an innovative solution to landslide stabilisation in an environmentally sensitive area on the Isle of Wight International Conference on ‘Landslides and Climate ChangeChallenges and Solutions’ Ventnor, Isle of Wight, UK. Gress, J.C. 1996. Dewatering a landslip through siphoning drain® . Ten years experiences. Proc 7th International Symposium on Landslide. Trondheim. Gress, J.C. 2002. Two sliding zones stabilized through siphon drains® . International conference on Landslide, slope stability of infrastructures. Singapor. Nagaraj, T.S. et al. 1993. Stress state—permeability relation for fine grained soils. Geotechnical (43): 333–336. Mitchell, J.K. 1993. Fundamentals of soil behavior. 2nd ed. John Wiley & Sons, Inc., New York. Pandian, N.S. et al. 1995. Permeability and compressibiliy behavior of bentonite—sand/soil mixes. Geotechnical Testing Journal (18): 86–93. Sivapullaiah, P.V. et al. 2000. Hydraulic conductivity of benton. Canadian Geotechnical Journal (37): 406–413. Tavenas, F. et al. 1987. State of the Art on Laboratory and in situ stress strain time behavior of soft clays. LAVAL University.

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Structure-controlled earth flows in the Campania Apennines (Southern Italy) F.M. Guadagno, P. Revellino, G. Grelle, G. Lupo & M. Bencardino Department of Geological and Environmental Studies, University of Sannio, Italy

ABSTRACT: Slow-velocity landslides predominate in the area of the Province of Benevento, due to the prevalent clay nature of its outcropping deposits. An analysis of these instabilities, via detailed inventory mapping, has shown a pervasive diffusion of earth flows, characterized by a reactivation tendency. Their evolution, in terms of activity and kinematic mechanisms, is structurally-controlled and can be generally connected to three principal controls: i) bedding; ii) stratigraphic or tectonic contact between lithologically differentiated sequences; and iii) zones of intense fracturing linked to folds and faults. Differences in pattern and controlling factors led to their grouping into recurrent types characterised by a different style of evolution.

1

INTRODUCTION

Like other sectors of the Italian Apennines, the area of the province of Benevento (Campania region, Southern Italy) is characterised by recurring events of slope instabilities, in time and space, determining conditions of high risk. The geological and structural setting of the Campania Apennines is the basis of landsliding processes. Moreover, recent occurrences demonstrate and confirm that human activities can have a determinant role upon the morphologic evolution of the slopes. The changes linked to increasing urbanization of morphologically complex settings and agricultural and forestry practices seem to have an important effect. Further problems connected to landslide and erosion phenomena are linked to the responses of natural slopes to rainfall regime modifications induced by climatic changes, which could produce changes in the spatial-temporal recurrence. The principal objective of this paper is to highlight the influence of the geological and structural settings on the occurrence of earth flows in the study area, by using data from a ‘‘Landslide Inventory Map’’ (Guadagno et al., 2006) of the province. These controls affect the evaluation of the susceptibility required in hazard assessment.

2

stony flysches, resulting from the deformation of paleogeographic domains (Patacca & Scandone 1989). As a consequence, lithologically differentiated sequences belonging to marine sedimentary rocks, ranging from Cretaceous to Pliocene, and continental deposits of Pleistocene, outcrop in the area. They can be grouped as follows (Pescatore et al. 2000): 1) Platform carbonate Units: limestone and dolomitic limestone forming the higher slopes; 2) Pre-orogen basin Units: basinal facies and clayey, quarzarenite and arenite deposits; 3) Synorogen and lateorogen Units: arenaceous and arenaceous-clayey flysch-like successions; 4) Pyroclastic deposits: a) generally lithoid and coherent, grey-yellowish flow deposits (the Campanian Ignimbrite); and b) incoherent or weakly cemented air-fall deposits; 5) Continental, fluvial and detrital deposits: debris and colluvial fans at the toe

THE GEOLOGICAL ENVIRONMENT

The territory of the province of Benevento is located in one of the most geologically complex areas in Italy. The Southern Apennines consists of thrustbelt structures of carbonate terrain and clayey and

Figure 1.

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DEM of the province of Benevento.

Figure 2. Map of the lithotechnical sequences outcropping in the province of Benevento, as shown in table 1. Legend: 1) Clayey-silty sequences; 2a) Clayey-marly sequences; 2b) Clayey sequences; 3a) Calcareous sequences; 3b) Conglomeratic sequences; 3c) Tuffaceous sequences; 4a) Calcareous sequences (calcareous s.s. and calcareous-clayey); 4b) Arenaceousclayey and arenaceous-conglomeratic sequences; 5a) Sandy- arenaceous sequences; 5b) Alluvial sequences; 5c) Fluvial sequences; 5d) Cemented detrital sequences; 5e) Incoherent detrital sequences.

of the carbonate slopes; generally terraced, Quaternary alluvial fan deposits along the river valleys. The structural setting, resulting from the tectonic phases, strongly influences the morphological configurations on the area. The western portion of the province is characterized by the presence of calcareous mountains, while the eastern sector is characterized by a hilly morphology (Figure 1). In contrast, the central area corresponds to the depression where marine and continental clastic Pleistocene deposits outcrop and along which the main rivers flow. Bearing in mind the aims of this paper, it was felt opportune to group the deposits outcropping in the province into successions, characterized by a likely lithotechnical homogeneity. In other words, these successions are constituted by similar lithotypes

according to geological-technical and geomechanical features (Figure 2). This procedure permitted the obtainment of data regarding the relationships between landslides and involved deposits and consequently on the principle causes of landsliding. The tectogenetic phases have induced complex tectonic settings in the area, testified by typical structures of a ductile-type and fragile-type tectonics, which completely deform and displace the sequences. Sets of joints are connected to these, which pervasively affect the masses and show high frequency near the most important tectonic lineations. Therefore, most of the above-indicated sequences can be defined as structurally complex formations (AA.VV. 1985, Picarelli 1986) as consequence of lithostratigraphical and tectonical features.

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Table 1. Geomechanical and lithological characteristics of the sequences outcropping in the province of Benevento. Group of sequences

Lithotechnical sequences

Competence

Setting

1

Prevalently pelitic—from low-degree to mediumdegree of tectonization

Clayey-silty (Ag-L)

Incoherent

Stratified, generally monoclinalic

2

Prevalently pelitic—highdegree of tectonization

Clayey-Marly (Ag-M) Clayey (Ag)

From mildly to intensely folded. Scaly clay.

3

Stony—from low-degree to medium-degree of tectonization

Calcareous (Ca)

Complex. Prevalently incoherent Stony

Conglomeratic (Cg) Tuffaceous (Tf)

Stony Weakly lithified

Stony and complex—highdegree of tectonization

Calcareous (Ca)

Calcareous s.s.

Stony

Calcareous-clayey

Complex, prevalently stony

4

5

Coarse clastic and/or non-homogeneously lithificated

Well stratified strata and banks (from 30–40 cm to 5–10 m) Sometimes stratified Up to 30 m banks. Columnar jointing Well stratified and highly tectonized Well stratified strata and banks Intensely jointed and folded

Arenaceous-calyey Arenaceous-conglomeratic (Ar-AgCg)

Complex

Arenaceous and conglomeratic bank, thin layers of marl, clay and conglomer ate Intensely jointed and folded.

Sandy- arenaceous (S-Ar)

Non-homogeneous lithification

Generally stratified and well stratified

Alluvial (Al) Fluvial (Fl) Cemented detrital (Dc)

Incoherent or weakly cemented Non-homogeneous lithification Incoherent or locally weakly cemented

Unclear bedding

Incoherent detrital (Ds)

On the basis of qualitative and quantitative observations, together with data collected from site and laboratory tests (Guadagno et al. 2006), the masses were classified from a geo-mechanical point of view. Table 1 shows the grouping of the various successions as a function of their common mechanical and lithological characteristics. The main groups identified can be listed as below: i) Group 1, Prevalently pelitic sequences (from low-degree to medium-degree of tectonization); ii) Group 2, Prevalently pelitic sequences (high-degree of tectonization); iii) Group 3, Stony sequences (from low-degree to medium-degree of tectonization); iv) Group 4, Stony and complex successions (high-degree of tectonization); v) Group 5, Coarse clastic and/or non-homogeneously lithificated sequences. The map in Figure 2 shows the areal distribution of the above described sequences. It is possible to observe that wide areas are characterised by the presence of both clayey and stony tectonization formations. These areas correspond to the sectors which are the most involved in erosion and landslide processes.

3

Irregular bedding Irregular bedding

LANDSLIDES OF THE BENEVENTO PROVINCE

In order to investigate the distribution and type of failures and their correlation with the lithostructural and morphological setting, a landslide inventory map of the province was compiled on a 1:75,000 scale (a copy of the map can be requested by e-mail—[email protected]—from the authors). The surveys were performed on the basis of a 1:25,000 scale topographic map, whereas, specific areas, characterized by the presence of typical landslides and towns or infrastructures, were surveyed on more detailed maps (greater than 1:10,000 scale). Analyses were carried out by means of an interpretation of historical aerial photos from different time periods, dating from 1954, together with geological surveys carried out between the years 2001 and 2005. It should be noted that, 2003 and 2005, two heavy rainfall events triggered the reactivation of hundreds of landslides. The classification criteria adopted refer to wellknow classification systems (Varnes 1978, Hutch-

367

inson 1988, Cruden & Varnes 1996, Hungr 2001). Furthermore, they are based on some specific characteristics of landsliding processes which are worth pointing out. 3.1 Classification criteria Six different typological groups (numbered from 1 to 6 below) were distinguished, taking into account the survey results: i) (1) falls and topples; ii) (2) rock and debris avalanches, and debris flows; iii) (3) translational, rotational and composite slides, and; iv) earth flows distinguished in (4) single (Figure 3), (5) multi-source (Figure 4) and (6) coalescent (Fig. 5). This latter typological choice derives from the necessity to provide morpho-evolutive elements for the earth flows, addressed, on the one hand, to understanding the landslide mechanisms and, on the other, to defining hazard management. The use of diversified typological contexts permits to stress the

Figure 3. Oblique aerial view of the Sant’Agata de’ Goti landslide of 11 January ’97.

Figure 5. flows.

Example of basin affected by coalescent earth

wide differences in the landsliding processes involving the slopes. In relation to coalescent earth flows, the surveys highlighted that different sectors, mainly characterized by the outcropping of pelitic and complex sequences from a medium to high degree of tectonization, have been and are still affected by several landslide generations that mobilise interacting among themselves. These groups of landslides, which can be considered as a unicum from a morphological point of view, are localized inside unitary sub-catchments (Figure 5). Multi-source earth flows reflect the complex lithostructural settings. In particular, in specific areas, litholologically differentiated sequences alternate or they are in contact, inducing differentiated erosion zones. The result is the formation of articulated source areas, composed of several branches. The structural setting may also influence the morphology of the channels, as well as that of the accumulation zones. These aspects are important in defining the distribution of activity, which contributes to the assessment of the evolution of the landslide body and, therefore, to the obtainment of data on landslide susceptibility evaluation. 3.2 Areal distribution of the earth flows

Figure 4. Multi-source earth flow from a segment of the Landslide Inventory Map (Guadagno et al. 2006). Legend: 1) Distribution of the activity in the landslide area: a. Retrogressing; b. Enlarging; c. Advancing; 2) Source area; 3) Movement direction; 4) Lithostructural element controlling the evolution of the landslide mass.

In the study area, 3,160 landslides were inventoried, covering an approximate area of 358 km2 , equal to about 18% of the whole provincial surface. This value refers to parts of the territory already affected by landslides and mappable on the scale of representation. It does not include those areas where predisposing conditions have been verified and which indicate possible future landslide evolutions. Moreover, the Landslide index, calculated as the percentage of area affected by landslide events per grid of 1 km2 , reaches values even greater than 75% where clayey sequences are outcropping.

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Figure 6. Percentage distribution of areal landslide phenomena, computed for number and area, in dimensional classes.

Figure 6 shows the percentage distribution of the landslides in dimensional classes. The data refers to the single events, even if they make up part of the coalescent groups. It is interesting to note that landslides 25) represent over 20% of the areas affected by landslides. 3.3 Controlled evolution of earth flows The litho-structural setting determines favourable conditions for the development of landslides. To verify from a quantitative point of view, analyses of data were carried out both for the 5 groups of sequences and for the single lithotechnical sequences (cf. Table 1). Considering the landslide type, analyses were carried out evaluating the formations specifically involved in the source areas. As shown in Figure 7, the percentage of outcropping surface affected by landslides is up to 46% of the total. The prevalently pelitic and complex sequences at a high-degree of tectonization are those mostly affected by landslides, totalling more than 80% of the inventoried phenomena. Additionally, over 40% of the areas where clayey formations crop out, are also affected by landslides. In particular, lithotechnical sequences named clayey, clayey-marly and arenaceous-clayey—arenaceousconglomeratic are those in which most of the landslides occur. Once the initial failure is identified, landslide development can be guided by the presence and orientation of the structural control elements, which influence the style and the distribution (cf. WP/WLI 1993,

Figure 7. Percentage ratio between area involved in landsliding (SF) and total outcropping area (ST) for each lithotechnical sequence. Refer to Table 1 for the symbology used.

Cruden & Varnes 1996) of the landslides and therefore their evolution in time (Figures. 8, 9). In particular, the evolution of earth flows, in terms of activity and kinematic mechanisms, can be generally explained by: i) bedding of the homogeneous and complex sequences; ii) stratigraphic or tectonic contact between deposits with a different competence; iii) zone of intense fracturing linked mainly to the presence of axes of folds and faults. These settings constitute fundamental geological elements in understanding the control mechanisms in the earth flow source areas and channels. The hazard analysis and, in particular, aspects connected to the spatial prediction, imply a careful evaluation of the geostructural conditions that become a fundamental element in the areas where first-order events are developing. Where geological bodies with different competence outcrop, the lithostructural control is recurrent. A typical example is the instability in Figure 9. The presence of a bedding plane (N125◦ /40◦ ) of a stony sequence directs the shape, the orientation and the evolution of both the source area and the channel of the landslide, inducing a lateral retrogression of the source in comparison with the flow channel. Areas of severe structural control are the source areas of multi-source earth flows (more than 150), where separation into source branches is a direct consequence of the local lithological and structural setting. Figures 4 and 10 show the morphological characters of multi-source instabilities and the distribution of the

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A Figure 10. Example of multi-source earth flow. For the legend see Figure 4.

B Figure 8. Earth flow structurally controlled by a marlyclayey hill. During motion, the landslide body affected some houses: A) Image of November 2003; B) Image of March 2005. Figure 11. Examples of basin affected by coalescent flows; Legend: 1) Coalescent earth flows and basin boundary; 2) Flow direction.

of the surface is involved in mass movements on average, with extreme values of 72%, whereas the number of events with interconnected evolution varies from 5 to 36.

4 Figure 9. Earth flow from a segment of the Landslide Inventory Map (Guadagno et al. 2006). For the legend see Figure 4.

activity by means of the empty triangles. The peculiar shaping of the source areas are fixed by the presence of stony sequences overlapping highly-tectonized, pelitic terrains. In contrast, conditions of relative lithological homogeneity can be attributed to the basins affected by coalescent earth flows. In particular, about 50 basins (Figure 11) have been identified that include over 390 events (size 2 mm count 49 %∼53 %; particles with its diameter 2∼0.075 mm, 43 %∼47 %. Powder cosmids with its diameter 80% of the vector data points.

QMS Quartz Mica Schist Unit GS Graphite Schist Unit ABC geological faults; T1 T2 T3 transverse cracks 1P1, 2A etc total station survey markers points used in the photogrammetric adjustment survey targets + survey check points

Figure 1. Outline of the landslide superimposed on a simplified geological map of the site; and survey points.

300mm

Figure 2. Several joint sets cutting the low-dipping foliation within the Quartz Mica Schist Unit.

less than 30 cm thick. The foliation strikes generally north and the orientation of the excavated face of the hillside is NNW-SSE (Figure 3). The foliation dips at shallow angles towards the east, i.e. into the slope. The rock sequence at site is cut by sets of preexisting faults. The most prominent fault set dips steeply towards the E to ESE and three of these faults can be traced across the landslide (Figure 1, faults A, B & C). The fault planes form counterscarps at outcrop on the landslide and have oblique and vertical striations, suggesting distinct phases of slip. The faults also

Figure 3. Stereographic projection of poles to joint planes and foliation and surface displacement vectors.

show signs of recent but pre-landslide movement and have been reactivated during landslide movements. The schists at outcrop are highly jointed with typical joint spacing less than 0.5 m (Figure 2). The poles of the joints form a girdle that is roughly orthogonal to the low-dipping foliation (Figure 3). Where unweathered, the schist is generally strong to very strong and most of the rock material currently exposed across the site can be classified as ‘slightly’ to ‘moderately weathered’ (British Standards Institution, 1999: Figure 19). The main surface features of the landslide are the main scarp, the head graben, the north and south flanks, the counterscarps of the oblique faults (A, B & C etc) and a low-angle push-out structure at the toe zone (Figure 1). Neither the northern flank, which is partly concealed beneath an earthflow, nor the toe of the basal slip surface can yet be fully delineated. The term ‘toe zone’ is therefore used here in preference to the word ‘toe’. The head graben is crossed by multiple high-angle internal shears with counterscarps at outcrop (T1, T2 & T3 etc). 3

DEFORMATION MONITORING BY TOTAL STATION

Monitoring of the landslide has been carried out by the road contractor since October 2003. The work

458

Plane of foliation

Displacement

Figure 4. Displacements and velocities from total station monitoring; and rainfall at Stesen Kajicuaca Cameron High-lands.

involves nominally weekly measurement, by total station (Sokkia SET5E), of distances and horizontal and vertical angles from base stations west of the road to reflective markers installed on the landslide. The plan co-ordinates and reduced levels of the markers are computed from these data. The magnitude, dip and dip direction of the displacement of each survey marker have been calculated and velocities of movement have been determined. Some of the monitoring data are presented in Figure 4. Uncertainty in the data may be assessed by examining the reported movements of a marker located above the crown of the landslide (Marker 2D). In contrast to the markers within the landslide, the reported changes in the position of Marker 2D are very small (save for an unexplained excursion in September and October 2006) and no systematic pattern is evident. The variance in horizontal position data to August 2006 (standard deviation of data = 45 mm) is about three times that expected from equipment error alone. The variance in height determination is as expected from equipment error alone. 4

TOPOGRAPHIC SURVEY AND PHOTOGRAMMETRIC MEASUREMENTS

A digital elevation model had been created from topographic survey in November 2003 and another was made photogrammetrically from aerial photographs taken from a helicopter in September 2005. Displacement vectors were constructed from the differences between the two digital elevation models at identifiable features such as the ends of drainage channels and berm edges. Many of the 150 displacement vectors are shown in Figure 5. Uncertainty in these measurements is associated with the coordinates of the ground control points used

Figure 5. 3D visualisation of the excavated hillside showing the surface displacement vectors (2003–5) to scale. Inset: surface movement of the central toe zone block relative to foliation.

for photogrammetric adjustment and errors in the digital camera system. The photogrammetric survey was compared to the November 2003 survey at 56 survey check points in areas thought not to have moved between 2003 and 2005 (see Figure 1). The error standard deviation is 0.2 m. Uncertainty in the dip and dip direction of the vectors is less than 1◦ for the longest vector (24.3 m), a possible maximum of 13◦ for the shortest vector (1.6 m), and an average of 2.7◦ for the mean vector (8.0 m).

5

POST-FAILURE LANDSLIDE MOVEMENTS

The surface displacement vector data (2003–5) advance our understanding of landslide behaviour. Viewed in plan the vectors are seen to be normal to the slope face contours with lateral extension revealed by radial divergence (north-south spreading), conforming to topography, which takes the form of a subdued ridge. Movements are greater at the head than in the toe zone (compression) and, on any slope face contour, displacements are greater in the north than in the south (rotation). Viewed in cross-section the vectors are seen to plunge at the head of the landslide, to generally lie sub-parallel to the slope in the upper main body and to emerge in the toe zone. It is instructive to examine the disposition of vectors along the centreline of the displaced mass by means of stereographic projection (Figure 3). The directions of vectors within 100 m wide blocks of ground at the highest part of the head (mean 238.5◦ ) and at the toe zone (mean 246◦ ) closely correspond to the dip direction of the face (mean 245◦ ). The disposition of vectors

459

at the highest part of the head coincides with a concentration of joint planes (Figure 3 – X) and at the central toe zone block corresponds to the attitude of the foliation (Figure 3 – Y and Figure 5 inset). The vectors reveal significant downslope compression. Compressive strain (defined as the displacement normalized against downslope length) measured on centreline between upper main body (at the elevation of Markers 2A-2C) and the toe zone is about 5% (2003–5). Such compression is evident in small-scale sliding on foliation seen as shear offsetting (‘kicking out’), especially in the southern part of the landslide, and by slip on the reactivated faults A, B & C etc. Observed fault slip movements are dextral, increase to the south and are greatest on fault B, where slip at the centreline is 3.5 m. The total station data give further insights into landslide behaviour. Whilst the overall rate of displacement is declining slightly year on year, for much of the time the displaced mass appears to be either accelerating or decelerating. Five surges are apparent (Figure 4) and comprise an accelerating phase (six to eight weeks) and a decelerating phase (two to three months). The velocity reached during surges at markers 2A-C is generally about 20 mm/day (greater in late 2004).

6

DISCUSSION

The nature of the basal sliding surface(s) is of interest. Evidence is given above of movement at surface stations which is parallel to joint planes at the highest part of the head and to foliation in the central toe zone; slip on foliation is also visible on the ground. It may be inferred, if the effects of non-parallel internal shear and change in landslide thickness are assumed insignificant, that the landslide is sliding on joint planes at the highest part of the head (i.e. at the main scarp ‘normal fault’) and sliding upwards on foliation in the central toe zone (but oblique to dip, Figure 5 inset). The vectors plunge steeply at the head and emerge sharply in the toe zone, the profile suggesting a non-circular basal slip surface (Figure 5). The presence of multiple counterscarps in the head graben (T1, T2, T3 etc. Figure 1) may signify curvature of the basal slip surface (Hutchinson, 1988). There are joints disposed to facilitate slip on such a curved surface (Figure 3 – dashed oval). The landslide is probably a compound slide. An educated guess was made about the geometry of the basal slip surface, using the surface station movements and crack patterns, and estimates were made of landslide volume.

It appears that the volume of the landslide is about 2 million m3 . After failure the landslide decelerated until March 2004 and it has since continued to move, for much of the time accelerating and then decelerating in surges. The timing of the surges generally coincides with peaking in the 30-day rolling rainfall (Figure 3), rainfall being measured at the Stesen Kajicuaca Cameron Highlands raingauge of the Malaysian Meteorological Service, 13 km SSE of the site. The bimodal rainfall pattern shown in Figure 4 is characteristic of an inland climatic regime in peninsular Malaysia. It may be that the landslide is responding to rainfall-induced seasonal rise and fall of groundwater levels. Such fluctuation is manifest by intermittent seepage from the southern toe zone. Other causal factors may have contributed to surges: a surge in late 2004 concurred with the removal of 100,000 m3 of ground from the northern toe zone of the landslide.

7

CONCLUSIONS

The landslide is a slow rock slide in schist. Failure occurred in September 2003 and by December 2006 the head had moved downwards more than 21 m. The rate of displacement is declining slightly year on year, but for much of the time the landslide mass is accelerating and then decelerating in surges. There is some correspondence between the timing of the surges and the seasonal rainfall pattern. It is likely that the surges are induced by groundwater fluctuations. It may be inferred from surface observations that the failure involves sliding at the head and in the upper main body of the landslide on joints roughly orthogonal to the foliation, which dips at a shallow angle into the slope; in the central toe zone the landslide is sliding up and out on the foliation. The failure, which is probably a compound slide of volume about 2 million m3 , has reactivated major pre-existing faults that run obliquely through the landslide mass.

ACKNOWLEDGEMENTS The study of the Pos Selim landslide was carried out on behalf of the Slope Engineering Branch of the Public Works Department of Malaysia and facilitated by the road contractor, MTD Construction Sdn Bhd., who supplied the total station data. Prof YQ Chen of the Department of Land Surveying and Geo-informatics, Hong Kong Polytechnic University checked the displacement and velocity calculations and assessed the errors in the total station surveying.

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REFERENCES Andrew Malone Ltd 2007. Landslide study at Ch 23+800 Simpang Pulai-Lojing Highway, Malaysia. Report to Minister of Works of Malaysia. British Standards Institution, 1999. Code of Practice for Site Investigations BS5930:1999. Hansen, A. 2007. Semi-automated geomorphological mapping applied to landslide hazard analysis. Ph.D. Thesis,

Department of Spatial Sciences, Curtin University of Technology, 281p. Hutchinson, J.N. 1988. General Report: Morphological and geotechnical parameters of landslides in relation to geology and hydrogeology. In Proc Fifth International Symposium on Landslides (C Bonnard ed.) Balkema Rotterdam v1 3–35.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Characteristics of rock failure in metamorphic rock areas, Korea W. Park & Y. Han Architectural Engineering Team, Civil Part, Samsung Corporation, Seoul, Korea

S. Jeon School of Civil, Urban and Geosystem Engineering, Seoul National University, Seoul, Korea

B. Roh Technical advisor Team, Samsung Corporation, Seoul, Korea

ABSTRACT: The metamorphic rock ranges in Korea very extensively, which has the unique characteristics through crustal movement and complex metamorphic processes for a long time. So the metamorphic rocks have discontinuities such as dislocation and fault as well as weak part such as fractured zone and fault gauge. For these reasons, there exist great potential of collapses, so it is required great caution to construction and design roads or tunnels in this area. This paper describes the past experiences of slope failure during the construction of roads at the area of metamorphic complex, especially Gyong-gy gneiss, and analyzes the characteristics of the failure patterns in order to prevent from loss of the life and property. 1

INTRODUCTION

Length of the Korean Peninsula is about 1, 000 km and the area is 223,000 km2 , especially 70% of the land consists of the mountain. But despite of small area, there exist various rock types and the complex geological structures from Pre-Cambrian to Cenozoic era. As the characteristics of distribution of the rock, metamorphic rock of Cambrian period and plutonic rocks of Mesozoic exist extensively in the middle of the Korean Peninsula. Sedimentary rock and igneous rock which include stratum of Paleozoic and after that period scatter on these bedrock. According to roughly compositions of the rock, the metamorphic rock is 40%, igneous rock is 35% and sedimentary rock is 25%, and in the report of slope failures, mostly rock collapses were happened in metamorphic and sedimentary rock area in the road constructions. Actually, many losses of life and property happened from repetitive slope failure. Therefore, it is important to understand geological characteristics and collect data of the failure cases of the each rock type in order to avoid natural disaster and failures of artificial structures of rock. For this purpose, this paper described characteristics of failures in metamorphic rock area.

the type of gneiss and schist in the middle region and sporadically puts in the whole country. As it is well known, metamorphic rocks have many complicated geological structures by the movement of crustal and metamorphism processes for a long time, so they have irregular discontinuities, rock fractured, various scale fault and weak gouge. In case of the fault in these rocks, the size is not large scale, but considering direction and location of fratures, it is not easy to determine and predict. Discontinuities such as joints and foliations are very complex and these crevices are infilled by gouge and weak deposit material. So on excavating slope or pit

Gyonggy Province. Seoul

Mokpo

1.1 Geological characteristics of Korea Figure 1 shows the geological map of South Korea. Most of metamorphic rock of Pre-Cambrian exists as

Figure 1.

463

The geological map of Korea.

Busan

and tunneling in rock, the failures were frequently happened by the disadvantage geological structures. The sedimentary rock which is mostly mud stone, shale and sand stone from Cenozoic era ranges in the south-east region, and most of the these rock, except the Pohang’s mud stone, are very hard and mass state. This rock sometimes has some problems when the excavation directions are the same with the dip direction of sedimentary rock. Many engineers know how to design in this rock type, so comparatively the damages are low, but if the failures have happened this area, the loss would be largely. In case of the igneous rock, even if the rock is intrusive rock or eruptive rock, it can be seen easily in Korea. Representative igneous rock is granite which ranges very largely in Seoul and Kyung sang Province and mostly eruptive rock such as the andesite and tuff exists in south region nearby Mokpo. The characteristics of this rock failure are that the many failures were happened after cutting the slope by erosion and weathering. As the rock types, there are unique characteristics of the movements, so for designing the slope or tunnel, it is very important to apply the appropriate geological structures.

1.2 Statistics of the failures About the slope failures, the important things, except geological problems, are related with the point time of excavation and elapsed time. Figure 2 shows the data of slope failures as the elapsed time of the slope excavation at highway construction site in Korea (You, 1997). As seen in the graph, though initial collapses were relatively rare, many slope failures occurred shortly after cutting slope and opening to public, totally 85% of the failures were these times.

Many failures in metamorphic rock area happened immediately after cutting, and then as the time goes, failures of sedimentary rock occurred much more. And in case of the igneous rock, the slope became weak by erosion of the rock and fast weathering, therefore swallow face-failure and the scour were easy to happen on the slope. On exposing the slope face, weathering velocity of some ingredients of the rock mass becomes much faster and the rainfall and underground water infiltrate into discontinuities. These unpredicted collapses brought many losses of the construction period and the additional cost. The important points of these accidents were that most of the failure could not predict in design. This paper is the case study about investigation of the collapse of soil and rock slope by cutting slope in the metamorphic rock site which is located at Gyong-gy Province and analyses the pattern of the failures.

2 2.1

Geological characteristics

This investigation area was located at the Gyong-gy Province near the Seoul and there was the Route 45 which was a new lying road in the South-North direction, especially most parts of the route passed the mountain area, so major works were tunneling and cutting slope. As characteristics of the geology, most of rock at this area was metamorphic rock, called Gyong-gy gneiss which is the representative metamorphic rock in South Korea. This rock has unique characteristics of the geological structures through crustal movements and metamorphism processes for a long time. Joints, dislocations, fault and weak part such as fractured zone and fault gauge were existed, and till some of depth, discontinuity surfaces were covered by ferrite oxide and very weak clay, also it is hard to predict the direction of the discontinuity. Figure 3 is the cut slope with containing the fault which parallels the strike of slope. Discontinuities of these structures were common in this site, because the fault is the same direction with the road. In the other side, upper and side of slope consisted of residual soil and weathered soil, and some of the complex dyke rock. Figure 4 shows the plane failure and wedge failure which was two persistent joint with the line of intersection of the joints daylighting at the rock face. 2.2

Figure 2. Ratios of slope failure in term of elapsed time from cutting slope for there different rock types in Highway construction sites in Korea (You, 1997).

CHARACTERISTIC OF THE SITE

Design and construction the slope

In order to design, designers need many data of the site such as geotechnical investigation data and the detailed geological structures. But it is hard to get the all data

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Table 1.

Standard Guide for slope’s dip.

Condition

Height

Dip

Remarks

Soil

0∼5 m Above 5 m

1:1.2 1:1.5

Establish the 1 m ramp per 5 m height

Rock

Weather rock Hard rock

1:1.0 1:0.5

Establish the 3 m ramp per 20 m height

Fault

a

b

Figure 3. The exposed fault in the slope and this fault is the same direction with the Yong-in fault.

Figure 5. Stereographic Analysis; a) Discontinuities at the first design. b) after cutting.

Figure 4. Exposed discontinuity which can be collapsed in the slope.

in case of the construction site of the road, owing to the limit of the investigation, so the standard of the slope design, which need not the precious data, is used very often. The standard of slope design is practically the same as related organizations of the civil construction, and most of the design criteria are based on the strength of the rock without the rock type. But if the designer used only these guides no concerning the geological information, many problems can be occurred, and a few organizations recommend that the characteristics of the geological structures are considered on the slope, if need. Table 1 shows the standard guide for design of slope at the Korea Expressway Corporation & Ministry of Construction and Transportation. This guide can be used comfortably in case of lack of the geotechnical investigation data, but there are many potentialities of the dangerous failure, but in case of being applied to importance cutting slope. As the briefly examples, the dips of the slope in this investigation site were determined by the standard guide such as Table 1 without considering direction of discontinuity and geological structures. The Figure 5a shows stereographic analysis of the slope at the initial design. As seen the result of the stereographic analysis, the dangerous block could not be found, but as the face mapping on slope after the

slope was finished the excavation, unpredicted fault and joint were existed in the slope. Actually this slope was happened the wedge failure and small scale plan failure by the discontinuities. So in case of design the slope with the standard guide, it is simple to use it, but it is always not safe. 3

CHARACTERISTIC OF FAILURE IN THIS SITE

3.1 Characteristic of the slope failure The rock failures are influenced on many factors such as the discontinuities, strength of rock and the time of construction. In case of the metamorphic rock, there are many collapses after finishing the excavation as result of the investigation. These results were the same of the investigation which was performed by Korea Expressway Corporation. Table 2 shows results of investigation of the relation rock type which was classified by the rock strength and elapsed time of the excavation. 66% of slope failure happened after finishing excavation, and 85% of the failure occurred in the rock masses which were above strength of weather rocks. These reasons that the more slope was cutting, the much free surface in slope face was enlarged. As it is well-known, the free surface gives the many influences in the slope. By the investigation, the causes of many failures were by the discontinuity of the slope, especially fault and joint. On the other side, the failures of residual soil or completely weathered soil, which had no geological

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Table 2. Number of slope failure. Rock R.S W.R S.R/H.R Total

On construction 3 2 7 12 (34%)

a

On complete

After project

Total

1 5 15

1 − 1

5 (15%) 7 (20%) 23 (65%)

21 (60%)

2 (6%)

35

∗ R.S

is residual soil; W.R is weathered rock; S.R is soft rock; H.R is Hard rock.

structure patterns, were caused by the soil weight, and most of failure shapes were circular type in this part. Figure 6a shows the circular failure, and it is typical circular failure by the weight of soil mass. However geological structures such as dyke rock and foliation are faintly reminded in slope on account of the different weathered speed, so failures in soil occasionally happened like the type of rock mass failures, especially in the metamorphic or sedimentary rock area. Figure 6b shows the circular failure by dyke rock. Dyke rock which did a role as discontinuity which divided into some soil layers, if this discontinuity was the same direction with the cut slope, circular or other shape failure would be occurred. As seen the result of the Table 2, soil failures did not often happen, but 2-case happened on the finished excavation and opening the public, the three failures happened the on the construction. In case of weak zone in the slope, the characteristic of the movement was same as soil rather than rock, so it is hard to predict the slope failure in the area where the fractured zone and mixed fault. As seen the table 2, collapse of 85% is related with rock, and large scaled failures were caused by the fault and gouge which is called Yong-in fault and this discontinuity was put to South-North direction. Even if the fault were so small scale, when the fault was created, nearby the crust was move at once, so short and dangerous discontinuities were extensively existed at this area. And as another reasons, there were fault gouge and infilled materials into the gap of joint. Figure 7 shows the representative large scale failure before starting to excavate in this area. The height of the mountain is about 70 m and width is 250 m. The main cause of the collapse was fault which existed in slope of lower part, and there was a thin thickness of the fault clay between discontinuities. The strike of fault paralleled with direction of slope, so mostly mode of failures were plan failure. Due to this failure, 5 m of settlements and over the 50 cm of tension crack were occurred for several weeks. Because this failure was large scale and the

b

Weathered Acidic D yke

Figure 6. Soil failure; a. The circular failure through the line of least resistance by soil weight. b. The circular failure by dyke.

Figure 7. This picture is large slope failure in the this site. This slope’s height is 56 m and the width is 120 m. a. view of the slope. b. Vertical discontinuity exposed by ground settlement at the failure. c. Tension crack on the ridge.

problem of compensation with inhabitants around the site, it was hard to do additional cutting and lots of monetary problems were happened. The sources of these problems were lack of properly geotechnical investigation, especially investigation of the fault and dangerous discontinuities is not easy and in usually case, small fault does not consider to design.

466

Table 3 shows classification of the slope failure as shape of the failures and number of the occurrences are referred to table 2. As above mentioned, because the discontinuity are the main cause, failures are happened the rock which related the discontinuity. According to the investigation of failures in the site, there were several cases of wedge and plane failure but fortunately the toppling failure did not happen. On the excavation, some of topping failure blocks were in the slope, but reinforcement methods such as the bolt and rock anchorage were used to prevent from failure through using. Figure 9 shows the plane failure and wedge failure after cutting and passed after 1year in this site. Height

of slope was 25 m and width that collapse occurs was about 50 m. The main cause was the unexpected fault which of dip of the fault was N40◦ E/63◦ SE. Figure 10 is the stereographic analysis of the Figure 9. As seen the results, the direction of the joint were changed as the construction, and could not reflect the fault at the first design because there were no data about the fault. As seen the Figure 10, although there were many dangerous blocks in slope at the initial condition, but designer used the standard guide, then the collapsed happened.

a

b

Figure 9. View of the failure in the slope: a: Slope shortly after cutting; b: Wedge failure in the slope; c: Fault gouge with 50 cm of width and slickenside in the slope; d: Tensile Crack in the ramp.

a

Figure 8. Rock face formed by persistent discontinuities: a. plane failure formed by bedding planes parallel to the face with continuous length over the slope (Gneiss on Route 45 near Yong-in); b. wedge failure formed by two intersecting planes dipping out of the face.

b

Table 3. Types of the slope failure. Type of failure

Circle failure

Planar failure

Wedge failure

Toppling failure

No. of occurrence

5

19

11

−

Figure 10. Stereographic analysis of the discontinuities plane at the construction site: a. is state of the discontinuity before digging and b. is result of the face mapping after 1st cutting. Dip of the fault was 63 degrees.

467

If more accurately investigation data of this site were existed in design, failure could be prevent by using a few rock bolts. 4

CONCLUSION

In this paper, we investigated the characteristics of slope failure in the metamorphic rock area which is called Gyong-gy gneiss. This rock mass has many unpredicted discontinuities from the metamorphism process and crustal movement for along time. Therefore most of the road and tunnel constructions of this region, many failures were happened by the geological structures. 85% of the slope failures were concerned discontinuities of the rock and most of the failures happened after cutting the slope or opening the public. As these reasons, the exposed discontinuities were revealed on the slope after excavation. Also in case of the soil parts, some of the failures were induced by the reminded geological structures such as fold and foliation, especially dyke rock. As the problems of the design, the designer depends on the standard guides of the slope without no considering the characteristic of the geology, even if there are many studies of slope stability.

Therefore, in order to prevent from the rock failure, it is important to collect the more data such as accurate geotechnical investigation which include the characteristic of the geological structures and to evaluate slope stability by the face mapping after cutting, and through the international research on the slope design, the more stable and rationally design method must be established. REFERENCES Lee Dae-Sung (edited). 1987. Geology of Korea. South Korea. Hoek, E. & Bray, J. 1977. Rock Slope Engineering, 3rd edn, IMM, London. Hoek, E. & Bray, J. 2004. Rock Slope Engineering, 4th edn, IMM, London. Korea Expressway Corporation. 1996. Highway slope manual. Park, W.S. 2002. The report of the slope stability in Yongin, Samsung Corp. You, B.O. 1997. A Study on Harzad Rating System and Protective Measures for Rock Slope, PH.D Dissertation, Dept. of Civil Eng., Hanyang University, Seoul, South Korea.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Shape and size effects of gravel grains on the shear behavior of sandy soils S.N. Salimi, V. Yazdanjou & A. Hamidi Tarbiat Moallem University, Tehran, Iran

ABSTRACT: The shear behavior of sandy soils containing gravel particles has been investigated by many researchers. However, the effects of the shape and size of gravel particles have not been particularly evaluated. In the present study the shear strength of sand-gravel mixtures with two different gravel grain sizes and shapes is studied in loose, medium and dense states using large direct shear test. The results of this study indicate that the gravel shape and size has a little effect on the shear strength of sand-gravel mixtures in low gravel contents. Increasing the gravel content to higher values makes this effect more clear. This is more obvious when the gravel particles are no more floating in the sandy soil matrix. The samples containing angular gravel particles generally show higher shear strength and dilation compared to the mixtures containing rounded to sub-rounded gravels. Also the samples containing larger gravel particles usually show more shear strength and dilation compared to the samples with smaller gravel particles in the same gravel content.

1

INTRODUCTION

Fragaszy et al (1990, 1992) introduced a new method to evaluate the shear strength of soils containing oversized particles. This method was based on the assumption that larger particles floating in a matrix of finer grained material do not significantly affect the strength and deformation characteristics of mixture. In other words, while oversized particles are floating in the finer matrix without any contact, the strength and deformation characteristics are controlled by the matrix part alone. However, for higher oversized contents, it is controlled by both the sand matrix and oversized particles. Therefore, in a floating state the behavior of the soil containing oversized particles can be simulated by testing the matrix portion alone, provided that the model specimen is prepared in near field density. The near field density is the density of matrix in vicinity of oversized particles (Fragaszy et al. (1992)). Based on this concept the shear strength of granular soils containing oversized particles reduces by increase in oversized content provided that the relative density of the mixture remains constant. Yagiz (2001) investigated the effects of the shape and content of gravel particles on the shear strength of fine sandy soils using direct shear tests. It was concluded that the shape and content of gravel particles have important effects on the friction angle of the mixture.

Simoni and Houlsby (2004) performed 87 large direct shear tests on sand-gravel mixtures with different gravel contents. They concluded that increase in gravel content enhances the dilatancy rate and the critical state friction angle. In the present study a regular set of 84 large direct shear tests were performed using a 300×300×170 mm direct shear box apparatus to investigate the effects of the shape and size of oversized particles on the shear behavior of sand-gravel mixtures. 2

SAMPLE PREPARATION

A bad graded fine sand with sub-rounded to rounded grains was used as the base soil. Also two gravel types were used as the oversized particles. The first type was river gravel with rounded to sub-rounded grains and the second type was an angular to sub-angular one. Two different gradations with maximum grain size of 12.5 mm and 25 mm were considered for each gravel type. Each one was mixed with the base sand in different contents to prepare sand-gravel mixtures. The maximum and minimum void ratios of the mixtures were measured according to the ASTM-D4253 and ASTM-D4254. Also the specific gravity of the sand and gravel grains was measured as 2.74 and 2.64 respectively according to ASTM-D854. The dry unit weights of different mixtures in desired relative densities were computed. The weight of the soil required

469

Figure 1. Shows the shear stress-shear displacement curve for two mixtures containing 20% gravel of maximum size of 12.5 and 25 mm.

Figure 2. Vertical displacement-shear displacement curves of mixtures containing 20% of different gravel sizes.

2 shear stress (kg/cm2)

to make samples in a specific relative density was selected. The gravel and sand portions were mixed based on the desired weight percents. The whole mixture was divided into three equal fractions and the soil sample was made in three layers by purring the first fraction and to compact it using a metal hammer in order to fill one third of the shear box height. The two other layers were placed using the same procedure.

1.5

1

0.5

0

3

EFFECT OF THE GRAVEL SIZE ON THE SHEAR BEHAVIOR OF SAND-GRAVEL MIXTURES

surcharge: 1.5kg/cm2 Dr: 60%, gravel content: 40%

max. size: 0.5 inch max. size: 1.0 inch 0

10 20 30 shear displacement (mm)

40

Figure 3. Shear stress-shear displacement curves of mixtures containing 40% of different gravel sizes.

The tests performed under an overburden pressure of 1.5 kg/cm2 on samples in a relative density of 60%. As it can be seen the shear stress-shear displacement curves are almost coincided for both mixtures. Although the peak shear strength of mixture containing larger gravel grains is a little higher, the residual strengths of both mixtures are the same. In fact in low gravel contents, the gravel grains are floated in the sand matrix and there is little contact between them. Therefore the shear behavior of the mixture is controlled mainly by the sand portion and the gravel grain size does not play an important role on the shear behavior of the mixture. The variation of vertical displacement with shear displacement for the above mentioned mixtures is shown in Figure 2. It can be observed that the vertical displacement of the mixture containing gravel grains with maximum size of 25 mm is more than the one containing gravel grains with maximum size of 12.5 mm. It seems that the increase in gravel grain size increases the vertical displacement or the dilation of mixture due to the more vertical movement and slip or topping of the larger gravel particles over each other. The more dilation of the mixtures containing larger gravel

grains is the main reason for higher shear strength in comparison to the mixtures with smaller gravel grains. Mixtures with 40% and 60% gravel contents were prepared to investigate the effect of gravel size on the shear behavior in higher gravel contents. Figure 3 shows the shear stress-shear displacement curve for samples with 40% gravel content. It can be seen that the difference between the curves has been increased for different gravel grain sizes compared to the curves for mixtures with 20% gravel content. This is mainly due to the increase in the contacts between gravel grains in the mixture in higher gravel content. The contact forces between particles are more for larger gravel grains compared to the smaller ones due to the more contact surface that leads to higher shear strength. The same comparison was made considering the vertical displacement-shear displacement curves. It was concluded that larger grains cause higher vertical displacement or dilation. The difference between dilation of mixtures with different gravel grain sizes increases with increase in gravel content. Figure 4 indicates this case for the mixtures with 40% gravel content. The differences of shear strengths and

470

dilation of the mixtures with different gravel grain sizes are more obvious when the gravel content is 60% in mixture. This is shown in Figures 5 and 6 which indicate the shear stress-shear displacement and vertical

displacement-shear displacement curves for samples with 60% gravel content. 4

surcharge: 1.5kg/cm2 Dr: 60%, gravel content: 40%

vertical displacement (mm)

2.5

EFFECT OF THE GRAVEL SIZE ON THE FRICTION ANGLE OF SAND-GRAVEL MIXTURES

The variation of friction angle for mixtures containing gravel grains of different size prepared in different relative densities is shown in Figure 7. It can be concluded

2

1.5 1

max. size: 0.5 inch

0.5

max. size: 1.0 inch 0

-0.5

0

10

20 30 shear displacement (mm)

40

Figure 4. Vertical displacement-shear displacement curves of mixtures containing 40% of different gravel sizes.

Figure 5. Shear stress-shear displacement curves of mixtures containing 60% of different gravel sizes.

Figure 6. Vertical displacement-shear displacement curves of mixtures containing 60% of different gravel sizes.

Figure 7. Friction angle of mixtures containing gravel grains with maximum size of 12.5 mm and 25 mm.

471

that the increase in gravel content leads to the increase of friction angle for all the mixtures. Besides the mixtures containing larger gravel grain particles show higher friction angle. Also the difference between the friction angles of mixtures containing different gravel size increases in higher gravel contents. 5

EFFECT OF THE GRAVEL SHAPE ON THE SHEAR BEHAVIOR OF SAND-GRAVEL MIXTURES

In order to study the influence of gravel shape on the shear behavior of sand-gravel mixtures, a set of direct shear tests have been conducted on mixtures containing angular and rounded gravel grains. It was observed that in low gravel content of 20%, the shear stress-shear displacement curves of mixtures containing angular and rounded gravel grains are nearly coincided as shown in Figure 8. However, the vertical displacement of the mixture containing angular gravel grains is a little more than the one containing

rounded gravel grains as shown in Figure 9. In fact the angularity of gravel grains results in more vertical displacement or dilation. This is due to the increase in overtopping of gravel grains during shearing. Figures 10 and 11 indicate the shear stress-shear displacement and vertical displacement-shear displacement curves for mixtures with 40% of different gravel shapes. It can be observed that the effect of gravel shape on the peak and residual shear strength and dilation of sand-gravel mixtures are more obvious in higher gravel contents. The more shear strengths of mixtures containing angular gravel grains are mainly due to the higher dilation occurs in these mixtures compared to the mixtures containing rounded gravel grains. The same trend can be seen when the gravel content increases to 60% as shown in Figures 12 and 13.

Figure 10. Shear stress-shear displacement curves of mixtures containing 40% of different gravel shapes.

Figure 8. Shear stress-shear displacement curves of mixtures containing 20% of different gravel shapes.

Figure 9. Vertical displacement-shear displacement curves of mixtures containing 20% of different gravel shapes.

Figure 11. Vertical displacement-shear displacement curves of mixtures containing 40% of different gravel shapes.

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Figure 12. Shear stress-shear displacement curves of mixtures containing 60% of different gravel shapes.

Figure 14. Friction angle of mixtures containing angular and rounded gravel grains.

For mixtures containing angular gravel with maximum grain size of 12.5 mm in a relative density of 60%: φ = 0.32Gc + 24.7 R2 = 0.99

(2)

In the above equations Gc is the gravel content in percent and φ is friction angle of the mixture in degrees. Figure 13. Vertical displacement-shear displacement curves of mixtures containing 60% of different gravel shapes.

6

EFFECT OF THE GRAVEL SHAPE ON THE FRICTION ANGLE OF SAND-GRAVEL MIXTURES

The variation of friction angle for mixtures containing angular and rounded gravel grains in different gravel contents is shown in Figure 14 in a relative density of 60%. The figure shows that increase of the angular gravel content in sandy soil increases the friction angle of the mixture more compared to the rounded gravel. A relationship can be determined for mixtures containing rounded and angular gravel grains to estimate the friction angle of the mixtures containing certain amount of gravel using a linear regression as shown in equations (1) and (2). These equations may be used to determine the friction angle of sand-gravel mixture with a relative accuracy. For mixtures containing rounded gravel with maximum grain size of 12.5 mm in a relative density of 60%: φ = 0.24Gc + 24.7

R2 = 0.99

(1)

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7

SUMMARY AND CONCLUSION

A regular set of direct shear tests performed on mixtures of sand and different types of gravel grains to investigate the effects of gravel grain shape and size on the shear behavior of sand-gravel mixtures. It was concluded that the addition of each type of gravel grains to the sandy soil increases the shear strength and dilation of the mixture intensively. However, the addition of larger gravel grains increases the shear strength and dilation of mixture more compared to the smaller ones. Also the difference between shear strengths of mixtures containing gravel grains of different size is more obvious in higher gravel contents. The same results obtained for mixtures containing gravel grains with different angularity. Addition of angular gravel grains increases the shear strength and dilation of the mixture more than rounded gravel grains. The effect of gravel grain angularity is more obvious in higher gravel contents. The friction angles of the mixtures containing different amounts of gravel content were also studied and it was understood that the increase in gravel content results in more friction angle. However, the friction angle of the mixtures containing angular gravel grains is higher compared to the one containing rounded gravel grains in equal gravel content. The same result obtained for friction angle of mixtures containing larger gravel grains compared to the mixtures

containing smaller gravel grains. For equal gravel content, the friction angle of the mixture containing larger gravel grains is higher than the one containing smaller gravel grains. REFERENCES Fragaszy, R.J., Su, W. & Siddiqi, F.H. 1990. Effect of oversized particles on the density of clean granular soils. Geotechnical Testing Journal 13(2): 106–114.

Fragaszy, R.J., Su, W., Siddiqi, F.H. & Ho, C.L. 1992. Modeling strength of sandy gravel. Journal of Geotechnical Engineering Division, ASCE 118(6): 920–935. Yagiz, S. 2001. Brief note on the influence of shape and percentage of gravel on the shear strength of sand and gravel mixture. Bulletin of Engineering Geology and the Environment 60(4): 321–323. Simoni, A. & Houlsby, G.T. 2006. The direct shear strength and dilatancy of sand-gravel mixtures. Geotechnical and Geological Engineering 24(3): 523–549.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Nonlinear failure envelope of a nonplastic compacted silty-sand D.D.B. Seely IGES, Inc., Salt Lake City, Utah, USA

A.C. Trandafir Department of Geology and Geophysics, University of Utah, Salt Lake City, Utah, USA

ABSTRACT: Accurate evaluation of shear strength of compacted fill materials represents a key issue in performing reliable stability analyses of embankment slopes. In engineering practice, the shear strength of compacted soils is expressed in terms of a linear failure envelope usually derived from consolidated-undrained (CU) triaxial tests with pore water pressure measurements. Experimental evidence indicates, however, that the failure envelope of many soils is not linear, particularly within the range of low effective normal stresses. The present study addresses the nonlinear character of failure envelope of a nonplastic compacted silty-sand. Linear and nonlinear strength functions were fitted to experimental results from CU triaxial tests to derive the strength parameters of linear and nonlinear failure envelopes. For the investigated soil and typical range of mean effective normal stresses of interest in embankment slope design, the nonlinear failure envelope appears to give a more accurate representation of the experimental information compared to the linear strength model.

INRODUCTION

The importance of incorporating the nonlinear character of soil failure envelope in slope stability computations has been demonstrated by two dimensional limit equilibrium (LEM) and finite element (FEM) analyses (Trandafir et al. 2000) as well as three dimensional slope stability studies (Jiang et al. 2003). Previous research has shown that failure envelopes of various soils (e.g., silt, sand, compacted clay, and compacted rockfill) exhibit nonlinearity especially within the range of small normal stresses (Penman 1953, Ponce & Bell 1971, Charles & Soares 1984, Day & Axten 1989). In this context, the present study is concerned with the nonlinear character of the failure envelope of a compacted silty-sand representing potential embankment material for earthworks throughout Utah. Experimental data from a series of consolidated-undrained (CU) triaxial tests with pore pressure measurement are utilized to derive both the linear and nonlinear failure envelopes of the analyzed soil, and the results are discussed. 2

D 422-63). A compaction test (ASTM D 698-00) was also conducted in order to obtain the optimum water content associated with the maximum dry unit weight of the material. The results of the grain size analysis are shown in Figure 1, whereas the proportion of clay, silt, and sand characterizing the soil are depicted in Figure 2. Based on the laboratory test results, the analyzed soil classifies as a nonplastic silty-sand (SM) according to the Unified Soil Classification System (USCS) (ASTM D 2487-06). The soil consists of 66.4% sand, 30% silt, and 3.6% clay (Figure 2). The compaction test yielded an optimum water content of 16.14% and maximum dry unit weight of 100

Percent finer by weight

1

80 60 40 20 0 0.001

PRELIMINARY SOIL TESTING

Preliminary laboratory tests were performed on the soil used in this study for the purpose of geotechnical classification, including liquid and plastic limits (ASTM D 4318-05), and grain size analysis (ASTM

0.01

0.1 Grain size (mm)

1

10

Figure 1. Grain size distribution curve of the silty-sand used in the experimental study.

475

Sand 0

100

20

80

40

60

60

40

80

20

100

Clay

0

0 20

40

60

80

100

Silt

Figure 2. Ternary diagram showing the percentages of sand, silt and clay in the analyzed soil.

Dry unit weight (kN/m3)

17.0 16.8 16.6 16.4 16.2 8

10

12 14 16 18 Water content (%)

20

22

Compaction curve ZAV Gs = 2.6 Triaxial test specimens

Figure 3. Results of the compaction test performed in accordance with ASTM D698-00. The water content-dry unit weight characteristics of the soil specimens used in the triaxial testing program are plotted on the same curve.

16.82 kN/m3 (Figure 3). These were the target compaction parameters used to prepare the soil samples for triaxial testing. 3

TRIAXIAL TESTING

A total of eight samples were prepared for triaxial testing. The soil was blended with water at a target water content equal to the optimum water content derived from the compaction test, and allowed to attain the steady moisture condition for a minimum of 16 hours prior to compaction. The soil specimens were hand compacted in the laboratory to approximately 100%

of the maximum dry unit weight obtained from the compaction test using 254 mm lifts, with the soil being scarified between lifts. The initial dimensions of the specimens were 61.37 mm in diameter and 152.4 mm in length. The samples were then trimmed to a height to diameter ratio between 2:1 and 2.5:1 prior to preparation in the triaxial cell. The moisture-dry unit weight data points of the compacted specimens for triaxial testing plot very close to the point of optimum on the compaction curve, as seen in Figure 3. The triaxial samples were percolated with CO2 to aid in saturation prior to the initialization of the triaxial tests as recommended by Rad (1984). The samples were backpressure saturated, ensuring a Skempton’s B pore pressure parameter of 0.95 prior to consolidation. A strain rate of 0.06%/min was used for undrained shearing, as derived from the time rate of consolidation data. The triaxial testing program involved CU axial compression (AC) and axial extension (AE) tests on the compacted soil specimens. The experiments were conducted using a state-of-the-art microprocessor controlled fully automated triaxial equipment manufactured by Geocomp Corporation (Dasenbrock 2006). The effective stress paths from triaxial tests are shown in Figure 4. Seven CU-AC samples were sheared up to 20–25% axial strain with average data sampling every 0.1% strain. The AC specimens demonstrated a dilative response yielding an effective confining stress at failure larger than the initial (consolidation) effective confining stress (Figure 4). Consequently, a CU-AE test (characterized by a contractive tendency of the soil before failure) was included in the experimental program to capture the soil strength behavior within the range of low effective normal stresses (Figure 4). The AE specimen was sheared up to approximately −10% axial strain with data sampling every 0.01% strain. 4

INTERPRETATION OF FAILURE IN TRIAXIAL TESTING OF SILTY SOILS

Failure in dense low-plasticity silty soils subjected to triaxial compression testing is difficult to define due to a continuous dilative behavior of these materials with increasing shear strain. Consequently, a variety of criteria used to define the onset of failure for triaxial tests can be found in the literature (Brandon et al. 2006). Three of these criteria have also been evaluated in the context of the triaxial test results shown in Figure 4, i.e., Skempton’s A pore pressure parameter = 0 (Brandon et al. 2006), peak deviator stress (σ1′ −σ3′ )max , and peak principal stress ratio (σ1′ /σ3′ )max . Figure 5 graphically demonstrates all three of the previously mentioned criteria and their applicability to triaxial test results in this study.

476

300 40 kPa AE 20 kPa AC 40 kPa AC 76 kPa AC 132 kPa AC 188 kPa AC 244 kPa AC 300 kPa AC

q (kPa)

200

100

0 0

100

200

300 p' (kPa)

400

500

600

Figure 4. Results of the CU-AC and CU-AE triaxial tests. The values in the legend represent the initial effective consolidation stresses for the triaxial tests.

300

q (kPa)

ESP 200 ESP

100

TSP

TSP

0 0

100

200

300 p' (kPa)

400

500

600

Peak deviator stress, ( '1- '3) max Peak principal stress ratio, ( '1 / '3)max A=0

Figure 5. Graphical representation of various criteria to define the onset of failure in triaxial tests. Both samples in the figure were sheared up to 20% axial strain. The A = 0 failure condition occurs where the effective stress path, ESP, and the total stress path, TSP, intersect.

Brandon et al. (2006) recommends the condition A = 0 to define the onset of failure in low-plasticity silts. As illustrated in Figure 5, the A = 0 condition seems an appropriate choice for the triaxial specimen with an initial effective consolidation stress of 132 kPa. It results in a large quantity of points on the failure envelope described by the effective stress path above the A = 0 stress point. However, the A = 0 condition is not achieved for some of the triaxial test results in this study (e.g., the triaxial specimen with an initial mean

effective consolidation stress of 300 kPa in Figure 5). For this reason, the A = 0 criterion was discarded. The disadvantage of the peak deviator stress (σ1′ −σ3′ )max criterion applied to the tested soil is that it results in only one point on the failure envelope, i.e., the strength value at the maximum shear strain attained at the end of the triaxial test (Figure 5). Compared to the other analyzed criteria, the (σ1′ −σ3′ )max criterion will provide the minimum number of experimental data points that can be used to derive the strength parameters characterizing the failure envelope. The criterion selected to define the onset of failure in this experimental study is the peak principal stress ratio (σ1′ /σ3′ )max . As seen in Figure 5, the stress point associated with (σ1′ /σ3′ )max is above the stress point corresponding to A = 0 failure condition for the sample with an initial effective consolidation of 132 kPa thus providing a smaller number of experimental data points characterizing the failure envelope in this case. However, unlike the A = 0 criterion, the (σ1′ /σ3′ )max failure condition is achieved in all of the performed triaxial tests (thus including the sample with an initial effective confining stress of 300 kPa in Figure 5 that did not achieve the A = 0 condition). Furthermore, the (σ1′ /σ3′ )max criterion appears to provide a significantly larger number of experimental data points describing the failure envelope compared to the (σ1′ −σ3′ )max criterion. Figure 6 demonstrates the range of experimental data points located on the failure envelope based on (σ1′ /σ3′ )max failure condition, in relation to axial strain, εa , for a triaxial test with an initial effective confining stress of 244 kPa.

477

4.0

300

3.0 2.5

q (kPa)

'1/ '3

3.5

Range of experimental data used in deriving failure envelope

2.0 1.5

200 100 0

1.0

0

0.0 2.5 5.0 7.5 10.0 12.5 15.0 17.5 20.0 a (%)

100

200

300

400

500

600

p' (kPa)

Figure 6. Principal stress ratio plot for sample with an effective confining stress of 244 kPa showing the range of data used in failure envelope regression analysis. The diamond point represents the peak principal stress ratio.

Power, q = 0.761p' 0.946 ; R2 = 0.999 Linear, q = 0.547p' + 2.70 ; R2 = 0.997

Figure 8. Linear and nonlinear failure envelopes obtained from fitting linear and nonlinear strength functions to the experimental data.

q (kPa)

300 200 100 0 0

100

200

300

400

500

600

p' (kPa)

Figure 7. Experimental stress points on the failure envelope according to the principal stress ratio failure definition.

By selecting the (σ1′ /σ3′ )max criterion to define the onset of failure in triaxial testing, a reasonably large number of experimental stress points describing the failure envelope was obtained. This will allow for a more accurate evaluation of the shear strength parameters characterizing the linear and nonlinear strength functions derived from regression analysis. Figure 7 displays the experimental failure envelope data from triaxial testing according to the (σ1′ /σ3′ )max failure condition.

5

LINEAR AND NONLINEAR FAILURE ENVELOPES FROM REGRESSION ANALYSIS

A regression analysis was performed using the experimental data set shown in Figure 7. Conventional least squares minimization methods were employed to derive the strength parameters for a typical linear kf function and a nonlinear empirical power law relationship. The resulting linear and nonlinear failure envelopes derived from the regression analyses

are shown in Figure 8. A high coefficient of determination R2 of 0.997 and 0.999 obtained for the linear and nonlinear regression analyses indicates that both strength functions provide a very good description of the experimental information which in fact covers a range of mean effective normal stresses, p′ , within 25 to 500 kPa (Figure 7). For the interval of very small mean effective normal stresses (i.e., p′ < 25 kPa), the nonlinear strength function predicts smaller strength values compared to the linear model (Figure 9). Pariseau (2007) has shown that both forward and backward extrapolation of a particular linear fit applied to the experimental data can overestimate the available shear strength of the material. This aspect is also demonstrated in this study by the q-axis intercept of the linear failure envelope, unlikely for a cohesionless nonplastic soil such as the analyzed silty-sand (Figure 9), and the deviation of the linear and nonlinear failure envelopes in the range of mean effective normal stresses above 350 kPa (Figure 8). Reliance on the R2 value alone to assess the accuracy of a strength function in describing the experimental information should be approached with caution. Pariseau (2007) pointed out that a common but erroneous procedure is to interpret the failure envelope with the greatest R2 value as the ‘‘best fit’’ failure criterion. Care must be taken in evaluating the ‘‘best fit’’ for the range of normal stresses available from laboratory testing as long as this range does not coincide with the interval of normal stresses for the design problem of interest. In case of compacted embankment slope applications, the mean effective normal stress typically does not exceed 200 kPa. However, the experimental results in Figure 7 are within a larger p′ interval (i.e., 25 to 500 kPa), thus the R2 parameter characterizing the linear and nonlinear strength functions in Figure 8 corresponds to this large range of mean effective normal

478

' '

Figure 9. Expanded view of the linear and nonlinear failure envelopes together with the experimental data in the range of small mean effective normal stresses.

stresses. Therefore, in order to investigate the accuracy of strength predictions within the interval of mean effective normal stresses of interest, an approach based on cumulative squares of residuals has been used. The residual for a given p′ value was calculated as the difference between the experimentally measured q value and q predicted by the linear or power strength functions in Figure 8. The cumulative squares of residuals was plotted against the mean effective normal stress (Figure 10) to demonstrate the ‘‘fit’’ of the two strength models within the range of stresses applicable to compacted embankment slopes (i.e., 0 to 200 kPa mean effective normal stress). As seen in Figure 10, for p′ within 0 to 200 kPa, the nonlinear strength model provides either a comparable or a much lower value of the cumulative squares of residuals compared to the linear strength function. Thus, it may be concluded that for the analyzed compacted silty-sand and for the range of mean effective normal stresses of interest in this study, the failure envelope described by the nonlinear strength function is a better representation of the actual strength of the soil.

6

CONCLUSIONS

Linear and nonlinear strength functions were fitted to laboratory triaxial test data to derive the failure envelope of a compacted silty-sand. For the range of experimental mean effective normal stresses, both linear and nonlinear strength models provided equally

Figure 10. Demonstration of the ‘‘best fit’’ failure criterion for the given range of stresses using the cumulative residual2 . The nonlinear failure criterion provides a more accurate representation of the experimental data, having an overall lower cumulative residual2 for the given range of stresses.

valid descriptions of the experimental information. However, a closer examination of the accuracy of predictions within the interval of mean effective normal stresses of interest for compacted embankment slopes revealed a better ‘‘fit’’ of the nonlinear strength function to the triaxial data. The derived nonlinear failure envelope also predicted lower strength values than the linear strength model within the range of very small normal stresses uncovered by the experimental information. This departure from linearity is usually

479

associated with smaller computed safety factors in a conventional slope stability analysis based on the nonlinear strength model, implying therefore a safer slope design. ACKNOWLEDGMENTS Funding for this research was provided by the Undergraduate Research Opportunity Program (UROP) at the University of Utah. The authors would like to thank IGES, Inc. for open access to their extensive laboratory facilities that were used in the present experimental investigation. REFERENCES ASTM 2007. Standard D 4318-05: Standard Test Methods for Liquid Limit, Plastic Limit, and Plasticity Index of Soils, Annual Book of Standards, Vol. 4, ASTM International, West Conshohocken, 2007. ASTM 2007. Standard D 422-63: Standard Test Method for Particle-Size Analysis of Soils, Annual Book of Standards, Vol. 4, ASTM International, West Conshohocken, 2007. ASTM 2007. Standard D 698-00: Standard Test Method for Laboratory Compaction Characteristics of Soil Using Standard Effort (12,400 ft-lbf/ft3 (600 kN-m/m3 )), Annual Book of Standards, Vol. 4, ASTM International, West Conshohocken, 2007. ASTM 2007. Standard D 2487-06: Standard Classification of Soils for Engineering Purposes (Unified Soil

Classification System), Annual Book of Standards, Vol. 4, ASTM International, West Conshohocken, 2007. Brandon, T.L., Rose, A.T. & Duncan, J.M. 2006. Drained and undrained strength interpretation for low-plasticity silts. J. Geotech. Geoenviron. Eng., 132(2): 250–257. Charles, J.A. & Soares, M.M. 1984. Stability of compacted rockfill slopes. Geotechnique 34(1): 61–70. Dasenbrock, D.D. & Hankour, R. 2006. Improved soil property classification through automated triaxial stress path testing. Proc. GeoCongress, Atlanta, Feb. 26–Mar. 1 2006. Day, R.W. & Axten, G.W. 1989. Surficial stability of compacted clay slopes. J. Geotech. Eng. 115(4): 577–580. Jiang, J.C., Baker, R. & Yamagami, T. 2003. The effect of strength envelope nonlinearity on slope stability computations. Can. Geotech. J. 40(2): 308–325. Pariseau, W.G. 2007. Fitting failure criteria to laboratory strength tests. J. Rock Mech. & Mining Sciences, 44: 637–646. Penman, A. 1953. Shear characteristics of saturated silt measured in triaxial compression. Geotechnique 15(1): 79–93. Ponce, V.M. & Bell, J.M. 1971. Sear strength of sand at extremely low pressures. J. Geotech. Eng. 9(4): 625–638. Rad, N.S. & Clough, G.W. 1984. New procedure for saturating sand specimens. J. Geotech. Eng. 110(9): 1205–1218. Trandafir, A., Popescu, M. & Ugai, K. 2001. Two dimensional slope stability analysis by LEM and FEM considering a non-linear failure envelope. Proc. 40th Annual Conf. of Japan Landslide Society, Maebashi, August 2001: 219–222.

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An investigation of a structurally-controlled rock cut instability at a metro station shaft in Esfahan, Iran Ali Taheri Zaminfanavaran Consulting Eng. (ZAFA), Esfahan, Iran

ABSTRACT: On July 29th 2006 at 1.30 A.M, an overall slope failure took place at eastern wall of a 28 m-depth metro station shaft in Esfahan, Iran. The failure caused some serious damage to the city utilities and loss of one man life. The instability occurrence was about 10 month after the shaft completion, where, the shaft walls had been supported by a regular grid of grouted dowels, steel mash and shotcrete. According to the geotechnical investigation results carried out after the failure, the main cause of the rock mass instability at the shaft wall was a major shear joint existence in the rock mass, trending almost parallel to the slope face and dipping towards the shaft bottom with about 60 degrees angle. This structural feature was unrecognized before the failure incidence. In this paper, by describing the induced failure mechanism, the dominant role of the existing natural plan of weakness on the rock instability is analyzed and the unsuccessful effect of the used rock supporting system is discussed. It has been concluded in this paper that the possibility of the structurally-controlled rock failures should be taken into account seriously in design and execution of any earth structures, especially in urban areas. 1

INTRODUCTION

The Esfahan Metro Project is a large scale infrastructure project which has attracted international interest form inception as it passed through the city of Esfahan with 2500 years history and architecture and home to a number of highly treasured world heritage sites. On the other hand, the geological and hydrogeological conditions of the project area is various and rather complicated. The line 1 of the Esfahan subway passes through bedrock at southern part of the city and through soft deposits at the middle and northern parts. The kargar Station shaft instability occurred in the discontinuous bedrock mass due to sliding on a major shear joint, 10 months after completion of the ground excavation and support installation. In this paper, by describing the geological conditions of the site and explanation of the installed support system of the station cut, the induced failure mechanism has been discussed and conclusions and recommendations are given in order to prevent such instabilities in similar projects. 2 2.1

GEOLOGY Lithology

The lower Jurassic (Lias) deposits, comprising shale and sandstone alternation, form the bedrock of

the project area. These deposits which are denoted as ‘‘Shemshak Series’’ are overlain by a slightly cemented, coarse grained alluvium, which its thickness is around 9 m at the proposed site. 2.2 Structural features The dominant structural features of the bedrock mass are bedding plane and three systematic joint sets, with the orientations given in Table 1. There is no visible geological discontinuity in the covering alluvial deposits. 2.3 Groundwater condition There is a shallow, low transmissible groundwater aquifer at the southern part of the Esfahan city, which has been formed in the alluvial deposits and shallow layers of the bedrock mass. The main recharge source Table 1. mass.

Orientations of discontinuity system in the rock Orientation (deg.)

Discontinuity type

Dip direction

Dip amount

Bedding plane Joint set 1, J1 Joint set 2, J2 Joint set 3, J3

185–200 310 110 010

35 85 75 60

481

of this aquifer is water infiltration form green areas, as well as, leakage from water and sewage lines. The groundwater flow direction is generally from south to north, with hydraulic gradient of about 3%. The original groundwater table at the proposed site lies in the range of 6.2 to 8.8 m. bellows the ground surface.

3

GEOTECHNICAL ASPECTS

In order to assess the effect of the geological discontinuities on stability of the proposed rock mass, the orientation of the four distinguished discontinuity sets have been plotted on a streonet, together with the orientation of the shaft cut face (Figure 1). It is evident from this streoplot that the existing discontinuities were unlikely to be involved in the slope failures. So, the potential slope failure had been considered as non-structurally controlled and the slope stability analyses were performed on the basis of this assumption. This led to design and execution of the following slope supporting system. In designing the executed supporting system, by assuming the rock mass as an equivalent continuum media, the shear strength of the rock mass had been determined on the basis of the Hoek-Brown failure criterion, as follows (Hoek et al. 2002). σ1′ = σ3′ + 49(0.0176σ3′ + 0.0003)0.508

(1)

where σ1′ and σ3′ are the major and minor effective principle stresses at failure, in Mpa.

Figure 1. Streoplot of the discontinuity system in the rock mass.

4

THE INSTALLED SUPPORT SYSTEM

In order to support the proposed station shaft walls, a passive supporting system has been designed and executed during the excavation process. In designing this system, the rock mass had been considered as a equivalent continuous media, with potential failure plane as non-structurally controlled (similar to soil). The wall supporting system was comprised of φ32 mm grouted dowels with 12 , 8 and 6 m. lengths, at 2 × 2 m. regular grid, and 30 cm. thickness shotcrete, reinforced by two 10 × 10 × 0.8 cm. weldmeshes (Figure 2). 5

THE FAILURE MECHANISM

When the overall failure of the shaft wall started to take place at 1:30 A.M., there were five persons working inside the shaft, which one of them had not chance to survive. Another worker who was witness of the incident reported that the failure started from bottom of the shaft by mass sliding towards the opposite wall with a loud voice, and followed by toppling of soil, rock and shotcrete blocks from high levels of the wall (Figure 3). It should be added that 10 days before failure incidence, some tension cracks had been formed at upper surface of the shaft, which nobody paid attention to it. According to the geotechnical investigation results, carried out after the failure incidence, a major shear joint (local fault plane), trending almost parallel to the slope face and dipping towards the shaft bottom with about 60 degrees angle, was the main cause for the failure occurrence. This fault plane which had not been recognized before the failure incidence was filled by a soft and saturated gouge of about 30 cm thickness.

Figure 2.

482

Wall supporting system of the station shaft.

Figure 3. General view of the failed shaft wall. Figure 4.

As shown in Figure 4, the induced failure had a dual mechanism; sliding of a discontinuous rock on the existing fault plane and then rotation and toppling of the soil blocks and shotcrete slabs from higher levels of the wall (slide head toppling).

6

THE MAIN REASONS OF THE INSTABILITY

In spite of the supporting system installation for stabilization of the shaft walls, about half of the eastern wall was failed in a few seconds, 10 months after completion of the shaft excavation. By consideration of the induced failure mechanism and taking into account all the parameters affecting the stability of the slope, the main reasons of the shaft wall instability were reported as follow. • Design of the wall supporting system by assuming the rock mass as a continuum media with equivalent geomechanical parameters. While, the existing fault plane caused the rock mass to behave as discontinuous, with structurally controlled failure incidence.

Plan and cross section of the failed shaft wall.

• Besides the insufficient supporting system, the poor installation of the grouted dowels intensified the instability problem of the shaft wall. For instance, by inspecting the failure plane after removal of the debris, it was found that just a limited number of the dowels were failed in tension, while, most of them were pulled out from failed or remained rock masses (Figure 5). • As the failure took place 10 months after completion of the shaft excavation, the sliding of the potentially unstable rock block can be related to increase of groundwater pressure on the fault plane, due to the high water infiltration rate from adjacent green areas in hot summer.

7

FAILURE MECHANISM ANALYSIS

In order to analyze the induced failure mechanism, a windows grogram called ‘‘ROCPLANE’’ has been used. This program is an interactive software tool for assessing the stability of planar sliding blocks in

483

rock slopes. It also allows users to estimate the support capacity required to achieve a specified factor of safety. To do this analysis, the Mohr-Coulomb shear strength parameters of the fault plane have been determined by performing the Consolidated-Undrained triaxial tests on a number of representative undisturbed samples of the fault gouge material in the laboratory. The mean shear strength parameters determined in this manner are as follows: • Cohesion, c′ = 0.5 kg/cm2 , • Angle of internal friction, φ′ = 30 degrees.

By assuming the groundwater pressure distribution as shown in Figure 6, the factor of safety of the so supported slope was found as 1.1, without taking into account the retaining effect of the reinforced shotcrete. In this case, the total retaining force of the installed grouted dowels was found 157.5 ton/m, assuming all the dowels were mobilized simultaneously. While, by factor of safety of 1 the retaining force of the dowels drops to 118.5 ton/m, i.e. 75% of the expected total force, and by taking into account the retaining effect of the reinforced shotcrete, it was found that around 50% of the installed grouted dowels were not mobilized simultaneously with the others, at the verge of the failure.

Figure 5. Pull out of the dowels from the failed rock mass.

8

CONCLUSIONS AND RECOMMENDATIONS

According to the induced failure mechanism and reasons, the following conclusions and recommendations can be given.

Figure 6. Geometry of the failed slope.

• Regarding the dominant role of the major geological discontinuities on stability of rock slopes, in preliminary designing of the wall supporting systems, the potentiality of structurally controlled failures should be taken into account, especially in urban areas. To do this, a good practice is to carry out some sensitivity analysis by assuming a range of possible orientations and shear strength parameters for the major structures in rock masses. • To confirm the geological structure of the slope, further geological mapping together with evidence on groundwater and tension cracks, would provide information for review of the situation to decide upon the best means of slope stabilization, in addition to the drainage measures. • Due to the considerable effect of the supporting systems installation quality on stability of rock masses, the proper execution of the rock supporting designs is highly recommended. In this sense, it is desirable that unlike the studied case, the passive supporting systems be mobilized simultaneously. Otherwise, the resisting capacity of the supporting systems will

484

be decreased and may not be able to sustain the destructive forces of the unstable rock blocks. • In order to control the stability of the slopes and efficiency of the ground stabilization systems, it is recommended that to perform some instrumentation and monitors surface movements and subsurface deformations of the ground, during execution and operation of the shafts in the urban areas. ACKNOWLEDGMENTS The author would like to record his appreciation of help from Mr. H. Mansoori Broojeni in collection and analysis of the failure data. REFERENCES

Haraz Rah Consulting Eng. Group. 2006. A study of the Kargar Station eastern wall failure. Technical report, in Persian, Tehran-Iran. Hoek, E. & Bray, J.W. 1981. Rock Slope Engineering, 3rd. ed. London: IMM. Hoek, E., Carranza-Torres, C.T. & Corkum, B. 2002. HoekBrown failure criterion-2002 ed., proc. North American Rock Mechanics Society meeting in Toronto in July 2002. Ortigao, J.A.R. & Sayao, A.S. 2004. Handbook of Slope Stabilization. Berlin: Springer. Simons, N., Menzies, B. & Matthews, M. 2001. Soil and Rock Slope Engineering. London: Thomas Telford. US Army Corps of Engineers. 2003. Engineering and Design Slope Stability, Manual No. 1110–2-1902. Wyllie, D.C. & Mah, C.W. 2004. Rock Slope Engineering, Civil and Mining, 4th ed., Based on the 3rd. ed, by Hoek, E. & Bray, J. , London: Spon Press. Zaminfanavaran Consulting Eng. (ZAFA). 2006. Concluding report of the instability at the Kargar Station eastern wall. in Persian, Esfahan-Iran.

Alamoot Bridge & Building Eng. Co. 2006. Geological report of the failure event at the Kargar Station eastern wall, Report No. EURO-ALMT-RO3-RPT-GEO-2003(Rev. 0), in Persian, Tehran-Iran.

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Yield acceleration of soil slopes with nonlinear strength envelope A.C. Trandafir Department of Geology and Geophysics, University of Utah, Salt Lake City, Utah, USA

M.E. Popescu Department of Civil and Architectural Engineering, Illinois Institute of Technology, Chicago, Illinois, USA

ABSTRACT: This paper presents and discusses the results of a pseudostatic slope stability analysis aiming to address the influence of strength envelope nonlinearity on the computed yield acceleration of a uniform slope in a homogeneous soil. Published linear (Mohr-Coulomb) and nonlinear strength envelopes derived from the same experimental database were employed in the analysis. For the variety of slope geometries analyzed in this study, the nonlinear strength envelope always resulted in a smaller yield acceleration compared to the traditional Mohr-Coulomb strength envelope. The difference between yield accelerations associated with the two failure criteria appears to increase with decreasing slope height and increasing slope inclination. For the same slope geometry, the critical sliding surface corresponding to the yield acceleration for the nonlinear strength envelope is shallower than the critical sliding surface associated with the Mohr-Coulomb strength envelope. 1

INTRODUCTION

Assessment of seismic displacements of slopes based on Newmark sliding block method (Newmark 1965) represents a routine engineering practice. Newmark model consists of one-block translational or rotational mechanism along a rigid-plastic sliding surface which is activated when the ground shaking acceleration exceeds the yield acceleration of the sliding mass. The yield acceleration is defined as the earthquake acceleration required to bring the sliding mass to the limit equilibrium condition corresponding to a safety factor of 1.0. The yield acceleration is the yield coefficient (ky ) multiplied by the gravitational acceleration (g). The yield coefficient (ky ) is obtained from conventional pseudostatic slope stability analyses and depends, among other factors, on the shear strength properties of the material along the sliding surface. Soil shear strength is generally expressed in terms of Mohr-Coulomb linear failure criterion τ = c + σ tan φ, with c (cohesion) and φ (angle of internal friction) representing the conventional shear strength parameters. There is however considerable experimental evidence showing that the strength envelope of many soils is not linear particularly within the range of small effective normal stresses (e.g., Bishop et al. 1965, Charles and Soares 1984, Atkinson and Farrar 1985, Maksimovic 1989). Furthermore, slope stability analyses conducted for materials exhibiting a nonlinear failure envelope resulted in lower

static safety factors compared to those provided by the linear (Mohr-Coulomb) strength envelope (Maksimovic 1979, Charles and Soares 1984, Popescu et al. 2000, Trandafir et al. 2001, Jiang et al. 2003). Given the interdependence between static safety factor and yield acceleration of a potential sliding mass, the importance of considering the strength envelope departure from linearity in seismic slope stability evaluations may be easily inferred; the result may be a smaller value of the computed yield acceleration and therefore a higher vulnerability of the analyzed slope to earthquake induced instability. This paper examines through a parametric study the influence of soil strength envelope nonlinearity on the computed yield acceleration of a uniform slope in a homogeneous soil. Published linear Mohr-Coulomb and nonlinear strength envelopes derived from the same experimental database are employed with pseudostatic slope stability computations to obtain the yield acceleration of the sliding mass.

2

SLOPE GEOMETRY AND SHEAR STRENGTH ENVELOPES

Figure 1 shows the geometric variables (i.e., slope height H and slope inclination parameter b) for the parametric study, and the forces acting on the sliding mass in a pseudostatic slope stability analysis to

487

the shear strength estimate in the range of small normal stresses obtained from the projection of the linear strength envelope in the range of larger normal stresses is unsafe.

Ground surface

H

1

b

kW =17.7 kN/m3

W

b=2 (NL) b=2.5 (NL) b=3 (NL)

Sliding surface

Figure 1. Slope geometry and forces acting on the slide mass.

0.55

Yield acceleration (g)

150

Mohr-Coulomb strength envelope Nonlinear strength envelope

(kPa)

b=2 (MC) b=2.5 (MC) b=3 (MC)

100

50

0.45

0.35

0.25

0 0

50

100

150

200

0.15

250

4

(kPa)

6

8

10

Slope height, H (m)

Figure 2. Linear and nonlinear strength envelopes considered in the analysis.

Figure 3. Yield acceleration versus slope height for various slope inclinations; NL—nonlinear strength envelope; MC—Mohr Coulomb strength envelope.

H=4m (NL) H=7m (NL) H=10m (NL)

H=4m (MC) H=7m (MC) H=10m (MC)

0.55

Yield acceleration (g)

determine the yield acceleration of a specific slope. The linear (Mohr-Coulomb) and nonlinear strength envelopes used in the analysis are depicted in Figure 2. These envelopes were derived by Jiang et al. (2003) using least-square estimates on experimental results from numerous (i.e., 103) laboratory triaxial tests on heavily compacted Israeli clay. The conventional shear strength parameters characterizing the linear (Mohr-Coulomb) failure envelope (Figure 2) are c = 11.7 kPa and φ = 24.7◦ . The nonlinear strength envelope is described by a power-type relationship τ = Pa A(σ/Pa )δ (Baker 2004), with Pa representing the atmospheric pressure, whereas A and δ are the dimensionless strength parameters of the nonlinear strength function. For the nonlinear failure envelope in Figure 2, A = 0.582 and δ = 0.857. By comparing the sums of squares of residuals, Jiang et al. (2003) noticed equally valid descriptions provided by both linear and nonlinear strength models of the available experimental data associated in fact with normal stress levels greater than 35 kPa. It is to be noted that triaxial tests reported by Atkinson and Farrar (1985) indicate that the nonlinear shear strength envelope apparent in the range of small stresses becomes linear in the range of larger normal stresses. Therefore

0.45 0.35 0.25 0.15 0.05 1.5

2

2.5

3

Slope parameter, b Figure 4. Yield acceleration versus slope inclination for various slope heights; NL—nonlinear strength envelope; MC—Mohr Coulomb strength envelope.

488

1.06

1.08

1.04

1.16

1.02

1.14

1.12

1.0 2

1.10

1.08

1.04

1.06

1.18

1.06

b)

15.0

15.0

12.5

12.5

10.0

10.0

Elevation (m)

Elevation (m)

a)

1.08

F s =1 F s =1

7.5

kyMC = 0.479

5.0 2.5

7.5

kyNL = 0.251

5.0 2.5

0.0 05

10

15

20

0.0

25

05

Horizontal distance (m)

10

15

20

25

Horizontal distance (m)

Figure 5. Critical sliding surface corresponding to the yield coefficient of a 5 m high slope with an inclination b = 2: (a) linear strength envelope model (MC); (b) nonlinear strength envelope model (NL).

For the particular soil considered in our analyses (Figure 2), the nonlinear strength model predicts smaller strength values than linear (Mohr-Coulomb) strength envelope for normal stresses below 35 kPa. The following section of this paper presents and discusses the results of a parametric study performed to illustrate the effect of the nonlinear failure envelope in the range of very low normal stresses (i.e., 3 m: Fs /Fs > 10%; Knowing the hazard thresholds with regards to different triggering phenomena, a hazard zonation can be drawn, based on the resultes of the 3D groundwater flow model.

W

Montegrassano

E S

Cavallerizzo %

Cerzeto ## % Torano castello #

#%

#

San Giacomo # #%

#

San Martino di F

##%# # #

6

CONCLUSIONS

Water is one of the main causes of landslide; as a results, the hazard assessment has to consider the groundwater setting and its effects on slope stability. These analysis are more and more widespread for the study of individual landslides; in spite of that, usually in hazard mapping they only concern the pluviometrical thresholds triggering soil slips, whereas groundwater effects on deeper landslides are not taken into account. Such a simplification comes from: – the complexity of the ruling hydrogeological processes, – the lack of suitable hydrogeological data.

N %

The Figure 10 shows a map of hydrogeological susceptibility to landslides based on the groundwater discharge, whereas the Figure 11 shows the susceptibility to groundwater boundary conditions (as recharge, up hill supply, etc.).

# #

The paper describes an example of parametrical modelling that provides a connection between the slope stability and some hydrogeological parameters easy to survey and to monitor (as for instance rainfall, piezometrical level and springs discharge). Moreover, hazard thresholds have been pointed out as regard to several triggering phenomena and an hydrogeological susceptibility map has been achieved for the test area. This map is a very important informative layer for both landslide hazard mapping and risk analysis.

#

REFERENCES

#

Rota Greca % #

Lattarico %

# ##

# #

#

# # #

#

#

##

# San Benedetto Ul % #

Rainfalls hydrogeological sensibility: Low Medium High Very high #

Figure 11.

Spring Town River

Rainfall hydrogeological sensitivity map.

Bartolomei, A., Brugioni, M., Canuti, P., Casagli, N., Catani, F., Ermini, L., Kukavicic, M., Menduni, G. & Tosoni, V. 2006. Analisi della suscettibilità da frana a scala di bacino (Bacino del F. Arno, Toscana, Italia). In AIGA (eds) Giornale di Geologia Applicata 3 (2006): 189–195. Chieti. Benedetti, A.I., Dapporto, S., Casagli, N. & Brugioni, M. 2006. Sviluppo di un modello di previsione di frana per il bacino del fiume Arno. Giornale di Geologia Applicata 3 (2006): 181–188. Corominas, J. 2001. Landslides and climate. 8th Int. Symp. On Landslides. No.4: 1–33. Cortopassi, P., D’Amato Avanzi, G., Guidotti, M., Marconi, C., Milano, P.F., Mussi, M. & Puccinelli, A. 2006. Approccio multidisciplinare per la valutazione della pericolosità da frana: indagini geologico-tecniche, idrogeologiche, geochimiche e isotopiche per la determinazione degli aspetti idrici sotterranei nella frana di Cassana in Val di Magra (Massa Carrara, Italia). In AIGA (eds) Giornale di Geologia Applicata 3: 196–204. Chieti. Farkas, J. 1992. Experiences from landslide investigations in Hungary. Proc. of the VIth International Symposium on Landslides 10–14 February 1992. Rotterdam: Balkema.

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Francani, V. & Gattinoni, P. 2000. Influenza della struttura geologica del versante sulla distribuzione delle pressioni neutre nei corpi franosi. Quaderni di Geologia Applicata 7 (3): 61–77. Bologna: Pitagora Ed. Gattinoni, P. & Trefiletti, P. 1999. Instabilità di sponda in seguito a variazioni piezometriche. Quaderni di Geologia Applicata 6 (2): 23–33. Bologna: Pitagora Ed. Gillon, M.D. & Hancox, G.T. 1992. Cromwell Gorge landslides—A general overview. Proc. of the VIth International Symposium on Landslides 10–14 February 1992. Rotterdam: Balkema. Iverson, R.M. 1992. Sensitivity of stability analysis to groundwater data. Proc. of the VIth International Symposium on Landslides 10–14 February 1992. Rotterdam: Balkema.

Kawabe, H. 1992. On the influence of pore pressure on land deformation of a landslide. Proc. of the Vith International Symposium on Landslides 10–14 February 1992. Rotterdam: Balkema. Lacerda, W.A. & Schilling, G.H. 1992. Rain induced creeprupture of Soberbo Road landslide. Proc. of the Vith International Symposium on Landslides 10–14 February 1992. Rotterdam: Balkema. McDonald, M.G. & Harbaugh, A.W. 1988. A modular three dimensional finite difference ground water flow model. U.S. Geological Survey Techniques of Water Resources Investigations, Book 6.

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Sustainable landslide stabilisation using deep wells installed with siphon drains and electro-pneumatic pumps A. Gillarduzzi High-Point Rendel, London, UK

ABSTRACT: In the winter of 1994/95, a major accelerated reactivation of a pre-existing landslide occurred at Castlehaven, Isle of Wight, UK. This event was a consequence of both coastal erosion and high groundwater levels. If left unchecked this landslide had the potential to evolve further causing direct and consequential financial losses estimated to be £20 million (at 2004 prices). To address the problem of landsliding and to reduce the risk to public safety, property and infrastructure, a coastal protection and landslide stabilisation scheme was implemented. The slope stabilisation scheme included the use of 151 deep pump drainage wells, installed with 125 siphon-drains and 35 electro-pneumatic pumps, to reduce winter groundwater levels to the equivalent summer levels where movement is substantially reduced, thus improving landslide stability and at the same time protecting valued environmental assets. This paper describes the substantial advantages, in terms of sustainability, provided by the adopted design approach and by the type of drainage equipment used. 1

INTRODUCTION

Castlehaven is located at Niton on the southern most point of the Isle of Wight, UK, centreed on coordinates 50◦ 34′ 47′′ N and 1◦ 17′ 14′′ W (Figure 1) The instability of the site is due to a combination of susceptible geology, loss of slope toe caused by marine erosion of the sea cliffs and high groundwater levels. The later is caused by particularly intense and prolonged wet periods occurring over the autumn/winter season. Castlehaven, since the latest post-glacial period is characterised by slow rate instability with periodic accelerated and catastrophic reactivations. A major reactivation, affecting an area up to 500 m inland from the coast, occurred in 1994/95 following a very wet winter. If left unchecked this landslide had the potential to further evolve causing direct and consequential losses estimated at £20 million in 2004 prices. The landslide is characterised by the widespread occurrence of springs and seepages forming areas of wet bare ground areas and scarps. These particular local settings, in combination with a site-specific microclimate, create an important and sensitive environment supporting rare fauna and flora. For this reason, the area, over the years, has been covered by both national and international designations including: Site of Special Scientific Interest (SSSI), candidate Special Area of Conservation (cSAC), Site of Importance for Nature Conservation (SINC), Area of Natural Outstanding Natural Beauty (ANOB) and Heritage Coast. To address the problem of landsliding and to reduce the risk to public safety, property and infrastructure

a coast protection and landslide stabilisation scheme was planned, designed and constructed. This had to be sustainable taking into account the environmental importance and quality of the area and simultaneously being effective. 2

GEOLOGICAL SETTINGS

The site includes (Figure 2) an unstable slope of approximately 500 m in length limited to the north by the inland vertical to sub-vertical inland cliff (∼20 m high) and to the south by the vertical sea cliffs (∼8 m high). The geology of the area consists of an interbedded sequence of high and low permeability (sub-horizontal) Cretaceous strata. The inland cliff is formed by Upper

Figure 1.

1547

Site location.

Inland cliff

N

Upper tier

Gault Clay scarp Lower tier

Sea cliff protected by the armourstone revetment

Electro-pneumatic pump line Siphon drain line Drainage connection Compressor chamber Outfall

Figure 2. Aerial view of Castlehaven at completion of the stabilisation works.

Greensand (UG) strata underlain by the Gault Clay (GC) and the Lower Greensand (Carstone and the Sandrock). The near vertical sea cliff is formed by intact Sandrock. The landslide body is complicated and highly heterogeneous. This includes an inland area of variable thickness of multi-rotational failures (Upper tier) separated by the seaward compound landslides (Lower tier) by exposure of the in situ GC scarp. The Upper tier is generally formed by a unit of GC debris up to 25 m in thickness overlaid by UG debris up to 10.5 m in thickness derived from weathering, erosion and rockfalls from the inland cliff.

The main source of groundwater feeding the unstable slope is from recharge of the upper and deep aquifers at the rear of the landslide from the inland UG aquifer perched on the underlying low permeability GC. Additional minor recharge is from direct rainfall (over the landslide area only) infiltrating in the soil and from artificial sources including leaking water supply, sewers, highway and building drainage, gardening, etc.

4 4.1

3

GROUNDWATER

The hydrogeological conditions at Castlehaven are anisotropic and variable but in general include an upper unconfined aquifer (perched over the GC strata) and a deep confined artesian aquifer in the Sandrock strata. The comparison of the groundwater and inclinometer monitoring data (together with direct observations, slope back- and sensitivity-analysis and historical records) indicates that high groundwater level, which occurs during the autumn-winter period, is the most important factor causing the recurrent reactivations and general instability of the site (Clark et al. 2007, Gillarduzzi et al. 2007).

DESIGN APPROACH General approach

A combined coastal protection and landslide stabilisation scheme was planned and designed during the period 2001–2003 to improve the stability of the site. The adopted design approach included the prevention of the sea erosion and landward retreat of the sea cliff and the improvement of the stability of the Upper tier landslide by reducing and stabilising the groundwater level (Clark et al. 2007). The minimum design life of the scheme was fifty years. 4.2

Coastal protection

The foreshore coast protection design included a rock armour buttress revetment with a 3.5 m wide buried toe. This was placed in front of the erodible 8 m

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high Sandrock sea cliff. The rock armour consisted of 45,000 t of Carboniferous limestone. The primary armour (imported by barge from northern France) ranged in size from 3 to 6 t placed on 5 to 500 kg sized core-stone. The revetment had a 1 in 1.5 front slope with a 3.5 m crest at an upper level of 5 m Ordnance Datum (mOD). The lateral extent and height of this structure were limited to reduce the impact on the coastal behaviour and for aesthetic reasons. 4.3

Groundwater level lowering

The slope stability analyses and monitoring clearly indicated that the most effective way of improving the landslide stability was to lower and maintain the groundwater within the Upper tier to a level comparable to that monitored during previous summers when no significant slope movement occurred. The main technical constraints in the selection of the dewatering system included: – Variable characteristics and thicknesses (maximum 35 m at the rear of the landslide) of the landslide debris; – Substantial length and width of the landslide and access problems due to property limits and topography; – Variable groundwater level depending on the position within the landslide, topography, local geology and period of the year; – Maximum recorded groundwater winter levels of approximately 6 and 1 m below ground level (mbgl) at the rear and the front of the Upper tier respectively; – The calculated groundwater drawdown to achieve the required improvement was typically between 7.5 and 15 mbgl; – The need to maintain partial activity of the spring line that occurs at the top of the GC scarp and causes localised slope movement, which is important for unstable slope habitat creation; – The final scheme had to have low running and maintenance costs following construction. A total of 151 vertical deep drainage wells (150 mm casing diameter) were constructed in 2003–2004. These wells are arranged in four different lines and the wells are generally spaced at 6 m centres. The wells depth is variable (range 15–25 mbgl) dependant on of their position within the landslide. A total of 116 wells were installed with Siphon Drains (SD) and 35 wells with Electro-Pneumatic Pumps (EPP). The wells were connected by a system of trench drains and ducts directly discharging to a sea outfall. The details of the SD and EPP are described in detail in Bomont et al. (2005) and the principles of both systems are summarised below.

Both EPP and SD systems are extremely flexible and are able to automatically respond to changing groundwater conditions. This may be required in the event of major climatic and significant ground water changes. Both systems were design to cope with particularly wet winters that typically trigger catastrophic landslides. EPP and SD, if required, could be easily retrofitted and modified to increase the discharge output of the drainage system. For example, the EPP and electrical switches can be lowered deeper in the wells to achieve a greater drawdown. Additional wells can be constructed and installed at the site due to the redundancy of the compressor system. The wells installed with SD can be fitted with multiple siphon tubes or larger diameter tubes (as already done in five wells) or with electro-pneumatic pumps to allow for supplementary drainage.

4.4 Electro-pneumatic pumps The main source of water feeding the landslide is due to infiltration from the large catchment located to the north of the inland cliff (Figure 2). Therefore the draw down required in this area had to be at least between 11 and 15 mbgl (depending on the location) to substantially reduce the effect of this source. The wells located in this area where installed with EPP. The adopted EPP have been specifically developed to stabilise landslides where the groundwater lever is deep and the abstraction volume is substantial. The electro-pneumatic system is similar in concept to other systems deployed to drain leachate from landfills and in industrial installations. The basic system includes an air compressor and compressed air tanks, an electropneumatic control system and the EPPs located in the wells. The system can be described in a simplified way as follows. Electric switches positioned in the well detect when the groundwater level is higher than the design requirements to maintain the slope stability. When the groundwater level is high, compressed air is released into pipes running from the air tanks (charged by two compressors) to the pumps installed in the wells. The compressed air release is controlled by the pneumatic system (i.e. solenoids). When the compressed air reaches the pumps, it displaces the groundwater contained in them, pushing it out of the well through an outlet pipe. The groundwater is then drained by gravity to a sea outfall trough a system of buried ducts and pipes. The electro-pneumatic pumps operate, only when required, in continual fill and drain cycles whose frequency is dictated by the groundwater level in the well, recharge of the aquifer and required drawdown level. This allows to save a substantial amount of energy.

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The EPP system was required to achieve a groundwater abstraction at a greater depth than that allowed by the siphons. Cost estimates over the entire life of the scheme indicated a clear economical advantage in the use of this system compared to a more standard installation of wells with submergible pumps. 4.5

Siphon drains

The 116 wells installed with SD were arranged in three longitudinal lines across the landslide (Figure 2) with the purpose of intersecting the groundwater around the middle area of the Upper tier slope. Five wells were installed with two SD each due to local high groundwater recharge. The wellheads were located at the base of large manholes between 1.5 m and 7.5 mbgl. The wells are dewatered using small diameter siphon tubes that allow groundwater to drain from the well through an intermediate point (i.e. wellhead) that is up to 8 m higher than the groundwater level in the well. The system operates in accordance to Bernoulli’s principle. The peculiarity of the adopted SD is that they are installed with two patented products (Gress 1996); a permanent reservoir located in the well and an accumulator/flushing system at the siphon tube outfall (Figure 3). These products prevent normal problems afflicting conventional siphons such as the break of siphon flow when one or both sides of the tubes are not immersed in water. Furthermore, the periodic automatic flushing of the siphon tubes creates a rapid turbulent flow. This removes impurities, looses concretions and gas bubbles accumulating in the siphon tubes therefore maintaining the system’s efficiency. When the groundwater level rises in the well, the siphon activates abstracting water out of the well in cycles (i.e. 2.9 to 24 m3 /day). The flow continues until the groundwater level in the well falls to the level of the

Siphon tubes Water flush Accumulator/Flushing system Figure 3. Siphon drain accumulator/flushing systems installed in an outlet manhole.

outlet. This level can be changed depending on future design requirements by simply rising or lowering the accumulator/flushing system level within the outfall manhole. One of the reasons for the selection of the SD system was its simplicity of operation and cost since the siphon does not require any source of power to operate. Field trials were carried out to confirm the practical operation and performance of the siphon well system before adopting it. 5

DISCUSSION

The stabilisation of Castlehaven area has been planned in an attempt to provide the best outcomes for the human and the natural environment both in the short and long term. A sustainability approach has been adopted in various aspects of the planning and design phase, in particular: 5.1 Planning – A lengthy process of planning and consultation was required to obtain approval of the scheme (Clark et al. 2002) and to fully address the requirements listed below; – The planning activities systematically identified and addressed the risks and uncertainty associated with stabilising a developed slope in an environmental sensitive area, therefore setting realistic targets; – The environmental aspects of the scheme were evaluated and appreciated through dedicated studies and monitoring. It was clearly identified that an absolute priority was to conserve the specific biodiversity and ecological integrity of the site; – Environmental, social, human and economic goals were clearly identified and integrated in the subsequent design; – The local community was involved and consulted during the planning and design phase to address specific needs and to secure approval of the scheme; – The consultation was also extended to the general public to take into account the global integration of the locality; – The design was planned with the Local Authorities, Governmental Agencies and stakeholders; – The benefits of the planned activities had to continue after the initial funding. The scheme had to be maintainable to ensure its long lasting effect (sustainable benefit); – The Client (Isle of Wight Council) for the scheme was involved in all the stages (planning, design, construction, maintenance and training) to promote its participation and a sense of ownership of the scheme. The Client and a local contractor were

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trained on site on all aspects of the scheme to allow for the proper maintenance of the scheme. In addition, a detailed procedure for scheme performance, monitoring and maintenance was prepared. 5.2

Design

– The scheme was designed with a clear commitment to best practice and innovation. New technology (some of which was used for the first time in the UK) were considered and adopted in the design; – The target was to improve the slope stability of the area to an acceptable level with no net loss of human or natural capital; – The design addressed and complied with all the environmental constraints of the site minimising the disturbance caused by the construction works (Figure 4) which were particularly severe in a developed area with poor access; – The slope stability was improved by modifying the slope drainage (using deep well drainage) but omitting works on the Lower tier of the slope. This approach maintains the natural integrity of the undeveloped slope Lower tier whilst improving the stability of the developed Upper tier; – The slope stability analysis also demonstrated that lowering of the groundwater below the equivalent summer level would have further substantially improved the stability of the slope. However a target slope stability factor of improvement of 1.10 was adopted to avoid the potential of causing damage or stress to the local flora and fauna; – The use of the pumping system was restricted to the shortest period possible over the year. This approach addresses the need to maintain the groundwater very close to the natural average level during summer when the flora and fauna are at their maximum annual activity; – The water discharge to sea was limited to 500 m3 /day to avoid any detrimental impact to the marine environment in proximity to the outfall;

– The drained water was discharged without preventive treatment. This was justified by a probabilistic assessment carried out using an approved methodology. It was considered that there was no economic justification in installing an oily water separator due to the remoteness of the hazard and that the limited funds would have been better prioritised; – To reduce the visual and noise impact of the scheme, the entire drainage system (including the electropneumatic compressor chamber) was located underground and was accessible through manholes; – To reduce the impact to the local community the wells and ducts were largely sited within (generally low traffic) public roads to reduce the need of acquisition and use of private land during construction and subsequent maintenance; – Climatic changes in the area, occurring during the design life of the scheme leading to more seasonal precipitations concentrated over the winter period and consequential changes in the groundwater levels were difficult to predict (Gillarduzzi 2008). Therefore, the selected system had to be flexible to cope with possible changing scenarios. The slope stability design philosophy included the ability to cope with increased rainfall and groundwater levels raised to existing ground levels (i.e. total soil saturation); – During the design, the need was identified of carrying out recurrent performance reviews and control visits. These are required to calibrate, improve, optimise and maintain the system to achieve the design groundwater drawdown and reduce the long-term maintenance cost. The design included the provision of monitoring equipment to verify the performance of the systems and to compare the results to targets. The drainage system and the slope were installed with monitoring instrumentations including borehole inclinometers, observation wells (standpipes and vibrating wire piezometers), flow meters, weather station, telemetric compressor monitoring and a surface-monitoring network. – The drainage was carried out using deep wells installed with siphon drains (SD) and electropneumatic pumps (EPP). The advantages of these systems, in terms of sustainability, are described above; – The design requirements were specified in a schedule of minimum drawdown or trigger levels to be achieved in each siphon and electro-pneumatic wells and these were included in the performance specification. 6

Figure 4. Construction of a temporary work platform to prevent excessive disturbance of the environment.

CONCLUSION

The quality of the environment along the southern coast of the Isle of Wight is the reason for a large

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part of the resident population choosing to live in this area and it is closely linked to the economy of the site and of the entire island. Therefore, the slope instability at Castlehaven is a threat to public safety, property and infrastructures but simultaneously it is also one of the factors strongly characterising the local natural environment. This scheme demonstrates how a sustainable approach can be successfully adopted in the planning and design of slope stabilisation schemes to provide the best outcome for both the human and the natural environment. The planning phase should correctly identifying the existing needs of all the parties directly and indirectly involved and influenced by the scheme in both the short and in the long term. This allows setting realistic and achievable targets and to correctly plan the following design phase. This has the purpose to define a robust design, which meets the planning requirement. The detailed design phase can bring substantial optimisation of a conceptual scheme in terms of sustainability. In the example illustrated above this was achieved by installing conventional dewatering deep wells with siphon drain and electro-pneumatic pumps and in various others less evident, but similarly significant, aspects. In this particular scheme, the use of an unconventional and innovative dewatering method was the key point to optimize the sustainability of the scheme and the effectiveness of the engineering solution. ACKNOWLEDGEMENTS The author would like to acknowledge the co-operation and assistance of the Isle of Wight Council and of the many colleagues at High-Point Rendel in particular Dr A.R. Clark and D.S. Fort.

REFERENCES Bomont, S., Fort, D.S., & Holliday, J.K., 2005. Two applications for deep drainage using siphon and electro pneumatic drains. Slope works for Castlehaven Coast Protection Scheme, Isle of Wight (UK) and slope stabilisation for the Railways Agency, France. In, Proceedings of the International Conference on Landslide Risk Management. 18th Annual Vancouver Geotechnical Society Symposium. Clark, A.R., Fort, D.S., Holliday, J.K., Gillarduzzi, A. & Bomont S., 2007. Allowing for climate change; an innovative solution to landslide stabilisation in an environmentally sensitive area on the Isle of Wight. In: McInnes, Jakeways, Fairbank & Mathie (Eds), Landslides and Climate Change. Challenges and Solutions. Proceedings of the International Conference on Land-slides and Climate Change, Ventor, Isle of Wight, UK, pp. 443–451. (c) 2007 Taylor & Francis Group. Clark, A.R., Storm, C.V, Fort, D.S & McInnes, R.G. 2002. The planning and development of a coast protection scheme in an environmentally sensitive area at Castlehaven, Isle of Wight. Procurement International Conference on Instability, Planning and Management. Pub. London: Thomas Telford. pp. 509–518 Gillarduzzi, A., 2008. Physical impact of climate change on coastal slope instability: the Undercliff landslide, Isle of Wight, UK, (under review of Quarterly Journal of Engineering Geology and Hydrogeology). Gillarduzzi, A., Clark A.R., Fort, D.S. & Houghton, J.E.M. 2007. Monitoring coastal slope instability within the western Undercliff landslide, Isle of Wight, UK. In: McInnes, Jakeways, Fairbank & Mathie (Eds.), Landslides and Climate Change. Challenges and Solutions. Proceedings of the International Conference on Landslides and Climate Change, Ventor, Isle of Wight, UK, pp. 345–354. (c) 2007 Taylor & Francis Group. Gress, J.C., 1996. Dewatering a landslip through siphoning drain—Ten years experiences. Proc.7th International Symposium on Landslides, Trondheim. Pub.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Biological and engineering impacts of climate on slopes – learning from full-scale S. Glendinning & P.N. Hughes Newcastle University, UK

D.A.B. Hughes Queens University, Belfast, UK

D. Clarke, J. Smethurst & W. Powrie Southampton University, UK

N. Dixon & T.A. Dijkstra Loughborough University, UK

D.G. Toll & J. Mendes Durham University, UK

ABSTRACT: Our climate is set to change significantly over the next century; future change is likely to have a serious effect on UK slopes. The scenario of hotter drier summers, followed by more intense periods of rainfall has the potential to reduce stability by increasing degradation mechanisms and/or increasing positive pore water pressure generation. There is evidence that the scenario of more intense rainfall is already having an impact on the UK slopes. However, there is also potential for stability to be improved through the generation of greater suctions during longer periods of drought. Newcastle, Southampton, Belfast, Durham and Loughborough Universities have all been carrying out research into the impacts of climate and vegetation on embankment and cut slope stability. These five Universities, along with international partners in Canada, Singapore, China, South Africa, France and Portugal, are conducting a collaboration programme the aim of which is to link research groups undertaking full-scale monitoring of slopes to improve the understanding of the complex interaction between climate, vegetation and clay soils. This paper presents results of current full scale infrastructure slope monitoring and model development at the involved universities and plans for future collaborations. 1

INTRODUCTION

The climate of the UK is set to change significantly over the next century, and is likely to have significant effects on the stability of both natural and man-made slopes. Future UK climate change scenarios predict consistent and significant increases in temperature of up to 3◦ C on average in south-east UK (Hulme et al., 2002). Predicted changes in rainfall are less consistent, but key aspects are: little change or small increase in annual rainfall; a general increase in winter rainfall and decrease in summer rainfall. This is likely to increase the rate at which certain clay soils degrade by increasing the amplitude of the seasonal shrink swell cycle. Thus failures governed

by progressive failure mechanisms have the potential to increase in frequency. Increased periods of summer drought will cause clay soils to shrink, leading to surface cracking, potentially increasing surface permeability. The more intense periods of Autumnal rainfall will then allow pore pressures to increase more rapidly within slopes, triggering more failures. There is evidence that the scenario of more intense rainfall is already having an impact on the UK including major landslides in Scotland (e.g. Stromeferry) and, in the winter of 2000/1, which was documented the wettest on record, over 100 slope failures occurred in the Southern Region of Railtrack alone. However, there is also evidence that the pore suctions generated by dry (summer) weather conditions control the long-term ultimate limit state stability

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of clay slopes (Loveridge & Anderson, 2007). It is therefore possible that drier summers will actually increase the stability of clay slopes. Additionally, if a similar amount of annual rainfall occurs as at present, but is experienced as shorter, more intense events, it will lead to less of the rain entering the soil as more is lost as run-off. This again could improve stability overall. In either case, the study of the interactions of climate and pore water pressures in slopes is key to determining their long-term future stability. This paper sets out to explain how this problem is being addressed through the combined efforts of a consortium of UK-based universities. 2

CLIFFS is managed by Loughborough University and is supported by a large core group of academic institutions (including the Authors’) and stakeholders. It currently has more than 150 members, mainly from the UK. It operates by organizing multi-disciplinary themed workshops and by providing a web-based information exchange facility. Workshop themes have included issues of risk and uncertainty, and aspects of the responses of natural and constructed slopes to changes in climate. Details of these workshops can be accessed at the network’s website on cliffs.lboro.ac.uk. Whilst current membership is mainly UK based, CLIFFS is seeking new international members in order to learn from the experience of both researchers and practitioners who deal with slopes in a wide range of soils and vegetation, subjected to different climates.

COMMUNICATION AND NETWORKING

The nature of the problem of climate impacts on slopes is such that it affects many different stakeholders and end-users. The problem is also being approached from many different angles and with different objectives in mind. It thus forms a very broad multi-disciplinary field in which geographers, mathematicians, statisticians, physicists, engineers, ecologists, hydrologists, etc. try to work out their own particular problem angles and seek to forge links to provide a broader solution than would be achieved individually. This is not always easy as specialists speak different (scientific) languages and do not always share the same philosophical approach to problem solving. With this in mind the network CLIFFS (climate impact forecasting for slopes) was funded by the UK Engineering and Physical Sciences Research Council (EPSRC) in 2005 to bring together academics, research and development agencies, stakeholders, consultants and climate specialists. The main aim of bringing these people together is to stimulate an integrated research response to address the intricately linked problem of forecasting, monitoring, design, management and remediation of climate change induced variations in slope instability. The size of the task and the complexity and multi-disciplinary nature of the problem requires active participation of a wide group to assess the magnitude of the resulting impact on UK society and to identify appropriate management and remediation strategies to achieve a better insight into the links between climate change and slope stability in the UK, firstly there is a need to determine the information requirements and, secondly, a need to focus research efforts on targeted assessments of long-term scenarios. Although detailed processes or individual site conditions are being addressed, general process-response issues are still not well understood or researched—a problem exacerbated by poor communication in this multi-disciplinary field (Dijkstra & Dixon 2007).

3

CURRENT UK RESEARCH

Five universities, Newcastle University, Queens University Belfast, University of Southampton, Loughborough University and Durham University have all been carrying out research into the impacts of climate and vegetation on embankment and cut slope stability. This has already included field instrumentation work to measure seasonal moisture and pore water pressure changes in a number of embankments and cut slopes, back analysis and numerical modelling. The research has started to give a more detailed picture of embankment response (lateral and vertical deflections) to seasonal variations of both moisture content and pore water pressure. The behaviour of these embankments is complex, and in terms of trying to model their behaviour there are still many challenges to be overcome. Recent work has shown that the numerical models are very sensitive to the values and distributions of parameters such as permeability, which in a clay embankment can vary considerably as a result of summer desiccation and cracking close to surface, and the nature and compaction of the clay fill. It is possible that a very dry summer followed by a wet winter is most critical for stability, as the summer cracking allows a path for rainfall infiltration. However, this is still not well understood. Future changes in climate in the UK are likely to lead to more extreme rainfall events with higher intensity storms. Such rainfall events are common in the tropical regions of the world and the team is drawing on collaborative work in Singapore, Thailand, China, Canada and Hong Kong. The five universities recently received a major travel grant from the Engineering and Physical Sciences Research Council (EPSRC). This will allow the team to visit and build better links with both UK and overseas infrastructure owners and research organizations.

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The Roads Service in Northern Ireland (Department for Regional Development in Northern Ireland) and Northern Ireland Rail are funding Queens University Belfast to develop a risk based method of assessment of the geotechnical infrastructure on the Northern Ireland road network following a major slope failure on the road network in 2000 (Hughes et al, 2007). As part of this research programme a cutting on the A1, 4 miles south of Dromore has been heavily instrumented (Clarke & Hughes et al 2005). Pore water changes were recorded during the excavation of the cutting and currently much data is being gathered on the pore water dynamics forced by rainfall and evapotranspiration effects (Figure 1). A transient predictive model incorporating climate events has been calibrated and verified against the field data using GeoStudio 2004. GeoSlope International (from Calgary, Alberta, Canada) have been supporting the project with technical assistance and the provision of the latest modelling software. Southampton University has been carrying out intensive monitoring of soil moisture and pore water pressures at a Highways Agency owned road cutting near Newbury in Southern England since 2002. The climate is temperate with average annual rainfall of 850 mm, summer temperatures of +20◦ C and winter temperatures of 0–3◦ C. An array of 40 sensors were inserted in five groups along a cut slope in the London Clay, vegetated with a mixture of short grass and small bushes up to 0.5 m tall. Data were recorded for soil moisture content using Time Domain Reflectrometry (TDR) in the upper layers of the soil (0–2.5 m) below the surface. Pore water pressures were also monitored using Vibrating Wire Piezometers. Readings have been made every 10 minutes since 2002. Hydrological inputs and losses at the site have been measured, including rainfall, surface runoff, depth to saturation together with climatic parameters to estimate potential evapotranspiration (temperature, humidity, wind speed, solar radiation). Soil characteristics have been determined using both field and Factor of Safety (FoS)

Depth of GWL below crest GL

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1.4

Factor of Safety (FoS)

3

2 1.3 1.5 1.2 1 1.1 1.0 Jan-05

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Depth of GWL below crest GL

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laboratory approaches and the soil moisture sensors were calibrated using gravimetric methods and backed up with regular neutron probe measurements Figure 2 shows the variation in volumetric soil moisture content (m3 /m3 ) between 0.3 and 1.5 m depth This long term monitoring clearly shows the cyclical changes between summers (warm and relatively dry) and winters (cool and relatively wet). Most drying occurs in the upper 1.0 m of the soil profile, where the roots from the vegetation are most active. The maximum drying usually occurs at the end of summer in September (month 9), followed by a rapid re-wetting of the profiles in November-January. Also apparent is the effect of the climatic patterns of different years; 2003, 2005 and 2006 were relatively dry in the summer whereas 2004 and 2007 had higher than average rainfall. Figure 3 shows the associated variations in pore water pressures for the same period. Near hydrostatic conditions occur in the winter months (NovemberMarch) but the seasonal growth of vegetation between April and September dries the soil and negative pore water pressures develop up to 2.5 below the surface. Suctions of up to −70 kPa are recorded at 1.0 m depth and we have recorded suctions as high as −400 kPa 30 cm using some temporary instruments. The magnitude and duration of the negative pore water pressures varies from year to year, again depending on the climatic conditions experienced. A series of hydrological and numerical models have been developed to describe and explain the behaviour of the processes at the site. Figure 4 shows the result of a Soil Moisture Deficit model based on the FAO CROPWAT methodology. The losses of moisture from the soil profile are calculated based on evapotranspiration and a root zone model and converted to an equivalent pore water pressure. These have been used to validate a FLAC model of the slope. This work has demonstrated the use of hydrological models in describing the surface boundary conditions and their impacts on pore water pressures.

0 Jul-05

Jan-06

Jul-06

Month

Figure 1. Monthly fluctuation of GWL.

Figure 2. surface.

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Moisture content (m3 /m3 ) 0.3–1.5 m below

embankment is constructed to modern highway specifications using modern compaction plant, and half has been constructed to poorer specification using as little compaction as possible in order to simulate older rail embankments. Data from the in-situ testing conducted on the embankment during its construction utilizing high suction tensiometers developed at Durham University (Lourenço et al, 2006) has demonstrated that high negative pore pressure were generated during the construction process. Figures 5 and 6 show suctions measured in the ‘‘poor’’ and ‘‘well’’ compacted zones respectively. These tests clearly indicate that modern construction techniques generate higher soil suctions than older methods. Piezometers installed after construction have shown that the initial soil suctions had in part dissipated six months after construction was complete. Now that vegetation has become more established on the embankment soil suctions of up to −30 kPa have been measured in both poor and well compacted sections at 4.5 m depth using fully flushable piezometers. An additional system for measuring soil suctions, developed by Durham University has been installed in the BIONICS embankment using

Figure 3. Pore water pressures 1.0–2.5 m below surface.

Figure 4. Soil moisture deficit and run-off. 0 -20 0

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400 Panel A, Layer 8

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Panel A, Layer 9

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Panel D, Layer 5

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Figure 5. Suctions measured during construction of the poorly compacted sections. Note: Section constructed in 1 m layers, beginning with layer 1.

0 -50 0

Suction (kPa)

These models are being used to explore the impact of many years repeated cycling of wetting and drying on the slope stability and UKCIP climate change scenarios are being using to investigate the long term performance of the slopes. In addition to the monitoring of ‘real’ infrastructure slopes, subjected to ‘real’ UK climate, a consortium of asset owners has been put together by Newcastle University to oversee the BIONICS (Biological and Engineering Impacts of Climate change on Slopes; www.ncl.ac.uk/bionics) research project. This is a four year programme that aims to establish a unique facility for engineering and biological research. This facility is in the form of a full-scale, fully instrumented embankment, with climate control over part of its length. Thus, the facility allows the control of the climate necessary to study the effects of future climates, coupled with a fully characterized engineering soil and vegetative cover. A database of embankment performance data is being compiled of the results of testing and monitoring during both the construction and the climate experiments. This unique set of data, describing the full history of the embankment will be available for all future research based at the facility. The embankment is 90 meters long and has been constructed in two distinct parts. Half of the

Suction (kPa)

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Figure 6. Suctions in the well compacted sections. Note: Sections constructed in 0.3 m layers.

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Wykeham Farrance—Durham University tensiometers (Mendes et al, 2008). The borehole locator system allows readings at different levels in a single borehole, permitting observations of the variation of suction with depth. It also allows tensiometers to be removed for resaturation whenever necessary. The wide measuring range of the tensiometers (down to −2 MPa) allows usage of such a system in most natural and manmade earth structures. Preliminary results (from three months of monitoring) show that there are different patterns of suction measurements from the tensiometers installed in the well compacted section of the embankment compared to those installed in the poorly compacted section (Mendes et al, 2008). It has been observed that tensiometers installed in the poorly compacted section of the embankment react rapidly to rainfall. The well compacted section instead shows a slower change of suction and does not respond rapidly to rainfall. Currently a climate control system is being constructed, consisting of flexible, roofing sections that can be pulled over the embankment when required (similar to those used to cover sports facilities), and arrays of computer controlled rainfall sprinklers mounted on poles. Automatic weather stations will monitor wind speed, net radiation, temperature, relative humidity and atmospheric pressure, with tipping bucket and storage rain gauges to measure rainfall rates and totals. The performance of the proposed arrangements will be measured to ensure that it provides the climatic conditions required. In particular, the heating effect provided by covering (and leaving covers in place over night to prevent heat loss) will be compared to the temperatures predicted by climate change. This system will then be used to study the response of the embankment and the vegetation to controlled patterns of rainfall and heating. Associated with the BIONICS project is an EPSRC funded research project at Loughborough University to develop novel instrumentation to detect slope instability (Dixon & Spriggs 2007). A real-time continuous slope monitoring system based on detection and quantification of acoustic emission generated by slope deformations is currently being trialed on the BIONICS embankment. Performance of the acoustic system is being compared to traditional deformation measurement techniques including in place inclinometers. Acoustic emissions are related to slope deformation rates. The system is sensitive to both small magnitudes and rates of displacements and the technique has potential to provide an early warning of instability. This too will be monitored closely during the controlled climate experiments. Durham University has been studying rainfallinduced slope failures in collaboration with Universities in Singapore and Thailand who currently

experience more extreme patterns of rainfall and temperature. Experience in both countries is that there has been an increase in landslide activity associated with increased rainfall events (Toll et al, 2008; Jotisankasa et al, 2008). Rainfall has been the dominant triggering event for landslides in Singapore and Thailand. Studies show spates of landslides occurring after unusually wet periods. Observations of past landslides in Singapore suggest that a total rainfall of 100 mm within a six day period is sufficient for minor landslides to take place (Toll, 2001). In Thailand, a total rainfall of 150–400 mm would tend to trigger major landslides (Jotisankasa et al, 2008). Measurements of pore-water pressures in slopes in Singapore and Thailand show that rainfall infiltration produces changes in pore-water pressure near to the surface. However, at greater depths (around 3 m) the pore-water pressures do not change significantly (Tsaparas et al, 2003). Numerical modeling shows that this is because water tends to flow down the slope within the zone of higher saturation (which has higher permeability) that develops near the surface (Tsaparas and Toll, 2002). As a result, rainfall-induced failures tend to occur within the near surface zone and are not usually deep-seated. Pore-water pressures measured within slopes in Singapore (Tsaparas et al, 2003) were, for a large part of the monitoring period, only slightly negative and at 3 m depth were generally positive. However, there were periods during the year when pore-water pressures reduced significantly following a drier period. Pore-water pressures dropped to as low as −70 kPa near the surface (0.5 m depth). Interestingly, this is similar to the values of suction measured at similar depths by Southampton University in very different climatic conditions in the UK. However, piezometer data in Singapore shows that there was little change in ground water table level (which remained below 15 m depth). Therefore, these suction changes were the result of infiltration and evapotranspiration occurring at the surface and were not due to changes in water table. Therefore, it is important that when studying climate effects on slopes that we do not always assume that rainfall will produce a rise in water table level. Infiltration of rainfall at the surface can produce significant changes in pore-water pressure without a change in water table (although a perched water table may be induced at the surface). Field measurements in Singapore suggest that porewater pressures do approach the hydrostatic condition near the surface due to infiltration (Toll et al, 2001). However, pore-water pressures remain significantly below the hydrostatic line, even at the wettest time of the year. Therefore, assuming that pore-water pressures were hydrostatic throughout the slope (as would

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often be assumed in a saturated soil analysis) would be over-conservative. Work is now underway with the National University of Singapore to investigate the impact that future climate change will have on Singapore, including slope stability problems (Toll et al, 2008). 4

CONCLUSIONS

There is compelling evidence that our climate is changing and that this change will have a significant impact on the behaviour of slopes globally. In the UK, there is sufficient concern for the owners and operators of its transport networks to be actively funding research to investigate the problem. However, it has been recognized that the problem of climate influences on slopes is a sufficiently complex problem that a much greater understanding of the problem can be gained by sharing the existing knowledge from a wide range of disciplines. The CLIFFS network has been funded to facilitate such an exchange of ideas. Additionally, there is much to be learned by exchanging ideas on an international scale where researchers necessarily have been determining the effects of a range of climates, vegetation and soil types. This exercise is currently being kick-started using a travel grant awarded to five UK-based Universities. The rewards, in terms of shared experience and improved understanding are only just beginning to be realized and it is anticipated that the shared experience will provide a more complete picture of the impacts of climate (and climate change) on slopes. REFERENCES Clarke, G.R.T., Hughes, D.A.B., Barbour, S.L. and Sivakumar, V. (2005). ‘Field Monitoring of a Deep Cutting in Glacial Till: Changes in Hydrogeology. GeoSask2005, the 58th Canadian Geotechnical Conference and 6th CGS & IAH-CNC Joint Groundwater Specialty Conference, Saskatoon, Canada, September 18–21. Dijkstra, T. and Dixon, N. (2007). Networking for the future—addressing climate change effects on slope stability. Proceedings Int. Conf. on Landslides and Climate Change: Challenges and Solutions, Isle of Wight, UK, May 2007 (CD). Dixon, N. and Spriggs, M. (2007). Quantification of slope displacement rates using acoustic emission monitoring. Canadian Geotechnical Journal, 44 (8), 966–976.

Hughes, D., Sivakumar, V., Glynn, G. and Clarke, (2007) ‘A Case Study: Delayed Failure of a Deep Cutting in Lodgement Till’, Journal of Geotechnical Engineering, ICE, Volume: 160-Issue: 4 Cover date: October 2007 (accepted) Page(s): 193–202 Print ISSN: 1353-2618. Hulme, M., Jenkins, G.J., Lu, X., Turnpenny, J.R., Mitchell, T.D., Jones, R.G., Lowe, J., Murphy, J.M., Hassell, D., Boorman, P., McDonald, R. and Hill, S. (2002). Climate Change Scenarios for the United Kingdom: the UKCIP02 Scientific Report, Tyndall Centre for Climate Change Research, School of Environmental Sciences, University of East Anglia, Norwich, UK. 120pp. Jotisankasa, A., Kulsawan, B., Toll, D.G. and Rahardjo, H. (2008). Studies of Rainfall-induced Landslides in Thailand and Singapore, 1st European Conference on Unsaturated Soils, Durham, UK, July 2008. Lourenço, S.D.N., Gallipoli, D., Toll, D.G. and Evans, F.D. (2006). Development of a commercial tensiometer for triaxial testing of unsaturated soils. Proc. 4th International Conference on Unsaturated Soils, Carefree, USA, Geotechnical Special Publication No. 147, ASCE, Reston. Vol. 2, 1875–1886. Loveridge, F. and Anderson, D. (2007). What to do with a vegetated clay embankment, Slope Engineering Conference, Thomas Telford, London, UK. Mendes, J., Toll, D.G., Augarde, C.E. and Gallipoli, D. (2008) A System for Field Measurement of Suction using High Capacity Tensiometers, 1st European Conference on Unsaturated Soils, Durham, UK, July 2008. O’Brien, A. (2001). Personal communication. Perry, J, Pedley, M. and Reid, M. (2001). Infrastructure Embankments—condition appraisal and remedial treatment. CIRIA publication C550. Toll, D.G. (2001). Rainfall-induced Landslides in Singapore, Proc. Institution of Civil Engineers: Geotechnical Engineering, Vol. 149, No. 4, pp. 211–216. Toll, D.G., Mendes, J., Karthikeyan, M., Gallipoli, D., Augarde, C.E., Phoon, K.K. and Lin, K.Q. (2008). Effects of Climate Change on Slopes for Transportation Infrastructure, 1st ISSMGE International Conference on Transportation Geotechnics, Nottingham, UK, September 2008. Toll, D.G., Tsaparas, I. and Rahardjo, H. (2001). The Influence of Rainfall Sequences on Negative Pore-water Pressures within Slopes, Proc. 15th International Conference on Soil Mechanics and Geotechnical Engineering, Istanbul, Rotterdam: Balkema, Vol. 2, pp. 1269–1272. Tsaparas, I. and Toll, D.G. (2002). Numerical Analysis of Infiltration into Unsaturated Residual Soil Slopes, in Proc. 3rd International Conference on Unsaturated Soils, Recife, Brazil, Lisse: Swets & Zeitlinger, Vol. 2, pp. 755–762. Tsaparas, I., Rahardjo, H., Toll, D. and Leong, E.C. (2003). Infiltration Characteristics of Two Instrumented Residual Soil Slopes, Canadian Geotechnical Journal, Vol. 40, No. 5, pp. 1012–1032.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Some attributes of road-slope failure caused by typhoons M.W. Gui, C.H. Chang & S.F. Chen Civil Eng. Dept., National Taipei Univ. of Technology, Taipei, China

ABSTRACT: Taiwan is an island with two-thirds of its area located in mountainous zones. Because of the scarcity of usable land, many developments and roads have been built on the hillsides that were formed by relatively weak geological materials. During heavy rainfalls, these materials will be weakened and hence led to debris flows or landslides. The main purpose of this study is to explore some of the attributes of the 1567 road (excluding express/motor-ways) slope failures occurred after the heavy rainfalls that were brought by four intense typhoons in 2004 (Mindulle, Aere, Haima and Nock-Ten). This preliminary study statistically examined the significance of the types of geological material, slope angles, distance to the nearest active faults, and rainfalls intensity on road slope failures.

Taiwan is an island where one-third of its total areas are plains with elevations below 100 m while the other two-thirds of the island are occupied by hills and mountains with elevations varying from 100 m to 4,000 m. Because of the scarcity of usable land, many developments and roads have been built at the hillsides that were formed by relatively weak geological materials. During heavy rainfalls, particularly those brought by typhoons, these materials will be weakened and led to natural hazards such as debris flows and landslides. According to the Central Weather Bureau (CWB 2004), the island’s average annual rainfall was about 2500 mm, in which approximately 70% of the precipitation occurred between May and September, i.e. during the typhoons season. Four intense typhoons: Mindulle, Aere, Haima and Nock-Ten struck the island and inflicted a total of 1567 road-slope failures along the island’s road networks. The road networks mentioned included all the roads on the island except the express/highways. The main purpose of this paper was to identify the attributes of landslides generated after the heavy rainfalls that were brought by the four intense typhoons in 2004. The method of study was via statistical approach. Hopefully, area prone to landslides after heavy rain may be identified so that mitigation and contingency plans could be derived to minimize life and property losses in the event of another typhoon. 2

number of typhoons—about 29.3% of all typhoons (Chen 2004). 75.2% of all the typhoons struck in the months of July, August and September. The annual precipitation/rainfall also concentrated between these few months (Figure 1) as these typhoons brought along with them heavy rainfalls. There were four intense typhoons struck the island in 2004: typhoon Mindulle, typhoon Aere, typhoon Haima, and typhoon Nock-Ten. 2.1 Typhoon Mindulle Typhoon Mindulle moved west after taking shape and landed on the east of Taiwan, about 20 km from the southeast of the island with a speed of 15 km/h on July 1, 2004 and left from the northwest of the island on July 2. The path of the typhoon Mindulle has been categorized as Path 4 (Figure 2). It brought heavy rains to the east, north, northeast and mountain regions of the island. The highest rainfall recorded was 1182 mm on Ali Mountain (usually called Alisan), the second 600

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Monthly and accumulated rainfalls through 2004.

Aere caused a total of fifteen deaths and fourteen people missing, three-hundred and ninetynine wounded, one-hundred and sixteen houses damaged, more than 1.4 millions electricity interruptions, and about 380,000 tap-water interruptions. Agricultural losses were estimated at more than NT$180 millions (US$5.6 millions).

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Figure 2. Possible paths of typhoon encountered between 1897 and 2002 (Chen 2004).

highest was 703 mm in Jade Mountain. Water levels at several reservoirs rose on July 2 and 3, with Jenwen Reservoir in Tainan County accumulated 419 mm of rain, while Nanhua Reservoir, which supplies water to Tainan, Kaohsiung and Chiayi, received 560 mm of rain. There were many rainfalls and landslides in the central regions of the island. The death toll as a result of the typhoon and the flood was thirty-two. In addition, there were twenty persons wounded, thirteen missing, and one-hundred and forty houses damaged and over 80,000 electricity interruptions. Agricultural losses were estimated at more than NT$8,900 millions (US$278 millions). 2.2 Typhoon Aere Typhoon Aere moved toward the northwest after taking shape on August 20, 2004 and landed on the northeast of the island with a speed of 15 km/h on August 25. The typhoon moved gradually southwest after entering the Strait of Taiwan and entered Fujian from northeastern side of Jinmen at 12:00 pm on August 25. Typhoon Aere weakened over the northern part of the island on August 25 after bringing heavy rainfall to the central and northern parts of the island. The path of typhoon Aere has been categorized as Path 1 (Figure 2). Thus, it rained a lot in the north and the central parts of the island. The highest rainfall encountered was 783 mm at Alishan and 774 mm at Zhuzihu, Taipei (CWB 2004).

Typhoon Haima

Typhoon Haima was formed in the east of Taiwan Sea on September 11 with its center about 60 km to the east of the island. It passed through the north of the island with a speed of 14 km/h on September 12 and moved to the east of China. The path of typhoon Haima is Path 4 (Figure 2). It brought a lot of rains to the north of Taiwan and the highest rainfall area was in Anbu with rains of 706 mm, the second highest was 516 mm in Taipei. Rock and earth were loosened by the continuous downpours There were four deaths, two missing, thirty-four electricity interruptions, and 44,400 tap-water interruptions (National Fire Agency, 2004). Agricultural losses were estimated at more than NT$15 millions (US$0.47 millions). 2.4

Typhoon Nock-Ten

It started raining on October 24 as a result of typhoon Nock-Ten. It was a mild typhoon between October 18 to 25. The typhoon landed in Yilan on October 25 and left Taiwan a few hours later from Danshui with a speed of 19 km/h. The path of typhoon Nock-Ten is Path 4 (Figure 2). It rained heavily in the north of the island with the highest rainfall in Anbu with 496 mm of rain, the second highest was in Zhuzihu with 372 mm of rain. Accumulated rainfall from midnight of October 24 to 5 pm of October 25 exceeded 300 mm in many places in the northern part of the island. In northern Taiwan, the water surface of Danshui, Dahan and Xindian Rivers exceeded the alarming levels. The worst situation was reported in Taipei County, where the Keelung River suddenly rose from 42.09 m (1:00 October 24 2004) to 47.23 m (12:00 October 25 2004) due to the heavy rains. Around the island, a total of four deaths, onehundred and four wounded, two missing, 383,752 electricity interruptions and 12,566 tap water interruptions were reported. In addition, many agricultural facilities were also ruined. Financial losses in the agricultural sector reached NT$35 million (US$1.1 millions). Among the four typhoons, the most serious disaster in 2004 was caused by typhoon Mindulle. The associated maximum accumulated rainfall was 1182 mm at Alishan between 29th of June and 2nd of July.

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3

4

TYPES OF SLOPE FAILURES

According to Varnes (1978), slope failures can be classified into: (1) fall, (2) topples, (3) slides, (4) lateral spreads, (5) flows, and (6) complex. As they were not specifically related to our study, we have defined the following five types of failure instead: 1. Type-1: the upper slope failed and caused damage to the retaining wall supporting it [Figure 3(a)]. Soil/ rock and broken retaining wall were seen on the road. This is the type with the least failure; only 86 failures, or 5.5% of all the 1567 failures, were reported compared to the other types of failures. 2. Type-2: the upper slope was stable but the retaining wall on the crest of the down slope was damaged due to the washing out of the down slope (Figures 3(b)). It was the second common failure among the five failures. The number of failures was 324 or 20.7% of all failures. 3. Type-3: Among all the types of failure, it was the greatest failure in terms of landslide volume because both the upper and lower slopes failed together, either with (Figure 3(c)) or without the retaining wall (Figure 3(d)). It accounted for 13.8% of all the failures. 4. Type-4: the natural upper slope failed with soil/rock debris fallen on to the road (Figure 3(e)). This was the type with the second fewest disasters, albeit only slightly less than the Type 3 failure. About 13.1% of the failures were belonging to this category. 5. Type-5: there were pits on the surface of the road together with the failure of the natural down slope (Figure 3(f)). This was the most common failure among all the five failure types. Close to half, or 46.9%, of all the failures reported were of this type of failure. The damaged of the upper slope drainage channels normally causes erosion on the down slopes, which may fail subsequently (Hung 1996).

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Figure 3. Types of slope failure defined for this study.

ATTRIBUTES OF SLOPE FAILURES

All the above road slope failures occurred after the heavy rainfalls brought along by the typhoons. However, it was noted that these failures occurred not necessarily at places with the highest accumulated rainfall or highest rainfall intensity. There should be some other possible attributes of road slope failures, for example topography, geology, slope angle, distance to the nearest active fault and so on. As a result, we have statistically linked the attributes of topography, geology, slope angle, and distance to the nearest active fault with the number of slope failures. The results are presented in the following subsections. 4.1 Rainfalls Landslides triggered by rainfall are the cause of thousands of deaths worldwide every year; and a large variety of landslides occur as a consequence of heavy rainfall in tropical and temperate climates (Jakob et al., 2003). There have been many studies on the relationship of rainfalls and landslides (Finlay et al. 1997, Lin and Jeng 2000, Collins & Znidarcic 2004, Alleotti 2004, Rahardjo et al. 2005, Cheng et al., 2005, Chang et al. 2005). According to Abramson et al (2002), during heavy rainfall, rain water seeps into the ground and travels across the top layer of the soil in the slope which is often unsaturated. Results of field measurements at various sites indicated that the unsaturated gradient was generally directed vertically downward during the steady rainfall and the water may perch on lower permeability materials or on a drainage barrier such as bedrock and highly impermeable clays, and so on, creating a temporary, localized saturated zone (Abramson et al., 2002). The water then continues downward and down-slope, eventually seeping into gullies or reaching the lower groundwater table and entire slope will be dangerous because of infiltration and erosion (Abramson et al., 2002). Finlay et al. (1997) studied the number of slope failures and the rainfall intensity in Hong Kong Island between 1984 and 1993. They used the relationship of accumulated rainfall and number of slope failures to predict the number of slope failures. Hence, the number of slope failures reported here could also be linked to the rainfalls. For this study, we tried to relate the average rainfall, maximum hourly-rainfall, 24-hour rainfall, and accumulated rainfall to the number of slope failures. 4.1.1 Average rainfall The average rainfall during each typhoon has been taken as the average rainfall value of 25 rain-stations throughout the island. The average value was then

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would not have had failed at all. This is clearly not the case and therefore the stability of a slope could not be correlated to a single rainfall threshold value.

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Figure 4. Relationship between the average rainfalls during each typhoon and the number of failures.

compared with the total number of slope failures during each typhoon. Figure 4 shows the relationship of the average rainfall during each typhoon and the number of slope failures. The number of slope failures during typhoon Mindulle was 462, during typhoon Aere was 961, during typhoon Haima was 113, and during typhoon Nock-Ten was 31. In general, the distribution indicated that higher average rainfall was associated with higher slope failures, except for the case of typhoon Nock-Ten. 4.1.2 Maximum hourly-rainfall From the records made by 133 rain-stations around the island, it was possible to derive a more representative value of the maximum hourly-rainfall for each failed slope. This method of study shows that the intensities that associated with the highest and second highest numbers of slope failures were 60 and 30 mm/hour, in which they accounted for 754 and 706, or 48.1% and 45.1%, of all failures, respectively. The intensities of 90 mm/hour and 120 mm/hour only accounted for 4.5% and 2.3% of all failures, respectively. 4.1.3 24-hour rainfall The 24-hour rainfall was obtained by summing the rainfall record measured from a consecutive 24 hour’s period during the typhoon. The maximum value was then used to compare with the number of road slope failures and it was found that 53.6% of the slopes failed when there were less than 300 mm of 24-hour rain while 38.4% failed between 300 and 600 mm of 24-hour rain. According to Hung (1992), each slope may be associated with a rainfall threshold, if exceeding it, the slope will fail. He suggested a rainfall threshold of 300 mm/day for some of the weak slope areas (crest of higher mountain, mudstone, lateritic terrace slope, residual slope, cataclinal slope, filled slope, over-used slope) in Taiwan. However, using the threshold suggested by Hung (1992), the 53.6% of failed slopes

4.1.4 Accumulated rainfall The relationship between the accumulated rainfall and number of slope failures was also examined. A total of 1316 failures occurred in places with less than 600 mm of accumulated rainfall, or equivalent to an accumulated percentage of 84.0%; the failures percentage for accumulated rainfall of over 1500 mm was only 0.5%. Presumably, most of the weak slopes have already failed at accumulated rainfall of 600 mm or less. Gabet et al. (2004), who studied the suspended sediment concentrations and discharge in a catchment in the High Himalayas of Nepal during the monsoon seasons of 2000 to 2002, found that landslides in Himalayas of Nepal were not triggered until 860 mm of accumulated rainfall have fallen. Their observations suggested that sufficient antecedent rainfall is necessary to bring the regolith up to field capacity (the soil moisture beyond which gravity drainage will ensue) such that future rainfall may produce positive pore pressures and trigger landslides (Campbell 1975, Crozier 1999). 4.2 Topography Taiwan mountain belt is located at the oblique convergent boundary of the Eurasian Plate and the Philippine Sea Plate. The Central Mountain Range (comprising the Hsuehshan Range and the Backbone Range), which was resulted from active mountain building process, is up-heaved with elevations exceeding 3000 m, and about 70% of the island is mountainous area (Lin et al. 2000) with steep slopes. Due to society developments, many of these slopes were excavated for the purpose of roads building. Figure 5 shows that 75.3% of all the slopes failures occurred on slopes located at elevation 600 m and below, in which 26.7% of all the failures concentrated between elevations 200 m and 400 m where most of the roads were built. 4.2.1 Geology These slope failures were seen concentrating on the west side of the island. Relationship between the period of geology and number of slope failures is shown in Figure 6 and the highest number of failures was distributed between the geological period of Pliocene and Pleistocene to Middle Miocene. The highest failures occurred in Late Miocene group, also called the SanSia Group, with 398 failures; the second is Pliocene group with 267 failures; the third is Pliocene and Pleistocne group with 235 failures; and the fourth is Middle Miocene group with 230 failures.

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noted that many slopes on the island had probably been severely disturbed by the September 21, 1999, Chi-Chi earthquake. The magnitude 7.5 September 21 Chi-Chi earthquake had triggered 25,845 landslides with a total area of 15,977 ha and severely disturbed mountainous slopes in central Taiwan, and setting the conditions for occurrence of more landslides and debris flows/debris floods in the future (Lin et al 2002). Since, tectonic activities caused by the collisions of the Philippine Sea Plate and the Eurasian Plate have resulted in a series of Quaternary thrust faults trending north-south and dipping towards the east (Ho 1982, Lu et al. 2000), thus, it was decided to examine the relationship between the degree of disturbances and the number of slope failures by assuming that the closer a slope to a fault the more severe it was disturbed; hence, the relationship between the nearest distance of active faults and the failure locations was studied. The finding shows that only 145 cases, or 9.3%, of failures occurred within 1 km of active faults. There were 1077 cases, or 68.7%, of the failures occurred within 10 km of active faults; while almost all the failures (96.6%) located within 30 km from active faults. The number of slope failures within 3, 5, and 20 km were 444, 726, and 1343, respectively. So, it could be concluded that earthquake had played an important role in destabilizing the stability of these slopes.

Figure 5. Elevation and number of failures.

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These four geological groups made up of 72.6% of all slope failures. The reason why there were so many failures in the SanSia group was probably because SanSia Group is the youngest Miocene sedimentary cycle in western Taiwan. The group is distributed more extensively than any other Miocene groups in western Taiwan. This group is divided into a lower coal-bearing formation (Nanchuang Formation) and an upper marine unit (Kueichulin Formation). Both formations are composed predominantly of thick sandstones, which are tertiary weak rocks (also included are shale and mudstone) that are not strong enough to be classified as hard rock as a result of a relatively short rock forming period with a typical strength ranges from 10 to 80 MPa (Jeng et al. 1994).

The relationship between slope angles and the number of slope failures obtained from this study is shown in Figure 7. The slope angles were obtained from a 40 m× 40 m digital terrain map. The result indicated that 293 cases, or 18.7%, of the failures occurred on slopes with gradient ranges between 25◦ and 30◦ . A total of 937, or 60%, of all the failures occurred on slopes with gradient ranges between 20◦ and 40◦ . It was also interesting to observe that 29%, or nearly one-third, of the failures had occurred on slopes with slope angle of 20◦ or less, but there were only 11.2% of the failures occurred on slopes with gradient greater than 40◦ .

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Figure 6. Relationship between geological periods and number of failures.

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Figure 7. failures.

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Relation between slope angles and number of

The reason why there were more failures observed on the gentle slopes than the steeper slopes was that during rainfalls, water was supposed to infiltrate into the slopes but the gradient of the slopes dictated the amount of the water infiltrating into these slopes. On one hand, a large amount of the rain water has sufficient time to infiltrate into the gentle slopes since it required a longer time for the rainfall to travel down the slopes. As the water content in the slope increases the shear strength of the slope reduces until a value equals to the mobilized shear stress of the slope, which is also increase due to the weight of water in the slope. At this point, the slope or a portion of the slope would fail. On the other hand, steeper slopes can not absorb all the rainfall quickly and therefore water travel as surface runoff down to the toe of slope by gravity. This has resulted that most of the steep slopes have failures near the toe of the slopes.

5

CONCLUSIONS

A statistical study has been carried to identify some of the attributes of road slope failures occurred after the heavy rainfalls brought by four intense typhoons that struck Taiwan in 2004. In total, there were 1567 road slope failures reported. The study shows that the number of slope failures is generally proportional to the average rainfall, and that 84% of all failures occurred under an accumulated rainfall of 600 mm or less. In terms of geology, the highest number of failures occurred in the late Miocene group, which is the youngest Miocene sedimentary cycle on the island. The study also shows that 96.6% of all failures were located within 30 km of active faults, which had weakened these slopes during tectonic activities in the past. In addition, 60% of all the failures occurred on slopes with gradient ranges between 20◦ and 40◦ .

ACKNOWLEDGEMENT The authors are grateful to the Public Construction Commission of Taiwan who has partially funded this study. REFERENCES Abramson, L.W., Lee, T.S., Sherma, S. & Boyce, G.M. 2002. Slope Stability and Stabilization Methods. Second Edition, Wiley-Interscience Publication. Alleotti, P. 2004. A warning system for rainfall-induced shallow failures. Engineering Geology, 73 (3–4), 247–265.

Collins, B.D. & Znidarcic, D. 2004. Stability analyses of rainfall induced landslides. J of Geotechnical and Geoenvironmental Engineering, ASCE, 130, 362–372. Campbell, R.H. 1975. Soil slips, debris flows, and rainstorms in the Santa Monica Mountains and vicinity, southern California: U.S. Geol. Survey Professional Paper, 851, 51p. Chen, I.C. 2004. The impacts of rainfall in typhoon on the slope failure along road of the Northern Coastal Range. Master of Science Dissertation, National Dong-Hwa University, Taiwan (in Mandarin). Cheng, J.D., Huang, Y.C., Wu, H.L., Yeh, J.L. & Chang, C.H. 2005. Hydrometeorological and landuse attributes of debris flows and debris floods during typhoon Toraji July 29–30, 2001 in central Taiwan. J of Hydrology, 306, 161–173. Chang, M.H., Chiu, Y.F., Lin, S.Y. & Ke, T.C. 2005. Preliminary study on the 2003 slope failure in Woo-wan-chai Area, Mt. Ali Road, Taiwan. Engineering Geology, 80 (1–2), 93–114. Crozier, M.J. 1999. Prediction of rainfall-triggered landslides: a test of the antecedent water status model. Earth Surface Processes and Landforms, 24, 825–833. CWB (2004). Central Weather Bureau of Taiwan: http://photino.cwb.gov.tw/ Finlay, P.J., Fell, R. & Maguire, P.K. 1997. The relationship between the probability of landslide occurrence and rainfall. Canadian Geotechnical Journal, 34, 811–824. Gabet, E.J., Burbank, D.W., Putkonen, J.K., Pratt-Sitaula, B.A. & Ojha, T. 2004. Rainfall threshold for landsliding in the Himalayas of Nepal. Geomorphology, 63 (3–4), 131–143. Ho, C.S. 1982. Tectonic evolution of Taiwan: Explanatory text of the geologic map of Taiwan. Central Geological Survey, Ministry of Economic Affairs, Taipei, 126p. Hung, J.J. 1996. Typhoon Herb, the New-Central-CrossIsland-Highway and slopeland failures in central Taiwan. Sino-Geotechnics, 57 (10), 25–30 (in Mandarin). Jeng, F.S., Ju, G.T. & Huang, T.H., 1994. Properties of some weak rock in Taiwan. Proc. of the 1994 Taiwan Rock Engineering Symposium, Taiwan, 259–267. Lin, M.L. & Jeng, F.S. 2000. Characteristics of hazards induced by extremely heavy rainfall in Central Taiwan—Typhoon Herb. Engineering Geology, 58, 191–207. Lin, P.S., Lin, J.Y., Hung, J.C. & Yang, M.D. 2002. Assessing debris-flow hazard in a watershed in Taiwan. Engineering Geology, 66, 295–313. Lu, C.Y., Chu, H.T., Lee, J.C., Chan, Y.C. & Chang, K.J. 2000. Impact of basement high on the structure and kinematics of western Taiwan thrust wedge-example from 1999 Chi-Chi Taiwan earthquakes, Int. Workshop on Annual Commemoration of Chi-Chi Earthquake, Taipei, 145–154. Rahardjo, H., Lee, T.T., Leong, E.C. & Rezaur, R.B. 2005. Response of a residual soil slope to rainfall. Can. Geotech. J., 42 (2), 340–351. Varnes, D. J. 1978. Slope movement types and processes. In Landslides: Analysis and Control. Edited by R.L. Schuster and R.J. Krizek. Transportation Research Board, National Academy of Science, Washington, Special Report, 176, 11–33.

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A small rock avalanche in toppled schist, Lake Wanaka, New Zealand G.S. Halliday Tonkin & Taylor Ltd., Wanaka, New Zealand

ABSTRACT: A rock avalanche of 100,000 m3 occurred in a glacial valley in the Southern Alps of New Zealand in 2002. It originated on a 35◦ slope and released debris over a steep bluff. The resulting rock avalanche travelled 300 m, coming to rest on a gently sloping glacial bench. Individual boulders continued downslope and a number hit the Haast Pass Highway 600 m below. The bedrock in the region is mica schist, dipping at 50◦ into the slope. Large scale toppling is evident in the source area, with dips reduced to 20–35◦ in fractured, dilated rock. Aerial photos taken several years before the rock avalanche show a fresh scarp around the head, indicating significant slope deformation prior to failure. It is inferred that the scarp was the result of incipient sliding that eventually led to a catastrophic failure through loss of strength by strain weakening. The residual friction angle on the sliding surface is believed to be significantly less than the 35◦ slope inclination, providing conditions for rapid sliding. 1

INTRODUCTION

Large-scale mountain slope instability is widespread in the tectonically active Southern Alps of New Zealand. In schist terrain, creeping landslides and slow bedrock deformation are common. The stability of schist landslides subject to reservoir flooding was intensively studied during the Clyde Hydroelectric Dam Project, located 70 km south of the current study area (Gillon & Hancox, 1992). Occasional rockfalls and rock avalanches occur in the schist, typically on slopes over-steepened by glacial or fluvial erosion. This study concerns a small rock avalanche of approximately 100,000 m3 that occurred on alpine slopes above the Haast Pass Highway (SH6) on the shores of Lake Wanaka in 2002. (Figure 1) A number of boulders hit the road, resulting in temporary closure of the highway. Investigation of the rock avalanche established that the source area was a spur weakened by large-scale toppling. The presence of widespread large-scale toppling in New Zealand alpine greywacke terrain has been previously noted (Prebble, 1992). There is less apparent recognition of similar toppling in schist, (the metamorphic equivalent) although toppling prior to a rapid rockslide-rockfall has been reported (Halliday & McKelvey, 2004). 2

GEOLOGICAL SETTING

The rock avalanche site is located in the Southern Alps 45 km east of the Alpine Fault, the boundary

between the Australian and Pacific tectonic plates. Uplift of >10 mm/yr is occurring due to the oblique plate collision, and there is a high horizontal tectonic stress regime. The area is seismically active, with major earthquakes of up to Richter Magnitude 8 on the Alpine Fault every 200–300 years. The Alps were heavily glaciated during the Pleistocene. Post-glacial ice retreat approximately 15,000 years ago left steep sided U-shaped valleys, subsequently subject to fluvial erosion. The bedrock in the region is typically mica schist, representing metamorphosed sandstones and mudstones of Permian-Jurassic age. It is tectonically deformed with moderate to steep dips. Extensive slow gravitational bedrock deformation and landsliding of schist debris has occurred on the steep valley sides. There is geological evidence of numerous past rapid rockslides and rock avalanches in the region, and in 2007 a large schist rock avalanche formed a landslide dam in the Young Valley, 20 km to the north.

3

GEOLOGY AND TOPOGRAPHY

The rock avalanche originated on a 35◦ sloping spur above the Haast Pass Highway (SH6), at Sheepskin Creek near the head of Lake Wanaka (Figures 1–3). Steep bluffs lie beneath the source area, with the slope grading out to a gently sloping glacial bench. Below the bench, steep slopes with numerous vertical bluffs extend down to the highway and lake.

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Figure 3.

Rock avalanche source area, side view.

Figure 1. Rock avalanche above the Haast Pass Highway.

Figure 2. Cross-section of the rock avalanche area.

In situ schist outcrops below the bench and dips into the slope at an average 50◦ (Figures 2 & 4). The schist in the source area and further upslope towards the crest of the range is extensively disturbed by largescale toppling, evidenced by reduced dips, fractured outcrops and ‘‘sagging’’ geomorphology. A large active landslide lies over the ridge crest to the north, and sliding has also occurred on the slopes to the south. The bedrock is composed of interlayered quartzrich schist and mica-rich schist, and the relative

proportions of the two rock types vary in the stratigraphic column. Predominantly quartz-rich schist units form prominent escarpments, and units dominated by weaker mica-rich schist have gentler profiles. The quartz-rich schist is typically moderately strong, and is composed of alternating laminae of quartz-feldspar and mica. Mica-rich schist is weak, well laminated and fissile, due to mica dominating over quartz-feldspar. Weathering of the rock mass is typically slight. The schist has a pronounced anisotropy, with UCS highly dependent on foliation orientation. Previous testing of similar schist in the Clyde dam reservoir gave USC values typically ranging from 5–105 MPa (Beetham et al, 1992). A similar range of strengths are considered likely in the site area, with the mica-rich lithologies at the lower end of the range, (approximately 5–20 MPa) and the quartz-rich at the upper. No faults or major crush zones were mapped in the rock avalanche area but foliation shears are common. The dominant joints of Set A dip at an average 80◦ out of the slope and are persistent for tens of meters forming the faces of steep bluffs. (Figure 4) They are rough and moderately-widely spaced. Set B joints are sub-vertical and strike at right angles into the slope. They are rough and moderately-widely spaced.

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Figure 4. Lower hemisphere, polar, equal area stereoplots of mean attitudes of rock defects. The great circle indicates average dip of slope surface in the rock avalanche source area.

4

LARGE-SCALE TOPPLING

The rock avalanche source area is located on a spur exhibiting undulating cross-slope ridges and troughs, typical of ‘sagging’ geomorphology (Figures 1 & 3). This landform extends upslope towards the crest of the range, with ridges apparently corresponding to quartzrich units and troughs to mica-rich units. Outcrops are fractured and dilated, and schistosity dips into the slope at 20–35◦ compared with 50◦ in undisturbed schist. The evidence indicates that the rock in the spur has undergone large-scale toppling. Exposures in the sides of the rock avalanche source area show no obvious increase in dip to a depth of 10 m, and it is inferred that the hinge zone of the topple lies at greater depth. Large-scale toppling in the slope is facilitated by low friction angles along micaceous foliation defects, and undercutting of the toe by glacial action. Strong seismic shaking from major earthquakes associated with the nearby plate boundary, and elevated porewater pressures during extreme rainfall events will also have promoted the toppling process. Flexural toppling appears the dominant mechanism in the mica-rich schist, and block-flexural or block toppling in the quartz-rich schist. 5

to frequent heavy rainfalls. There were no significant seismic events at the time of failure. The rock avalanche was not directly observed, but a motorist driving along the Haast Pass Highway saw boulders falling onto the road and raised the alarm. Because of the risk of further rockfalls, the highway was immediately closed. Investigations were initiated to determine the source of the rockfall and the risk to the highway. The slopes above the highway were examined by helicopter and on foot. A lobe of rock avalanche debris was visible on a glacial bench high above the road, and the travel path could be traced back to the source area on the hillside above (Figure 5). The bulk of the debris had stopped on the bench, but some boulders continued rolling and bouncing over the bluffs towards the lake, with a small proportion hitting the highway. Their routes could be traced by tracks left in the scrubby vegetation (Figure 6). The debris volume was estimated to be of the order of 100,000 m3 . The travel distance was 300 m, but individual boulders continued a further 600 m to reach the road and lake.

Figure 5.

View of the rock avalanche from the toe.

ROCK AVALANCHE OBSERVATIONS

The failure followed a period of rainfall, but it was not an exceptional event for the region, which is subject

Figure 6. Traces left by rolling/bouncing boulders below the rock avalanche toe.

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6

SOURCE AREA DESCRIPTION

7

An aerial photograph taken several years before the rock avalanche (Figure 7) shows a pre-existing fresh scarp around the head of the source area. The rock types in the source area were closely examined after the failure (Figure 2). A prominent horizon of quartz-rich schist occupies the base, and forms the steep bluffs below. Above is a thick unit of mica-rich schist that underlies most of the failure, with a further quartz-rich unit at the head. The evacuated source area was roughly triangular in shape and sloped at 35◦ (Figure 3). A scarp up to 10 m high, faced by toppled schist, extended around the head and followed the southern margin. The scarp around the northern margin was generally lower and less defined, and lay in toppled schist or schist debris. Toppled schist exposed in the southern scarp was closely fractured and dilated along schistosity and other rock mass defects. Very rough bending fractures across schistosity were also evident. A layer of loose schist debris covered the floor and proved difficult to traverse on foot, due to the slippery micaceous surfaces of the slabs. Only limited field work was possible due to danger from falling debris. Of particular interest was a large semi-intact ‘raft’ of predominantly quartz-rich schist, still bearing vegetation, that had moved a short distance downslope from the head region. (Figure 3). A fresh scarp about 10 m high extended across the lower face. The ‘raft’ had clearly been displaced by translational sliding. A small exposure of a smooth micaceous silt coated plane underlain by schist was seen on the southern lateral margin adjacent to the ‘raft’, and is thought to be a section of the sliding surface. An apparently similar feature was seen from the air on the northern margin.

Figure 7. Source area prior to failure showing fresh scarp around head region, remaining boundary indicated.

FAILURE MECHANISM

The available field data, coupled with the properties of the rock mass, suggest the failure mechanism was translational sliding. Strong evidence to support this mechanism was found at the top of the source area, where the large semi-intact raft of quartz-rich schist remaining on the slope had undergone translational sliding. Smooth rock surfaces exposed at the adjacent lateral margins appear to be sliding planes. The remaining slope below (apart from the toe region) was underlain by weak, mica-rich schist. There was no field evidence to indicate the failure mechanism in this unit. However, studies of toppling mechanics (Nichol et al, 2002) indicate that catastrophic toppling failure is highly unlikely in such weak rock, and sliding is thus considered the probable mechanism. The quartz-rich schist in the toe region probably failed by sliding in a similar manner to the quartz-rich unit at the head. How the incipient slide plane developed through the toppled schist is unknown. No obvious persistent slope-parallel joints or other defect sets were detected, and the toppling hinge zone appears to lie too deep to form the failure surface. Possibly progressive failure along minor slope parallel fractures generated by the toppling process, or along stress relief surfaces, may have been involved. Translational slides are likely to be rapid if strain weakening on the rupture surface results in a residual friction angle significantly less than the rupture surface inclination (Fell et al, 2000). In slowly creeping schist landslides in the Cromwell Gorge, effective field strengths along failure surfaces range from 21–29◦ (Macfarlane et al, 1992). This range is considered indicative of residual strength values on well developed schist sliding surfaces in the region. The residual strength on the failure surface in the predominantly mica-rich schist at Sheepskin Creek would be expected to lie at the lower end of this range ie significantly less than the slope inclination of 35◦ . Thus rapid translational sliding would appear possible as a result of strain weakening. Following the main failure, a slow retrogressive slide of the quartz-rich unit at the head is thought to have occurred, translating the ‘raft’ a short distance downslope. Sliding may have been slower due to a higher residual friction angle in the quartz-rich schist. The fundamental watertable in the source area appears to lie beneath the sliding plane, but infiltrating rainfall may have perched on the incipient failure

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surface. Elevated porewater pressure from the period of rainfall before failure was probably the final trigger of movement.

8

DISCUSSIONS AND CONCLUSIONS

The rock avalanche is thought to have resulted from a rapid sliding failure of a weakened mass of toppled schist. Rapid translational sliding is thought to have occurred as a result of strain weakening, with the residual friction angle in the weak, mica-rich schist being significantly lower than the slope inclination. Large-scale toppling is widespread in the schist rock adjacent to the failure area and above other sections of the Haast Pass Highway, and rockfalls and rock avalanches present a hazard to the road. It may be possible to gain warning of future failures by aerial examination of the slopes for precursory scarps and tension cracks. There is high probability of a major earthquake on the Alpine Fault in the next 50 years. It is likely to trigger widespread rock avalanching in weak rock masses such as the toppled schist, and could result in major blockages of the Haast Pass Highway.

REFERENCES Beetham, R.D., Moody, K.E., Fergusson, D.A., Jennings, D.N. & Waugh, P.J. 1992. Landslide development in schist by toe buckling. Sixth International Symposium on Landslides. Christchurch, New Zealand. Fell, R., Hungr, O., Leroueil, S. & Reimer, W. 2000. Geotechnical Engineering of the Stability of Natural Slopes, and Cuts and Fills in Soil. Proc. GeoEng 2000, Melbourne, Australia. Gillon, M.D. & Hancox, G.T. 1992. Cromwell Gorge Landslides—A General Overview. Sixth International Symposium on Landslides. Christchurch, New Zealand. Halliday, G.S. & Mc Kelvey, R.J. 2004. Video-analysis of an extremely rapid rockslope failure at Nevis Bluff, New Zealand. Ninth International Symposium on Landslides. Rio de Janeiro, Brazil. Macfarlane, D.F., Riddolls, B.W., Crampton, N.A. & Foley, M.R., 1992. Engineering geology of schist landslides, Cromwell, New Zealand. Sixth International Symposium on Landslides, Christchurch, New Zealand. Nichol, S.L., Hungr, O. & Evans, S.G. 2002. Large scale brittle and ductile toppling of slopes. Canadian Geotechnical Journal 39:773–788. Prebble, W.M. 1992. Landslides in New Zealand. Sixth International Symposium on Landslides. Christchurch, New Zealand.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

NRCS-based groundwater level analysis of sloping ground L.I. Ju, O.T. Suk, M.Y. Il & L.S. Gon University of Seoul, Korea

ABSTRACT: Slope-related disasters have been happened between July and September which is rainy reason in Korea. It is known that the rainfall is one of the most important that lead to slope-related disasters. Based on slope analysis, rainfall intensity was applied as a factor of rainfall. It was calculated groundwater level in case of 100% rainfall infiltration. This paper calculated the amount of infiltration by using slope inclination, soil type and executed comparative analysis. NRCS model has the highest suitability level among analytic infiltration models. As a result, soil type occurred greatly permeation by the biggest permeability coefficient, and the small amount of infiltration in hydraulic conductivity happened in occasion of small D soil type. Also, this paper concludes the increasing in the amount of infiltration as slope and soil condition to calculate according to soil condition of slope range from D type to A type, and that infiltration occurs little and the slope is steep.

1

INTRODUCTION

Slope-related disasters have been happened in very short time and it could generate big damage. When it rains heavily, the collapse of slope was caused by rising pore water pressure of underground infiltration and erosion by surface flowing water, specific weight was increased activity of soil layer by saturated soil etc. Resisting powers for collapse reduced and stability of slope was fallen extremely. Therefore, if rainfall characteristics are applied rationally for stability analysis of slope, slope-related disasters can be forecasted and considered in advance. Rainfall intensity was applied as a rainfall factor for slope and this is permeated all of the rainfall in slope, excessive groundwater level can be lead to wrong estimation. It is a result which does not consider an outflow. In this paper, NRCS model was applied to calculate the amount of infiltration by analysis sections and it is necessary to study groundwater level based on time changes. 2 2.1

RESEARCH BACKGROUND AND THEORY Research background

The amount of infiltration, computed by rainfall, has achieved in different aspects. According to oversea research, D.M. Fox (1997) applied to the inclination slope and final infiltration capacity through the experiments. The final infiltration capacity was decreased by the inclination slope. Surendra K.M. and Vijay

(2004) suggested stable estimation method regarding the amount of infiltration based on NRCS model, using 55 infiltration observed data which were consisted of 4 soil groups by Mein and Larson (1971). S, NRCS-CN’s parameter, was presented dynamic behavior of Horton infiltration. λ was offered by using Horton parameter k and ponding time as well. Li Chen and Michael H. Young (2006) calculated the amount of infiltration for slope by using Richards equation and Green-Ampt equation. Shin (2003) suggested the correlation between slope safety and regulation of train operation considering change of slope safety ratio through the indoor experiments. Kim (2003) developed equation regarding rainfall infiltration depth which was achieved from slope related infiltration experiments for rainfall and the inclination slope. This empirical equation presented that rainfall infiltration depth is related to rainfall intensity. 2.2 Theoretical background NRCS (National Resources Conservation Service) developed for estimation of the effective rainfall from soil characteristics and land use condition in the ungaged basin. The effective rainfall in NRCS model included soil type and land use or vegetation type, hydrologic soil condition and AMC. NRCS the model is based on two assumptions. First, the maximum latency retention water(S) exists through the principal part and soil storage capacity in the basin. Second, actual storage capacity(F) and ratio of the maximum latency retention water(S) are equal to the amount of

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rainfall P which was inducted from outflow Q and initially the loss Ia .

3

Q F = S P − Ia

3.1 Computation of infiltration

(1)

Ia and S have the linear relationship as following. (2)

Ia = aS

Where, a is constant. When it is the heavy rain, a is applied to 0.2 or 0.1.

THE AMOUNT OF INFILTRATION AND GROUNDWATER LEVEL COMPUTATION

KMA (Korea Meteorological Administration), Seoul data (1961∼2005) were applied to calculate the amount of infiltration and the most suitable distribution was selected to GEV dist. For 24 hour and 50 years return period by frequency analysis. The determinated probable rainfall is 434.4 mm and the amount of effective infiltration is decided by Huff method. Fig 1 and 2 display accumulated infiltration slope 1:1.0 and 1:1.2, these were analyzed according

(3)

F = P − Ia − Q

Eq. (3) is the water balance equation and Eq. (4) is inducted from Eq. (1) and Eq. (3). F can be induced as following. S(P − Ia ) P − Ia + S

F=

(4)

Eq. (5) shows a cumulative infiltration with P, S and Ia. The differential Eq. (4) is as following. f =i

S2 (P − Ia + S)2

(5)

Where, f = dF/dt, I = dP/dt. Parameter S is calculated from CN (Curve Number). S can be changed from infiltration progress condition and the infiltration is changed soil type and overland condition. The decision of S is as following Eq. (6). S=

25, 400 − 254 CN

Figure 1.

Accumulated infiltration(1:1.0).

Figure 2.

Accumulated infiltration(1:1.2).

(6)

where, CN has the range from 1 to 100 and the relationship with CN and S is shown as Table 1. Table 1. Relations of CN and S. CN

S

P vs. Q

Note

1

25,146

S is in inverse proportion to CN.

100

0

All rainfall can be loss and the direct runoff is equal to 0. There is no rainfall loss and total rainfall is equal to a direct out flow.

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Height(m)

to soil specific characteristics. As a result, soil A type was occurred great permeation by the biggest permeability coefficient, and the small amount of infiltration was happened in occasion of soil D type. Table 2 displays the accumulated infiltration ratio for total rainfall.

Groundwater Level

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0

Infiltration Flow

0

1

2

3

4

5

6

7

8

9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35 36 37 38 39 40

Distance(m)

a)

Slope

1:1.0

1:1.2

CN

A B C D

0.603 0.337 0.223 0.166

0.637 0.347 0.227 0.168

30 58 71 78

Height(m)

Table 2. Total rainfall ratio and accumulated infiltration (NRCS model).

Groundwater Level

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0

Infiltration Flow

0

1

2

3

4

5

6

7

8

9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35 36 37 38 39 40

Distance(m)

b)

Groundwater Level

Height(m)

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0

Infiltration Flow

0

1

2

3

4

5

6

7

8

9

10

11

12

13

14

15

16

17

18

19

20

21

22

23

24

25

26

27

28

29

30

31

32

33

34

35

36

37

38

39

40

c)

Infiltration Flow

Height(m)

Height(m)

Distance(m)

Groundwater Level

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0 0

1

2

3

4

5

6

7

8

9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35 36 37 38 39 40

Distance(m)

a)

Groundwater Level

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0

Infiltration Flow

0

1

2

3

4

5

6

7

8

9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35 36 37 38 39 40

Distance(m)

Height(m)

d) Groundwater Level

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0

Infiltration Flow

0

1

2

3

4

5

6

7

8

Figure 4. Groundwater level change by slope 1:1.2 (120 hours) a) A soil type, b) B soil type, c) C soil type, d) D soil type.

The amount of infiltration has increased from soil D type to soil A type according to slope and soil condition. Infiltration was occurred little in the steep slope.

9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35 36 37 38 39 40

Distance(m)

Height(m)

b) Groundwater Level

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0

Infiltration Flow

3.2 0

1

2

3

4

5

6

7

8

The change of groundwater level was observed be using the amount of infiltration for NRCS model after 120 hours. Fig 3 shows that groundwater level was changed by soil A type (CN = 30), soil B type (CN = 58), soil C type (CN = 71), soil D type (CN = 78) based on 1:1 slope. Fig 4 shows groundwater level at 1:1.2 slope.

9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35 36 37 38 39 40

Distance(m)

Height(m)

c) Groundwater Level

20 19 18 17 16 15 14 13 12 11 10 9 8 7 6 5 4 3 2 1 0

Infiltration Flow

0

1

2

3

4

5

6

7

8

Computation of groundwater level

9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34 35 36 37 38 39 40

Distance(m)

4

d)

Figure 3. Groundwater level change by slope 1:1.0 (120 hours). a) A soil type, b) B soil type, c) C soil type, d) D soil type.

CONCLUSIONS

In this study, the amount of the effective infiltration was estimated by slope using NRCS model. It was applied to calculate the groundwater level at Sand

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Load of which slope 1:1.0 and 1:1.2. When the groundwater level is located near the slope, the slope is unstable because of the amount of excess infiltration in case of soil A type and slope 1:1.0 and 1:1.2 as well. However, soil B, C, D types have slight effect on groundwater level. In addition, the ground water level has been increased by 3% due to the amount of infiltration in the case of 1:1.2 slope compared to 1:1.0 slope. Finally, in further study, it is necessary to evaluate slope stability and consider stability by soil type on the groundwater as well. ACKNOWLEDGMENT This research was supported by a grant (NEMA06-NH-05) from the Natural Hazard Mitigation Research Group, National Emergency Management Agency. REFERENCES Bujang, B.K., Huat, Faisal Hj. Ali. & T.H. Low. 2006. Water infiltration characteristics of unsaturated soil slope and its effect on suction and stability. Geotechnical and Geological Engineering. 24, pp1293–1306. Fox, D.M., Bryan, R.B. & Price, A.G. 1997. The infiltration of slope angle on final infiltration rate for interrill coditions, GEODERMA 80 pp181–194. Kim Hyeon-Ki, Lee Jin-Wook & Shin Min-Ho. 2003. The Development of Rail-Transport Operation Control using the Variation of Slope Stability under Rainfall. Korean Society For Railway.

Li Chen & Michael H. Young. 2006. Green-Ampt infiltration model for sloping sufaces. Water Resources Research, Vol. 42, W07420. Moon Young-Il, Shin Dong Jun, Oh Keun-Tak, Shin Heung—Keon & Lee Su-Gon. 2007. Analysis of the Groundwater level and Characteristic of the Sloperelated Disasters according the Infiltration. Korea Water Resources Association. Nanee Chahinian, Roger Moussa, Patrick Andrieux & Marc Voltz. 2005. Comparison of infiltration models to simulate flood events at the field scale. Journal of Hydrology, 306, pp191–214. SEEP/W. 2002. User’s Guide, GEO-SLOPE International Ltd. Surendra, K.M., Shashi, R.K. & Vijay, P. Singh. 1999. Calibration and validation of a generral infiltration model. Hydrol. Process. 13, pp1691–1718. Surendra Kumar Mishra & Vijay P. Singh. 2004. Validity and extension of SCS-CN method for computing infiltration and rainfall-excess rates. Hydrol. Process. 18, pp3323–3345. Tingwu Lei, Yinghua Pan, Han Liu, Weihua Zhan & Jianping Yuan. 2006. A run off-on-ponding method and models for the transient infiltration capability process of sloped soil surface under rainfall and erosion impacts. Journal of Hydrology 319, pp216–226. William, L. Crosson, Charles A. Laymon, Ramarao Inguva, Marius P. & Schamschula. 2002. Assimilating remote sensing data in a surface flux-soil moisture model. Hydrol. Process. 16, pp1645–1662. Zhou, Z.C. & Shangguan, Z.P. 2007. The effects of ryegrass roots and shoots on loess erosion under simulated rainfall. CATENA 70, pp350–355.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

A numerical case study on load developments along soil nails installed in cut slope subjected to high groundwater table A.K.L. Kwong & C.F. Lee Department of Civil Engineering, The University of Hong Kong, Hong Kong

ABSTRACT: Soil nailing is a technique routinely used in Hong Kong whereby closely spaced steel bars are installed into a slope so that the stability conditions of slopes comprising either in-situ or compacted soils can be improved. A full-scale field test has been carried out by The Department of Civil Engineering of The University of Hong Kong to study the development of passive load along the full length of soil nails when subjected to induced rise in groundwater table. The cut slope was formed to a very steep angle of 55◦ and 10 m high in completely decomposed volcanic material. Grouted curtain was also formed behind, bottom and on both ends of the slope in order to form an impermeable barrier that would allow groundwater table to increase artificially by injecting water into slotted PVC inlet pipes. Nine number of soil nails (in regular 2 m c/c spacing of 3 rows and 3 columns) of 6 m long high yield steel bar were installed at 15◦ from horizontal into the formed cut slope. Instrumentation included strain gauges along the nails, inclinometers, piezometers, moisture probe and settlement prisms. This paper describes the load developed along the instrumented soil nails when the groundwater table was raised to the ground surface. It was found that the measured passive load along the soil nail was smaller than the commonly assumed design parameters, an indication that substantial savings can be achieved if mobilization of shearing resistance along the full length of the soil nail was considered in routine design. Finite element analysis has also been carried out to compare the stability factor with the factor of safety. A comparison of factor of safety of this soil nailed system using conventional limit equilibrium and that from strength reduction technique is also described in this paper.

1

INTRODUCTION

Soil nailing is a technique whereby closely spaced steel bars are installed into a slope to improve its stability conditions. This method has been routinely used in Hong Kong and overseas, and the analysis and design of the soil nails are mostly based on limit equilibrium approach which compares the resisting moment and force with the driving moment and force on the most critical slip surface. A safety factor is usually applied in the design so that the resisting moment and force will be larger than the driving moment and force. Limit equilibrium approach does not consider how much deformation will be generated in the soil mass before the passive force can be fully mobilized in the soil nail. It has also not considered the relative stiffness between the soil and the steel bars. To date, there has not been any field tests in Hong Kong conducted on slope comprises in-situ soil where load development along the steel bars are monitored during each stage of rise in groundwater table to the point of failure of the whole soil mass. Thousands of slope in Hong Kong have been upgraded using soil nails but they are not monitored. The only available

field measurement of soil nail forces in Hong Kong was in a report given by Yim and Yuen (1998) where a 7 m high slope in completely decomposed volcanic was cut to 70◦ and the working loads developed in the 4 rows of nails were only 7% (10 kN) to 52% (70 kN) of the design working load (134 kN) of the nails. No movements were observed. Li (2003) summarized the current design and installation practice of soil nails in Hong Kong and described the construction and instrumentation of a loosely compacted soil-nailed slope. Li (2003) concluded that although the loads developed in the nails were small and at the working stage (40 to 70 kN), it had significantly reduced the slope deformation by preventing the formation of shear band or distinctive slip surface. Failure of the slope was governed by excessive movement (both lateral and downward due to surcharge at the top) rather than due to failure of the nail or failure of the bond between the nail and the soil. Using a strength reduction technique, Cai and Ugai (2003) modeled an idealized slope reinforced with a single row of nail and compared the safety factors calculated by Bishop’s method. A range of nail orientation, spacing, and shear strength of the soil-grout

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interface was analyzed and reasonable agreement was achieved on the factor of safety (FOS). However, field measurements are not available to compare the load developed in the nail and the deformation of the soil-nailed slope at failure with that predicted by the finite element method. It is imperative that the load and deformation predicted in the model is checked and calibrated against field data before it could be adopted for use in the practice. In this regard, a research involving a full-scale field test has been carried out with instrumentation mounted at strategic locations in order that the failure mechanism and the soil-nail interaction can be fully understood This paper describes the full-scale field test carried out at the Kadoorie Agricultural Research Centre of The University of Hong Kong, where groundwater table was artificially raised to the surface to study the development of loads in the soil nails and compare with that obtained using a finite element approach. The objective of the research is to develop an in-depth understanding of the failure mechanism and load development along the nail at each stage of the loading step, based on a full-scale test with dimensions compatible to common slope upgrading work in Hong Kong. Based on the instrumented results and the calibrated numerical model, it is hoped that savings can be achieved in the future design of soil nails in Hong Kong.

2

Figure 1. General site location plan at Kadoorie Agricultural Research Center.

Figure 2. Overall view of cut slope with completed instruments and water inlet pipes installed.

CONSTRUCTION OF FULL-SCALE FIELD LOAD TEST

2.1 Construction procedure At the Kadoorie Agricultural Research Centre of The University of Hong Kong, a cut slope was formed to a very steep angle of 55◦ and 10 m high in completely decomposed volcanic material (see Figure 1 for location). Two boreholes were drilled at the crest and the slope comprised 1.5 m of fill overlain 1 m of residual soil above the completely decomposed volcanic (CDV) layer. The residual soil can be described as firm, moist, brown, grayish brown, slightly sandy silt with some angular to sub-angular fine to coarse gravel, whereas the CDV can be described as extremely weak, light grey, completely decomposed fine ash crystal TUFF (slightly sandy silt with occasional angular to sub-angular fine gravel). Grouted curtain was formed behind, at the bottom and on both side ends of the slope in order to form an impermeable barrier that would allow groundwater table to increase artificially by injecting water into slotted PVC inlet pipes. Nine number of soil nails (in regular 2 m c/c spacing of 3 rows and 3 columns) of 6 m long high yield steel bar were installed at 15◦ from horizontal into the formed cut slope. Figure 2 shows the slope formed before the

Figure 3.

Plan view of instruments installed.

test began in October 2005. A detailed construction procedure is being published (Kwong and Lee, 2008). Instruments including strain gauges along the nails, inclinometers, piezometers, moisture probe and settlement prisms were installed with locations shown on Figure 3 and cross sectional view shown on Figure 4.

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small diameter grout pipes (25 mm diameter perforated pipes) cannot provide sufficient water inflow into the slope to induce a rise in groundwater table when the test was first carried out in September 2005. Subsequently, larger diameter grout pipes (50 mm diameter perforated pipes in 100 mm diameter hole) were installed in October 2005 and with a 24-hour supply of water (with an average flow rate of about 30 l/min) continuously in 10 days, the groundwater table was finally raised to the bottom of the fill layer, approximately 1.5 m below the surface. Figures 5 and 6 show the maximum pore water pressure recorded in the piezometers and maximum load developed in the nails between 20 Oct. and 31 Oct. 2005. 3

NUMBERICAL MODELLING

Figure 4. Sectional view of instruments installed.

3.1 Parameters and procedures

Figure 5. Pore pressure recorded during induced rise in groundwater table.

Consolidated triaxial tests were carried out on the Mazier samples of residual soil and CDV and their shear strength properties are shown in Table 1. For a 10 m high slope with a steep angle of 55◦ , the use of these typical shear strength parameters under an 8 m pressure head (see Figure 4) would result in a factor of safety (FOS, conventional limit equilibrium method) equal to 0.689. The design of this experiment is to induce groundwater table so that the slope would reach or close to failure. Therefore, 9 number of 25 mm diameter high yield steel bars with 2 m c/c horizontal spacing were installed and the load monitored throughout the test. The resulting factor of factor is about 1.104 with the design load of 17 kN, 20 kN and 25 kN in the upper, middle and lower nails respectively, as shown in Figure 7. A finite element study is then carried out to investigate whether the measured loads are close to the design loads. The following steps were employed in the modeling sequence so that it follows closely the stress path in the field. 1. Set up the geometry and boundary of the problem and turn on gravity to the soil mass. A high elastic property is assigned to all elements to ensure numerical stability.

Figure 6. Load recorded in soil nails during induced rise in groundwater table.

Table 1.

2.2

Materials

Cohesion kPa

Frictional angle degree

Fill Residual soil CDV

0 5 6

35 30 33

Measurements of pore pressure, loads and movements

Although an impermeable barrier was formed by grouting on 4 sides of the slope (1 back face, 2 side faces and 1 bottom), the use of four number of

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Shear strength properties of residual soils and CDV.

of 8 mm, which is very close to the measurement and all the analyses were repeated using this value. The thickness of the fill and residual soil is small and the same Young’s Modulus as that of CDV is assumed. Poisson’s ratio of CDV is taken as 0.25. Young’s Modulus of steel is taken as 2 × 108 kPa. 3.2 Comparisons of field measurements with numerical simulations Figure 8 shows a comparison of the measured load with that generated from the finite element method at three groundwater levels (highest being 145.5 mPD shown in Figure 7). Figure 7. Minimum factor of safety (from limit equilibrium) at highest groundwater level.

2. Excavate the slope (unloading) and change the elastic model to elastic-plastic model. Steps to equilibrium while the groundwater table is maintained at the toe level. 3. Assign structural elements to the grids, beyond the potential slip surface, to represent the installation of the passive soil nails in an existing slope. 4. Raise the groundwater table gradually and steps to equilibrium at each step. 5. Record the development of loads along the soil nails at each stage of groundwater rise and also the maximum movement of the slope. 6. Calculate the stability factor (SF) of the slope. The stability factor of a slope by finite element stress method is defined as the ratio of the summation of the available resisting shear force to the summation of the mobilized shear force along a slip surface. They are different in different points on the slip surface. 7. Calculate the stress reduction factor (SRF) by repeating the above steps with a new shear strength parameter (divide the design shear strength parameter by a factor of safety), steps to equilibrium until numerical instability occurs. 8. Repeat the above steps by varying the Young’s Modulus of Completely Decomposed Volcanic from 2 × 104 kPa, 5 × 104 kPa, 5 × 105 kPa to 10 × 105 kPa). 9. For each above step, plot the development of loads along each nail and compare with that measured in the field. The inclinometer readings show that under the highest water level, the outward rotation in the middle section of the slope is about 0.3◦ or 10 mm movement. Based on the movement at the toe of the slope from the finite element analysis, the use of the Young’s Modulus equal to 5 × 104 kPa would predict a movement

Figure 8. Comparisons of measured load with the finite element results.

Table 2. Comparison of maximum and average load measured in three rows of nails with that from finite element results. Maximum measured load kN Upper nail 19.6 Middle nail 24.4 Lower nail 36.2

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Maximum simulated load kN

Average measured load kN

Average simulated load kN

29.2 24.6 31.4

13.2 13.5 20.1

15.9 11.4 19.5

Table 2 shows a comparison of the maximum and average load measured with that from the finite element results at the three rows of the nails. Very close agreement can be found between the ones generated from the finite element with those measured in the field.

Figure 9. Plot of factor of safety and stability factor against rise in water level.

Figure 10. Critical slip surface location and maximum shear strain contour at highest water level of 145.5 mPD.

Figure 11.

4

FINDINGS FROM NUMERICAL SIMULATIONS

With the proper selection of soil parameters that have been calibrated with field measurement, and numerical simulation procedure following closely to that in the actual construction sequence, it has been demonstrated that the load simulated in the finite element model can replicate those in the field with close agreement. However, there is significant implication in practice if we then assess the stability condition of the slope with the developed nail forces. Is the slope approaching failure when the load developed is less than that assumed in the design and when the factor of safety and stability factor are so close to 1.0? Figure 9 shows the plot of factor of safety and stability factor as the water level increases without nails. It shows that the difference between FOS and SF is very small and at the highest water level of 145.5 mPD, the FOS and SF are 0.689 and 0.688 respectively. If the average loads measured in the field (13 kN, 14 kN and 20 kN for the upper, middle and lower nails respectively) are used, the corresponding FOS is equal to 0.966. After nails are installed and the loads matched very well with the field measurement, the SF increased to 1.030 at the highest water level of 145.5 mPD. Figure 10 shows the critical slip surface is very close to the maximum shear strain contour, an indication that the mechanism developing may be physically approaching unstable condition. Stress reduction factor is also calculated at each stage when the water level increases, and it is 1.110 at the highest water level of 145.5 mPD. It is noted that no convergence problem occurred in the numerical

Plot of FOS and SF with an increase in water level for the cases without soil nail and with nail.

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procedure because of the very small loading steps and a simple elastic perfectly plastic model adopted. In the SRF method, the limiting load that can be developed at the nail is assumed to be 100 kN, based on 4 mm movement from the pullout test. The above findings are summarized in Figure 11. It shows the change of FOS and SF as a function of water level increases on the cases without soil nail and with soil nail. The following observations are made: 1. If the measured nail forces are used in the limit equilibrium method, the FOS is equal to 0.966. Theoretically, the slope should have failed but it had not. 2. If the commonly assumed design nail forces are used in the limit equilibrium method, the FOS is equal to 1.104. The slope should have been very close to unstable condition, but less than 10 mm movement was observed. There were no signs of instability or distresses observed. The magnitude of the load developed was very small in the order of 30 to 40 kPa, indicating that the slope was very stable even though water level was raised almost to the top surface. 3. When nails are installed, the decrease in SF is less than that of FOS, indicating that if there was a further rise in water level, the corresponding decrease in SF is still small and may still be above 1.0. 4. Although SRF is close to FOS at the highest water level, the developed load from the finite element method was 100 kN, which is much higher than the measured. This indicates that it may not be an appropriate method for this system where soil structural interactions and relative stiffness are involved. 5

CONCLUSIONS

2. Measured movement and load allowed proper calibration of material properties for numerical works. 3. The average loads at different rows of the nails from the numerical simulation compares very well with that measured, although some differences appear in the distribution along the entire length. 4. Although the FOS (based on limit equilibrium method) was close to 1.0, there was no sign of significant movement or distress observed. 5. SF may be a good approach to use because its changes are gradual and the method considered the relative stiffness and movement between the steel and the soil. ACKNOWLEDGEMENT The financial support by the Research Grant No. HKU7109/04E provided by the Research Grants Council of the Hong Kong Special Administrative Region, China is greatly appreciated. REFERENCES Cai, F. and Ugai, K. 2003. Reinforcing mechanism of anchors in slopes: a numerical comparison of results of LEM and FEM. International Journal for Numerical and Analytical Methods in Geomechanics, 27, pp. 549–564. Kwong, A.K.L. and Lee, C.F. 2008. A field test study on instrumented soil nail installed in cut slope. Sixth International Conference on Case histories in Geotechnical Engineering and Symposium, Arlington, VA (USA)-August 11–16, 2008 (accepted). Li, J. 2003. Field study of a soil nailed loose fill slope. Ph.D thesis, The University of Hong Kong, 218p. Yim, K.P. and Yuen, K.S. 1998. Design, construction and monitoring of a soil nailed slope in decomposed volcanics. Slope Engineering in Hong Kong, Balkema, pp. 67–73.

The following conclusions can be made based on the field measurements and numerical works. 1. Rise in groundwater level was successfully induced, thus allowing the loads developed in the soil nails accurately measured.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Landslides at active construction sites in Hong Kong T.M.F. Lau, H.W. Sun, H.M. Tsui & K.K.S. Ho Geotechnical Engineering Office, Civil Engineering and Development Department, Hong Kong SAR, China

ABSTRACT: Notable landslides have occurred at construction sites involving slope works in Hong Kong. Apart from posing a risk to the site workers and causing delays and cost overrun to projects, slope failures at construction sites can also threaten public safety as the failures can affect facilities beyond the site boundary. There have been cases resulting in casualties as well as near-misses. This paper presents some case studies of selected notable landslides incidents and the key findings from the post-failure investigations. The review has identified some areas that warrant attention in order to mitigate the risk of slope failure during execution of slope works.

1

INTRODUCTON

Following the disastrous landslides in the 1970’s which killed more than 150 people in Hong Kong, the Government established the Geotechnical Control Office (renamed as Geotechnical Engineering Office (GEO) in 1991) in 1977 as a central authority to regulate geotechnical practice and manage slope safety in Hong Kong. Since 1977, the design and construction of all slope works are subject to geotechnical control by the GEO to ensure public safety. The GEO provides roundthe-clock emergency services in respect of landslides. Professional geotechnical engineers are dispatched to inspect significant landslides and provide advice on the need for road closure, building evacuation and the necessary emergency works to remove the immediate danger and avoid casualties. In addition, systematic landslide statistics have been compiled since the mid-1980’s. Notable landslides are studied by the GEO to improve the understanding of the causes and mechanisms of slope failures. 2

LANDSLIDES DURING SLOPE WORKS

Notable landslides have occurred at construction sites involving slope works in Hong Kong. Apart from posing a risk to the site workers and causing delay and cost overrun to projects, slope failures at construction sites can also threaten public safety as the failures can affect facilities beyond the site boundary. There have been cases resulting in casualties as well as near-misses whereby casualties were fortunately avoided. Based on the available landslide records, there are a total of 76 incidents that occurred on slopes within

Figure 1. Annual landslide incidents during slope works (1984–2006).

Figure 2. Consequence of landslide incidents during slope works (1984–2006).

active construction sites between 1984 and 2006. The statistics of slope failures within construction sites are presented in Figure 1. Of the total 76 reported incidents, 25 (about 33%) are major landslides (defined as a landslide with a volume ≥50 m3 ).The consequences of the corresponding landslides are summarized in Figure 2. During the period of 1984 to 2006, slope failures within construction sites had

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resulted in one fatality and nine injuries in six separate incidents. Other consequences have included blockages or closures of roads (representing 42% of the 76 landslides), temporary evacuation of buildings (13%), and evacuation of squatter structures (11%). Thirty-six cases had comparatively more serious consequences, in which 6 occurred on private works sites and 30 on government works sites.

3

TYPES OF FAILURES

Twenty-two of the above 36 more serious cases were classified as ‘landslides’, three as ‘rock/boulder falls’, one as ‘flyrock incident’, five as ‘retaining wall failures’ and five as ‘washout incidents’. About 70% of the cases involved insitu soil/rock and the other 30% involved fill material. About two-thirds of the 36 cases were associated with the formation of new slopes in site formation projects and the remaining third occurred during upgrading works on existing slopes. Some notable incidents are highlighted below.

4

CASE STUDIES

4.1 The 18 August 1995 rockfall at Tuen Mun highway, Siu Lam The rockfall occurred during scaling works which were in progress on a rock cut slope some 30 m above Tuen Mun Highway (Figures 3 and 4). The excavation works were part of a road improvement project which was being carried out under a design-and-build contract. The rockfall involved the detachment of a rock block of about 2 m by 2 m by 2.5 m in size from the overhanging rock outcrop at the slope crest onto the highway below. The detached rock block bounced and rolled across the road lanes, landing on the middle

Figure 4.

Cross-section of the slope at Tuen Mun Highway.

eastbound lane of the highway. A van travelling along the road drove into the large rock block as a result of which the driver was killed and the passenger in the van was injured. The incident also resulted in the closure of three eastbound lanes and the suspension of the works on the rock cuts for that section of the project for about 11 months. The scaling works involved rock splitting using drilling, feathering and wedging. No precautionary measures such as cabling or dowelling were implemented before the detachment caused by the site operations. The incident has highlighted some major lessons from a risk management point of view. The Coroner noted that the rock breaking works were not adequately supervised. The Coroner also noted some apparent confusion between the contractor and its Independent Checking Engineer regarding the consultant’s role under the contract. As a result, the design of the temporary works was not checked properly.

4.2 The 4 December 1997 rockfall at Sau Mau Ping Road

Figure 3. The 18 August 1995 rockfall at Tuen Mun Highway.

The failure occurred on a 25 m high rock slope adjacent to Sau Mau Ping Road (Figures 5 and 6). At the time of the incident, the top portion of the rock cut was being excavated under the on-going site formation works. The failure happened a few seconds after blasting had taken place behind the slope crest. As a normal safety precautionary measure, the road section was closed to vehicular and pedestrian traffic during blasting. The volume of the failure debris was about 1000 m3 and contained some large angular rock blocks, the largest being about 150 m3 in volume. The debris destroyed a section of the protective steel fence erected along the slope toe, completely blocked a 25 m long section of Sau Mau Ping Road and punched through the metal hoarding on the far side of the road at several locations.

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The failure resulted in the complete closure of Sau Mau Ping Road for 17 days. According to the detailed landslide study by the GEO (Leung et al, 1999), blasting was carried out at a close distance to the crest of the failed slope and the amount of explosives used was found to have exceeded the permitted value. The slope failure was probably triggered by the shock waves and gas pressures generated by the blasting which took place close to the slope crest.

4.3 The 9 and 10 June 1998 landslides at Junction of Sai Sha Road and Tai Mong Tsai Road The two landslides (with failure volumes of about 900 m3 and 1700 m3 respectively) occurred on a soil cut slope where slope upgrading works were in progress. The works comprised the construction of a retaining wall at the slope toe with rockfill up to about two-thirds of the slope height (Figures 7 and 8). Permanent soil nails were provided to support the upper third of the slope and temporary soil nails were proposed to support a 50◦ temporary cut at the lower portion for wall construction. Both landslides occurred when the soil nails were being installed during the temporary works stage. A detailed investigation (FSWJV, 1999) of the incidents revealed that the ground conditions at the site were complex with persistent infilled relict discontinuities, high and complex groundwater regimes and possible subsurface drainage concentration due to the presence of a depression in the rockhead, a history of major failures, etc. The landslides were likely caused

Figure 5. The 4 December 1997 rockfall at Sau Mau Ping Road.

Figure 7. Road.

The 9 and 10 June 1998 landslides at Sai Sha

Figure 6. Cross-section of the site at Sau Mau Ping Road.

Figure 8.

Cross-section of the failed slope at Sai Sha Road.

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by groundwater conditions that were more severe than that assumed in the temporary works design, which relied on the works being carried out during the dry season. The construction was delayed and temporary slope works had to be carried out during the wet season, hence exposing the temporary cuts to conditions not previously allowed for in the design. Some of the nail heads had not been constructed at the time of failure and the detailing of the nail heads, which incorporated small (150 mm square) steel plates, were probably not effective in providing support to the temporary cuts. Figure 10. Cross-section of the landslide site at Lung Ha Wan Road.

4.4 The 24 August 2000 incident below Lung Ha Wan Road The incident occurred during heavy rainfall at a road widening project. A 20 m long section of a temporary sheet pile wall failed with an estimated failure volume of about 200 m3 . The failure resulted in the temporary closure of Lung Ha Wan Road for two days (Figures 9 and 10). The sheet pile wall was being constructed to provide temporary support to Lung Ha Wan Road during the construction of a new retaining wall. The subsequent landslide study (HCL, 2001) concluded that the failure of the wall was probably due to the build-up of groundwater pressure behind the wall together with inadequate lateral support to the wall. The build-up of groundwater pressure was likely to have been partly associated with the driving of the sheet pile wall through a culvert that drained a fairly large catchment area and causing a blockage to the culvert. The temporary sheet pile wall system was designed by the contractor and certified by an Independent Checking Engineer. The design was also accepted by

the Engineer. However, it was observed from postlandslide inspections that the shoring for most of the failed sheet pile wall was not in place and that the construction of the shoring system in the adjacent section did not conform to the approved drawings. Despite repeated warnings from the Resident Engineer over a 6-month period preceding the failure with regard to the non-conformities, no action was taken by the contractor to rectify the non-conformities.

5

DISCUSSIONS

Slopes can be vulnerable to failure during site formation or slope upgrading works, especially when the construction is carried out during the wet season. Adequate planning, scheduling and supervision of the works are of the essence. Based on an overall review of the previous incidents involving landslides at active construction sites, key areas that warrant attention are presented in the following.

5.1 Buildability of the design

Figure 9. The 24 August 2000 incident below Lung Ha Wan Road.

The buildability of slope designs should be thoroughly considered by the designers at the option assessment stage as well as the detailed design stage to avoid conditions whereby slopes are rendered particularly vulnerable to failure during construction. It is also important to ensure that the method statements should include detailed procedures on the proper use of explosives for blasting. Also, the Blasting Assessment submitted by the contractors should contain sufficient details including, where necessary, contingency measures to cater for possible changes in the site conditions or scenarios whereby the actual ground conditions are more adverse than that assumed for the design.

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5.2

Management of site supervisory staff and auditing of site works

The designer should: i. ensure that the site supervisory staff are suitably qualified and experienced; ii. formulate a plan to manage the site supervisory staff and upkeep the standard of site supervision; iii. provide site supervisory staff with sufficient information for their appreciation of the geotechnical content of the works, key design assumptions and potential anomalies that could be encountered; iv. review regularly the adequacy of the site staffing level and make necessary adjustments to suit the prevailing circumstances; v. carry out site audits without prior warning, particularly for critical site activities, to check the standard of site supervision (including site staff’s familiarity and knowledge of the site activities and progress); and vi. ensure that prompt action is taken to rectify the situation where non-compliances are identified by the audits and undertake follow-up audits to check and ensure the effectiveness of the rectification measures. 5.3

Site supervision

An adequate site supervisory staffing level commensurate with the scale and complexity of the works should be provided. All critical works that will be buried in the ground whereby the quality of the asbuilt works will not be readily visible (e.g. insertion of steel reinforcement and grouting of soil nail drill holes, placement of drainage filter material and construction of subsoil drain, etc.) should be supervised on a fulltime basis with detailed records kept. The competence of the site staff and their familiarity with the technical and contractual requirements of the project should be continually reviewed by the designer. The site staff should be alert to the range of potential anomalies and prepare comprehensive records of relevant observations, such as signs of ground distress, suspected deviations from the design assumptions (e.g. high level seepage) or discovery of unexpected water-carrying services, and report promptly to the designers for advice. In particular, they should be alert to identifying non-compliances with the specifications or agreed method statements and initiate prompt action to rectify the situation. 5.4

Periodic supervision by geotechnical professionals

Suitably qualified geotechnical professionals should undertake periodic supervision and liaise closely with the resident site staff to ensure that the works are not

carried out in such a manner as to invalidate the design assumptions. Arrangement should be made for independent checking of any specific recommendations involving major changes in, or refinement of, the design. The possible need to seek further specialist advice from an experienced engineering geologist in geologically complex sites where appropriate should also be borne in mind. 5.5 Temporary support and slope protection Large and steep unsupported temporary cuts should be avoided as far as possible. Adequate support must be provided to temporary cuts to maintain their stability. For sizeable temporary cuts in fill material or geologically complex or potentially problematic sites, the designer should carry out a stability assessment of the temporary cut and the design of the temporary support system including consideration of the construction sequence. Apart from the stability of the cut face, consideration should also be given to the possibility of excessive ground movement to the adjacent facilities caused by the formation of a temporary unsupported cut. Sufficient instrumentation of the adjacent facilities should be provided and suitable pre-determined trigger levels for action and pre-defined action plans and action parties should be established. During the wet season, the extent of temporary cuts or stripping of existing hard surfacing/vegetated cover on slopes should be kept to a minimum as required for the execution of the works. Adequate temporary protection (e.g. shotcrete, tarpaulin, etc.) should be provided to any exposed slope face. Where an impermeable cover is to be applied to a slope face, in particular a temporary cut, adequate provisions should be made to maintain the functions of all existing subsurface drainage measures behind the slope face (e.g. drainage blankets and raking drains). 5.6 Temporary drainage provisions Slope works can be particularly vulnerable to washout failure due to uncontrolled surface runoff and susceptible to instability due to concentrated water ingress. Adequate temporary surface drainage must be provided at all times, especially during the wet season, to avoid the adverse effects of uncontrolled surface water flow. The temporary drainage system should be maintained and cleared of any blockage on a regular basis to ensure that the drains remain functional especially at times of heavy rainfall. Consideration should be given to requiring the contractor to construct part of the permanent drainage measures at an early stage of the works (e.g. crest drain and the associated discharge points) to enhance the drainage provision during construction.

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The assessment of the adequacy of temporary drainage should take account of the overall site setting in an integrated manner, with due consideration given to the possible influence of adverse topography and impact of adjacent works where appropriate. 5.7

Management of earthworks

to check the length of the reinforcement, exposing a section of constructed soil nail (e.g. by an inspection pit near the slope surface or through sinking a trial pit from a slope berm) to inspect the integrity of the grout annulus. 6

CONCLUSIONS

For sites involving earthworks, particularly in a congested works area such as fill slope re-compaction sites, due attention should be given to the proper management of the excavated spoil or imported fill through proper planning and good house-keeping. Significant stockpiles of loose materials that are unprotected against infiltration must be avoided, as these can be vulnerable to sudden and mobile failure upon water ingress.

A holistic review of the reported cases with notable failures during slope works in Hong Kong has been carried out. The review has identified some areas that warrant particular attention in order to mitigate the risk of slope failures during construction viz. temporary site drainage, temporary slope support and protection, non-compliance or poor workmanship, management of earthworks and site supervision.

5.8

ACKNOWLEDGEMENTS

Non-compliance/poor workmanship

Any discovery of non-compliances of the site works by the resident site staff should be reported to the designer/project office as soon as possible. Follow-up actions should be stepped up progressively by the site supervisory staff against repeated non-compliances by the contractor, such as ordering the rectification to non-compliant works to be completed within a short period of time, suspension of works, or undertaking emergency measures, following consultation with the designer/project office as appropriate. Apparent temporary stability of works that do not comply with the specifications or with the agreed method statements during dry weather can give a false sense of security, as failures are liable to occur without much prior warning, or with insufficient time to take rectification actions, during intense rainstorms. Apart from the routine compliance tests stipulated in the specification for quality control, further compliance tests (e.g. on the finished works) may be included as part of the site audits as appropriate in order to detect, and deter, non-compliances and poor workmanship. For instance, for soil nailing works, these may involve non-destructive tests (e.g. Time Domain Reflectometry tests) on randomly selected soil nails

This paper is published with the permission of the Head of the Geotechnical Engineering Office and the Director of Civil Engineering and Development, Government of the Hong Kong Special Administrative Region. REFERENCES Fugro Scott Wilson Joint Venture 1999. Detailed Study of the Landslides at the Junction of Sai Sha Road and Tai Mong Tsai Road in June 1998. Investigation of Some Selected Landslides in 1998 (Volume 4). Geotechnical Engineering Office, Hong Kong, 159 p. (GEO report No. 111). Halcrow China Limited 2001. Review of the 24 August 2000 Sheet Pile Wall Failure at Lung Ha Wan Road, Clear Water Bay. Geotechnical Engineering Office, Hong Kong, 29 p. Leung, B.N., Leung, S.C. & Franks, C.A.M. 1999. Report on the Rock Slope Failure at Cut Slope 11 NE-D/C7 along Sau Mau Ping Road on 4 December 1997. Geotechnical Engineering Office, Hong Kong, 69 p. (GEO Report No. 94). Sun, H.W. & Tsui, H.M. 2003. Review of Notable Landslide Incidents during slope works. Geotechnical Engineering Office, Hong Kong, 137 p. (GEO Report No. 177).

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Landslide ‘‘Granice’’ in Zagreb (Croatia) Z. Mihalinec Civil Engineering Institute of Croatia, Zagreb, Croatia

Ž. Ortolan Faculty of Civil Engineering, Osijek, Croatia

ABSTRACT: Landslide ‘‘Granice’’in Zagreb is an example of a shallow translational landslide formed on a gentle slope (6–7◦ ). It is extremely elongated in shape, i.e. 300 m in length and 50 m in an average width. Although most of this landslide is on an undeveloped land, its lower part passes a developed area with some ten housing units. First movements registered in this locality date back to 1963/64. Investigations for landslide improvement were undertaken in 1995. Landslide improvement activities were completed in 1998. During ten years after the improvement works no displacements were observed and the life for the people on the landslide area has normalized.

1

INTRODUCTION

The north part of the Croatian capital Zagreb is settled on the lower part of mountain Medvednica, which highest peak Sljeme is 1033 m in height. In this, for living very attractive part of town, a number of landslides were formed. It was the result of urbanization and natural conditions. Some of these landslides have been known for a long time and some of them have activated recently. One of the landslides which has been active for a long period of time is the subject of this paper. 2

GEOTECHNICAL INVESTIGATIONS AND INTERPRETATIONS

Landslide ‘‘Granice’’ was first recognized during the basic geotechnical investigations of a wider area, which were performed in 1987 to define terms for urbanization planning. One borehole with double piezometer was made on the landslide. In contact with local citizens it was found that the first movements in the area took place about years 1963/64. The contour of the landslide was defined and it was established that its length is about 300 m, the average width is about 50 m and the area of landslide is about 1.4 hectares (Fig. 1). It was formed on a gentle slope (6–7◦ ). General direction of movement was north to south. North part of the landslide was on an undeveloped land, while its lower part was on a developed area with some ten housing units. By comparing older topographic plans, from 1962, and the situation in the field, it was estimated that the horizontal movements have already

reached about 3.5 m. The amounts of movements were visible on passages between the houses. The passages and stairs leading from the street Granice down to the houses were originally built along straight lines (Fig. 1). The lower parts of the stairs and passages were obviously moved south. All houses built on the lower south part of landslide suffered damages as the result of landslide activities and were ‘‘traveling’’ together with the landslide. It was concluded that thorough geotechnical investigations and improvement measures are necessary. But a lack of funding delayed further actions. More intensive activity of the landslide was observed after the rainy season in 1989. Finally, in 1995 detailed geotechnical investigations of the landslide with the objective to establish a landslide improvement proposal were conducted. New topographic plan and eleven boreholes, most of them with piezometers, were made to establish a correct geotechnical model. Comparisons of the old topographic plans from 1962 (Fig. 1) and the new plan from 1995 (Fig. 2) were made for estimating horizontal displacements of landslide. It was established that during thirty-three years the displacements of most points that could be compared on the landslide exceeded approximately 5 m. The house which is positioned the most south, on the steeper part of landslide, was moved about 10 m. As a result of geotechnical investigations (field and laboratory), a geotechnical model of the landslide was formed. The RNK-method was used for forming a landslide model of good quality based on relatively small quantity of investigations.

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LANDSLIDE "GRANICE"

E

S

19

0

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18

5

190

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0

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17

5

Figure 1. Landslide ‘‘Granice’’—the contour on the old topographic plan from 1962. The passages and stairs by the east side are in straight lines. 175

0

17

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165

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The RNK-method, or the Reference Level of Correlation Method (Ortolan, 1996, 2000), is a fully developed method for engineering-geological and/or geotechnical modeling. It can be used for both soils and soft rock formations. The RNK is defined as an unequivocally recognizable and visually identifiable (or graphically defined) bedding plane or any other reference plane within a structural feature, in relation to which the altitude of all studied profiles can be unambiguously defined, with individual point analysis of any material property. Such a plane is a part of a single vertical geotechnical correlation column. The geotechnical correlation column is a consistent engineering-geological or geotechnical soil model (design cross section) in which adequate parameters (defined in the laboratory or in situ, either by the point method or continuously), can be reasonably allocated to every defined layer (and portions of such layers) along the entire height of the vertical sequence of formations covered by the study. From such a geotechnical correlation column one may in principle distinguish zones of minimum residual shear resistance, with their thicknesses and continuities, and also layers with different moisture, permeability, natural compaction, compressibility, etc. Geotechnical column of landslide ‘‘Granice’’ is shown in figures 3 and 4. The hanging wall of a fat

LEVEL LINES: 190

SLIP SURFACE 190

MAXIMUM PIEZOMETRIC PRESSURES AT SLIP SURFACE LEVEL

195

0

50 m

100

Figure 2. Landslide ‘‘Granice’’ on topographic plan made in 1995. The passages and stairs beside east side are not in straight lines any more.

clay layer of high plasticity, soft to firm, light-blue in colour, was selected as the reference level of correlation (RNK) for the geotechnical correlation column. In this case the RNK coincided with the slip surface position.

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Figure 3. Geotechnical correlation column of the landslide Granice (Ortolan, 1996) established using the visually recognizable reference level of correlation.

Figure 4. Some geotechnical characteristics of materials in geotechnical correlation column of the landslide ‘‘Granice’’ (Ortolan, 1996).

All boreholes were positioned in the correlation column. The absolute height of the RNK was defined at the position of every borehole. The interpolation of the absolute height of the RNK level between the boreholes was made and lines connecting equal absolute heights of the RNK were constructed (Fig. 2). The depth of the RNK and the geotechnical correlation column enable the extrapolation of geotechnical characteristics of materials established for one position on the landslide on any other position on the landslide. This interpolation also yields the absolute height of the slip plane on the complete area of the landslide. The slip plane is almost parallel with the ground surface in the direction of movements, i.e. north to south. It is 3.5–4.5 m deep along the central profile (Fig. 2). In the perpendicular direction, i.e. west to east, the depth increases from 2 to 7 m. The interpolation of the maximum absolute heights of piezometric pressures at the level of slip surface was also made, and the level lines were constructed (Fig. 2) on the basis of monitoring of installed piezometers.

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Established geometry of the slip plane and the maximum piezometric pressures at the slip surface level, are the basic data for stability and seepage analyses.

Table 1. zone. Borehole

Depth (m)

LL (%)

PL (%)

PI

SF (%)

CF (%)

ϕR (◦ )

3

S-2 S-3 S-8

3.1–3.2 6.8–7.1 6.4–6.7

73.5 73.1 71.7

22.0 23.7 26.6

51.5 49.4 45.1

19 18 15

36 34 36

13.9 13.7 13.5

3.1

STABILITY AND SEEPAGE ANALYSES Stability analyses

After defining the geometry of the slip plane and the maximum piezometric pressures at the slip surface level stability analyses were performed. It was adopted that the strength of the material along the slip surface is represented by its residual strength, because the landslide was active for a long period and movements that already occurred were large (Skempton 1985). The assumptions for the stability analyses were: the strength of the material along the slip surface is equal at all points and when piezometric pressures exceed maximum measured values factor of safety falls to FS = 1. Stability analyses were performed on the central profile, whose position is shown in figure 2. Only the part from the north edge to the break point of the profile was considered as relevant for landslide behavior. After the break point, sliding mass was just pushed by movements in the upper part in the south-west direction as a result of surface configuration. Analyses were performed using Spencer’s method (Spencer 1967). Factor of safety FS ≈ 1 (0.984) was found for the residual friction angle ϕR = 11◦ (with c = 0, γ = 19 kN/m3 ). This result is considered realistic on the basis of the plasticity index established for the material on the slip surface (Fig. 4) and known correlations between the plasticity index and the residual friction angle (Ortolan & Mihalinec 1998). It was concluded that the best improvement measure for the landslide would be to install drain trenches. Further stability analyses were performed to determine how much piezometric pressures should be lowered for assuring an acceptable factor of safety. The installation of the drain trenches was only possible on the north undeveloped part, so in the analyses the lowering of piezometric pressures was assumed in the first 200 m. A satisfactory factor of safety FS = 1.319 was obtained for lowering the piezometric pressures for 2 m. After the improvement design was already completed three more samples from the slip zone were examined in the laboratory (Table 1). The established residual friction angle, with an average value ϕR = 13.7◦ (Table 1), was slightly greater than calculated in previous analyses. It was concluded that the difference appeared because the samples were contaminated with sand which underlies the fat clay layer (Table 1, 15–19% sand particles).

Laboratory results for the samples from the slip

On the other hand, the difference could mean that the piezometric pressures were underestimated (particularly along the east side which was under the influence of unsolved sewerage in the Granice street), but if that was the case the improvement measures were designed on safe side. 3.2

Seepage analyses

Seepage analyses were performed to define the distance between the drain trenches which would provide the necessary lowering of piezometric pressures. The fact that conditions along the whole central profile are very similar allowed that analyses could be performed on a smaller simplified model. The seepage analyses model is shown in figure 5. Seepage analyses were made with computer program which enables 3D analysis by defining seepage conditions in parallel planes (Jovi´c & Radelja, 1981). For this problem four planes were adopted: the first was through the drain trench and the fourth was on half distance between the trenches, while the second and the third were positioned so the distances between successive planes were equal. The results were the pore water pressures in the points of four planes, and the average pressures from which new PL was constructed (Fig. 5). The analyses have shown that the distance between the drain trenches should be s ≤ 6 m for the adopted model, or generally s ≤ 1.8 h (where h is the height from the slip surface to the PL). In that case average piezometric pressures at the slip surface level would be 2 m lower than established maximum piezometric pressures. After these calculations, the positions of drain trenches were adopted as shown in figure 5. A factor of safety was then controlled for the adopted disposition of drain trenches for local conditions on different parts of the landslide. It was done because the depths of the slip surface and the pore water pressures were different from the west to the east side (Fig. 2). Factors of safety for infinite slope models and the influence of drain trenches on different parts of the landslide were calculated using diagrams (Stani´c, 1984). These simple calculations proved that local factors of safety on all

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SEEPAGE ANALYSES

192

Model

GROUND SURFACE

190

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PIEZOMETRIC LINE

188

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M1

186 M3

184

SLIP SURFACE

M2

S

182 180 0

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SEEPAGE ANALYSES

192

Equipotential lines before drainage

190

186

187

188

189

188

184 182 180 5

0

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15

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SEEPAGE ANALYSES

192

Equipotential lines after drainage

190 189

188

188 187

186

186

186 185

184

185 184

182 180 0

5

10

194 192

GROUND SURFACE

190

15

20

25

30

SEEPAGE ANALYSES Piezometric lines before (PL1) and after drainage (PL2)

PL1

188

PL2

186 SLIP SURFACE

184 182 180 0

5

10

15

20

25

30

Coefficients of filtration for materials M1, M2, M3: k1 = 10-8 m/s, k2 = 5x10-10 m/s, k3 = 5x10-9 m/s LANDSLIDE "GRANICE"

Figure 5. Seepage analyses. Model was formed from central profile, between 100 and 130 m from its north edge (Fig. 2).

parts of the landslide would be satisfactory (FS >1.3) and that the improvement in the FS would be 20–30% compared with those before the drainage. 4

IMPROVEMENT MEASURES AND MONITORING

0

50 m

100

Figure 6. Landslide ‘‘Granice’’ — the positions of the drain trenches realized as main improvement measure.

The landslide was improved by the drain trenches, which seemed to be an appropriate solution already at the initial stage of investigation works, because of the landslide geometry. The system of seven parallel drains was installed in the north undeveloped portion of the landslide. The distances between the drain trenches were, from the

west to the east side: 4, 4.5, 5, 6, 8 and 8 m. The lengths of the drains were from 50 m for drain 1 to 182 m for drain 3. The drainage system ended north of the residential structures. Three additional drains were installed to the south of the residential structures. This constructive measure was needed because the material in this part of the

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sliding body was very soft and wet. Indeed it was so soft that it was partly replaced with drainage material on a large part of this surface. According to the landslide improvement design, the problem of sewerage for residential structures situated in the Granice street, which is positioned above the landslide surface along the east side of the landslide, was to be solved simultaneously with the landslide improvement activities. Unfortunately, construction of the sewerage system was prevented by administrative and financial problems. Even so, the improvement was successful which was confirmed by monitoring. Landslide monitoring activities were performed during one year after the landslide improvement. They revealed that the improvement works were successful. This was established by displacement measurements using inclinometers and geodetic benchmarks, which showed that no displacement occurred on the landslide after the improvement works. Piezometric monitoring revealed that the piezometric pressures were significantly reduced with respect to the prior situation on a greater portion of the zone covered by the drainage system. Next to the eastern edge of the landslide, which remained affected by the unsolved sewerage in the Granice street and in the populated portion of the landslide, the piezometric level did not change significantly when compared to the situation prior to the improvement works. Improvement and monitoring activities were used to collect additional geotechnical data on the landslide in order to verify the adopted geotechnical model (Jurak et al. 2004). The data gathered in this way (Fig. 7) confirmed that the geotechnical correlation column based on 1995 data (Fig. 3) was fully reliable.

5

CONCLUSION

Landslide ‘‘Granice’’ in Zagreb was active more than 30 years. The displacements of the objects on the landslide exceeded approximately 5 m and one house positioned the most south, on steeper part of landslide was moved about 10 m. Geotechnical investigations were conducted in 1995 and the result was a landslide improvement project. The RNK method (Ortolan, 1996) was used to define a geotechnical model of the landslide. Contoured map of the slip plane with the sliding surface clearly delimited, contoured map of the maximum piezometric pressures at the slip surface level and the geotechnical column were provided so appropriate stability and seepage analyses could be done and appropriate improvement measures defined. Geotechnical correlation column constructed on the basis of geotechnical investigations was verified by

Figure 7. Verification of the geotechnical correlation column presented in Figures 3 and 4, based on plasticity indexes from 1998 (samples from boreholes drilled for the installation of monitoring equipment and from drain 6).

the data collected during improvement and monitoring activities (Jurak et al. 2004). The landslide was improved in 1998 by installing the drain trenches. Landslide monitoring activities revealed that the improvement works were successful.

REFERENCES Joviæ, V. & Radelja, T. 1981. Manual for computer program ELIPTI, for solving filtration problems in saturated media (in Croatian). Zagreb: Civil Eng. Institute of Croatia. Jurak, V., Ortolan, Ž., Slovenec D. & Mihalinec, Z. 2004. Verification of Engineering-Geological/Geotechnical Correlation Column and Reference Level of Correlation (RNK) Method by Observations in the Slip-Plane Zone. Geologia Croatica 57/2, 191–203, Zagreb: Croatian Geological Survey and Croatian Geological Society. Ortolan, Ž. 1996. The creation of a spatial geologic engineering model of deep multilayered landslide (on an example of the Podsused landslide in Zagreb). Ph. D. Thesis (in Croatian). Zagreb: Faculty of Mining, Geology and Petroleum Engineering, University of Zagreb, 245 p.

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Ortolan, Ž. 2000. A Novel Approach to the Modeling of Deep Complex Landslides with Several Sliding Planes. Landslides in Research, Theory and Practice: Proc. 8th Int. Symp. on Landslides, Cardiff 26–30 June 2000, Vol.3, 1153–1158, London: Thomas Telford. Ortolan, Ž. & Mihalinec, Z. 1998. Plasticity index—Indicator of shear strength and a major axis of geotechnical modeling. Geotechnical hazards: Proc. of the XI-th Danube-European conference on soil mechanics and geotechnical engineering, Porecˇc (Croatia), 25–29 May 1998, 743–750, Rotterdam: Balkema.

Skempton, A.W. 1985. Residual strength of clays in landslides, folded strata and the laboratory. Geotechnique, Vol.35, No.1, 3–18. Spencer, E. 1967. A method of analysis of the stability of embankments assuming parallel inter-slice forces. Geotechnique, Vol.17, No.1, 11–26. Staniæ, B. 1984. Drainage cuts and slope stability (in Croatian). Graðevinar, Vol.36, No.3, 105–112.

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Improvement of subsurface drainage provisions for recompacted soil fill slopes in Hong Kong K.K. Pang & J.M. Shen Fugro Scott Wilson Joint Venture, Hong Kong SAR, China

K.K.S. Ho & T.M.F. Lau Geotechnical Engineering Office, Civil Engineering and Development Department, Hong Kong SAR, China

ABSTRACT: Distress has been observed on some recompacted soil fill slopes in Hong Kong. A detailed investigation of the distress at the recompacted fill slope behind the Hong Kong Sanatorium and Hospital revealed that the distress might have been caused by adverse transient groundwater conditions. This paper presents the results of a review of the sub-surface drainage provisions for recompacted fill slopes in Hong Kong and proposes means to enhance their robustness and effectiveness. Plausible measures that may enhance the conventional sub-surface drainage detailing have been identified and seepage analyses carried out to examine the effectiveness of the proposed measures. One effective means is to provide prescriptive filter pipes at the upstream end of the sub-horizontal drainage blankets to help rapid dissipation of any build-up of groundwater pressure behind the recompacted fill cap.

1

INTRODUCTION

Soil fill slopes formed by end-tipping without proper compaction were common in Hong Kong prior to the establishment of the Geotechnical Engineering Office (GEO) in 1977. The potential instability problem of loose fill slopes was highlighted when two large soil fill slopes failed catastrophically at Sau Mau Ping in 1972 and 1976 respectively, causing multiple fatalities. Following the recommendations of the Independent Review Panel for Fill Slopes after the investigation of the 1976 incident (Government of Hong Kong, 1977), upgrading of loose fill slopes in Hong Kong is generally done via the recompaction of the top 3 m of the fill material to at least 95% maximum dry density. Inclined and sub-horizontal drainage blankets are provided at the base of the recompacted fill cap for each of the batters of the recompacted fill slope. The function of the drainage blankets is to quickly dissipate any build-up of groundwater pressure behind the recompacted fill cap. Law et al. (1998) carried out a review of 128 fill slopes upgraded by recompaction in Hong Kong and concluded that there was a general improvement in the performance of fill slopes after recompaction of the top 3 m, in which major distress occurred on only one recompacted fill slope, which was caused by leakage from a water main. Law et al. (1998) noted that

in some cases, cracks continued to develop especially at the transition zone between the uncompacted fill and the 3 m thick recompacted fill layer. They also noted that there was no improvement in tackling or treating the problems of blocked drainage and groundwater seepage (including leakage from water-carrying services), which was suspected as being "an indication of the poor performance of the drainage layer behind the recompacted fill". Bolton et al. (2003) carried out a number of centrifuge tests to examine the behaviour of loose granular fill slopes, formed of materials derived from completely decomposed granite (CDG), subject to infiltration conditions. Based on a number of centrifuge tests, they observed that model slopes formed using uniform CDG materials in a deep profile over bedrock, irrespective of whether they are in a loose or dense state, did not fail. This was attributed to the CDG being too permeable to achieve saturation and develop significant water pressure and turn into a flowslide as water can drain out from the soil pores. They also noted that it was necessary to constrain the groundwater flow, either by raising the bedrock, or by inserting a ‘tongue of permeable material’ (Figure 1) in order to promote a mobile failure caused by hydraulic blowout due to the elevated groundwater pressure. For model slopes built of fine CDG over coarse CDG (which acts as the tongue of permeable material), densification of

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Figure 1. Hydrological consideration of a layered slope analyzed by Bolton et al. (2003) (reproduced from Figure 11 of Bolton et al. (2003)).

the fine CDG (to a relative compaction of 91%) was found to be not of much benefit in preventing a failure according to centrifuge tests. Lee & Bolton (2006) described two further centrifuge tests and noted that a deep-seated failure with significant distortion was induced when a layered fill slope was subjected to seepage flow from a more permeable layer. The results indicate that "shear failure can be developed when transient pore water pressure is allowed to build up in a blind layer underneath a fill slope, even though the top fill layer is compacted to a high degree of compaction". It is noted that the heterogeneity of the in-situ condition of loose fill slopes may not have been adequately modelled in the above centrifuge tests and hence observations made from the laboratory tests should be treated with caution. Notwithstanding this, the tests have re-affirmed the critical importance of ensuring sufficient sub-surface drainage provisions at the base of the compacted fill layer.

2

DISTRESS RECENTLY NOTED ON A RECOMPACTED FILL SLOPE

The fill slope (registered as Slope No. 11SW-D/FR1 in the Government’s Slope Catalogue) is situated above the Hong Kong Sanatorium and Hospital at Happy Valley. The slope has an overall height of about 65 m and comprises eight batters inclined at around 32◦ , separated by 1 m to 2 m wide berms with a 7 m high retaining wall at its toe (Figure 2). The slope was upgraded between 1977 and 1979 with at least 3 m soil fill compacted to a minimum 95% relative compaction. Major distress of the slope was first observed in March 2004 by the slope maintenance department, which was subsequently mapped in detail by FSWJV (2005). The distress included major cracking on berms

Figure 2. Typical cross-section of slope No. 11SW-D/FR1 at Hong Kong Sanatorium and Hospital.

and the lower two batters of the slope, displaced shotcrete cover beyond the edge of the berm drainage channels, crushed concrete drainage channels and blocked drainage outlet pipes. The detailed study of the incident (FSWJV, 2005) concluded that the distress was mainly related to water ingress into the old loose fill body underlying the compacted fill cap within the upper half of the slope, and the build-up of groundwater pressure in the compacted fill within the lower half of the slope. Piezometers installed within the compacted fill in the lower half of the slope indicated a piezometric head between 1 m and 2 m below the ground surface. Some piezometers exhibited storm responses of 0.5 m rise with a time lag of 3 to 6 hours. Tensiometers installed within the compacted fill in the lower slope also indicated positive groundwater pressure during rainstorms. The main source of water ingress was probably from the catchment in the upslope areas (e.g. leakage from stormwater drains within private lots cannot be precluded), although direct surface infiltration through the vegetated areas of the slope and leakage from a sewer beneath the road at the crest of slope might also have played a contributory role. The setting of the site (which is within a buried valley) could have led to concentrated sub-surface groundwater flow along preferential paths, resulting in the slope distress. The ineffectiveness of the sub-surface drainage provision probably played a key role in the observed slope distress.

3

SUB-SURFACE DRAINAGE PROVISIONS FOR FILL SLOPES IN HONG KONG

3.1 Previous practice in sub-surface drainage provisions The fill slope above the Hong Kong Sanatorium and Hospital was one of the early slopes upgraded by the

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prescriptive recompaction treatment in Hong Kong. The sub-surface drainage provisions to this slope are representative of the prevailing practice in the late 1970’s. The sub-surface drainage for slope No. 11SWD/FR1 comprises drainage blankets at the base of each batter of the compacted fill slope (Figure 2). Only ‘Filter A’ (i.e. fine granular filter) of 750 mm thick was provided to the upper batters, whereas a sandwich-type ‘Filter A/B/A’ (i.e. fine/coarse/fine granular filter), with a total thickness of 1, 000 mm (Filter ‘B’ of 500 mm thick), was provided to the lower two batters. A reinforced concrete cover was provided on the slope surface where the drainage blankets daylight. Discharge from drainage blankets was through a row of drainage outlet pipes of 75 mm diameter and 750 mm long, at 1.5 m horizontal spacing.

practice of 75 mm. The intercepting length has been marginally increased to 800 mm, as compared to the earlier practice of 750 mm.

4

SUB-SURFACE DRAINAGE PROVISIONS FOR RECOMPACTED FILL SLOPES IN OTHER COUNTRIES

Figure 3 gives the details of the current practice in respect of sub-surface drainage provisions to soil fill slopes upgraded by recompaction. A notable difference with the practice adopted in the late 1970’s is that the current practice requires a sandwich-type fine/coarse/fine granular filter to be provided to each of the compacted slope batters. The thickness of the sandwich-type filter is reduced to 900 mm (coarse granular filter of 300 mm thick), as compared with the earlier practice of 1000 mm or 1500 mm. The drainage blankets do not daylight but stop short at about 500 mm from the slope surface. Discharge from the drainage blanket is through a row of drainage outlet pipes of 50 mm diameter at 1.5 m horizontal spacing with the perforated section penetrating to a length of 800 mm into the coarse granular filter. It is noteworthy that the diameter of the outlet pipe has been reduced to 50 mm, as compared with the earlier

There is limited information in the literature on the detailing of sub-surface drainage provisions for recompacted fill slopes formed against a sloping ground, a site setting that is common in Hong Kong. In the Seattle Landslide Study Project (Seattle Department of Planning and Development, 2000), some typical details of the drainage blanket for the construction of ‘compacted earth buttress fill’ against sloping ground are proposed (Figure 4). A drainage blanket of minimum 18 inches (approximately 450 mm) thick is proposed at the base of the ‘compacted earth buttress fill’, which does not daylight at the slope surface. Discharge from drainage blanket is from ‘subdrains’, comprising a perforated drain pipe surrounded by drainage materials in a trough. The main ‘subdrain’ is placed at the toe of the slope and the intermediate ‘subdrains’ are provided at higher levels. In 2002, the Department of Transportation of the State of New York issued Standard Drawing No. M203–3R1, which depicts the details of granular fill slope protection installation (Figure 5). Where the granular fill is to be constructed against a sloping ground, no drainage blanket provisions would be required at the base of the granular fill. Instead, at the locations of ‘seepage planes’, ‘pipe drains’ (surrounded with filter) are to be provided to convey sub-surface water issuing from the slope. These ‘pipe drains’ are to be designed to ensure that the intercepted water is carried to a drainage system.

Figure 3. Current practice on sub-surface drainage provisions for fill slopes in Hong Kong (extracted from CEDD Standard Drawing C2302 (Rev F).

Figure 4. Typical drainage blanket details (reproduced from Seattle Department of Planning and Development (2000)).

3.2

Current practice in sub-surface drainage provisions

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Figure 5. Granular fill slope protection details of Department of Transportation, State of New York (reproduced from Standard Drawing M203–3R1).

credible option in increasing the discharge capacity of the drainage blankets. The purpose of providing laterally persistent filter pipes across the slope at the up-stream end of the sub-horizontal drainage blankets is to intercept the transient groundwater earlier, thus releasing any buildup of hydraulic pressure sooner. An added advantage of filter pipes is that they would avoid over-reliance on the effectiveness of the drainage blankets and the outlet pipes for discharging any transient groundwater, which could be prone to blockages (Law & Thorn, 2001). These filter pipes need to be fabricated from relatively flexible material in order to cater for any potential small differential settlement that might occur in the long run arising from the variable loose fill below. 6

5

SCOPE FOR ENHANCING ROBUSTNESS AND EFFECTIVENESS OF SUB-SURFACE DRAINAGE PROVISONS

In the present review, four plausible measures have been identified which may enhance the robustness and effectiveness of conventional sub-surface drainage provisions for loose fill slopes subjected to recompaction treatment. These include (a) lengthening the slotted section of outlet pipes; (b) increasing the diameter of outlet pipes; (c) increasing the permeability and/or the thickness of the filter materials; and (d) providing sub-soil pipes (viz. filter pipes) across the slope at the up-stream end of the sub-horizontal drainage blanket. Lengthening the slotted section of the outlet pipes will permit transient groundwater to be tapped earlier to permit release of hydraulic pressure. It also has the added advantage of increasing the intercepting capacity of the outlet pipes as well as reducing the likelihood of pipe clogging, although the latter is difficult to quantify. Increasing the diameter of the outlet pipes would increase the intercepting and discharge capacity of the outlet pipes. This would be an effective measure if the outlet pipe is the ‘ bottle-neck’ in the drainage system but would have little effect if the ‘bottle-neck’ is in the drainage blankets. Increasing the permeability of the filter materials would increase the discharge capacity of the drainage blankets. However, since filters are designed to satisfy the requirements of stability, permeability and segregation (GEO, 1993), there appears to be little room in readily increasing the permeability of the filter materials unless a very permeable material (such as rockfill) is entrapped within ‘Filter B’. However, such provision can be complicated and costly. On the other hand, increasing the thickness of the filter materials (in particular the coarse filter) could be a more

THEORETICAL ANALYSES

The effects of the four plausible enhancement measures identified above have been examined by reference to theoretical analyses, including hand calculations as well as finite element seepage analyses using the computer program SEEP/W. The seepage analyses carried out were not aimed at giving a definitive assessment of what might happen on site. On the contrary, the seepage analyses were intended to provide a sensitivity analysis of the likely differing behaviour of the sub-surface drainage provisions upon implementation of the various enhancement measures, i.e. to examine the potential changes in the behaviour of the drainage provisions. For this review, slope No. 11SW-D/FR1 has been modelled in the seepage analyses, since major distress has been observed at this site and comparatively more information, in terms of material properties and groundwater conditions, is available. The specific site setting comprises a fill slope formed over a well-defined old drainage line with concentrated subsurface groundwater flow and the compacted fill at the lower half of the slope was probably underlain by a tongue of fairly permeable loose fill. The results of the seepage analyses prior to the provision of enhancement measures indicate that drainage blanket with the provision of ‘Filter A’ alone would not be effective in discharging the build-up of transient groundwater pressure. The drainage blanket with the provision of ‘Filter A/B/A’ proves to be more effective, but it would still take a fairly long time to release any build-up of water pressure. The potential ‘bottleneck’ in terms of discharge of transient groundwater discharge appears to be with the drainage blankets. 6.1 Lengthening the slotted section of outlet pipes The results of seepage analyses indicate that the benefit in the release of transient groundwater is not very

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obvious when the length of the slotted outlet pipes is increased from 0.75 m to 3 m. It still takes a fairly long time for the transient groundwater to reach this extended portion of the outlet pipes, prior to releasing the hydraulic pressure. With an increased length of 12 m for the slotted outlet pipes, the benefit becomes more obvious. The results of simplified theoretical hydraulic analyses also indicate a similar conclusion that there is only a marginal (about 10%) increase in the discharge capacity of the outlet pipe when it is extended by 3 m into the drainage blanket. 6.2

Figure 6a. Transient groundwater pressures before provision of filter pipes.

Increasing the diameter of outlet pipes

The results of theoretical hydraulic analyses indicate that there is little merit in increasing the diameter of the outlet pipes since their current capacity (with a diameter of 50 mm) is generally sufficient to discharge the transient groundwater collected in the drainage blankets which comprise a fine/coarse/fine granular filter. The slow response in releasing transient groundwater pressure was primarily a result of the insufficient transmissivity (i.e. permeability x thickness) of the drainage blankets. Increasing the diameter of the outlet pipes alone will not help in relieving the transient groundwater pressure. However, in the case of discharging groundwater collected by the filter pipes (see Section 6.4), increasing the diameter of the outlet pipes (such as from 50 mm to 75 mm) will have a direct benefit in the release of transient groundwater flow. 6.3

Increasing the thickness of ‘Filter B’

The results of theoretical hydraulic analyses indicate that there is little merit in increasing the thickness of the ‘Filter B’ layer from 0.5 m to 2 m. The 2 m thick ‘Filter B’ layer is still insufficient in quickly dissipating the hydraulic pressure built up at the far end of the ‘tongue of permeable material. 6.4

Providing filter pipes at the up-stream end of sub-horizontal drainage blanket

The provision of lateral filter pipes in the drainage blankets has been promulgated in some other countries as presented previously. However, this practice has not been commonly adopted in Hong Kong for upgrading loose fill slopes by recompaction. In this review, a 75 mm diameter flexible slotted plastic pipe was assumed to have been placed laterally across the slope at the intersection of the inclined and sub-horizontal drainage blankets. To model the effect of this filter pipe, a 75 mm diameter hole was assumed to be open at that location, which permitted water to enter from its circumference and be drained away immediately.

Figure 6b. Transient groundwater pressures after provision of filter pipes.

Figure 6 presents a comparison of the transient groundwater condition in the slope before and after provision of filter pipes. As can be seen, the highest hydraulic pressure was 20 kPa at the far end of the tongue of permeable loose fill after 36 hours since the start of heavy rainfall when a 75 mm diameter filter pipe was provided within the drainage blanket, whereas a hydraulic pressure of 80 kPa would build up at the same location after 36 hours when no filter pipe was provided. The results indicate the effectiveness of the filter pipes in enhancing the robustness of subsurface drainage provisions for fill slopes upgraded by recompaction. The results of simplified theoretical hydraulic analyses indicate an improvement of about 60% in the discharge capacity of the system when a filter pipe is provided at the up-stream end of the sub-horizontal drainage blanket.

7

DISCUSSIONS

The performance review by Law et al. (1998) indicates a marked improvement in the performance of loose fill

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slopes upgraded by recompaction in that no liquefaction failure has occurred on these slopes. Since then slope distress and signs of inadequate drainage provisions have been observed in certain site settings and slope detailing. The present review has examined various means of further enhancing the effectiveness of sub-surface drainage provisions which may be considered for the more vulnerable site settings, e.g. a deep layer of loose fill over a natural drainage line with a fairly large catchment in the upslope area. The results of theoretical analyses suggest that a drainage blanket with only a fine granular filter may be limited in its effectiveness in discharging significant transient groundwater flow. The current practice in Hong Kong requires the installation of a drainage blanket of sandwich-type comprising fine/coarse/fine granular filter to each of the slope batters, which is an improvement in this regard. Theoretical analyses suggest that the drainage blanket itself, even with a fine/coarse/fine granular filter, might be a ‘bottle-neck’ in the sub-surface recharge provisions during heavy rainfall and rapid sub-surface recharge from the catchment. At times of significant inflow of transient groundwater, the prescriptive drainage blanket alone may not be sufficient in quickly dissipating the build-up of water pressure within it. The diameter of the outlet pipes plays a relatively minor role in improving the effectiveness of the subsurface drainage provisions if the outlet pipes are short. The provision of very long outlet pipes (of the order of 12 m) is more effective in intercepting the transient groundwater table at the up-stream end of the subhorizontal drainage blankets. The installation of such drains is relatively cheap and simple. An even more effective way to enhance the effectiveness of the sub-surface drainage system is to provide continuous flexible filter pipes at the upstream end of the sub-horizontal drainage blankets across the slope. The installation of such filter pipes is relatively simple and cheap for slopes to be upgraded by recompaction. Any water collected in these filter pipes can be discharged to the berm U-channels. A possible arrangement is presented in Figure 7. It should be noted that the fill around the filter pipes could settle unevenly and compaction of fill above the pipes may affect their vertical/horizontal alignment. In the event of highly distorted pipes, they are liable to locally form a concentrated source of water infiltrating into the loose fill behind/below the pipes. In view of this potential concern, the suggested arrangement will need to incorporate a steeper fall for the drainage blankets as well as the filter pipes than that adopted in the current practice. Should there be an undue risk of major uneven settlement occurring around the pipes, designers should carefully review whether the provision of prescriptive filter pipes is warranted.

Figure 7a. Suggested improvement to details of sub-surface drainage provisions for soil fill slope recompaction.

Figure 7b. Suggested improvement to details of sub-surface drainage provisions for soil fill slope recompaction.

It should also be noted that the sub-horizontal drainage blankets, as well as the filter pipes, might sometimes be designed to fall laterally across a slope in order to suit the actual site conditions. Designers should exercise discretion in considering whether there is a need to provide outlet pipes with a larger diameter or outlet pipes at a closer spacing at the location of the lowest point with a view to facilitating the release of any potential concentration of groundwater flow. ACKNOWLEDGEMENTS The paper is published with the permission of the Head of the Geotechnical Engineering Office and the Director of Civil Engineering and Development, Government of the Hong Kong SAR. Contributions by colleagues, especially Ir S M Tam and Mr. W H Lu, are gratefully acknowledged. REFERENCES Bolton, M.D., Take, W.A., Wong, P.C.P. & Yeung, F.J. 2003. Mechanisms of failure in fill slopes after intense rainfall.

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Int. Conf. on Slope Engineering, Hong Kong vol. 1: pp 1–25. Fugro Scott Wilson Joint Venture 2005. Study of Distress on Slope No. 11SW-D/FR1 below Stubbs Road Happy Valley (Landslide Study Report LSR 3/2005). Geotechnical Engineering Office, Civil Engineering and Development Department, Hong Kong SAR Government. Geotechnical Engineering Office 1993. Review of Granular and Geotextile Filters (GEO Publication No. 1/93). Geotechnical Engineering Office, Civil Engineering Department, Hong Kong Government. Government of Hong Kong. 1997. Report of the Independent Review Panel on Fill Slopes, Government of Hong Kong. Law, K.T., Lee, C.F., Luan, M.T., Chen, H. & Ma, X. 1998. Appraisal of Performance of Recompacted Loose

Fill Slopes (GEO Report No. 58). Geotechnical Engineering Office, Civil Engineering Department, Hong Kong SAR Government. Law, K.T. & Thorn, M.R. 2001. The acceptability and longterm performance of general fill in slope construction. Geotechnical Engineering: Meeting Society’s Needs: Proceedings of the Fourteenth Southeast Asian Geotechnical Conference, Hong Kong. Lee, Y.S. & Bolton, M.D. 2006. Centrifugal modelling of the landslides triggering mechanism in layered fill slopes. Proceedings of the Sixth International Conference on Physical Modelling in Geotechnics, Hong Kong. Seattle Department of Planning and Development 2000. Seattle Landslide Study. Department of Planning and Development, City of Seattle.

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Biotechnical slope stabilization and using Spyder Hoe to control steep slope failure P. Raymond Terra Erosion Control Ltd., Nelson, BC, Canada

ABSTRACT: A fill slope failure on a public road initiated a landslide in February of 2002. The British Columbia Ministry of Highways and Transportation contracted Terra Erosion Control Ltd. to design and implement soil bioengineering/biotechnical slope stabilization techniques in order to reduce surface erosion and increase slope stabilization. The landslide was approximately 18 m (59′ ) in width and 70 m (230′ ) in length, and was composed of mostly sand, with a small component of silt, gravel, and cobbles to small boulders. The slope gradient of the landslide ranges from 35 to 40 degrees. Seepage was noticed throughout the site, as well as a tension crack below the recent failure head scarp during a field visit on March 12, 2003. A soil bioengineering prescription was developed and the site was then treated in May of 2003. A combination of the following techniques and treatments were used on the site; brush layers, live pole drains, drain fascines, live staking, vegetated lift, straw wattles, erosion matting, the planting of native seedlings (Alnus viridis ssp. sinuata) and broadcast seeding. The following species of cuttings were used; Salix sp. (scoulieriana, bebbiana) and Populus balsamifera ssp.trichocarpa. The cuttings were harvested dormant, placed in cold storage and then soaked. A Spyder Hoe was utilized to install most of the structures and to re-grade the contour of the slope. On this site the use of a Spyder Hoe demonstrates how cuttings can be planted at a greater depth than using manual excavation, resulting in deeper rooting and higher survival rate. The Spyder Hoe also moved large obstacles such as logs and boulders. These soil bioengineering techniques help stabilize the surface layer of the site by addressing surface erosion. The live pole drain system directs surface and seepage water to the toe of the slope. By establishing a woody shrub component, in conjunction with grasses and legumes growing on the site, rilling and gullying is reduced. Live stakes placed near the terminus of the landslide anchor the soil by providing an inter-twined root mass. The seedlings and legumes provide a deep-rooted nitrogen fixing species. The site was monitored during the summer and fall of 2003 and again in the spring and summer of 2004 and fall of 2007. The survival rate is currently >85% with an average growing height of up to 2.4 meters. Good growth rates were attributed to a long soaking period and organic soil amendment including local mycorrhiza fungus inoculation. 1

INTRODUCTION

This paper describes the soil bioengineering work undertaken to stabilize the Walker’s Landing site. The treatments were intended to slow and eventually stop the debris slide, which took place twice within a one year period below the road. As well, the treatments were intended to establish vegetation to reduce surface erosion and address seepage flow from the site. One other attempt to re-habilitate the site was carried out before the soil bioengineering treatment described in this paper. The treatment consisted of filling and compacting the failure area with gravel pit material followed by a hydro seeding application. 1.1

(see location map below, Figure 1). The initial slide event occurred in February 2002. Rainfall events resulted in erosion at the face of the Powerline Gravel Pit above the road from the site and the creation of a deep vertical headwall scarp on the downstream end of

Site description and history

The project area is located on the east shore of Kootenay Lake in southeastern British Columbia, Canada

Figure 1.

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Site location map.

a culvert that crosses Walker’s Landing Road onto the site (Source MOTH Creston). The site failure was due to excess and concentrated drainage water increasing pore water pressure to the point of failure on an over steepened slope comprised of deep sandy lacustrine soils. The culvert was removed and the area was filled in with gravel, cobble, and small boulder material. The fill materials were compacted from the bottom up with an excavator using approximately 3000 m3 (2294 yard3 ) of material. The site was then hydro seeded using a road side erosion control mix. A second failure of approximately 90 m3 (70 yard3 ) of material occurred in December 2003. The site has an area of approximately 0.13 ha (0.03 acre), with an average of 18 m (59′ ) in width by 70 m (230′ ) slope distance. It is composed of sand, coarse gravel, and cobbles to small boulders with a small component of silt. Slope angle ranges from 1.25:1 to 1.4:1 (∼35 to 40 degrees). The elevation ranges from approximately 581 m (1906′ ) at the top to 536 m (1758′ ) at the bottom with a north/west aspect. Seepage was noticed throughout the site, as well as a tension crack below the recent failure head scarp, during a field visit carried out on March 12, 2003 by the author. 1.2

Climate

Total annual precipitation is in the order of 886 mm/year (35 inches), most of which falls as snow. Snowmelt takes place mostly in April and in some seasons is accelerated by warm spring rains. Summer drought periods of up to a month occur in July and August with temperatures to 35◦ Celsius (95 Fahrenheit, Canadian Climatic Normals 1971–2000, Environment Canada weather station for Kaslo, B.C.) Kaslo is located at the same elevation and to the north of the project area. 2

REMEDIATION MEASURES

The prescription for remediation measures was completed by P. Raymond of Terra Erosion Control Ltd. and W.H. Wells, P.Ag. of William H. Wells Consulting based on observations in March of 2003. The following remediation measures were applied to the site: The brush layer type structures (see Figure 2) will enhance the survival of the living materials by increasing the root to shoot ratio, providing moisture to the cuttings and deeper rooting development. The layout of the brush layers will allow the seepage water to percolate toward the middle of the site into the live pole drain system. The rows of brush layers were combined with drain fascines and straw wattles, to help in draining surface water and act as sediment traps. The use of drain fascines in combination with the central live pole drain system (see Figure 3) will help control the excess surface water and seepage on the

Figure 2. 1996).

Brush layer installation (from Schiechtl & Stern.

site. As these structures when constructed are partially buried they will be resistant to the summer drought. These structures will also act as sediment traps to catch flowing sediment as the woody vegetation sprouts from the fascines (see Figure 4). In the area directly below the road vegetated lifts were installed using cuttings of 1.5 m (4′ 11′′ ) and 2.0 m (6′ 7′′ ) in length to provide deep rooting. In the area directly above the debris pile at the bottom of the slope, vegetated lifts were also installed to help in providing toe buttress (see Figure 5). Live stakes of 1.0 m (3′ 3′′ ) to 1.5 m (4′ 11′′ ) in length (600) were planted in the sandy debris in the slide run out zone. The roots from the live stakes will intergrow and help stabilize the accumulated materials (see Figure 6). Broadcast seeding of grasses and legumes was carried out to reduce surface erosion. Container grown Sitka Mountain alder (Alnus viridis ssp. sinuata) seedlings were planted in between structures to speed up initial growth of a stabilizing root mass and to provide a source of nitrogen fixation. 2.1

Machine work

A Spyder Hoe was utilized to assist in the installation of the various structure types such as brush layers and vegetated lifts. The Spyder Hoe was also used to re-contour the slope where needed and to place stumps and logs below the debris pile at the toe of the slope (see Figure 5). The Spyder Hoe worked the slope from the bottom up, efficiently digging the trenches for each structure while a crew manually installed the cuttings.

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Figure 5.

Prescribed treatment (Author 2003).

Figure 3. Live pole drain/fascines (from Donald H. Gray, Robbin B. Sotir 1996).

Figure 4. Brush layer/fascine (from H.M. Schiechtl & R. Stern. 1997).

Figure 6. Live staking (from Donald H. Gray, Robbin B. Sotir. 1996).

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location within the same biogeoclimatic zone. A mix of approximately 80% cottonwood and 20% willow was used. Portions of the cuttings were kept in cold storage prior to installation and portions of the cuttings were harvested just before installation. All cuttings were soaked in Kootenay Lake prior to installation for a period of 10 to 15 days. This enhanced the growth hormones within the cuttings. The cuttings were enclosed within a metal fence supported by rebar to protect them from beaver damage. The brush layer cuttings and the cuttings used for live stakes were treated as follows: The exposed portion of the cuttings was painted with a mix of 50% latex paint and 50% water. This mix helps prevent desiccation and entry of disease. Figure 7. Spyder Hoe.

The versatility of this machine to operate on such steep terrain allowed for cuttings to be planted at a greater depth than manual labour while increasing production and resulting in an overall better end product. A regular 200 series excavator was also used to install vegetated lift directly below the road and to place a concrete barrier next to the road.

4.2

Growing medium

In order to enhance the moisture retaining capacity during the summer drought and provide an adequate growing medium, a layer of approximately 5 cm (2′′ ) of Sunshine mix #4 added to organic fertilizer 4-4-4 and locally gathered mycorrhiza fungus was mixed, applied and soaked within the installed structures and during planting of native seedlings.

2.2 Spyder Hoe description The Spyder Hoe has four articulated legs, two of which have claws and two that have wide rubber tires. The Spyder Hoe also has a telescopic boom that is used to push the machine up slope using back wheels. It has hydraulic commands and each leg is operated independently. Some Spyder Hoe’s have an electric winch to tether themselves. Conventionally, the machine would push itself up the slope by raising the front claws, and pushing with the telescopic boom until it reaches the full extent of the boom. The Spyder Hoe can then stabilize itself by placing the front claw on the slope and leveling the cab. It then carries out the slope grading and structure placement through excavating (see Figure 7 below, May 2003).

3

4.3

A crew of five people installed the structures listed above; broadcast seeded and planted native seedlings over nine working days between May and June, 2003. In most locations the ground was moist at the time of installation. 4.4

Broadcast seeding

The seeding prescription was developed by Polster Environmental Services Ltd. of Duncan, B.C. Seeding was carried out at 25 kg/ha (23 lbs/acre). Bluebunch wheatgrass—Pseudoroegneria spicata (Pursh) A. Löve ssp. spicata Durar Hard Fescue—Festuca trachyphylla (Hack.) Krajina Sodar Streambank Wheatgrass—Agropyron riparium Scribn. & J. G. Sm. Primar Slender Wheatgrass—Elymus trachycaulus (Link) Gould ex Shinners ClimaxTimothy—Phleum pratense L. Rambler Alfalfa—Medicago sativa L. Aurora Alsike Clover—Trifolium hybridum L.

SITE PREPARATION

All wildlife/dangerous trees were assessed by a certified assessor and felled as per Worker’s Compensation Board (safety) regulations. The site was rock scaled to remove all loose rock and debris and to ensure safe work sites.

4

Installation of structures

IMPLEMENTATION AND TREATMENT 4.5

4.1 Collection and preparation of materials Willow spp. (Salix scouleriana and bebbiana) and black cottonwood (Populus balsamifera ssp.trichocarpa) were gathered in the vicinity of the project

Planting of native seedlings

A total of 200 Sitka mountain alder (Alnus viridis ssp. sinuata) PSB 412A 1 + 0 were planted the on site. The Sitka mountain alder added a deep-rooted and nitrogen fixing component to the sites.

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5

SUMMARY OF TREATMENT TYPE Work description/Linear meter/Structure Type. Scaling of loose rocks and debris. On-site layout of structures. Brush Layer 1 m (3′ 3′′ ) 113 m (371′ ). Brush Layer 1 m (3′ 3′′ )/Fascines 20 cm (8′′ ) diameter 107 m (351′ ). Live Pole Drain 30 cm (12′′ ) diameter 77 m (253′ ). Live Stakes 1.0 m (3′ 3′′ ) to 1.5 m (4′ 11′′ ) at bottom. Vegetated lift using coconut matting and 2 m (6′ 6′′ ) cuttings. 28 m (92′ ) Vegetated lift using coconut matting and 1.5 m (4′ 11′′ ) cuttings. 149 m (489′ ) 745. Seedlings planted (Sitka Mountain alder) 200. Broadcast seeding. 25 Kg/ha (23 lbs/acre). Machine work (Spyder Hoe and 200 series excavator) including low bed. Straw wattle 22 cm (9′′ ) dia. 140 m (460′ ). Growing medium.

6

PROJECT COST

Total cost for the initial prescription, project implementation, broadcast seeding, monitoring, and maintenance in 2003 was $40,000. 7

MONITORING, MAINTENANCE AND REPAIR

June 27, 2003—Early growth of soil bioengineering structures as well as the Sitka Mountain alders, grasses and legumes was satisfactory. About 90% of the structures had 10 to 20 cm (4′′ to 8′′ ) of growth and root development was noted. The live pole drain was functioning. Light watering of the site was carried out from the upper road. October 25, 2003—Average growth of cuttings was estimated at 60 to 150 cm. (24′′ to 59′′ ). There was a good take of the grasses and legumes. Weather conditions were very dry for most of the summer of 2003 which caused lots of wildfires in British Columbia. This resulted in about 10% mortality of the cuttings and some dieback of growth on the brush layers. A minor tension crack was noticed on the road edge most likely caused by earth settling. Minor fungus was noticed on leafs of the black cottonwood cuttings. April 30, 2004—The growth from 2003 and the new growth was noted as up to 150 cm (59′′ ) combined with the survival remaining over 90%. The live pole drain was functioning very well and controlled runoff water. Minor browsing was noticed on some of the structures. Small failures of brush layer/fascine structures approximately 2 meters (6.6′ ) in length occurred, from a large animal walking above it.

July 16, 2004—Desiccation of approximately 30% upper brush layer row and live stakes located on the north portion of lower debris pile was noted. Average growth on brush layers ranged from 0.6 meters to 2.4 meters (24′′ to 95′′ ) on live pole drains from 0.9 meters to 2.0 meters (35′′ to 79′′ ), live stakes from 0.4 meters to 1.4 meters (16′′ to 55′′ ) and the Sitka Mountain alder growth averaged 0.8 meters (32′′ ). The grasses and legumes were getting established and were growing through the erosion control matting. Seepage was noticed approximately half way down the site. Minor repair of the spring failure was carried out. September 9, 2007—Weather conditions were very dry for most of the summer of 2007 (highest temperatures ever recorded for July >40 C), which caused several wildfires in British Columbia, resulting in about 75% mortality of the live stakes planted in the sandy debris in the slide run out zone. The remainder of the site was not affected so much by the drought. Average growth on brush layers ranged from 2.8 meters to 4.3 meters (110′′ to 169′′ ) on live pole drains from 1.3 meters to 3.5 meters (51′′ to 138′′ ), live stakes from 1.1 meters to 3.5 meters (43′′ to 138′′ ) and the Sitka Mountain alder growth averaged 1.5 meters (59′′ ). The grasses and legumes were very well established; native herbaceous such as fire weed (Epilobium angustifolium), thimbleberry (Rubus parviflorus) and woody shrubs such as wood rose (Rosa gymnocarpa) and mock orange (Philadelphus lewisii) were getting colonized within the site. Erosion control matting netting was still present with no remaining fibers. 8

MARKETING

The Walker’s Landing Road case study shows a cost effective solution to permanently restoring a slope failure and the aesthetics of a hillside and riparian habitat by stabilizing the soil and re-establishing vegetation on a steep slope site. This case study also showed the efficiency and results of working with a Spyder Hoe on a site with extremes of slope steepness and unstable glaciolacustrine materials. REFERENCES Donald H. Gray & Robbin B. Sotir. 1996. Biotechnical and Soil Bioengineering Slope Stabilization. A Practical Guide for Erosion Control. A Wiley-Interscience Publication, 378p. Schiechtl H.M. & Stern R. 1996. Ground Bioengineering Techniques for Slope Protection and Erosion Control. Blackwell Science Ltd.146p.

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Rapid landslides threatening roads: Three case histories of risk mitigation in the Umbria region of Central Italy D. Salciarini & P. Conversini Department of Civil and Environmental Engineering, University of Perugia, Perugia, Italy

E. Martini, P. Tamburi & L. Tortoioli Department of Environment, Territory, and Infrastructures, Umbria Region, Perugia, Italy

ABSTRACT: The geomorphologic and geologic characteristics of the Umbria Region of central Italy, along with its urbanization, are the main cause of the widespread landslide risk. In the region, landslide risk is classified in a ranking between 1 and 4, where 4 is for the highest level. Shallow landslides, debris flows and rock-falls are included in this highest level, since they are extremely rapid and intense. Many of these typologies of landslide recently occurred in the Umbria Region of Central Italy affecting the major road network. Pilot study areas were selected within the Region where important transportation corridors lie in the zone threatened by rapid landslides. This paper describes the risk mitigation strategies adopted for the selected roads and introduces the crucial issue of the residual risk management. By presenting different situations, this paper shows how the regional authorities faced the problem, and tries to summarize the answer that they gave to the issue by combining countermeasure, information and maintenance actions. 1

INTRODUCTION

In the Umbria Region of central Italy rapid landslides (e.g. rock-falls and debris flows) are very common and may cause severe damages to public and private properties (e.g. Felicioni et al. 1995, Guzzetti et al. 2003, Salciarini et al. 2006a). The most frequent triggering factors are represented by rainfall and earthquake (e.g. Guzzetti et al. 2004, Salciarini et al. 2006b). Most of these kinds of landslide threats roads and life-lines, generating a significant risk for people safety. Before 1998, regulations on the identification, assessment, and mitigation of landslide risk did not exist in the Italian normative, although the national and international scientific community had already elaborated many proposals. A set of laws was emanated urgently only after the disaster generated by the landslides occurred in May 1998 in the Campania Region of southern Italy. Masses of mud and debris triggered from the mountain above the villages of Quindici, Sarno, Siano and Braciliano (Salerno) destroying buildings and infrastructures. Beside the serious economical damage, these landslides also provoked 159 deaths. After the disaster, the Italian Government emanated an urgent law that included obligations both for the Regions and the River Basin Authorities. The law required both the delineation of extraordinary plans of

works for the areas prone to high hydro-geologic risk, and plans of hydro-geologic order at basin scale, for the other landslide and flooding risk areas. Also, the technical normative imposed to individuate the localization and delimitation of areas at risk, subdivided into 4 classes, as showed in Table 1. Table 1.

Classes of risk from the national regulation.

R4—Very high Risk

R3—High Risk

R2—Medium Risk

R1—Low Risk

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There is the possibility of: human loss and serious injuries to the people; serious damages to the buildings, infrastructures and the environmental patrimony; destruction of socialeconomic activities. There is the possibility of problems for the people safety; functional damages to buildings and infrastructures; interruptions of functionality of social-economic activities and damages to the environmental patrimony. There is the possibility of minor damages to buildings, infrastructures and the environmental patrimony that may impair people safety, the functionality of the buildings and of the economic activities. The social, economic and environmental damages are marginal.

The Umbria region is nearly entirely included in the River Tevere basin. It extends for 8457 km2 , of which 80% is hilly and mountainous territory. There are about 815,000 inhabitants, 5 regional parks and 1 national park. About 615 km2 of the region is indicated as potentially unstable by the landslide inventory map by CNR (Guzzetti & Cardinali 1990) that is about 9% of the territory. There are 179 areas prone to landslide risk, of which 69 are in the R4 class and 110 in the R3 class. For these areas, landslide risk mitigation is planned by funding specific programs that comprise the work designing and execution, and the ordinary maintenance and monitoring actions (Tamburi 2006). Other than the areas classified as R3 and R4, the regional regulation also individuated the priority of the consolidation works of numerous historical centres classified as unstable, as for the towns of Orvieto, Todi, Montone and Massa Martana, and the natural heritage of Marmore Falls (Conversini et al. 1995; Felicioni et al. 1995). For monitoring and maintaining the consolidation works completed in these areas, a regional system of ‘‘Observatories’’ were instituted, (e.g. the Orvieto Observatory) (Martini 1995, Pane & Martini 1996). Currently, analogous observatories are in preparing for the risk mitigation works regarding the infrastructures. In what follows, we briefly introduce the theoretical definition for risk, describe three pilot areas where different mitigation strategies were adopted, and conclude with a discussion on the acceptability of the residual risk after the works, underpinning the challenging of the issue.

One of the challenge currently faced by the international scientific community is the feasibility of formally adopting a ‘‘risk acceptance criteria’’. Acceptability decisions are based both on consideration of the characteristics and probabilities of hazard, both on the potential losses of the elements at risk. The hazard level can be theoretically measurable both at basin scale and at regional scale, and so it can be for the economical assessment of damages or losses of structures; while it is very difficult to handle the ‘‘risk acceptance criteria’’ when the indicators for the acceptability are human lives. The issue becomes even more complex when the discussion is referred to infrastructures. Indeed, a great difference exists whether, for equal hazards, the element at risk is static or dynamic. For example static elements at risk are villages, and their inhabitants. In such cases, theoretically, the occurrence of the hazard produces a partial or total loss of the elements a risk. Thus, the permanence of human beings at risk requires mandatory actions from the authorities. Dynamic elements at risk are, for example, people travelling along a road. In such a case the possibility of occurrence has to be combined with the possibility of the presence of such elements at risk in the same time. Therefore, the acceptability of the risk is measurable only with probabilistic methods, and for the authorities is not trivial to decide how strength the mitigation actions have to be. The following study cases that we present constitute a basis for discussing the complexity of this subject. 3

2

TERMINOLOGY AND DEFINITIONS FOR RISK

Risk is defined as the value of potential loss, times the probability of loss. Losses depend on the presence of elements at risk, such as people, infrastructures, social and environmental values (e.g. Hungr 2006). The hazard assessment and the risk assessment could be regarded as two successive steps of a procedure. Hazard assessment includes the analysis of the landslide hazard, characterizing its probabilities of occurrence, intensities, etc. Risk assessment comprises the identification of existing or potential elements at risk (Varnes 1984). Mitigation strategies aim to reduce the risk for people and environment. However, although all theoretically possible safety measures would be applied, a certain level of risk cannot be completely eliminated, because of the necessity of balance between economical and environmental sustainable actions. This portion of risk, defining as ‘‘residual risk’’, should be evaluated and managed.

MANAGEMENT AND MITIGATION OF LANDSLIDE RISK IN UMBRIA REGION

In the Umbria Region, the management and mitigation of the risk produced by rapid landslides that threat life-lines have been a crucial challenge from several years, due to the widespread landslide areas in the region, and to the intensity of each event (e.g: Guzzetti et al. 2003, Conversini et al. 2005, Tamburi 2006). In add, the morphologic characteristics of the territory often lead the transportation paths to be nearly obliged, particularly in some areas where the bottoms of the narrow valley are the only alternative for the roads. In the Umbria Region the landslide inventory map by CNR (Guzzetti & Cardinali, 1990) is an important tool to individuate the critical areas but, for its definition, it cannot provide support to the engineering design phase. Indeed, a conceptual model for the effective mitigation of landslide hazard requires a detailed characterization of the landslide deposits that includes information on water condition, material properties, shear strength parameters, topography, and material thickness. For rapid landslides the characterization is

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even more challenging, as the interval between the triggering phase and the propagation is typically very short. The mitigation strategies for such a kind of phenomena, combine the realization of active protections and passive defences, and are mostly delineated after a catastrophic event. The priority order of the funds financed by the Government necessarily affects the scheme of the management and mitigation planning. Indeed, typically, the works to mitigate landslide risk are subsequent to landslide occurrences, as shown in the flow chart of Figure 1. Therefore, although the landslide inventory maps are widely used to identify the potential unstable areas, the economical resource is provided, principally, after that the events actually occurred. In such context, also the works defined as ‘‘preventive’’ are often realized successively to the actual occurrence of a landslide. Extending the investigation to the adjacent areas, these works aim to reduce the possibility that potential analogous situations occur for a second time in the same zone. As already introduced, every time that a mitigation strategy is approved, a residual risk is accepted, derived from the necessity of balance between the economical and environmental sustainability. Up to now, a welldefined technical methodology to quantitatively assess such a residual risk is still lacking. Thus, the regional authorities have tended to use a very low threshold for the ‘‘residual risk acceptability’’. Also, they have managed the residual risk challenge investing part of the Government funds in monitoring and maintenance actions. Finally, they have tried to strengthen the information phase, by which people are made aware and conscious of the presence of the risk. In what follow, we show study cases where this kind of management is adopted.

4

PILOT AREAS DESCRIPTION

Figure 1. Typical successive actions adopted for the mitigation of the risk from rapid landslides.

Figure 2. Location of the three pilot areas in the Umbria Region of central Italy.

We selected three pilot areas located in the Umbria Region (Figure 2), where rapid landslides (rock-falls and debris flows) impacted the major interchange lines (state and local roads). The mitigation strategies were different, depending on the specific case. Although the residual risk was not quantitatively evaluated, the study cases considered in the planning actions the importance of its presence. 4.1 Case study 1: debris flows in the Terria basin During the summer 2005, two debris flow events occurred in the Terria basin (Terni Province, in the southern Umbria). The first happened on 30 June 2005 and was triggered by a heavy rainstorm that was concentrated over a small area, including the basin. The second debris flow occurred in the early evening of 31 August, after a short but heavy rainfall. In both cases the triggered deposit impacted the local road, damaging the bridge (Figure 3). The loss was exclusively economic, both direct for the structural damage and indirect for the temporary closing of the viability. The mitigation actions were subdivided into two stages of work. The emergency actions were focused in

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reducing the possibility which other successive events could reach the road. They comprised the installation of selective steel checkdams (Figure 4) to protect the roads. Further mitigation actions followed, along with a detailed study of the hydrologic, morphologic and geologic characteristic of the basin, with a general consolidation of the watercourse and the hillslopes, including the realization of concrete checkdams and vegetation of the slopes in erosion.

In the basin, the loose material volumes that can be triggered again are very relevant. For this reason a residual risk still persists for the community that uses the road to reach the village, although the mitigation works have been completed. The acceptability of this risk has been considered insufficient by the authorities. Thus, they have decided to integrate the structural actions (check dams and consolidations) with a warning system to real-time inform the community about the safety condition of the road. With this aim, a rain gauge to alert the population when critical rainfall threshold that can initiate the landslide are exceeded was installed in the basin. When rainstorms are characterized by intensity and duration that can produce unsafe conditions for the road users, the road traffic is temporarily interrupted. The residual risk is then managed combining the effectiveness of the structural works (by ordinary maintenance) with informative action derived by at-site monitoring. 4.2

Figure 3. Mechanical means at work, few hours later the event, to move the deposit that obstructed and damaged the bridge of the local road.

Figure 4. Selective steel checkdams installed few weeks after the debris flow events to protect the road from the sediment.

Case study 2: rock-falls along the state road n. 298 ‘‘Eugubina’’

The state road n. 298 is located in the northern Umbria (Perugia Province) (Figure 5). Following the heavy rainfalls of the winter season, in December 1996, a block triggered from a steep cliff and after destroying two orders of fences along the hillslope, reached this road. With emergency works the fences destroyed by the block were substituted and consolidated. Then, from the single rock-fall event, the regional authorities decided to extend the study to the entire hill slope, due to the importance of the road that is an interchange line between two regions. After detailed geo-structural analyses, helicopter surveys and geotechnical investigations, the sites with higher propensity to landslides were zoned and the following

Figure 5. Fences damaged by the rocks that impacted the state road.

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set of synergic works with active and passive function were funded: – elastic fences with dissipation systems; – cortical consolidation; – rock bolts. The residual risk management is highly correlated to the effectiveness of the passive defence systems. To preserve it unaltered, the regional authorities planned an efficient framework of maintenance works that include ordinary and extraordinary action during the year. 4.3

Figure 6. The seventy-tons block that detached from the cliff impacting the state road.

Case study 3: rock-falls along the state road n. 209 ‘‘di Cascia’’ and n. 320 ‘‘Valnerina’’

The state roads n. 209 and 320 are located in the eastern Umbria (Perugia Province). After the earthquake of 1997 several rock-falls occurred along these roads, impacting sectors very dangerous for the people safety. Following these events, the Umbria Region funded active and passive actions to mitigate the landslide risk. The steep hill slopes along the roads were consolidated by: – blasting of the block in evident prone to collapse conditions; – cortical consolidation of the hill slope with steel netting; – rock-fall barrier systems with high energyabsorbing capability; – rock bolts and cables. Due to the urgency of restoring the path, the works started while the residual seismic activity was still present. During the work progress, a secondary earthquake triggered a huge block from the cliff that destroyed the defence works just installed. In every countermeasure action the acceptation of a predictable or unpredictable residual risk is always implicit, and although every action was carried out with detail and analysis, two orders of barriers could not blocked a rock of about 70 tons that detached from an altitude of about 150 meters (Figs. 6–7). Following this event, beside the described structural works, an action of monitoring and maintenance was planned. It comprised the monitoring of: 1. the effectiveness of the barriers, measuring the stress variations along the longitudinal axis of the fence, by means of an electric cable connected to the acquisition system; 2. the effectiveness of the cortical netting, by means of pressure and displacement transducers; 3. the stability conditions of critical rock masses that could not be blasted, by means of extensimeter systems.

Figure 7. Mechanical mean removing the block few hours later the event.

The total cost of the countermeasures, including the active and passive actions, and the monitoring was related to the actual length of the mitigated road sector, obtaining an assessment of the ‘‘works cost’’ per ‘‘defended km’’ of about 250 thousands euros per km. This very relevant cost was justified by the importance of restoring the route of the road that could not have any alternative paths.

5

CONCLUDING DISCUSSION

After presenting the three study cases, is evident that the management of the residual risk after that the countermeasure works are concluded is a relevant issue. Warning system, monitoring, and maintenance actions seems to be a correct answer to the problem although sometime the unpredictability of the natural event can overcome the solutions. For a correct management of the countermeasure planning it is important to consider both the cost/effect ratio and the environmental sustainability of the work.

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In such a context, a certain level of residual risk has always to be accepted. The increasing imbalance between requirements and resources for landslide risk mitigation seems to turn into a management of the risk that has to admit higher threshold of acceptable residual risk. To face this tendency, regional authorities decided to associate the structural works both with communication and information actions to keep people aware, and with maintaining efforts to preserve the effectiveness of the installed protections. REFERENCES Conversini, P., Martini, E., Pane, V., Pialli, P., Tacconi, P., Tortoioli, L. & Ubertini, L. 1995. La rupe di Orvieto e il colle di Todi: due casi di città fragili (in italian). Geologia Applicata e Idrogeologia XXX:211–224. Conversini, P., Salciarini, D., Felicioni, G. & Boscherini, A. 2005. The debris flow hazard in the Lagarelle Creek in the eastern Umbria region, central Italy. Natural Hazards and Earth System Sciences 5: 275–283. Felicioni, G., Martini, E. & Ribaldi, C. 1995. Studio dei centri abitati instabili in Umbria (in italian). Regional Atlas CNR-Umbria Region. Rubbettino eds., Catanzaro, Italy. pp. 418. Guzzetti, F. & Cardinali, M. 1990. Landslide inventory map of the Umbria region, Central Italy. Proceedings in: 6th ICFL-ALPS 90, Milan, Italy: 273–284. Guzzetti, F., Reichenbach, P., Cardinali, M., Ardizzone, F. & Galli, M. 2003. Impact of landslides in the Umbria Region, Central Italy. Natural Hazards and Earth System Sciences 3(5): 469–486.

Guzzetti, F., Reichenbach, P. & Ghigi, S. 2004. Rock-fall hazard and risk assessment in the Nera River Valley, Umbria Region, central Italy. Environmental Management 34(2):191–208, DOI: 10.1007/s00267–003–0021–6. Hungr, O. 2006. Terminology of hazard and risk, suggested definitions. Abstract in: Workshop on Guidelines for hazard and risk mapping, 18–21 September 2006, Barcelona, Spain. Martini, E., 1995. Osservatorio: strumento per la prevenzione dei rischi e la manutenzione degli interventi (in italian). Proceedings in: Orvieto e Todi, due città da salvaguardare, 10–11 February 1995, Orvieto, Italy. Pane, V. & Martini, E. 1996. The preservation of historical towns in Umbria. The Orvieto case and its observatory. Proceeding in: Int. Symp. on Geothecnical Engineering for the preservation of monuments and historic sites, 4–5 October 1996, Naples, Italy: 489–498. Salciarini, D., Conversini, P. & Godt, J.W. 2006a. Characteristics of debris flow events in eastern Umbria, central Italy. Proceedings in Geological Society of London (eds): 10th International Congress of the IAEG 6–10 September 2006, Nottingham, UK: IAEG-285. Salciarini, D., Godt, J.W., Savage, W.Z., Conversini, P., Baum, R.L. & Michael, J.A. 2006b. Modeling regional initiation of rainfall-induced shallow landslides in the eastern Umbria region of central Italy. Landslide, 3(3): 181–194. Tamburi, P. 2006. Dal risanamento delle frane alla manutenzione, conservazione, valorizzazione e sviluppo: l’esperienza Umbra (in italian). In Regione Umbria (ed.): Pianificazione, prevenzione, mitigazione e controllo del dissesto idrogeologico in Umbria, 14–19 November 2006, Perugia, Italy. Varnes, D.J, 1984, Landslide hazard zonation: a review of principle and practice. Unesco Natural Hazards Series n. 3, Paris, 63 pp.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Assessment of the slope stabilisation measures at the Cadas Pangeran road section, Sumedang, West Java D. Sarah, A. Tohari & M.R. Daryono Research Centre for Geotechnology, Indonesian Institute of Sciences, Indonesia

ABSTRACT: This paper presents the results of a study conducted to assess the stability of the slope above a provincial road at Cadas Pangeran section which had undergone slope modification and reinforcement after the failure in April 2005. The study involved geotechnical investigation and slope stability modeling to elucidate the present stability of the slope. Subsurface investigation showed that the sliding surface is located between loose silt and dense sandy silt layers at depth of 5–6 m below the slope surface. Slope stability analysis showed that the slope is still in critical condition with a factor of safety close to 1.0. This indicated that slope failure is likely to re-activate under the effect of rainfall infiltration. Recent field examination found the development of new tension cracks on the slope surface. Therefore the stabilization measures employed have not been thoroughly successful in mitigating the slope failure in the area. Inappropriate design of bored pile could be the reason for the present instability. Keywords: 1

Slope stability, failure mechanism, residual soil slope, factor of safety, slope stabilization.

INTRODUCTION

West Java province is the area with the highest landslide vulnerability in Indonesia. Factors like, intensed weathering, the existence of microstructures, the steep morphology (slopes with inclination over 25◦ ), uncontrollable land use and high intensity of rainfall (up to 100 mm/day) during wet months seem to control this susceptibility. The occurrences of landslides in this area have claimed a number of human lives and caused great economic losses every year.

The hilly area in Cadas Pangeran road section, Sumedang regency, West Java is one of the areas highly prone to landslide. On April 4 2005, a slope failure occurred in a natural slope above this road section. Local witnesses revealed that the slope failure was initiated by small failure at the toe of the slope and tension cracking along 100 m along the slope surface. Based on an investigation carried out following the failure event, the failure was classified as a nonrotational multiple slumping, with a sliding surface being located at the depth of 10 meters above the

Figure 1. Slope stabilization method employed in Cadas Pangeran slope.

Figure 2. Cracks shown by white dashed line along the slope surface indicating slope instability.

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shows the topographical map of the study site and the locations of the geotechnical investigations. Mechanical drilling was aimed to reveal the soil stratification. For this purpose, one borehole is located at the upper and middle slope, while the other two are at lower portion of slope. The drilling program also included the N-SPT tests at an interval of 1.50 m and undisturbed sampling. The depth of groundwater table was measured following the completion of drilling program. Disturbed and undisturbed samples recovered from the mechanical and hand borings were subjected to a series of laboratory tests to determine the index and shear strength properties. These properties were then used in the slope stability analysis. The analysis was carried out using Slope W/package (Geoslope, 2002a) under the change of groundwater level and seismic loading condition for the existing slope stabilization and the proposed alternative of slope stabilization. The analysis was focused on the stability of the lower portion of the slope. Non-circular failure was considered in the analysis with the failure surface cutting through the slope toe.

ground surface (Tohari et al, 2006). This slope failure involved total debris volume of 100.000 m3 which could have blocked the road and possibly dammed the Cipeles River below the road. To prevent further hazard, the Public Works Division of West Java provincial government had improved the slope by trimming it to 1:1 (H:V) slope and later installed a row of bore piles of diameter 0.6 m at 6.0 m depth at the slope toe. The bore piles were installed at 1.0 m interval and were tied up with a cap beam of 1.2 m width and 0.8 m thick. The configuration of stabilisation method is shown in Figure 1. Despite the completion of stabilization works, recent field investigation revealed that cracks were still found along the slope surface, indicating that the slope is still in the critical condition (Fig. 2). This paper presents the result of a series of geotechnical investigations carried out to evaluate the effectiveness of the existing slope stabilisation method employed. The investigations included topographical mapping, mechanical drilling, mechanical cone penetration tests, a series of laboratory tests and slope stability modelling. 2

3

METHODS OF INVESTIGATION

RESULTS OF FIELD INVESTIGATION

3.1 Soil stratification

The field investigation began with local geological and topographical mapping, followed by four mechanical drilling up to a maximum depth of 22 m, twelve Dutch cone penetration tests, and eight shallow hand borings up to a maximum depth of 6.0 meters. Figure 3

9.238.100

The results of the field geotechnical investigations showed that the slope is composed of three layers of soils originated from volcanic tuff and breccia of approximately 28 meters thick. The stratification of

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TOPOGRAPHICAL MAP OF LANDSLIDE STUDY AREA

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Figure 3. Detailed topographical map of Cadas Pangeran landslide and the locations of the subsurface investigations.

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Elevation (m) 790

1 Clayey Silt 785

2 Loose Sand 2

qc (kg/cm ) 0

780

0

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3 Dense Sand 4 Volcanic Breccia

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Figure 4. Cross section of Cadas Pangeran landslide along line A-A′ illustrating the soil stratification.

the slope is divided into four different layers (Fig. 4) as the following: 1. Reddish brown clayey silt of 3.0–4.0 m thick with N-SPT value of 13–22 and cone resistance value, qc of 30 and qc >250 kg/cm2 .

Table 1.

Depth of groundwater table.

Bore no.

Borehole depth (m)

Depth of groundwater table (m)

Aquifer layer

DH-01 DH-02 DH-03

22,0 16,0 13,5

16,8 9,9 4,7

Dense fine sand Dense fine sand Loose silty sand

Table 2.

Geotechnical properties of soils.

Soil type

Unit weight (kN/m3 )

Cohesion, cu (kPa)

Friction angle (◦ )

Clayey silt Loose silty sand Dense sand Volcanic breccia

18,0 17,6 15,0 19,0

19,0 5,0 33,0 250,0

30,0 20,0 32,0 0,0

the slope toe. At the slope toe, the groundwater table presents within the loose sand layer. 3.3 Soil properties

3.2

Groundwater table

The results of measurement showed in Table 1 indicate that the groundwater table becomes shallower towards

Results of laboratory tests on undisturbed samples of loose silty sand showed that the soil has porosity of 55–60% and degree of saturation of higher than 90%. The shear strength of the silty sand was determined

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by back-analysis (Tohari et al, 2006). The geotechnical properties of other soil layers are summarised in Table 2.

4

ANALYSIS OF THE STABILITY OF SLOPES

4.1 Existing slope stability To examine the effectiveness of the existing slope stabilisation method, slope stability modelling was performed using the parameters in Table 1. Figures 5 and 6 showed the results of slope stability calculation for the existing slope stabilisation configuration without and with seismic loading, respectively. The safety factor of the slope without seismic load is slightly higher than minimum acceptable factor of safety (Duncan and Buchignani, 1975), while when seismic load of 0.08 g is included, the factor of safety dropped below 1.0. Therefore, under the earthquake loading, the slope is in critical condition. To understand the change of the slope factor of safety due to rainfall effect, slope stability calculation was conducted under the increase in the water table of 1.0 meter above the existing condition. This analysis was carried out without incorporating seismic load, and the result is presented in Figure 7.

Figure 5. Slope stabilization with one row of bore piles without seismic load (FS = 1.26).

Figure 6. Slope stabilization with one row of bore piles with seismic load (FS = 0.98).

Analysis result pointed out that the slope is in critical condition with safety factor of 1.15. The failure would involve the loose sand layer at the slope toe, and yet the critical sliding plane does not intersect the bore piles. Hence the installation of one row of bore piles at the toe of the slope is not effective to increase the safety factor to minimum requirement of 1.25, especially under the increase of groundwater able due to rainfall infiltration. 4.2

Alternative of slope stabilization

Considering that the present slope modification and bore pile reinforcement are not effective, alternatives of stabilization method are required. Reinforcement of one row of bore piles of 0.6 m in diameter at 1.5 m interval at DH-03 with pile embedment of 11.0 meters is proposed. This bore pile reinforcement is intended to increase the resisting force at the loose sand layer. The positioning of the bore piles can be seen in Figure 8. To check the effectiveness of the proposed alternative, slope stability calculations were performed at critical level of water table. Figure 9 displays the result of calculation without seismic load in which the factor of safety had increased to 1.98. While the under

Figure 7. Slope stabilization with one row of bore piles with 1.0 meter elevated water table (FS = 1.15).

Figure 8. section.

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Position and dimension of bore pile at A-B cross

Figure 9. Slope stabilization with one row of bore piles at DH-03 at critical water level without seismic load (FS = 1.98).

passing through the dense sand layer is more effective compared to the existing bore piles reinforcement at the slope toe. Lack of a comprehensive geotechnical investigation would likely to be the reason for the inappropriate design of the present slope stabilization measure. Detailed assessment of field geotechnical investigation and laboratory tests results revealed the existence of thick and loose silty sand layer from the ground surface, and high water table within the loose sand layer at the lower portion of the slope. These two factors are responsible for the present instability of the study slope. The nature of loose sand layer and high initial groundwater level would tend to accelerate the rise groundwater table, and hence, the critical pore-water pressure during rainfall infiltration. Thus, these factors are likely not to have been considered during the design of the present slope stabilization measure. 6

Figure 10. Slope stabilization with one row of bore piles at DH-03 at critical water level with seismic load (FS = 1.53).

seismic loading, the calculation resulted in a factor of safety of 1.53, as shown in Figure 10. Those values are well above the required minimum factor of safety. 5

DISCUSSION

Slope stability assessment of the existing condition revealed that the lower portion of the slope is in critical condition, and failure is likely to occur under elevated water table condition. Therefore installation of one row of bore piles at the slope toe could not guarantee the stability of the slope especially during rainy season. The current study shows that the installation of one row of bore piles with diameter of 0.6 meter at 1.5 m interval at the lower portion of the slope (DH-03)

CONCLUSION

The development of new tension cracks in Cadas Pangeran slope is detrimental to the stability of the slope especially during rainfall infiltration. Assessment of slope stability indicated that the existing stabilization methods employed in Cadas Pangeran landslide location is currently ineffective. Proposed alternative of bore piles reinforcement at the lower slope provides better stability during critical condition of elevated water table and earthquake loading. It can be concluded from the current study that inappropriate design of slope stabilization measure is responsible for the present instability of the slope. Comprehensive geotechnical investigation could have lead to an effective design of stabilization measure to prevent further slope failure hazard in this area. REFERENCES Duncan, J.M. & Buchignani, A.L. 1975. An Engineering Manual for Stability Studies. Civil Engineering 270B, University of California, Berkeley, CA. Geo-Slope 2002a. Slope/W Version 5.12 User’s Manual, GEO-SLOPE International Ltd., Calgary, Alberta, Canada. Tohari, A., Soebowo, E. & Sarah, D. 2006. Geological and Geotechnical Investigation of a Slow-Moving Landslide in Volcanic Residual Soil Slope for the Purpose of Hazard Assessment. Proceedings of International Symposium on Geotechnical Hazard: Prevention, Mitigation and Engineering Response, Yogyakarta, 24–27 April 2006, 167–175.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Analysis of control factors on landslides in the Taiwan area K. Shou, B. Wu & H. Hsu Department of Civil Engineering, National Chung-Hsing University, Taiwan, China

ABSTRACT: The main purpose of this study is to investigate the critical control factors to the landslides in the Western Foothill area. The study comprises two major parts, i.e., investigation and analysis. The study includes GIS investigation, field investigation, and testing on the geomaterial of different geological area. Then GIS is used to study the correlation of the control factors, including rainfall, geomorphology, geology and geomaterial properties and their relation with landslide. Hazard mitigation suggestions can be given based on the results of this study. 1

INTRODUCTION

It is generally accepted that the high frequency of landslide in Taiwan area is strongly affected by its geography and geology background. Taiwan is located in the subtropical area with high precipitation, especially during typhoon season in the summer. Besides, the geomaterials are highly fractured due to the tectonic activity of collision of Phillipine sea plate and Eurosian plate. As landslides are quite common and frequent in Taiwan area, a large proportion of those landslides are resliding in colluvial geomaterials. However, detailed theoretical studies on the control factors are still limited which motivates this study. The main purpose of this study is to investigate the critical control factors to the landslides in the Western Foothill area. The western foothill of Taiwan is mainly covered by sedimentary rock and metamorphic rock formations, such as the Pliocene Chilan formation, Kueichulin formation, the Plio-Pleistocene Toukoshan formation, and the Miocene Lushan formation. Those formations are generally poorly cemented or with intensive weak planes (joints or foliations), with high porosity and low weathering resistance. In order to prevent or reduce the landslide hazard, it is essential to have a more detailed understanding of landslide related behaviour of the geomaterial. A landslide prone area in Guo-Shing was adopted as the study area (see Fig. 1). The study comprises two major parts, i.e., investigation and analysis. The former part includes field investigation and GIS investigation on the DTM with focus on geology, geomorphology and weathering factors. And the later part includes in situ and lab testing on the geomaterial of different geological area and the study on the correlation of the control factors on the landslides. GIS is also used to study the correlation of the control factors.

Figure 1.

Study area in Central Taiwan.

Hazard mitigation suggestions can be given based on the results of this study. 2

LANDSLIDE IMAGE INTERPRETATION

The satellite images in this paper are from SPOT-4 (France) provided by the CSRSR, NCU. After Normalized Difference Vegetation Index (NDVI) was

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MINDULLE,2004/6/28

4

3

2

Figure 2. Time of earthquake and typhoon events.

2005/7/1

2004/7/1

2003/7/1

2002/7/1

Date

0

Earthquake magnitude

2004/7/23

5

2004/4/20

2002/1/6

2000/1/8

6

Typhoon Rainfall

Earthquake

1

2001/7/1

0

BILIS,2000/8/21 XANGSANE ,2000/10/31 2001/1/2

100

2000/7/1

200

611Earthquake, 2000/6/11

1999/1/5

300

1999/7/1

400

1 Ci-Ci Earthquake 1999/9/2

500

NARI ,2001/9/16

600

TORAJI,2001/7/28

700

1998/7/1

Precipitation at Sun-Moon Lake station(mm)

calculated (Rouse et al. 1973; Lillesand, 2000), the satellite images were divided by binary classification method into bare land of NDVI < 0.3 and nonbare land of NDVI ≥ 0.3. The landslide distribution and area were set up by combining DEM data and

slope data obtained from spatial analysis model in Arc View 9.1. In this study, a proper and efficient image processing procedure with threshold of NDVI can be built up through trial and error method. And, a database of landslide and slope distribution was established by the above image processing procedure on satellite images and DEM data. With a reference to Jeng et al. (2005), this research takes slope 30% as the threshold of landslide classification. In addition to direct interpretation and digitalization of landslide range on the screen, semiautomatically classified data and topographic map are overlapped for confirmation of landslide, and then landslide interpretation range is modified to improve efficiently the accuracy of landslide identification. After semi-automatically classified images are evaluated, it becomes possible to provide necessary data via threshold of NDVI. Therefore, landslide identification and distribution require combination of classified image results, NDVI, DEM, topographic map and GIS.

(a) before 1999 Ch-Chi earthquake

(b) after 1999 Chi-Chi earthquake

(c) after Typhoon Toraji Figure 3.

(d) after Typhoon Mindulle

Landslide distributions before and after major events.

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Conversely, SPOT satellite images for image analysis are determined by considering the change of landslide from Chi Chi Earthquake, TORAJI Typhoon (July 28, 2001) to Mindulle Typhoon (July 2 Flood, 2004), as shown in Figure 2. 3

RAINFALL INDEX FOR LANDSLIDE

Typhoons in summer season are the major sources of heavy rains in Taiwan areas. Those heavy rains frequently induce landslides in areas with landslide prone characteristics, such as weak geomaterials, steep topography, poor drainage, etc. To understand the correlation of rainfall characteristics and lanslides, Landslides induced by the typhoon Mindulle in 2004 was studied. The distributions of maximum precipitation and accumulate precipitation were established and reviewed. This study introduces the rainfall index for landslide to correlate the rainfall and landslide. Other control factors such as geology, topography, etc. are also analyzed. As basic data for analyses, precipitation records of 12 stations during June 2 and July 10, 2004 were collected from Central Weather Bureau (Figure 4). 3.1

Distributions of maximum rainfall density and cumulative rainfall

Although Typhoon Mindulle affected Taiwan on July 1, 2004, heavy rain hitted central Taiwan since July 2 and diminished after July 11. Precipitation records during July 2 and July 10 were adopted to calculate the cumulative rainfall. The distribution of cumulative rainfall in Figure 5 was established by applying the kriging function of GIS to the cumulative rainfall record at the precipitation stations in the study area. Maximum hourly precipitation is defined as maximum rainfall density in this study. The maximum

1800

1600

R1

1400

R2 R3

1200

R4 R5

1000

R6 800

R7 R8

600

R9 R10

400

Figure 6. The distribution of maximum rainfall intensity obtained by kriging.

rainfall density for each grid in the study area should be determined by adopting the maximum from the 216 (24 hours ∗ 9 days) hourly rainfall distribution. However, this approach is very time consuming and sensitive to comparatively lower values at precipitation stations. This study adopted a simple approach, i.e., kring the maximum hourly precipitation of each precipitation station (see Figure 6). Since only one precipitation distribution need to be calculated, this approach is more efficient in calculation.

2000

cumulative rainfall(mm)

Figure 5. The distribution of cumulative rainfall obtained by kriging.

R11 R12

200

3.2 Determination of rainfall index for landslide

0 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 1 2 3 4 5 6 7 8 9 10

Time(days)

Figure 4. The records at different precipitation stations within or adjacent to the study area.

In this study, to correlate the rainfall and landslide, the Typhoon Mindulle induced landslides determined by the satellite image interpretation was mapped with the distributions of cumulative rainfall and maximum

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cumulative rainfall(mm)

1800

L1

Landslide

1600 1400 1200 1000

L2

800 600 60

70

80

90

100

110

120

max-ranifall intensity(mm/hr)

Figure 7. The cumulative rainfall-maximum rainfall density distribution of landslides and its boundaries.

rainfall density. As shown in Figure 7, the rainfall characteristics of landslide possesses a range of distribution. The upper and lower boundaries can be defined as L1 : a1 x + b1 y + c1 = 0

(1)

L2 : a2 x + b2 y + c2 = 0

(2)

where L1 is upper boundary and L2 is lower boundary. For a certain landslide point, the distances to the upper boundary line and lower boundary line can be expressed as:    ax0 + by0 + c    d1 & d2 =  √  a2 + b2

(3)

where d1 and d2 are the distances to L1 and L2. The Rainfall Index for Landslide, Id is defined as

Id =

d2 d1 + d2

(4)

The value of Id can be between 0 and 1, where value 0 means landslide will not occur, value 1 means landslide will occur. For value between 0 and 1, the more close to 1 the more probability of landslide occurrence. However, for the exceptional cases with L1 < 0 and L2 < 0, the Id is set to be 0; for the exceptional cases with L1 > 0 and L2 > 0, the Id is set to be 1. According to the rainfall characteristics of landslides, the upper boundary line and lower boundary line for the study area were defined as L1 : 4.17x + y − 2080.47 = 0

(5)

L2 : 26.75x + y − 2827.75 = 0

(6)

Figure 8.

The distribution of rainfall index for landslide.

Based on the boundary lines, the distribution of rainfall index for landslide Id can be calculated and shown in Figure 8. The results show that the east side of the study area has higher Id distribution which reflects the higher maximum rainfall density and accumulative rainfall in the higher elevation area.

4

INSTABILITY INDEX METHOD

Instability Index method was commonly applied to the analyses of landslide potential. This method statistically establishes an empirical formula in terms of control factors, which can quantify the hazard potential of landslides. The larger the Instability Index is, the more the possibility the hazard might occur. For comparison, the instability index analyses are performed separately for regions of Plio-Pleistocene, Miocene, and Oligocene formations. Through preliminary studies, this study adopts nine influence control factors: slope angle, slope direction, elevation, distance to road, distance to fault, distance to river, Dipe Slope Index, NDVI, Rainfall Index for Landslide. To reflect the influence of dip slope, the Dip Slope Index is introduced. The dip slope index is used to quantify the correlation between dip direction of bedding plane and dip direction of slope, and is defined by angle between two dip directions: 1 for 0◦ ∼30◦ , 2 for 30◦ ∼60◦ , 3 for 60◦ ∼120◦ , 4 for 120◦ ∼150◦ , and 5 for 150◦ ∼180◦ . 4.1 Evaluation of control factors For a specific control factor, the evaluation score is determined by the sliding ratio, which is defined as Gi =

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Gi,landslide Gi,total

(7)

whereGi,landslide is total grid number with landslide and class i control factor, Gi,total is total grid number with class i control factor. If we classify the sliding ratio to 1∼10 degrees, the evaluation score of a control factor can be defined as

Di =

9(Si − Si,min ) +1 (Si,max − Si,min )

(8)

Where Si is the normalized sliding ratio of the control factor, Si, max and Si, min represent the maximum and minimum values of sliding ratio.

4.2 Weighting of the control factors

Vi V1 + V2 + V3 + · · · + Vn

(9)

where Vi is coefficient of variance.

Calculation of instability index

The potential of landslide can be quantified by the Instability Index, which is a function of control factors and their weighting factors:

Dtotal = D1W1 × D2W2 × · · · × DnWn

(10)

where Dtotal is the instability index, D1 ∼Dn are the evaluation scores of the control factors, and W1 ∼Wn are the weighting factors.

5

Coefficient of variance of the sliding ratio reflects the sensitivity of the specific factor. So the coefficient of variance can be applied to obtain a weighting. The weighting factor of the i-th control factor can be defined as

Wi =

4.3

RESULTS OF ANALYSES

The Instability Index analyses are performed separately for regions of Plio-Pleistocene, Miocene, and Oligocene formation. Through the test of significance, minor control factors can be removed. For the study area, there are 6, 7 and 6 major control factors for the Plio-Pleistocene, Miocene, and Oligocene formation regions. The results of the Instability Index analyses are shown in Table 1, Table 2, and Table 3. The results reveal slope angle, NDVI, and Id are the top three control factors for the landslide, although the rank is different due to the difference in geomaterial and precipitation.

Table 1. Results of instability index analysis for Plio-Pleistocene area.

Coefficient of variation (%) Weight Rank

Slope

Dis. to river

Dis. to road

NDVI

Dip slope index

Id

94.7 0.19 2

49.1 0.10 6

76.2 0.15 5

94.1 0.19 3

82.8 0.17 4

96.8 0.20 1

Table 2. Results of instability index analysis for Miocene area.

Coefficient of variation (%) Weight Rank

Slope

Dis. to fault

Dis. to river

Dis. to road

NDVI

Dip slope Index

Id

83.1 0.16 3

51.1 0.10 6

71.4 0.14 5

51.0 0.10 7

96.5 0.19 1

75.4 0.15 4

92.6 0.18 2

Table 3. Results of instability index analysis for Oligocene area.

Coefficient of variation (%) Weight Rank

Slope

Dis. to river

Dis. to road

NDVI

Dip slope index

Id

79.2 0.17 1

73.3 0.16 5

72.9 0.16 6

77.8 0.17 2

75.4 0.17 4

76.0 0.17 3

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6

REFERENCES

CONCLUSION

In this study, the landslide control factors were studied separately for regions with different geomaterials. The Instability Index method was applied with introduction of two new control factors, i.e., dip slope index and landslide rainfall index. And the results of analyses reveal that slope angle, NDVI, and Id are the top three important control factors.

Rouse, J.W. et al. 1973. Monitoring vegetation systems in the Great Plains with ERTS. Third earth resources technology satellite-1 symposium. Lillesand, T.M. & Kiefer, R.W. 2000, Remote sensing and image interpretation. John Willey & Sons, New York, 736p. Jeng, C.T. et al. 2005. Automatic Interpretation of Satellite Images for Landslide Study—Study on the Tachashi Catchment due to Chi Chi Earthquake and Subsequent Typhoons. Symposium CGS 1994 Annual Meeting.

ACKNOWLEDGEMENT This paper was made possible by the support of National Science Council, Taiwan. (NSC94-2211-E005-022).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Inclined free face riverbank collapse by river scouring Ji-chao Sun & Guang-qian Wang State Key Laboratory of Hydroscience and Engineering, Tsinghua University, Beijing, China

ABSTRACT: Inclined free face riverbank collapse by the river scouring is a new type of collapse, and is little involved in previous researches. In the paper, we researched this type of collapse, obtained the stress expression of the inclined free face riverbank by the stress function, and analyzed the anti-force source. The anti-force of collapse is from the anti-tensile strength of the riverbank, which increases with the water content decreasing, namely it increases with water level falling. The X-stress is the tensile stress on the top surface of the inclined free surface riverbank, of which the maximum tensile stress increases with the exposure surface increasing. With scour being kept on, the slope angle of the inclined free surface riverbank becomes less, and at same time the maximum tensile stress becomes more. When the tensile stress is more than the maximum tensile strength of the soils of the riverbank, the crack appears, and at this time the portrait crack of the riverbank appears, at last even to collapse.

1

INSTRUCTION

Under the river affection, the shore dike is scoured by the river, so the river becomes wider and wider with the riverbank collapse, which are many collapse types mentioned in literatures (Wang, 2004). Scientists and researchers did some researches on the collapses by the mechanics analysis, and made great achievements. But a kind of collapse type shown in Figure 1 is not concerned in previous researches, which needed to be studied. By the research in the paper, the new collapse type can be added into the whole collapse research, and it is probably that the mechanism opened out can solve some problems. In this research, many advices and suggestions can be given in the riverbank reinforced and collapse prevention. The riverbank collapse generally exists in Yangtze River, Yellow River and Han River etc. The riverbank collapse is treated as a typical disaster, including (a) collapse threatening the river dike safety, (b) collapse threatening the safety of building and farmland, (c) sediment from collapse leading to the riverbed evolution, (d) effecting to the shipping. The riverbank collapse is result from the combined affection by the incoming runoff and sediment, the river channel scour and siltation evolution, the riverbank soil, the geological structure and so on. The river bank soils are the internal factor, while the conditions of river water are the external factor. Nowadays, researches about the mechanism of many factors affecting the collapse are little, but many achievements are about the mechanism of dif-

ferent factors effecting on the collapse, including the flow intensity, the changing of river water level, the river channel scour and siltation evolution, the soils conditions and so on. From the size of collapse body and collapse form, the river collapses can be divided into Pit-type Slide, strip slide, slip slide and wash slide as shown in Figure 2 (Wang, 2003). And from the river channel form, river slope, disaster form and so on, the river collapse can be divided into the top eroded collapse, the lateral eroded collapse, the deep eroded collapse and the local eroded collapse (Wang, 2003). But a kind of collapse shown in Fig. 3 is a new type. When the river foot is scoured, the top river bank is in cantilever state and its body is big. The body is large enough to landslide. The drop collapse is dived into three types, as shown in Fig. 3. a. Shearing collapse When the weight of the soil body in cantilever state is more than the anti-shear strength, the soil body slides from the shear surface AB (Fig. 3a) b. Rotating collapse When the weight moment of the soil body in cantilever state is more than the anti-tense moment, the soil body rotates and collapses (Fig. 3b), which is shown in experiments and natural rivers. c. Tensing collapse When the tensile stress from weight of the soil body in cantilever state is more than the anti-tense strength, the soil body collapses (Figure 3c).

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Figure 1. Riverbank crack/collapse from scouring.

Figure 3.

Figure 2. Typical collapse.

But the previous researches and literatures have not given the mechanics analysis. In geotechnical engineering field, many serious of great achievements are made, including the calculating method about slope stability, such as circular slip algorithm (Chen, 2003), strength reduction algorithm (Deng, 2007), back-analysis (Aydan, 2002), Monte Carlo-grads method (Sun, 2005) and so on; including many factors, such as earthquake (Ulusay, 2007,

Drop collapse.

Ingles J, 2006, Pradel, 2005), rain storm (Krejci, 2002, Qiu, 2007), current (Stegmann, 2007; Miller, 1998), and so on; Slope monitor and test method, such as remote sensing (Yang, 2004), GIS (Muthu, 2007, Ayalew, 2005), map comparison (Ayenew, 2005) and so on. In Fig. 1, the slope toe is scoured leading to the slope in cantilever state. Some part of the slope collapses with river going on scouring, and a concave appears, about which few researches are, and which leads to researchers difficult in understanding. And in slope slide fields, the phenomena in the Fig. 1 is not concerned, especial about the mechanics analysis, so the research is needed, which can be added into the systemic researches.

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2

MECHANICS ANALYSIS

With water level descending in river, some parts of the riverbank appear from the water, shown in Fig. 4. Water scours the lateral part of the riverbank, and gives a supporting force to the riverbank from the high water level. While the level falls, the supporting force decreases and even to zero. The exposed riverbank is in the cantilever state. With the exposed surface enlarging, the crack on the top of riverbank appears, and at last the portrait crack appears, shown in Fig. 1. The ketch map for mechanics analysis is shown in Fig. 5. 2.1

Stress function

With the dimension analysis, the research obtains the stress function ϕ. The stress is connected with the x, y and α. The dimension of the stress is [Force] [Length]−2 , the dimension of ρg is [Force][Length]−3 , α is dimensionless, and x, y have [Length] dimension]. So the expression of the stress component is the connection of the aρgx and bρgy. But a, b are dimensionless and only connected with α. The stress

components are pure once polynomial and all the stress components are the second order partial derivative of stress function to the coordinates, so the stress function is pure cubic polynomial, ϕ = ax3 + bx2 y + cxy2 + dy3

(1)

2.2 Body forces Body forces are X = 0 and Y = ρg, so ⎧ 2 ⎪ σx = ∂∂yϕ2 − Xx = 2cx + 6dy ⎪ ⎪ ⎨ 2 σy = ∂∂xϕ2 − Yy = 6ax + 2by − ρgy ⎪ ⎪ ⎪ ⎩ ∂2ϕ τxy = − ∂x∂y = −2bx − 2cy

2.3

(2)

Boundary conditions

The top boundary on y = 0 is σy = 0, τxy = 0, so from Equation 2,  a=0 (3) b=0 The incline boundary y = −x tan α, x ∈ [−A, 0], and the follow equation is satisfied,  l(σx ) + m(τxy ) = 0 l(τxy ) + m(σy ) = 0

Figure 4. Riverbank force evolution with water level.

where A is the horizontal distance between the spire point and inner point of the research riverbank, l = sin α and m = cos α And from the Equation 3,  sin α(2cx + 6dy) + cos α(−2cy) = 0 sin α(−2cy) + cos α(−ρgy) = 0 And results are obtained, as follow, ρg 3 tan2 α ⎪ ⎩c = − ρg 2 tan α ⎧ ⎪ ⎨d = −

(4)

2.4 Stress expression From Equation 1, 2, 3 and 4, ϕ=−

ρg ρg xy2 − y3 2 tan α 3 tan2 α

⎧ ρg ⎪ ⎨σx = − tan α x − σy = −ρgy ⎪ ⎩τ = ρg y xy tan α

Figure 5. Research riverbank coordinates.

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2ρg y tan2 α

(5)

where ‘‘+’’ shows it is the tensile stress, and ‘‘−’’ shows it is the compressive stress. The Equation 5 is the riverbank soil stress. 3

ANTI-SHEAR SOURCE

Literature (Stegmann, 2007) believes that mineralogical composition and pore fluid pressure are the crucial controls for mechanical stability of water-saturated sediments. And some researches about unsaturated soils are rainfall (Sun, 2006), which leads to landslide (Salciarini, 2006; Sun, 2006). Bishop gave a effective stress expression,

Figure 6.

Stress state in different part.

Figure 7.

X-stress changing with the slope angle.

σ ′ = (σ − ua ) + χ(ua − uw ) Strength expression is as follow, τf = c′ + (σ − ua ) tan ϕ ′ + χ (ua − uw ) tan ϕ ′ In the theory, χ has no clear physical meaning. Blight in 1967 made experiments to research about χ characteristics in strength and the volume. He obtained that χ was different in strength and volume problem, the value of which can more than 1 (Wu, 2003; Fredlund, 1993). Fredlund gave the two stress state variables, and obtained the anti-shear strength formula, ′

′

τf = c + (σ − ua ) tan ϕ + (ua − uw ) tan ϕ

b

(6)

where tan ϕ b is the friction coefficient of suction. And in the last years, the experiment results show ϕ b is not constant and ϕ b has the unclear meaning and is difficult in being measured. The above two theories absorbed two key parameters χ and ϕ b , which have unclear meaning, and it is difficult in measuring them. From the two strength equations, tan ϕ b = χ tan ϕ ′ is obtained, so they are accordant each other. In literature (Li, 2000), a relationship between tensile strength and dry unit weight of soils is established. And the anti-tensile strength in soils with different water content is different (Lu, 1993; Lu, 1997; Tan, 2005; Miao, 1999). The anti-tensile strength is the soils cohesion (Wang, 2003), which is close connected with the water content (Liu, 2006; Li, 2006; Liu, 2006). By studying the saturated-unsaturated soils antistrength and the soil-water characteristic curves, results show that the strength of the unsaturated soils is more than one of the saturated soils. From the soilwater characteristic curves, the suction increases with the water content reduced and the anti-shear strength increases with the suction increasing.

4

INFLUENCING FACTORS

The parameters are chosen, in Fig. 5 ρg = 20000 N/m3 From the Equation 5, the X-stress σx is shown in Fig. 6, which is tensile stress and increases with the |x| increasing. Connected with the Fig. 4, with the water level falling, the exposure riverbank is more and more, the A becomes more and more, and the tensile stress becomes more and more. At last, when σx > [σt ], the surface has the crack, which well explains the portrait crack of the riverbank, even to collapse. We choose A = 0.4 m, so we obtained the result shown in Fig. 7. With the slope angle decreasing, the X-stress σx is increasing, which is the result from the river scouring. 5

CONCLUSIONS

Our object is to obtain the reasonable explanations about the incline free surface collapse by river scouring. The incline free surface collapse is a new type of

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the collapse, which is added into the existing classification. We obtain the stress expression of incline free surface river bank the by the stress function, the support force source, which well explains that incline free surface riverbank can stand, and the influencing factors. Conclusions are in detail as follow, First, a new type of collapse/incline free surface riverbank is defined. By the river scouring, the riverbank toe is hollowed out, but some part of riverbank can stand. With the river being kept on scouring, the slope angle is reduced, at last to collapses. In the previous research, the new type of collapse/stand incline free surface riverbank is not clear; in the paper, the new type of collapse is defined as one of the drop collapse (Wang, 2003). Second, we obtain the stress expression and the influencing factors of the incline free surface riverbank. By the stress function, we obtain the stress expression of the incline free surface riverbank. X-stress of the top surface of incline free surface riverbank is the tensile stress. When the tensile is more than the tensile strength of soils, the crack appears on the top of the riverbank. The maximum tensile stress of the top exposure riverbank increases with the exposure riverbank increasing. With river keeping on scouring, the slope angle becomes less and less, and the tensile stress becomes more and more. Third, the tensile stress source is analyzed. The anti force is little when it is saturated. And with the water content decreasing, the anti force of riverbank soils becomes more and more. Namely, with the river level decreasing, the riverbank gradually exposes from water, so the anti force of soils increases, which well explains the riverbank does not collapse justly when the river level decreases.

ACKNOWLEDGEMENTS The study is supported by the National Natural Science Foundation of China (No.50709015), by Post-Doctoral Science Foundation of China (No. 20070410531), Science Fund for Creative Research Group of China (Project No.50221903), by the National Key Basic Research and Development Program (973 Program No.2007CB714101).

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Ayenew T, Barbieri G. 2005. Inventory of landslides and susceptibility mapping in the Dessie area, northern Ethiopia. Engineering Geology 77 (1–2): 1–15. Chen Zuyu. Stability analysis of soil slopes-theory, methods and programs. Bejing: China Water Power Press, 2003. (in Chinese). Deng JH, Tham LG, Lee CF, et al. 2007. Threedimensional stability evaluation of a preexisting landslide with multiple sliding directions by the strength-reduction technique. Canadian Geotechnical Journal 44 (3): 343–354. Fredlund, D.G. & Rahardjo, H. 1993. Soil mechanics for unsaturated soils. New York: John Wiley & Sons. Ingles J, Darrozes J & Soula JC. 2006. Effects of the vertical component of ground shaking on earthquake-induced landslide displacements using generalized Newmark analysis. Engineering Geology 86 (2–3): 134–147. Krejci O, Baron I, Bil M, et al. 2002. Slope movements in the Flysch Carpathians of Eastern Czech Republic triggered by extreme rainfalls in 1997: a case study. PHYSICS AND Chemistry of the Earth 27 (36): 1567–1576. Li XiaoJun, Zhang Dengliang & Ren Yufang. 2000. Experimental studies on determination of tensile strength of road foundation soils. Journal of Xi’an Highway University, 20 (2): 20–21. Li, Yongle, Zhang, Hongfen, She, Xiaoguang, Hou, Jinkai & Cui, Xiangyu. 2006. Analysis on intensity characteristic and stability of unsaturated soil of the yellow river embankments. Yellow River 28 (1): 51–52. Liu, Cuiran, Yang, Mingqing & Liu, Zhenchao. 2006. Research on Experiment of Unsaturated Soil’s Shear Strength of Yellow River. Journal of Yellow River Conseruancy Technical Institute 18 (4): 24–26. Liu, Xiaowen & Geng, Xiaomu. 2006. Stability analysis of soil slope under rainfall seepage influence. Hydrogeology and Engineering Geology 6: 40–42, 47. Lu Zhao-jun, Wu Xiao-jing, Sun Yu-zhen, et al. 1997. The role of swelling pressure in the shear strenth theory unsaturated soils. Chinese Journal of Geotechnical Engineering 19 (5): 20–27. Lu Zhao-jun, Zhang Hui-ming, Chen Jian-hua & Feng Man. 1993. Shear Strength and Swelling Pressure of Unsaturated Soil. Chinese Jounal of Geotechnical Engineering 14 (3): 1–8. Miao Lin-chang, Zhong Xiao-chen & Yin Zong-ze. 1999. The relationship between strength and water content of expa-sive soil. Rock and Soil Mechanics 20 (2): 71–75. Miller DJ, Sias J. 1998. Deciphering large landslides: linking hydrological, groundwater and slope stability models through GIS. Hydrological Processes 12 (6): 923–941. Muthu K & Petrou M. 2007. Landslide-hazard mapping using an expert system and a GIS. Ieee Transactions on Geoscience and Remote Sensing 45 (2): 522–531. Pradel D, Smith PM, Stewart JP, et al. 2005. Case history of landslide movement during the Northridge earthquake. Journal of Geotechnical and Geoenvironmental Engineering 131 (11): 1360–1369. Qiu C, Esaki T, Xie MW, et al. 2007. Spatio-temporal estimation of shallow landslide hazard triggered by rainfall using a three-dimensional model. Environmental Geology 52 (8): 1569–1579.

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Salciarini D, Godt JW, Savage WZ, et al. 2006. Modeling regional initiation of rainfall-induced shallow landslides in the eastern Umbria Region of central Italy. Landslides 3 (3): 181–194. Stegmann S, Strasser M, Anselmetti F, et al. 2007. Geotechnical in situ characterization of subaquatic slopes: The role of pore pressure transients versus frictional strength in landslide initiation. Geophysical Research Letters 34 (7): Art. No. L07607. Sun jichao. 2006. Effects and Numerical Simulation of Rain Infiltration on Soil-Rock Aggregate Slope Stability. China University of Mining and Technology, Beijing. Sun Jichao, Gao Quanchen & Wang Haibiao. 2006. Numerical Simulation of Coupled Rainfall and Temperature of Unsaturated Soils. Key Engineering Materials 306–308: 1433–1438. Sun Jichao, Gao Quanchen, Zhao Yongcai. 2005. Calculating Safety Factor of Slope by the Monte Carlo-Grads Method. China Mining Magazine 14 (1): 67–69. Tan Luo-rong & Kong Ling-wei. 2005. Study on strength bebavior of expansive soil. Rock and Soil Mechanics 26 (7): 1009–1013.

Thorne C R and Tovey N K. 1981. Stability of Composite Riverbanks. Earth Surface Proeesses and Landforms 6: 469–484. Ulusay R, Aydan O & Kilic R. 2007. Geotechnical assessment of the 2005 Kuzulu landslide (Turkey). Engineering Geology 89 (1–2): 112–128. Wang GQ, Xia JQ & Wu BS. 2004. Two-dimensional composite mathematical alluvial model for the braided reach in the Yellow River. Water International 29 (4): 455–466. Wang Yangui. 2003. Study on mechanism of bank failure in the alluvial river. China Institute of Water Resources and Hydropower Research, [doctor thesis]. Wu Jianmin, Li Guangxin & Wang Chenghua. 2003. Effect of matric suction of unsaturated soil on the estimation of the load and internal force on supporting structure for deep excavation. Industrial Construction 33 (7): 6–10. Yang MD, Yang YF & Hsu SC. 2004. Application of remotely sensed data to the assessment of terrain factors affecting the Tsao-Ling landslide. Canadian journal of remote sensing 30 (4): 593–603.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Drainage control and slope stability at an open pit mine: A GIS-based hydrological modeling C. Sunwoo Geotechnical Engineering Division, Korea Institute of Geoscience & Mineral Resources (KIGAM)

Y.S. Choi & H.D. Park Department of Energy Systems Engineering, Seoul National University (SNU)

Y.B. Jung Geotechnical Engineering Division, Korea Institute of Geoscience & Mineral Resources (KIGAM)

ABSTRACT: Slope erosion and flooding of the open-pit mines are typical problems which need proper drainage controls at the heavy rainfall region where surface drainage pattern is dominated by the fast change of landform due to mining operation. This paper presents an application of GIS-based hydrological modeling, spatial analysis and slope stability in a region of Pasir coal mine, Indonesia. A detailed topographical survey was performed at the study area to generate a reliable DEM. By using GIS, the more reproducible information about hydrological characteristic of drainage system could be extracted from DEM. The results from drainage analysis and rasterbased spatial analysis showed that current arrangement of pumping facility is not suitable and some vulnerable places to erosion exist on the bench face due to concentrated surface runoff. Finally, some practical measures were suggested to optimize the design of drainage system and to monitor the slope stability by the surface water management at the study region during heavy rainfall. 1

INTRODUCTION

The need of a more sophisticated strategy for water management in open-pit mines has increased in recent years, because the laws passed and regulations implemented in most countries insist that open-pit mine planning includes a detailed discussion dealing with water resource management for mine reclamation (Hustrulid & Kuchta, 1995). Proper management and conservation of water resources in open-pit mines have always received some consideration in order to address the productivity and safety of mining operation (Meek, 1990). The strategy for water management in open-pit mines is usually dependent on the scale of the mining operation, the geological conditions, the characteristics of the ore deposits being mined, and especially the amount of rainfall. In this study, we focus on problems related to the drainage controls and slope safety in open-pit mines located on the high humidity region. In open-pit mines, drainage flow tends to concentrate in the lowest parts of pit through benches, ramps and haul roads. The steep gradient of the pit can significantly accelerate the velocity of drainage flow. The gullies can be created on the bench slopes owing to the concentrated and accelerated drainage

flows. Moreover, rapid erosion process on bench faces with concentrated drainage flow can become the cause of slope failures which endanger the safety of mining operations. Controlling the concentrated drainage flows is the most important problem to be solved in open-pit mines. Although some artificial ponds in the pit have roles as storage of the accumulated rainwater, their storage capacity can be insufficient to prevent the flooding during heavy rainfall. Therefore, pumping up rainwater to outside the pit is necessary in open-pit mines to prevent the flooding. Rainfall is known to have induced major slope failures in many countries. Runoff is the proportion of rainfall that flows from catchment area. Surface runoff depends on some of the following factors: rainfall intensity, area shape of catchment area, gradient and length of the slopes, condition of surface and nature of the subsuface soils. How much rainfall permeates into a slope depends on the above factors. Because these factors vary considerably between slopes, it is not possible to draw general relationships between rainfall and groundwater response with accuracy (Ambramson et al, 2002). Water plays a very important role in the slope stability. Water influences the strength of slope forming materials by increase in pore water pressure,

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and subsequent decrease in shear strength, reduction of apparent cohesion softening of stiff fissured clays and shales and chemical alteration and solution. In this study, we attempt to use Geographic Information Systems (GIS) as a decision making tool for the local-scale drainage controls in open-pit mines. This paper presents a case study to the Pasir open-pit coal mine in Indonesia which focuses on two problem solving processes for drainage controls: the identification of problematic benches for surface protection and the calculation of priority order to allocate the pumping facilities in the pit.

2

STUDY AREA

2.1 General and geology of Roto South area The Pasir coal mine is situated in the east side of Kalimantan, Indonesia (Figure 1). The mine area covers 504 km2 and has approximately a billion tons of recoverable reserves of bituminous coal. This mine is divided into Roto North, Middle and South in the north—south direction. The study area has a tropical climate with heavy rainfall, as much as 3055 mm of rainfall was recorded in 2005. The highest elevation in the study area is 180 ML and the lowest is currently—46 ML. Average slope angle is about 30 degrees. Slope erosions due to the accelerated and concentrated drainage flows have produced various scales of eroded gullies at the bench slopes (Figure 2). On Kalimantan Island, Tertiary sedimentary rocks are widely distributed. The basement to the sedimentary rocks is composed of Paleozoic gneisses and schists of the Subda shield and Mesozoic igneous rocks containing ophiolite. The Pasir coal field is located in the Warukin Formation which occupies large basin structure (Chung et al., 2003). Roto South area is divided into A, B, C (1–4), D (1–2), E (1–2) and F sectors as shown in Figure 1. The predominant rock in this area is mudstone comprising about 80% of the total area. Sandstone comprises from 15% and the rest 5% is coal. Rocks are very soft and compressive strength is very low. Due to this reason, the overall pit angle of about 28◦ is found be suitable for stable slope geometry. The dip angle of the coal seams varies from 75◦ to 85◦ . There are at least 15 different coal seams laid side by side with varying thickness of 1 m to about 25 m. 2.2

Mining method adopted at Roto South area

At the present time, Roto south area is already being mined with Strip Mining system. The proposed plan for coal extraction remains the same with only exception of introducing haul back mining system. It means that the overburden will be placed in the mined out

Figure 1.

Location and plan view of the Pasir coal mine.

Figure 2.

Erosion and failure of slope face by the water.

pit behind the current face of working. The overburden is loosened by drilling and blasting method and then it is removed using shovel/backhoe and 100 tonner dumpers. After the overburden is removed, coal is extracted using backhoe and trucks. The management proposes to initiate haul back mining system from the south end of the Roto South deposit and then advances towards northern side along the strike of the deposit.

3

GIS BASED DRAINAGE ANALYSIS

Although the software packages which have been widely used in the mining industry provide many convenient, they have no function to analyze the surface drainage patterns and to support the rapid decision for water management in open-pit mines where the landform is rapidly changed. Consequently, mining engineers depend on manual methods for drainage analysis, which are tedious, time-consuming, errorprone and highly subjective. It is also very difficult to analyze the surface drainage patterns in complicated configuration of the topography in open pit

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mine (Figure 3). For drainage analysis and rapid decision making, GIS can be used effectively, since most GIS software packages provide tools for hydrological modeling of drainage system and spatial analysis (Garbrecht & Martz, 2000, Burrough et al., 2005, Maidment et al., 2005). Recently in the mining industry some authors started to use GIS to assess the mineral potential (Jusmady, 1999) and the environmental impact of open-pit quarries (Berry & Pistocchi, 2003) and to support stream remediation by prioritizing acid mine drainage (Ayad, 2006). There have been many attempts to develop efficient algorithms for automated extraction of hydrological characteristics of drainage system from Digital Elevation Model (DEM) (Mark, 1984, Jenson & Domingue, 1988, Martz & deJong, 1988, Tribe, 1992; Martz & Garbrecht, 1999, Wang & Liu, 2006). These algorithms, especially the ones developed by Jenson & Domingue (1988), have been implemented by nearly all GIS software packages (Wang & Liu, 2006); consequently, GIS has been used as an effective and convenient tool for drainage analysis in various fields (Garbrecht & Martz, 2000; Burrough et al., 2005; Maidment et al., 2005). Figure 4 shows example of drainage analysis using DEM. Firstly, the flow direction of drainage is calculated from the elevation value at each cell in DEM. The D8 method (Mark, 1984; Jenson & Domingue, 1988), which specifies the steepest descent for each grid in eight possible directions, is generally used in this stage. Based on the flow direction of drainage, accumulated number of all cells into each down-slope is calculated as a flow accumulation of drainage at each cell in DEM. Cells have a flow accumulation value of zero generally correspond to the pattern of ridges (Jenson & Domingue, 1988) and cells have large flow accumulation value correspond to channels or streams. Finally, catchments can be delineated from flow direction of drainage by identifying ridge lines between reverse direction flows of drainage. In case of the example, there are two outlets (the grayshaded cells) in the DEM. Therefore, two catchments are delineated from the flow direction of drainage. For drainage analysis in open-pit mines, it should be noted that many cave-in terrains can be existed in the pit by mining operations, and most of them can be resolved in the DEM created from the detailed topographical survey. Therefore, filling depressions, which is a kind of depression removal technique from DEM by increasing the elevation of cave-in grids to the level of the surrounding grids (Jenson & Domingue, 1988; Wang & Liu, 2006), is required to ensure continuous flow patterns of drainage from upper to lowest parts of pit. The procedures of this study involved a collection of data from a number of thematic maps to construct a spatial database including DEM, coal seam, pond and haul road networks. The data except for DEM were

Figure 3. Complicated configuration of the topography at open pit mine.

(a) Filled DEM

Figure 4.

(b) flow direction

Example of drainage analysis.

available in the mining site as digital maps; therefore data processing was performed to convert the digital maps to the Shape file formats available in ArcGIS 9.1 (GIS software developed by ESRI). For generating a reliable DEM in the spatial database, a detailed topographical survey using a combination of Total Station and Differential Global Positioning System (DGPS) was also performed in the study area. We made a digital topographical map with a contour interval of 2 m from the detailed topographical survey result, and then created a Triangulated Irregular Network (TIN). A high accuracy DEM with 20 m grid spacing was generated from the TIN surface using the TIN to Raster function in ArcGIS. The flow direction, flow accumulation of drainage, catchment areas of ponds in the pit and slope gradient at each cell in DEM are required as factors for

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raster-based spatial analysis to identify the problematic bench slopes and to calculate the priority order for pump allocation. Firstly, we modified most of cave-in cells in the DEM by the filling depressions technique developed by Jenson and Domingue (1988), because we assumed that the small-size cave-in terrains in mining site can be completely filled with rainwater and can show overflowing patterns to ensure the continuous drainage flow during heavy rainfall. Only cave-in cells, which represent ponds in the study area, were filtered by threshold value (Wang & Liu, 2006) and were preserved in DEM, because ponds have roles of outlet as well as storage of rainwater if allocated facilities start to pump the rainwater up. It should be noted that most algorithms for drainage analysis consider the boundary cells of DEM to be outlets if all cave-in grids are modified from DEM by the filling depressions technique. A region of the Roto South that covers about 11.9 km2 (2900 m × 4100 m) was selected as a study area to apply the GIS-based hydrological modeling and spatial analysis. The five artificial ponds at the lowest parts of pit have the roles of storing rainwater. To pump the rainwater, several units of pumping facilities have been allocated at the three ponds (Figure 5). The flow direction and flow accumulation of drainage were analyzed at each cell in the modified DEM (Figure 6). The flow direction shows a mutually symmetric pattern on the west and east side of the coal seam strike inside the pit (Figure 6b). The accumulative amount of drainage is gradually increasing as the drainage flows down from the upper pit to the five ponds and many times the flow paths are concentrated along the haul road networks, especially ramp-ways (Figure 6c). Figure 6.d shows the catchment areas extracted from the simulated flow direction of drainage. Because we focused on five ponds in the pit as suitable places for pump allocation, only catchment areas which correspond to five ponds were considered. The ponds C3, C4 and D1 have relatively

Figure 5. Thematic map of the Roto South in Pasir coal mine.

a) filled DEM

b) flow direction

c) flow accumulation

d) catchment area

Figure 6.

Results of drainage analysis.

large catchment area, therefore most of the drainage could be gathered into these three in-pit ponds during rainfall (Table 1). The slope gradient at each cell in the DEM was calculated by searching the maximum change in elevation over the distance between the cell and its eight neighbors.

4

PREDICTION OF SLOPE TO BE ERODED

The spatial relationship between the flow accumulation of drainage and the slope angle was analyzed to identify the problematic benches due to fast surface erosion (Figure 7). Figure 7.a shows the rated value map derived from flow accumulation of drainage in the study area. The distribution of bench slopes in the study area with 30–45 degrees of gradient. Because the steep bench slopes with more concentrated drainage flow can be approximately assumed to be more vulnerable to slope erosions, we conducted a raster-based map algebra multiplying the rated value from flow accumulation and the rated value from slope gradient to identify the locations having high priority for surface protection. Figure 7b shows the locations of problematic benches. In order to prevent the erosion and failures of bench slopes due to concentrated drainage flow, it is necessary to class benches vulnerable to surface erosion. And if any abnormal symptoms are discovered at these benches through periodic monitoring, the

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a)

b)

Figure 7. Distribution of flow accumulation (a) and erosion possibility on the bench faces (b) at the Roto South. Figure 9. Optimal locations for constructing road safety berm and ditch to divert the concentrated surface runoff.

Figure 8. Suggested monitoring places on the bench faces at the Roto South.

immediate measures must be taken. The optimal locations for monitoring slope stability can be suggested as Figure 8. The number of monitoring locations can be adjusted according to managerial condition in mining site by changing the range assigning the rated value to both the flow accumulation of drainage and the slope gradient. Moreover, having analyzed the spatial relationship between the problematic benches and the concentrated flow paths of drainage, the optimal locations for constructing road safety berm and ditch to divert the concentrated surface runoff also can be suggested for preventing the risk of erosion on the slopes. 5

ARRANGEMENT OF PUMPING FACILITY

To optimize the allocation of pumping facilities in the study area, the priority order was calculated using

the catchment area at each pond and pond area as presented in Table 1. If the pond area is small, the possibility of flooding occurrence can get higher. The amount of incoming water at each pond can be approximately in proportion to its catchment area. We can assume the higher CP-value (i.e. the catchment area divided by the Pond area) has a higher priority order for pump allocation. The pond C4 has the first priority order, since the catchment area which represents the amount of incoming drainage is significantly large pond area. Although in the pond C3-A the largest capacity of pumping facilities is currently allocated, it will not be able to take effect during rainfall. Since the amount of incoming drainage will not be large (Table 1). Considering that pumping facilities are currently allocated at three ponds (C4, C3-A, D1), the arrangement of pumping facilities for pond C3-A is not appropriate. Therefore instead pumping facilities needs to be placed for the ponds C4, C3-B and D1. To make a decision for allocating the units of pumping facilities to ponds, we multiply the Normalized NCP-values by total amount of capacity of available pumping facilities (7,300 m3 /day). Based on the calculated value at each pond, the optimal allocation of pumping facilities could be determined as shown in Table 1. 6

DISCUSSION AND CONCLUSION

For the drainage controls and slope stability, drainage analysis and raster-based spatial analysis were applied to the large scale open pit coal mine in Indonesia, By using the GIS, faster, objective and more reproducible information about hydrological characteristics

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Table 1. Simulated results about suitability of pumping facilities and rise of water-level at the five in-pit ponds. Pumping capacity (m3 /hour)

Pond

Catchment area (m2 )

Pond area (m2 )

CP

NCP

Current

Suggested

C4 C3-A C3-B C1 D1 Total

1,493,913 277,088 1,005,607 259,039 1,182,065 4,396,373

12,582 34,725 14,740 10,377 39,877 112,301

122.5 6.2 84.4 18.0 30.3 261.4

0.47 0.02 0.32 0.07 0.12 1.00

900 3,200 − − 3,200 7,300

900 0 2,300 900 900 7,300

of drainage system could be obtained. And spatial analysis for finding rational solutions could be performed more accurately. Because the raster-based spatial analysis was based on quantitatively rated values, we could define slope stability classes for slope erosions and suggest the optimal arrangement of pump facilities. This indicates that mining engineers can find optimal solutions for drainage controls with alternating the rated values according to the conditions of their mining site. Since we used only two variables (catchment area and pond area) easily acquired through drainage analysis for the pump allocation, a rapid decision making could be possible without both a long-term monitoring and a detailed hydrological investigation which are not quite possible with a fast changing landform. Therefore, the presented approach in this paper can be more applicable in operating open pit mines, because the change of terrains in open pit mines does not allow enough time to perform detailed hydrological investigations based on field survey.

ACKNOWLEDGEMENTS This study was supported by the Research Project of the KIGAM funded by the Ministry of Commerce, Industry and Energy of Korea.

REFERENCES Ayad, Y. 2006. Prioritizing acid mine drainage stream remediation. In: ESRI (Eds.). GIS Best Practices: Mining. http://www.esri.com/library/bestpractices/mining.pdf. Ambramson, L.W., Lee, T.M., Sharma, S. & Boyce, G.M. 2002. slope stability and stabilization methods. John Wiley & Sons, Inc. 712. Berry, P. & Pistocchi, A. 2003. A multicriterial geographical approach for the environmental impact assessment of open-pit quarries. Int. J. of Surface Mining, Reclamation and Environment, Vol.17. 213–226. Burrough, P.A., Karssenberg, D. & van Deursen, W. 2005. Environmental modeling with PC Raster. In: Maguire, D.J.,

Batty, M., Goodchild, M.F. (Eds.). GIS, spatial analysis and modeling. 333–356. Redland: ESRI Press. Chung, S.K., Sunwoo, C., Han, K.C., Shin, H.S. & Park, Y.J. 2000. Stability analysis of open pit slopes in the Pasir coal field, Indonesia. Tunnel & Underground. Vol.10, 430–440 (in Korean with English abstract). Garbrecht, J. & Martz, L.W. 2000. Digital elevation model issues in water resource modeling. In: Maidment, D.R. & Djokic, D. (Eds.). Hydrologic and hydraulic modeling support with geographic information systems. 1–28. Redland: ESRI Press. Hustrulid, W. & Kuchta, M. 1995. Open pit mine planning and design: fundamentals. 636. Rotterdam: Balkema. Jenson, S.K. & Domingue, J.O. 1988. Extracting topographic structure from digital elevation data for geographic information system analysis. Photogrammetric Engineering & Remote Sensing. Vol.54. 1593–1600. Jusmady. 1999. Strategy of geographic information system spatial modeling for mapping mineral potential. Indonesian Mining Journal. Vol.5. 36–46. Maidment, D.R., Robayo, O. & Merwade, V. 2005. Hydrologic modeling. In: Maguire, D.J., Batty, M., Goodchild, M.F. (Eds.). GIS, spatial analysis and modeling. 319–332. Redland: ESRI Press. Mark, D.M. 1984. Automated detection of drainage networks from digital elevation models. Cartographica. Vol.21. 168–177. Martz, L.W. & DeJong, E. 1988. CATCH: A FORTRAN program for measuring catchment area from digital elevation models. Computers & Geosciences. Vol.14. 627–640. Martz, L.W. & Garbrecht, J. 1999. An outlet breaching algorithm for the treatment of closed depressions in a raster DEM. Computers & Geosciences. Vol.25. 835–844. Meek, F.A. 1990. Water and air management. In: Kennedy, B. A. (Eds.). Surface mining. Society for Mining, Metallurgy, and Exploration. Littleton. 819–840. Tribe, A. 1992. Automated recognition of valley lines and drainage networks from grid digital elevation models: a review and a new method. Journal of Hydrology. Vol.139. 263–293. Wang, L. & Liu, H. 2006. An efficient method for identifying and filling surface depressions in digital elevation models for hydrologic analysis and modeling. International Journal of Geographical Information Science. Vol.20. 193–213.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Assessment of regional rainfall-induced landslides using 3S-based hydro-geological model C.H. Tan, C.Y. Ku, S.Y. Chi & Y.H. Chen Sinotech Engineering Consultants, Inc., Taipei, Taiwan, China

L.Y. Fei, J.F. Lee & T.W. Su Central Geological Survey, MOEA, Taipei, Taiwan, China

ABSTRACT: An effective assessment of regional rainfall-induced landslides using 3S-based hydro-geological model was presented to investigate the most common shallow landslide in the Ta-Chia river watershed of central Taiwan. Several typhoon events have indicated that most shallow landslides in the study area occurred as a result of heavy rainfall. To consider the regional rainfall-induced shallow landslides, this study adopted a deterministic approach, the Transient Rainfall Infiltration and Grid-based Slope-stability (TRIGRS) model that couples an infinite-slope stability analysis with a one-dimensional analytical solution for rainfall infiltration. We examine rainfall-induced development of shallow landslides in Ta-Chia river watershed through the following steps. First, the physical, mechanical, geological and hydraulic properties were established using the GIS, GPS, and Remote Sensing framework (3S). Next, the distribution of rainfall intensity were analyzed by performing the Kriging interpolation method. Then, the response of transient pore-water pressure during a rainfall event was analyzed by TRIGRS. Finally, the applicability of model for characterizing shallow landslide susceptibility was addressed. The results demonstrated that good agreement was found between predicted shallow landslide susceptibility and the inventory.

1

INTRODUCTION

Taiwan is located in an active mountain belt created by the oblique collision between the northern Luzon arc and the Asian continental margin in which twothirds of this island is occupied by mountainous areas. The landslides are common natural hazards in this island for a long history. The landslide hazards were insignificant until the Chi-Chi earthquake hit the central Taiwan on September 21, 1999. This earthquake triggered numerous landslides and severely disturbed montane slopes which reduced the shear strength of the slopes, thus setting the conditions for occurrence of more landslides as well as debris flows. Accordingly, frequently landslides were occurred and caused severe property damage and inflicted heavy casualties in the following years especially during several typhoon events with intense rainfall, such as Toraji in 2001, Mindulle in 2004, Airi in 2004, and Hytarng in 2005. This stimulated the interests in studying the rainfall-induced landslide at Ta-Chia river watershed in central Taiwan. Shallow landsliding is the most common landslide type on steep natural hillslopes in the Ta-Chia river watershed. The location of the Ta-Chia river watershed

is shown in Figure 1, where the landslide area indicate the landslides occurred between the 921 Chi-Chi earthquake and the Mindulle typhoon. In this study area, it is found that shallow landslides are often triggered by rainfall as shown in Figure 2. The huge amount of debris sediment from landslides may easily mobilize into destructive debris

Figure 1. Taiwan.

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Location of the Ta-Chia river watershed in

Figure 2. Numerous landslides and debris flows in the study area after the Toraji typhoon (2001).

Figure 3. (a) Destructive debris flow occurred. (b) Hydropower facilities damaged by debris flow during the Mindulle typhoon.

response to rainfall over broad regions. In this paper, we examine rainfall-induced development of shallow landslides in Ta-Chia river watershed of central Taiwan through the following steps. First, the data of physical, mechanical, geological and hydraulic properties of the study area were established using the GIS, GPS, and Remote Sensing framework (3S). Next, the distribution of rainfall intensity were analyzed by the Kriging spatial interpolation method. Then, the response of transient pore-water pressure during a rainfall event was analyzed using TRIGRS model. Finally, the applicability of TRIGRS for characterizing shallow landslide susceptibility in the study area was addressed. Several typhoon events have indicated that most shallow landslides in the study area occurred as a result of heavy rainfall and consequent pore pressure increases in the near subsurface. To consider the rainfall-induced initiation of shallow landslides over a broad region, this study adopted a deterministic approach, the Transient Rainfall Infiltration and Grid-based Slope-stability (TRIGRS) model that couples an infinite-slope stability analysis with a one-dimensional analytical solution for transient pore pressure response to rainfall infiltration (Iverson 2000, Baum et al. 2002, Savage et al. 2003, 2004, Godt 2004, Chen et al. 2005). 2

(a)

30 m

(b)

Figure 4. Dramatic alteration of Ta-Chia riverbed (a) before the Chi-Chi earthquake (b) after the Mindulle typhoon.

flows as shown in Figures 3 and 4. The rainfallinduced landslides may be caused by the infiltration of rainfall which leads to increases in pore water pressures in the near subsurface that reduces the shear strength of the colluvial mass. A variety of approaches have been used to estimate the hazard from shallow, rainfall-triggered landslides, such as empirical rainfall threshold methods or probabilistic methods based on historical records. The multivariate statistical analysis and artificial neural network analysis are also often used to evaluate the landslide susceptibility in regional scale. In most cases of studying regional landslide susceptibility, the Geographic Information Systems (GIS) is common adopted for slope stability. TRIGRS model can be used to investigate both the timing and location of shallow landslides in

STUDY AREA

The Ta-Chia River watershed with a watershed area of 1,236 km2 is located in central western Taiwan. The elevation of highest mountain in the watershed is around 3,875 m. The river stretch extends 124 km from upstream to the sea. This river valley is notable because it incorporates the Central Cross Island road that links the east and west coasts of Taiwan across the Central Mountains. This very important infrastructure was constructed in the early 1960s, and was followed by a series of significant hydroelectric schemes that extend along the length of the river, consisting of one high, concrete arch dam at Techi, and a series of dams and hydropower stations. 2.1

Geological setting

The study area consists of a steeply incised valley orientated approximately east-west in the eastern part of the Central Range. The geology consists of a series of interbedded Tertiary sandstones and slates, with occasional limestone bands. The geologic strata of Ta-Chia river are argillite, slate, quartzite, sandstone, siltstone and shale including Lushan Formation, Tatungshan Formation, Kankou Formation, Chiayang Formation, Szeleng Sandstone, Tachien Sandstone, Kuohsing Formation, Guandaoshan Formation, Jinshuei Formation

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and Jhuolan Formation. The rock mass is extensively tectonically disrupted, with a high density of fractures and joints. The slope inclination in the watershed is range from 40 to 80 degrees.

×

∞

m=1

ierfc

+ ierfc 2.2

Climatic setting

The rainfall of Ta-Chia river watershed is abundant. There are more than 20 rainfall gauging stations installed in this area, but few of them were located at the upstream mountainous area. With the difference in location, the range of annual average rainfall is 3000–3500 mm in the upstream area, 2000–2500 mm in the midstream area, and 1500–2000 mm in the downstream, respectively. The long-term statistical data show that the main rainfall usually concentrates between May and September which was accounted for 75 percent of annual accumulated rainfall. Furthermore, the typhoons hit Taiwan frequently during summer season (June-August) when the daily accumulated rainfall could exceed more than 500 mm.

3

THEORETICAL BASIS OF TRIGRS MODEL

This paper uses the Transient Rainfall Infiltration and Grid-based Regional Slope-Stability (TRIGRS) model (Baum, 2002) to estimate the regional watershed landslide susceptibility. The TRIGRS model is based on the Iverson’s research results (Iverson, 2000) to assess the time-varying slope safety of each slope unit with transient pore-water pressure during a rainfall event. First, We divide the watershed area into discrete grids. The initial water table, hydrogeological properties (e.g. infiltration rate, hydraulic conductivity and hydraulic diffusivity and mechanical parameters) and rainfall intensities for each grid were assigned by values according to its characteristics in space and time. Next, the transient pore-water pressure can be obtained by solving the unsaturated flow equation (i.e. Richards’ equation). Then, the limit equilibrium method is used to estimate the safety of slope with the transient pore-water pressure and the hydro-geological properties. In the TRIGRS model, the calculation of the transient pore-water pressure distributed with a finite depth for the case of an impermeable boundary could be described as following formula:

ϕ(Z, t) = [Z − dz ]β

n 1 InZ +2 H (t − tn )[D1 (t − tn )] 2 K n=1 Z

−2 ×



n InZ

KZ

n=1

∞

m=1

(2m − 1)dLZ − (dLZ − Z) 1

2[D1 (t − tn )] 2

(2m − 1)dLZ + (dLZ − Z) 1

2[D1 (t − tn )] 2



 1

H (t − tn+1 )[D1 (t − tn+1 )] 2

ierfc

+ ierfc







(2m − 1)dLZ − (dLZ − Z) 1

2[D1 (t − tn )] 2

(2m − 1)dLZ + (dLZ − Z) 1

2[D1 (t − tn )] 2



 (1)

where ϕ is the groundwater pressure head, t is the elapsed time, Z = z/ cos α is the vertical coordinate direction, z is the slope-normal coordinate direction, and α is the slope angle; dZ is the steady-state depth of the water table measured in the Z direction, β = λ cos α, λ = cos α − (IZ /KZ )LT , KZ is the hydraulic conductivity in the Z direction, IZ is the steady surface flux, and InZ is the surface flux of a given intensity for the nth time interval and dLZ is the soil layer depth measured in the Z direction. D1 = D0 cos2 α, where D0 is the saturated hydraulic diffusivity, N is the total number of time intervals, and H (t−tn ) is the Heavyside step function. The function ierfc (η) is the complementary error function. The TRIGRS model imposes the physical limitation that pore-water pressure cannot exceed that which would result from having the water table at the ground surface during rainfall events, that is: ϕ(Z, t) ≤ Zβ

(2)

The infinite slope analysis is based on the limit equilibrium method, in which slope angles, unit weights of soil and water, shear strength parameters, and transient pore-water pressures are combined to estimate the safety of each grid as following equation:

Fs =

c − ϕ(Z, t)γw tan φ tan φ + tan α γs Z sin α cos α

(3)

where c and φ are cohesion and friction angle of weathered soil, γs and γw are unit weights of soil and water respectively. ϕ (Z, t) is the pore-water pressure calculated by equation 1 and limited by equation 2.

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4

APPLICATION OF 3S TECHNIQUES

Before we apply the TRIGRS model to analyze the regional rainfall-induced landslide susceptibility for the study area, the mechanical and hydro-geological parameters and their distribution in space and time must be established in advance. In order to apply the TRIGRS model for modeling the regional scale problem, the currently wide-implemented Geographic Information System (GIS) was considered for the preparation of the data input as well as the result displaying. A GIS could combine graphical features with tabular property data and could perform extensive spatial and statistical analyses. The results of analyses could also be displayed directly and visually. So we selected the ArcGIS, which was developed by ESRI, as a work platform for digitizing, storing, interpolating, overlaying and displaying our input data for the TRIGRS model. First, we use the Global Positioning System (GPS) to link the results of geological surveys, field tests, and topography surveys together by the three dimensional coordinates. Next, ArcGIS is used to establish specific layers involved geographical information (i.e., surface elevations, slope gradients and aspects), geological information (i.e., geological zones, formations and structures), hydrological information (i.e., spatial distribution of groundwater elevation and rainfall intensity) and geotechnical information (i.e., physical, mechanical and hydraulic properties). Then, we use the Kriging method for spatial analysis in ArcGIS to interpolate the above-mentioned information and export the results with the ASCII format to TRIGRS. Finally, the high-resolution RS imageries including the satellite images and aerial photos are adopted to verify the model results by comparing the predicted landslide locations and the inventory from RS imageries. The input-parameter layers created by ArcGIS are introduced as follows. 4.1 Slope angle The terrain of Ta-Chia river watershed is very steep and its elevation drops rapidly within a short distance. According to the digital terrain model (DTM) with 40 m × 40 m scale, the range of the slope angle is from 28 to 45 degrees, and the maximum slope is over 75 degree (see Figure 5). 4.2 Physical, mechanical and hydrologic properties The geological units in the Ta-Chia river watershed include Lushan Formation, Tatungshan Formation, Kankou Formation, Chiayang Formation, Szeleng Sandstone, Tachien Sandstone. Figure 6 is

Figure 5.

Slope angles in the study area.

Figure 6. Spatial distribution of geological formations along the midstream of Ta-Chia river.

the 1:250,000 scale geological map of the study area that shows the plane distribution of above formations. The adopted physical, mechanical and hydraulic properties for each zone are summed up in Table 1. We assume the physical properties of weathered layers is strongly related to properties of their fresh intact rock. Based on the assumption, the initial values of parameters are assigned to each geological zone. 4.3 Soil depth Although the weathered soil thickness is related to many effect factors including the vegetation cover, the underlying lithology, the climate, the angle and curvature of slope, the land use and so on, it is more convenient to simply assume that the soil thickness decreases with the increasing slope angle for the engineering objectives. We assume that there is a function relationship between the soil thickness and the slope angle. Figure 7(a) shows the correlation between soil depth and slope angle from our field surveys and

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Table 1. Physical, mechanical and hydraulic parameters of the geological units in the study area. Formation

c (KPa)

φ (◦ )

γt (KN/m3 )

Ks (10−6 m/s)

D0 (10−4 m2 /s)

Iz (10−8 m/s)

Kankou Lushan Sicun Szeleng Tatungshan

16–21 18–23 16–22 13–19 19–24

26 29 30 28 28

22 22 21 21 21

5 10 7 100 10

10 20 14 200 20

1 10 5 50 10

10 9 8

Soil Thickness(m)

7 6 5

H = 10.911e 2 R = 0.885

4

-0.0515 a

3 2 1 0 0

10

20

30

40

50

60

70

80

90

Slope angle(˚)

Figure 7(a). Relationship between weathered soil thickness and slope angle.

Figure 8. Spatial distribution of initial groundwater table in the midstream of Ta-Chia river.

mountainous areas are rare. This study collects the observed data of groundwater table from other engineering cases in the southern mountain areas of Taiwan and proposes the correlation between groundwater level (hw ) and surface elevation (h) as follows: hw = 0.9672h − 4.781

(4)

The spatial distribution of initial (steady-state) groundwater table in the study area from equation 4 is as shown in Figure 8. Figure 7(b). Weathered soil thickness in the study area.

4.5 Rainfall intensity

the other researches related to this issue (Delmonaco et al., 2003; Salciarini et al., 2006). The correlation is used to determine the distribution of soil thickness in the study area (Figure 7(b)). 4.4

Initial groundwater table

There has been few exploitation and development in the mountainous areas of Taiwan. Therefore, the available data related to field surveys and tests in the

The rainfall intensity data during the Toraji typhoon (July 2001) are used to investigate the effects of rainfall-infiltration on the slope susceptibility in the Ta-Chia river watershed. There are more than 20 rainfall stations in the watershed. The difference of the rainfall intensity and duration recorded at above stations obviously depends on the location of typhoon center at that time. The spatial distribution of rainfall intensity in the watershed also varies with time. Figure 9 shows the distribution of rainfall intensity at the moment of the peak rainfall intensity during the Toraji typhoon. Therefore, the difference in spatial and

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(a)

(b )

(c)

(d)

(e)

(f)

Figure 9. The spatial distribution of rainfall intensity at the moment of the peak rainfall intensity during the Toraji typhoon.

5

ASSESSMENT OF LANDSLIDE SUSCEPTIBILITY

Figure 10. Spatial distribution of landslide in the study area during the Toraji typhoon. (a) before the typhoon, (b) at the 1st hour, (c) at the 2nd hour, (d) at the 3rd hour, (e) at the 6th hour and (f) at the 12th hour. 4.0

100

3.5 80

Stable Slope Unit

3.0

Safety Factor

The results of landslide susceptibility estimated by the TRIGRS model are illustrated in Figure 10. Figure 10(a) represents the modeling results for the case of the steady-state groundwater table and shows that the safety factor of all slope units in the study area were almost greater than 1.0 before the Toraji typhoon. Figure 10(b)−(f) display the spatial distributions of landslide susceptibility at the 1st, 2nd, 3rd, 6th and 12th hour respectively. The results show that the landslide area spreads with the increase of rainfall intensity and the landslide locations develop following with the typhoon route. Figure 11 represents the safety factor of two slope units near the Shangkukuan rainfall station varying with time during the typhoon event. One is gradually to slide (FS < 1.0) and the other is safe (FS > 1.0), which were analyzed by TRIGRS model based on the recorded rainfall data of Shangkukuan station. The results also show that the failure slope unit begins to slide at the 10th hour, and the safety factor of the stable one is decreasing with time and reach the minimum value at the 14th hour (i.e., still greater than 1.0). The safety factor has no variation after 14 hours because the transient groundwater table has been raised to the ground surface. The multi-temporal RS imageries before and after the typhoon Toraji were used for landslide interpretation. The newly landslide area caused by the Toraji typhoon is 10,463,400 m2 . The predicted landslide area by the TRIGRS model is 8,789,830 m2 ,

60

2.5 2.0

40

1.5

Rainfall (mm)

time distribution must be considered in the analysis of regional rainfall-induced landslide susceptibility.

Unstable Slope Unit 20

1.0 0.5

0 0

5

10

15

20

25

Duration of Rainfall (hours)

Figure 11. Respective history of safety factor for the stable unit and the unstable one with the rainfall during typhoon Toraji.

which are less than the landslide area interpreted from RS imageries. The difference is mainly due to the underestimation of initial groundwater table or the overestimation of shear strength of weathered soil layer. Farther analyses may be needed to find the proper input values through the calibration process using more rainfall-triggered landslide events.

6

CONCLUSIONS

An effective assessment of regional rainfall-induced landslides using 3S-based hydro-geological model

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was proposed to investigate the most common shallow landslide in Ta-Chia river watershed of central Taiwan. This study depicts that the TRIGRS model is generally useful for the assessments of slope stability over regional scale. Although TRIGRS can accommodate spatially varying soil strength and hydraulic properties, it is often a paucity of the physical properties or the input data may vary significantly over a typical study area. Since detailed investigation of the physical properties for a regional scale problem is usually impractical, reasonable assumptions are necessary to made regarding input values. In our case, parameter calibration plays a crucial role for the accuracy of the predicted results. The use of remote sensing data such as multi-temporal satellite imaginary or aerial photographs can provide a useful solution for estimating the input data through the calibration process. Nevertheless, our preliminary results using estimated parameters appear to be useful for shallow landslide hazard assessments in the study area. REFERENCES

Chen, C.Y., Chen, T.C., Yu, F.C. & Lin, S.C. 2005. Analysis of time-varying rainfall infiltration induced landslide, Environmental Geology. 48: 466–479. Delmonaco, G., Leoni, G., Margottini, C., Puglisi, C. & Spizzichino, D. 2003. Large scale debris-flow hazard assessment: a geotechnical approach and GIS modeling. Natural Hazards and Earth System Sciences. 3: 443–455. Godt, J.W. 2004. Observed and modeled conditions for shallow landsliding in the Seattle, Washington area. Ph.D. dissertation, University of Colorado. Iverson, R.M. 2000. Landslide triggering by rain infiltration. Water Resour. Res. 36 (7): 1897–1910. Salciarini, D. et al. 2006. Modeling regional initiation of rainfall-induced shallow landslides in the eastern Umbria Region of central Italy. Landslides 3: 181–194. Savage, W.Z., Godt, J.W. & Baum, R.L. 2003. A model for spatially and temporally distributed shallow landslide initiation by rainfall infiltration. Proceedings of 3rd international conference on debris flow hazards mitigation: mechanics, prediction, and assessment: 179–187, 10–12 September 2003. Davos, Switzerland. Savage, W.Z., Godt, J.W. & Baum, R.L. 2004. Modeling timedependent slope stability. Proceedings of 9th international symposium on landslide: 23–28, 28 June–2 July 2004. Rio de Janeiro, Brazil.

Baum, R.L., Savage, W.Z. & Godt, J.W. 2002. TRIGRS— A fortran program for transient rainfall infiltration and grid-based regional slope-stability analysis. U.S. Geological Survey Open-File Report 02-0424.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Investigation of a landslide along a natural gas pipeline (Karacabey-Turkey) T. Topal & M. Akin Department of Geological Engineering, Middle East Technical University (METU), Ankara, Turkey

ABSTRACT: A new natural gas pipeline is under construction to provide the gas interconnection between Turkey and Greece, as well as to create gas ring for southern Europe. The new pipeline 0.90 m in diameter will be constructed next to an existing small diameter pipeline. However, the existing pipeline was broken by a landslide occurred in February 2006 near Karacabey (Bursa). In order to investigate the causes of the landslide and suggest possible remedial measures, geological and geotechnical investigations including surface geological mapping, trial pitting, drilling with field tests, laboratory testing, and limit equilibrium analysis were carried out. Based on the gathered data, possible remedial measures are suggested to prevent the reactivation of the landslide.

1 1.1

GENERAL INFORMATION Introduction

A new natural gas pipeline with a diameter of 0.90 m is under construction to provide gas interconnection between Turkey and Greece, as well as to create gas ring for southern Europe (Figure 1). The pipeline system will carry 750 million m3 natural gas at the first stage, and it will rise up to 11 billion m3 in 2012. The new pipeline route is next to an existing small diameter pipeline near Karacabey (Bursa). However, the existing pipeline was broken by a landslide occurred on February 20, 2006 in Keslik (Karacabey) area (Figure 2). The landslide has a length of 96 m and width of 48 m. Although the existing pipeline was temporarily repaired, it has been deformed by the landslide. The new pipeline system is totally completed except the problematic landslide zone, and the system should start to work within a very short period of time. Therefore, either the pipeline route should be relocated or the landslide should be stabilized urgently due to the fact that the other parts of the pipeline have already been completed and the system should work immediately after solving the landslide problem. In this study, a geological survey is carried out in the close vicinity of the landslide for site appraisal related to other existing landslides and possible relocation of the pipeline route if feasible. In order to investigate the causes of the landslide and suggest possible remedial measures, geotechnical investigations including geological mapping, trial pitting, drilling with field tests, inclinometer measurements, laboratory testing, and limit equilibrium analysis were carried out. Based on obtained data, possible alternatives on remedial measures are suggested.

For this purpose, firstly field geological mapping was performed. In the landslide area, eight trial pits and eight boreholes were opened. At each meter, standard penetration test (SPT) was performed, and undisturbed samples with Shelby tubes were taken. Constant head permeability tests corresponding to 3–7 different levels of each borehole were carried out in all boreholes. Inclinometer measurements were taken in six boreholes. Sieve and hydrometer analysis, Atterberg limits, soil classification on disturbed samples, and unit weight, water content, sieve and hydrometer analysis, soil classification on undisturbed samples were performed. Additionally, consolidated drained (CD) triaxial tests were carried out on undisturbed samples taken from a level corresponding to landslide failure surface. Borehole and trial pit logs, results

Figure 1. The route of the new natural gas pipeline (modified from www.tbys.org and Google Earth).

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Figure 2. Location map of the study area.

Figure 4. A view of the southern part of the site where several landslides exists.

Figure 3. Geological map of the study area and its close vicinity (modified from Ergül et al., 1980).

of laboratory tests and back analysis were all utilized to perform stability analysis and to suggest remedial measures for solving the landslide problem.

1.2 Site geology The landslide occurred in the study area is located in Neogene deposits (Ergül et al., 1980) (Figure 3). This unit covers a very large area in the region, and consists mainly of conglomerate, sandstone, siltstone, claystone, marl, clayey limestone, limestone, tuff, and volcanics. The unit with the age of Miocene-Pliocene, dominantly contains clayey limestone. This unit can only be observed at road cuts and valleys because the areas where this unit is exposed are generally covered with vegetation. At the bottom of the unit,

basal conglomerate exists. Above this level, semiconsolidated sandstone can be seen. The sandstone grades to claystone, clayey limestone and limestone with intercalation. At the top, tuff, agglomerate and andesite occasionally crop out. As a bedrock, the claystone of the Neogene deposits can be observed in the landslide area and its close vicinity. Above this unit, soft-firm clay with little amount of sand and gravel exists. This soft-firm clay has a vertical thickness of 28 m and it is overlain by 1 m of silicified claystone and 0.5–1 m of top soil at the upper part of the slope. The geological survey performed in the study area shows that there are other landslides already occurred in close vicinity of the pipeline (Figure 4). Soft-firm clayey level slides above the claystone. These landslides have a curved surface at crown. The movement type is rotational at the upper part of the landslide body, and becomes translational where the failure surface reaches the claystone level. The toe parts of the landslides are near a stream. Therefore, the stream undercutting seems to play an important role for triggering the landslides. The landslides formed near the stream enlarge backward creating retrogressive movements. Based on the field observations, relocation of the pipeline is not considered to be feasible due to the existence of several other landslides in the region. 2

SITE INVESTIGATION

2.1 Trial pits and boreholes The site investigation of the landslide which damaged the pipeline includes opening of eight trial pits, eight boreholes, sampling during SPT tests and using Shelby

1648

Figure 5. The location of the trial pits and the boreholes.

tube, constant head permeability tests, inclinometer measurements, and laboratory tests. The trial pits were excavated using back-hoe down to 2.10–6.50 m (Figure 5). The trial pitting was ended wherever the bedrock (claystone) was reached because it was very difficult to excavate the claystone with the back-hoe. The bedrock was observed in all of the trial pits, except two of them (TP-4 and TP-6). In one of the pits (TP-2), small volume of groundwater inflow was observed at the contact between soft-firm clay and claystone. In order to identify the failure surface, the boreholes (Figure 5) were drilled down to 10–15 m based on the field survey and trial pit data. In each borehole, SPT tests were performed at every 1 m interval and followed by undisturbed sampling. Disturbed samples were taken during the tests. Undisturbed samples using thin wall Shelby tubes were taken at every 1 m. Thus, continuous sampling for each borehole was achieved. Static groundwater level was recorded 48 hours after the completion of each borehole. Evaluation of the borehole data reveals that softfirm clay (landslide material) at the zone of accumulation of the landslide where the old pipeline was damaged, has a thickness of approximately 1.90– 12.40 m. SPT-N values of this material vary from 6 to 22. However, the clay with organic material observed near the stream has SPT-N value as low as 1. The clay is underlain by claystone with very high SPT-N value of more than 50 or refusal.

Groundwater is observed only in BH-7 and BH-8 at depths of 2.90 m and 2.60 m, respectively. At five to seven different levels of all boreholes, a total of 45 constant head permeability tests were performed in accordance with BSI (1999). The test results reveal that hydraulic conductivity values of the clay generally range from 10−4 to 10−7 m/sec. Although both soft-firm clay and claystone have low hydraulic conductivity values, one can find relatively high permeability values where gravel and sand contents of both materials increase. Therefore, the clayey zones are practically impervious whereas sandy and gravelly clay is slightly pervious. Since the distributions of sandy and gravelly levels are heterogeneous, long lasting rainfall and snow melt may thus infiltrate the ground as validated in TP-2 with groundwater inflow at the contact between clay and claystone. In six boreholes (BH-1, 2, 3, 4, 5, and 6), a total of 60 m inclinometer casing were installed with grooves oriented in such a way that the measurements can be taken both parallel and perpendicular to the direction of landslide movement in order to detect the failure surface of the landslide. The inclinometer measurements were taken twice after installing the casing due to a very limited time schedule of the project. The second inclinometer measurements indicate that there exist very small movements at the contact between soft-firm clay and claystone. On the basis of the field survey, trial pit and borehole data, and inclinometer measurements, it is concluded that the failure surface

1649

of the landslide is located at the boundary between the soft-firm clay and the claystone, and it is in the form of non-circular slide. 2.2 Laboratory tests The laboratory test results indicate that the landslide material is mainly clay with high plasticity (CH) according to unified soil classification of ASTM (1992). At some levels, gravelly clay, low plasticity clay, sand clay, clayey sand, silt with low plasticity and silty gravel also exist. The liquid and plastic limits of the landslide material are 30–74 and 19–33, respectively. The water content of the samples ranges between 11% and 47% and the clayey zones have higher water contents. The average unit weight of the soft-firm clay above water table is 19.05 kN/m3 and its saturated unit weight is 20.20 kN/m3 . However, the claystone has slightly higher unit weight (20 kN/m3 for dry and 21 kN/m3 for saturated samples). The uniaxial compressive strength of the claystone is 110–550 kPa. Consolidated drained (CD) triaxial tests performed on two undisturbed samples of the landslide material near the failure surface indicated that the residual effective cohesion (c′ ) and effective internal friction angle (φ ′ ) of the clay are 18 kPa and 8◦ , respectively.

3

ANALYSIS OF THE LANDSLIDE

The site investigation carried out in this study shows that the landslide is shallow one with failure surface located 1.90–12.40 m below the ground level. Therefore, it is decided to take remedial measures for the landslide rather than re-routing of the pipeline. For the analysis of the landslide, firstly, back analysis was carried out to assess reliable shear strength parameters of the landslide material. The shear strength parameters obtained from CD test and back analysis were compared, and suitable shear strength values were selected for the analysis. These values were then used to perform long-term stability analyses of the landslide. 3.1

determined for different shear strength pairs. The analysis was carried out for non-circular slide, using Janbu method by means of SLIDE (4.0) software of Rocscience (2000). The results obtained from the analysis reveal that shear strength parameters (c′ and φ ′ ) of the clay at the time of failure were 0.15 kPa and 9◦ , respectively (Figure 6). Although φ ′ values of the clay obtained from the back analysis and triaxial test results are identical, significantly higher c′ value is obtained from the laboratory. Considering the fact that the landslide material is moving very slowly, the residual cohesion (c′ ) value of the clay should be very low or zero. For this reason, the authors of this study consider that the shear strength parameters of the clay obtained from the back analysis are more realistic. 3.2

Stability analysis of the landslide

Long term stability analysis of the landslide (Table 1) was carried out using new topographical map of area after the landslide. The study area is located within first degree earthquake zone of Turkey. Based on a recent study performed by Erdik et al. (2005) in the close vicinity of the region, the maximum horizontal ground acceleration is expected to be 0.45 g. Seismic acceleration coefficient of 0.2 g which is in the range of 1/3–1/2 PGA (Peak Ground Acceleration) of the region as suggested by Marcuson and Franklin (1983) is employed for the analysis. Additionally, a pore pressure ratio of 0.2 was selected to incorporate long-term groundwater effect. The slope stability analysis indicates that the long-term factor of safety of the landslide

Back analysis of the landslide

Topographic map of the landslide area at a scale of 1/100 prior to its movement was used for assessing the shear strength parameters governing the landslide activity. Back analysis of the landslide (Sancio, 1981, Chandler, 1977; Turner and Schuster, 1996; Teoman et al., 2004) was performed using three parallel profiles in the direction of landslide movement. For the analysis, variation of the shear strength parameters (c′ and φ ′ ) of the clay satisfying factor of safety (FS) of 1 corresponding to limit equilibrium condition was

Figure 6. Variation of shear strength parameters of the clay layer obtained from the back analysis along three sections. Table 1.

Data used in the stability analysis.

Material

γnatural (kN/m3 )

γsaturated (kN/m3 )

c (kPa)

φ (◦ )

Clay Claystone

19.05 20

20.20 21

0.15 43

9 20

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Figure 7. Stability analysis of the landslide.

is 0.67 (Figure 7) and that’s why remedial measures should be taken into account.

3.3

Remedial measures

In the literature, there are various methods of remedial measures that can be adopted to overcome landslide problem. These methods include re-routing, unloading, flattening, buttressing, surface and subsurface drainage, reinforcement, retaining walls, vegetation, surface slope protection, soil hardening, thermal treatment, bridging, etc. (Bromhead, 1992, Turner and Schuster, 1996, Abramson et al., 2001, Cevik and Topal, 2004, Ontigao and Sayao, 2004, Ducan and Wright, 2005, Cornforth, 2005). Since re-routing (avoiding the problem) is not an alternative in our case, stabilization methods need to be considered. In this study, partial removal of landslide material, construction of toe buttress, slope flattening, surface drainage, and lowering the pipeline are considered as remedial measures.

4

CONCLUSIONS AND RECOMMENDATIONS

In this study, geotechnical aspects of a landslide along a natural gas pipeline near Karacabey (Bursa-Turkey) were investigated. The field studies revealed that there exist other landslides in the close vicinity of the landslide. Therefore, the re-routing of the pipeline is not preferred. The landslide damaged the existing pipeline has non-circular failure surface mainly following the boundary between clay and claystone. The failure surface is located at a depth of 1.90–12.40 m. Based on the long term analysis of the landslide, the remedial measures including partial removal of the landslide material, construction of toe buttress, slope flattening, surface drainage and lowering the pipeline are suggested with design guidelines to prevent the reactivation of the landslide. REFERENCES Abramson, L.W., Lee, T.S., Sharma, S. & Boyce, G.M. 2001. Slope stability and stabilization methods. 2e , Wiley, 736 pp.

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ASTM. 1992. Classification of soils for engineering purposes. D 2487, Annual Book of ASTM Standards, American Society for Testing and Materials, 325–335. Bromhead, E.N. 1992. The stability of slopes. 2e , Blackie, 411 pp. BSI, 1999. Code of practice for site investigations. British Standards Institution, 192 pp. Cevik, E. & Topal, T. 2004. Relocation of a problematic segment of a natural gas pipeline using GIS-based landslide susceptibility mapping, Hendek (Turkey). In: Hack R., Azzam R., and Charlier R. (eds.) Proceedings of the 1st European Regional IAEG Conference on Engineering Geology for Infrastructure Planning in Europe: A European Perspective, (Springer), 265–274. Chandler, R.J. 1977. Back analysis techniques for slope stabilization works: a case record. Geotechnique, 27 (4): 479–495. Cornforth, D.H. 2005. Landslides in practice, Wiley, 596 pp. Duncan, J.M. & Wright, S.G. 2005. Soil strength and slope stability. Wiley, 297 pp. Erdik, M., Sesetyan, K., Demircio˘glu, M.B. & Durukal, E. 2005. Assessment of Earthquake Hazard for Bakirköy, Gemlik, Bandurma, Tekirda˘g and Körfez. Prime Ministry Project Implementation Unit, Marmara Earthquake

Emergency Reconstruction (MEER) Project: A3 Component, Ankara, 61 pp. Ergül, E., Öztürk, Z., Akcören, F. & Gözler, M.Z. 1980. Balıkesir ili-Marmara Denizi arasının jeolojisi. MTA Rapor No: 6760, 57 s. Marcuson, W.F. & Franklin, A.G. 1983. Analysis and remedial measures to improve the stability of existing dams, Seismic Design of Embankments and Caverns, T.R. Howard (ed.), New York, ASCE. Ontigao, J.A.R. & Sayao, A.S.F.J. 2004. Handbook of slope stabilization. Springer, 478 pp. Rocscience. 2000. SLIDE 4.0-2D slope stability analysis for soil and rock slopes, Rocscience Inc., Canada. Sancio, R.T. 1981. The use of back-calculations to obtain shear and tensile strength of weathered rocks. Proc. Int. Symp. on Weak Rock, Tokyo, pp. 647–652. Teoman, M.B., Topal, T. & Isik, N.S. 2004. Assessment of slope stability in Ankara clay: A case study along E90 highway. Environmental Geology, 45 (7): 963–977. Turner, A.K. & Schuster, R.L. 1996. Landslides-investigation and mitigation. Transportation Research Board, National Research Council, Special Report 247, National Academy Press, Washington C.D., 673 pp.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Influence of extreme rainfall on the stability of spoil heaps I. Vanicek & S. Chamra Geotechnical Department, Czech Technical University, Prague, Czech

ABSTRACT: In the north part of the Czech Republic there are open pit mines of brown coal, stratum of which is overlaid by clayey tertiary deposits. Each year a huge amount of these clayey materials are excavated (about 200 mil. m3 ) and deposited in the form of spoil heaps, height of which is reaching 150 m. Because of such large affected area the new construction on these spoil heaps is a necessity. Nevertheless the stability of these spoil heaps is the first condition for such a decision. Therefore the problem of stability, especially from the view of extreme rainfalls is investigated with the help of laboratory and numerical models, to be able to define long term stability even for extreme hydrological conditions.

1

INTRODUCTION

Human being is significantly changing existing environment in particular during the last century. On one side it is the result of excavation on the surface e.g. open pit mines, quarries, borrow pits, but also on the other side the deposition on the surface, in particular different waste material, Sembenelli & Ueshita (1981). Our attention is focused on large volume waste material as mining waste, waste after burning of coal, industrial and building wastes as well as excavated soil coming from different underground structures such as construction of metro systems, urban tunnels, excavation for deep foundations and the like. Environmental geotechnics, environmental geology, engineering geology are solving many problems connected with these different waste materials, their storage in the form of spoil heaps, tailing dams, sanitary landfills. First of all it is stability and deformation of these new artificial earth structures that have to be solved as well as their influence on our environment, especially from the view point of different leachates. In some regions these new earth structures situated on the earth surface occupy large areas and by significant extent affecting the landscape pattern. This new geological structures created by human being can be indicated as new geological period—post quaternary period, or even ‘‘quinternary period’’. In these areas strongly affected by human activity the utilization of their surfaces is nearly necessity. Therefore the stability of these new structures is the priority question. For the Czech Republic spoil heaps composed from tertiary clayey deposits which overlay seams of brown coal in the northern part of the country are typical example of such situation. Therefore our attention is further devoted to this specificity, firstly on the

stability as a function of time, but counting also with extreme hydrological situation, namely with extreme rain falls. The problem is complicated by the fact that the properties of deposited clayey material is changing with time. 2

PROPERTIES OF SPOIL HEAP MATERIAL

Roughly 200 mil. m3 per year of clayey material overlaying brown coal are deposited into spoil heaps in the Czech Republic even when volume of brown coal is decreasing. Stripping ratio—volume of clay to the volume of coal is steadily increasing with time, now reaching in average a value of 6:1. If in average 40 m of material is stored on 1 m2 , after that each year spoil heaps are covering the area of 5 km2 . It means that in this region large part of territory is affected not only by mining activity but also by spoil heaps construction. The thickness of the tertiary sediments reaches 150 m in average in the North-Bohemian Brown-Coal Basin. Besides of brown coal this layer is composed mainly of clays and of claystones. But sand, underclay, slate coal, sandy clay and sandy claystone are also present. Clays and claystones have the following characteristics: – Plasticity limit wp = 30–35% – Liquidity limit wL = 70–80% – Plasticity index Ip = 35–50

From clay minerals mainly the kaolinite and illite are in abundance, montmorillonite is also present but with variable low percentage. The proportion of clay particles is ranging from 10 to 40% and silt particles from 20 to 60%, Vanicek & Vanicek (2008).

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From the place of excavation to the place of deposition into spoil heap body the clayey material is transported by train or by belt, recently preferably by belt. Before the belt transport (width of belt roughly 2–3 m) the excavated great clods are in some cases crushed down. During this transport the individual clods are rounded off and during wet weather their moisture content is increased. At the end of the transport the individual clods are partly compacted by free fall (for overburden conveyor bridge it is up to 20 m), see Figures 1–2. By free fall individual clods are partly crushed and compacted but their bulk density is approximately 1500–1600 kg · m−3 and so the macro porosity is around 30%. The individual macro pores between the individual clods are interconnected and air is in continuous form and so the permeability of soil for air is relatively high. Character of the fill is close to rock fill—see Figure 3. Under this condition an air pore pressure is relatively quickly equalized to the atmospheric pressure.

Figure 3.

Initial character of clay clods.

On the contrary the pore water pressure inside of individual clods is negative due to great unloading. Under this condition such soil will easily absorb water—free water or water from saturated air —even from air inside of the spoil heap’s body. In the extreme conditions the character of clay clods is changing up to the soft clay. The orientation on this type of spoil heap is not only the result of huge volume of such new earth structures but also due to the fact that the change of properties is extreme—the properties can change from one side of soil mechanics spectrum, from properties of rockfill, up to the other side of this spectrum, up to the soft clay. On the spoil heap surface the character of clay clods after significant unloading can lead to the shape shown in Figure 4. But generally the properties of the deposited clayey soils are changing with time due to two basic contradiction aspects, Vanicek (1995):

Figure 1. General view on open pit mine.

– Process of softening as a result of weathering, moisture content increase and kneading, – Process of hardening as a result of surcharge by new deposited layers.

Figure 2. Transport of clayey clods and spoil heap filling.

The result of these contradictory processes is a significant heterogeneity of the deposited material. These two processes are schematically described in Figure 5.

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Velocity of mass movement was maximally few meters per day. Therefore it was possible to stop the movement in one case by huge loading berm with volume of 0.75 mil. m3 composed from coarse material at the moment when central part of landslide was in average inclination 1:18. Therefore the inclination can vary in extremely large range, from roughly 1:1 (1:2) for newly deposited clay clods if the total height is not exceeding 20 m up to above mentioned 1:18. Mildest inclination is associated with old slip surfaces in the spoil heap body, when the residual angle of internal friction φr′ is about 6–8◦ . Effective angle of internal friction φ ′ is most often in the range of 13–17◦ . The progressive decrease of slope inclination is connected with landslides which can have either deep slip surfaces or shallow ones. To decrease the possibility of slip failures occurrence it is now recommended to construct spoil heaps with general inclination roughly 1:6, where steeper lower slopes are combined with benches. The slope stability during filling can be performed by two basic approaches:

Figure 4. Wealthered clay clod.

– to use total parameters of shear strength (cu , su ), – to use effective parameters of shear strength {φ ′ (φef ), c′ (cef )} and to control the pore pressure development inside of spoil heap body.

Figure 5. Two basic changes of clay fill—process of softening and process of hardening. Typical problem of double porosity.

3

APPROACHES TO THE SHORT TERM STABILITY

Slope stability of clayey fill calculation for spoil heaps is more complicated than for classically compacted fill due to following aspects: – Extreme height of spoil heaps, – Decrease of shear strength as a result of moisture content increase, – Possibility of development of local zones with significantly decreased strength. In eighties of the last century 4 large landslides of spoil heaps occurred in north part of Bohemia each with volume of moving mass exceeding 50 mil. m3 .

In the first case undrained shear strength can be estimated with the help of penetrometer tests—scatter is however very large but on the other side probabilistic approach can be used. Nevertheless the discussion is about the correlation between results of penetration tests and undrained shear strength especially for unsaturated zones. In the second case the biggest problem is connected with pore pressure measurement or with its estimation. During the last period pore pressure measurement is not the exception, using not only classical piezometers, but also special measuring devices described e.g. by Feda et al (1994). The observation proved that first signals about positive pore pressure starts at depth hi at which macro-pores are closed. Further increase is nearly linear, so that the pore water pressure can be expressed by equation: u = C · γfill (h − hi )

(1)

where h is depth below surface; C—coefficient of pore pressure, varying from 0 to 1.0 for fully saturated parts of the spoil heap. Average values C = 0.5 − 0.7. With the help of this approach the speed of filling can be checked to prevent slope failure development.

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4

THE INFLUENCE OF RAIN ON LONG TERM STABILITY OF SPOIL HEAPS

At the end of filling the upper part of spoil heap is permeable for air, and pore pressure is therefore zero. Middle part is partly saturated (pore pressure coefficient C is close to 0.5–0.7) and finally lowest part is fully saturated (C = 1). But with time the thickness of fully saturated layer is increasing and the final shape will depend also on the horizontal drainage layer at the toe of spoil heap. The influence of heavy rainfall was observed many times and some correlation between rainfall and number of slope instability is similar as observed correlation between rainfall and rate of spoil heap settlement; see Figure 6. This relation was also very sensitive during heavy rainfall in 2002, the period which is also connected with heavy floods. After that there is a tendency to calculate individual examples taking into account that during heavy rainfall the top layer, with thickness about 2–2.5 m, can be fully saturated and additional rain water is flowing along the spoil heap surface. So it means that the top layer is now divided into two parts, upper which is fully saturated and lower one, in which air bubbles are closed and pore air pressure is directed by pore water pressure at the bottom of the upper layer, see Figures 7a–7b, Bartozela (2007). At this moment the problem of stability can be divided into stability

Upper layer saturated by rainfall Layer with interconnected air pores – ua > uw Partly saturated layer, C = 0.5 - 0.7 Fully saturated layer, C = 1.0

Figure 7a. Modelled layers distribution after the end of spoil heap construction and for extreme rainfall.

– along shallow planar slip surface, – along deeper general slip surface. Deeper slip surfaces are by nature more dangerous from the point of view of prospective utilization of the spoil heap surface for new construction. However the stability along shallow slip surfaces has direct linkage on the effective regeneration of spoil heap surface—first of all on its revegetation.

Figure 6. Correlation between annual rainfall and monthly settlement rate of spoil heap body.

Figure 7b.

Expected pore pressure distribution.

Figure 8. surface.

Modelled case for shallow translational slip

From the view of shallow planar slip surfaces it is possible to proceed from the classical assumptions presented in Figure 8. The stability along this shallow slip surface is insufficient when neglecting cohesion and counting only with angle of internal friction in the range of 13 and 17◦ . Indifferent stability is reached for much lower slope inclination, in the range of

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1:9 to 1:10. However any cohesion is playing very significant role. In principle there are two ways leading to some cohesion: – primarily it is the result of weathering, when surface layer as consequence of weathering (firstly by change of moisture content—moistening and consequently drying connected with shrinkage) is getting cohesive character without macropores; – secondly it is the result of new vegetation system, especially of the root system of different grasses, bushes or small trees—in fact there are even first good positive results with vineyards or with apple orchards. Because the cohesion in the range of 5 kPa can guarantee the stability, the revegetation is recommended as soon as possible. From the view of deeper slip surfaces the first condition is connected with estimation of steady groundwater level in the spoil heap body. However this level is determined by the drainage layer at the bottom of the spoil heap. Therefore its construction and protection against clogging is playing very significant role. For long term steady state conditions and for heavy rainfalls the situation is displayed in Figure 9a and the distribution of expected pore pressures in Figure 9b. The limiting situation for pore pressure distribution exists under heavy rainfalls when additional water is flowing in the spoil heap body by some preferential ways. Potentially the phreatic line is going up and air in pores above is compressed. Air pressure can be higher then the pore water pressure at the bottom of the upper layer —air bubbles have tendency go up but not easily due to high homogenization —the result of the above described process of weathering. Therefore the critical air pore pressure is defined by vertical geostatic pressure at the bottom of this upper layer—ua,crit = γ . hzv . This critical assumption is shown in Figure 9c. But at the same time this condition is also critical for shallow planar slip surface—theoretically this upper layer will be lifted. However this theoretical case was not

Figure 9b. Expected pore pressure distribution for long term stability and heavy rainfall.

Figure 9c. Critical scenario of pore pressure distribution for long term stability and heavy rainfall. Upper layer saturated by rainfall Layer with interconnected air pores – ua > uw Fully saturated layer, C = 1.0

Figure 9a. bility.

Modelled layers distribution for long term sta-

observed up to now, probably the air bubbles are going up along some preferential paths, along some local zones of weakening. Nevertheless this situation represents most critical scenario so for long term stability so for extreme rainfalls.

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Stability was calculated by software GEO5, using conventional methods of slope stability. For slope with height H = 100 m, inclination 1:6 and soil shear strength parameters: φ ′ = 15, c′ = 10 kPa, the factor of safety was F = 1.27 for situation shown in Figure 9b. For situation shown in Figure 9c the factor of safety F was still higher than 1.15. 5

pressure of the upper layer (in the extreme case by geostatic vertical pressure) and is therefore increasing pore water pressure in the lower layer. Probability of failure along shallow slip surface is also high but the potential risk of this fact is not so sensitive from the construction on the surface of the spoil heap point of view. Partly this risk can be eliminated by revegetation of spoil heap surface.

CONCLUSIONS

Spoil heaps composed from the excavated tertiary clays which overlay brown coal are new artificial earth structures strongly influencing our landscape. Regarding the necessity to use the surface of these spoil heaps for new construction the long term slope stability is a primordial question. Clayey clods are untypical material, the character of which is changing from one side of the soil mechanics spectrum (rockfill) to the second one (soft clay). Therefore the technology of filling, construction of the drainage layer at the bottom of spoil heap and measures to prevent development of the local slip surfaces, along which the shear strength can fall down to the residual strength, are so important. The recommendation is to construct the spoil heaps with general inclination roughly in the range of 1:6–1:7 to prevent development of these local slip surfaces. This inclination after that was controlled for long term stability taking into account also extreme rainfall. For this case we assumed that intermediate layer which was up to now partly saturated is closed between two layers which are fully saturated and therefore the air pore pressure inside of this layer is directed by pore water

ACKNOWLEDGEMENTS The paper was written with support from the research project MSM 6840770005 ‘‘Sustainable construction’’, which is supported by the Czech Ministry of Education. REFERENCES Bartozela, J. 2007. The influence of extreme rainfall on the stability of spoil heaps. In Czech. MSc degree. CTU Prague. Feda, J., Herstus, J., Herle, I. & Stastny, J. 1994. Landfills of waste clayey material. In: Proc. 13th IC SMFE New Delhi, Oxford and IBH Publ. Co, New Delhi, vol 4, pp 1623–1628. Sembenelli, P. & Ueshita, K. 1981. Environmental Geotechnics. State of the Art Report. In: Proc. 10th ICSMFE Stockholm, Balkema, vol 4, pp 335–394. Vanicek, I. 1995. Ways of utilization of waste clayey material. In: Proc. 10th DEC SMFE, Mamaia, pp 967–974. Vanicek, I. & Vanicek, M. 2008. Earth structures. Springer, Dordrecht, in press.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Behavior of expansive soil slope reinforced with geo-grids M.Y. Wang & X.N. Gong Institute of Geotechnical Engineering, Zhejiang University, Hangzhou, China

M.Y. Wang, J.T. Cai & H. Xu Yangtze River Scientific Research Institute, Key Laboratory of Geotechnical Mechanics and Engineering of Ministry of Water Resources, Wuhan, China

ABSTRACT: In recent years, ‘‘expansive soil slope reinforced with geogrids’’ has been successfully applied in some projects in China, but there is lack of study on its engineering behavior. An ideal elastoplastic constitutive law is utilized to model the interaction between geogrid and expansive soil based on the pullout test. Behavior of slope reinforced with geogrids under moisture absorption condition of unsaturated expansive soils is analyzed adopting the finite difference method. The strength reduction method based on stress state is used to analyze the stability of slope reinforced with geogrids and study the restraint effect of geogrids on the shallow failure of expansive soil slope due to long-term weathering. Results show that the greater the elastic modulus of geogrids, the more obvious its restraint effect on the deformation due to moisture absorption of unsaturated expansive soil slope. The shallow failure of expansive soil slope can be effectively restrained by reinforced geogrids. 1

INTRODUCTION

Expansive soil has high clay content and strong hydrophilic minerals rich in bentonites and illites, as well as exchangeable cations. Its liquid limit and plasticity index are generally great, and its engineering characteristics are greatly affected by moisture content, drainage condition and weathering. Geo-hazards caused by expansive soil occur frequently worldwide and the resulting harms are huge. A large number of engineering show that failure of expansive soil slope appears as shallow slide and progressive failure. The failure mechanism and treatment measures are always the hotspot problems that scholars have studied. In China, geogrids has been used successfully to reinforce expansive soil slope of some expressways, railways and canals in recent years, and it broken through the relevant provisions of existing specifications. These projects include: ChangJiang BuJingMen Railway, ChuXiong-DaLi Expressway, XiangTan-ShaoYang Expressway, Xi’an-NanJing Railway, Inner Mongolia-KunMing Railway, Xi’anAnKang Railway, ZhengZhou-ShoLin Temple Expressway, and canals built in the irrigation district of water transfer from DanJiangKou reservoir. As for the South-to-North Water Transfer Project (Middle Route Scheme) in China under construction, measures like expansive soil reinforced with geogrids will be adopted to strengthen the channel slope, and have been

listed one of the Key Schemes of the ‘‘Eleven-Fifth National Plan’’ in China. Whereas, scholars in the world have little studied behavior of expansive soil slope reinforced with geogrids because of the rare application before. In this paper, aiming at expansive soil slope reinforced with geogrids, the stress and deformation is analyzed by the finite difference method (FDM), and the stability is analyzed by the strength reduction method based on stress state. Behavior of expansive soil slope reinforced with geogrids under moisture absorption condition of unsaturated expansive soil and under the action of long term weathering influence is investigated.

2

NUMERICAL ANALYSIS METHOD OF EXPANSIVE SOIL SLOPE REINFORCED WITH GEOGRIDS

2.1 Simulation of deformation due to moisture absorption of unsaturated expansive soil Recent practices indicate that infiltration of rainwater can cause the stress state and deformation model of unsaturated expansive soil slope to change (C.w.w. NG et al. 2003). Zhu (2003) states that the change in moisture content will make the structure expandshrink, leading to the redistribution of stress, and this problem is similar to the governing equation of thermal

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stress, so the same numerical solution method can be adopted. Strain increment due to moisture absorption of unconstrained isotropic expansive soil can be expressed as Equation (1): where α is coefficient of linear expansion, δij is Kronecker symbol, dω is moisture content increment, dεijω is strain increment due to moisture absorption. dεijω = αδij dω

(1)

Figure 1.

Mechanical model of geogrids.

Introducing the side limit condition and generalized Hook’s law, linear expansion coefficient can be calculated from the swelling ratio test without load as shown in expression (3), where ω is moisture content increment, δ is the swelling ratio without load; ν is the Poisson’s ratio. α=

δ(1 − υ) ω(1 + υ)

(2)

Total strain increment of expansive soil due to moisture absorption under load action is obtained by superposing the elasto-plastic strain increment and strain increment due to moisture increment, as shown in expression (3), where dεij is total strain increment, dεije is elastoplastic strain increment. dεij = dεije + dεijω

(3)

The initial strain method is adopted to analyze the stress state of expansive soil slope due to moisture absorption, and the strain caused by change in moisture content is regarded as initial strain. 2.2

(a) Relationship between the shear stress and relative displacement; (b) Shear strength criterion

Figure 2. sive soil.

Interface model between geogrids and expan-

is equal to its shear strength, the interface becomes at friction slip state as shown in Figure 2a. The shear strength at the interface is shown in Figure 2b, Where c is interface adhesion, φ is frictional angle at the interface. 2.3 Analysis of stability

Analysis of stress and deformation

Geogrid is a typical ductile material with a small flexural rigidity, and calculations show that under an ordinary operating condition, tension stress of geogrids in soil is generally smaller than its tensile strength, thus geogrids is modeled as linear elastic thin shell element. On the other hand, geogrid produces friction and shear actions with the fills in tangent plane direction, while it is restricted with fills in normal direction. Interaction between geogrids and soils is shown in Figure 1, of which σm is confining stress of geogrids, τ is shear stress at the interface, N is the resultant force of section stresses produced from geogrids. An ideal elastoplastic model is utilized to model the behavior of interaction at the interface. When shear stress at the interface is less than its shear strength, the interface is at elastic and binding state, ratio of shear stress to relative displacement at the interface is the tangential stiffness κ. Once shear stress at the interface

The limit equilibrium method is usually adopted to analyze the stability of reinforced slope in existing specifications, only the frictional action at the interface between geogrids and the fills is considered. Safety factor of slope obtained by this method is greatly smaller than its actual magnitude (Wang et al. 2000). The authors hold that first of all the change in stress state of slope caused by reinforced geogrids and the variance in strength of the fills within reinforced region are analyzed, and then stability are analyzed based on the stress state of reinforced slope. The strength reduction method based on stress field is adopted to calculate the safety factor of slope, and transfixion of equivalent plastic strain is taken as the indicator of slope instability. Strength parameter of soils is reduced at the same time as shown in Expression (4): c′ = c/w,

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tan ϕ ′ = tan ϕ/w

(4)

Where ω is reduction factor, namely safety factor; c′ , φ ′ are cohesion and internal friction angle of soils after reduction, respectively. 3

BEHAVIOR OF EXPANSIVE SOIL SLOPE REINFORCED WITH GEOGRIDS

Mitchell (1996) pointed out that some factors such as tensile strength and rigidity of geogrid, creep characteristics, durability and frictional characteristics of the interface are important characteristics affecting the reinforcing effect. Hatami and Bathurst (2005) carried out model test on retaining wall reinforced with geogrids with different vertical spaces. The authors hold that the reasonable vertical space between geogrids should be made from the strength parameter at the interface, thickness of atmospheric influence, and the geometrical character of slope etc. Expansive soil slope is of obvious shallow nature in failure, but it doesn’t fails along the slope surface, and position of failure plane is related to the slope height, both of which are different from that of sandy soil slope. Chandler and Skempton (1974) hold that failure of cut slope in stiff clay with multi-fissures during the period of several years to several decades is attributed to the reduction in effective shear strength parameters of clay due to long term weathering etc. Shen (2004) summarized the effect of weathering on geotechnical engineering and proposed the design concept of resistance to weathering. In this paper, considering effect of the parameters at the interface between geogrids and compacted expansive soils and the elastic modulus of geogrids, restraint effect of geogrids on the deformation of slope due to moisture absorption of unsaturated expansive soil is investigated. Taking the decrease of expansive soil’s cohesion caused by long-term weathering action, the space of geogrids vertically-arranged and the parameters of the interface into account, the restraint effect of geogrids on the shallow failure of expansive soil slope is studied. Computation model is shown in Figure 3, height of expansive soil slope is 10.0 m with a slope of 1:2,

of which 2.0 m thick surface layer is backfilling soil, geogrids are lay down horizontally. Mohr-Coulomb ideal elasto-plastic constitutive model and the associative flow rule are utilized to simulate expansive soil. Consolidation and undrained shear strength parameters on natural condition is taken, as shown in Table 1. 3.1 Restraint effect of geogrids on the deformation of reinforced slope due to moisture absorption of unsaturated expansive soil Space vertically-arranged between geogrids is taken as 0.8 m, and horizontal embedded length is 6.0 m. According to parameters at the interface between geogrids and expansive soils obtained by pull-out test, considering the influences of such factors as geogrid materials and moisture content of the fills, four research schemes as shown in Table 2 is proposed. According to observed data, initial moisture content of compacted expansive soil in reinforced zone is assumed to be 16.3%, the optimum moisture content; moisture content at slope surface after it fully rains is 25.8%; moisture content is constant when depth under slope surface exceeds 2 m; moisture content increment in 0 to 2 m range under the slope surface assumes to be linear relationship along the depth. Coefficient of linear expansion of expansive soil obtained via swelling ratio test without load is 20.2%. Taking the slope shoulder as an initial point, horizontal and vertical displacements at slope surface point at the middle of two adjacent layers of geogrids (called as slope surface points in this paper) are shown in Figs. 4 and 5. It can be seen that horizontal and Table 1.

Mechanical parameters of expansive soils.

Soils category

E (N/m2 )

c (kPa)

 (degree)

γ (kg/m3 )

μ

Undisturbed Backfill

5 × 107 5 × 107

20 15

17 15

2000 1850

0.35 0.35

Table 2. Parameters of geogrids and the interface between geogrids and expansive soils.

25m

15m

10m

20m

backfill soils geogrids

Scheme

60m

Figure 3. Section of expansive soils slope reinforced with geogrids.

1 2 3 4

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Elastic modulus of geogrids (N/m2 ) 1.2 × 1010 1.2 × 1010 1.2 × 1010 1.2 × 109

c (kPa)

 (degree)

16 10 16 16

10 10 16 10

Tangential rigidity (N/m3 ) 2.3 × 106 2.3 × 106 2.3 × 106 2.3 × 106

0.16

Relative displacement /m

Horizontal displacement /m

0.20 0.16 0.12 scheme1 Scheme2 scheme3

0.08 0.04 0.00 0.4

2.0

3.6

5.2

6.8

0.12

0.08

scheme1 scheme2 scheme3

0.04

0.00

8.4

0.8 2.4 4.0 5.6 7.2 Vertical distance from slope shoulder/m

Vertical distance from slope shoulder /m

Figure 4. Horizontal displacement at slope surface points when different interface parameters are used.

8.8

Figure 6. Relative displacement at the interface between geogrids and fills when different interface parameters are used. 17.6

0.30

17.2 0.20

0.10

Shear stress /kpa

Vertical displacement /m

0.40

scheme1 scheme2 scheme3

0.00 0.4 2.0 3.6 5.2 6.8 Vertical distance from slope shoulder /m

16.8 16.4

scheme1 scheme4

16.0 15.6

8.4

0.8

2.4 4.0 5.6 7.2 8.8 Vertical distance from slope shoulder/m a Shear stress at the interface at slope surface

Figure 5. Vertical displacement at slope surface points when different interface parameters are used.

20.0

vertical displacements for Scheme 1 and 3 are small, and for these two schemes effect of friction angle at the interface is quite small, which indicates that shear stress at the interface is all smaller than its shear strength. In Scheme 2, considering the influence of increase in moisture content of expansive soil, interface adhesion takes 10 kPa, friction angle at the interface takes 10 degree, horizontal and vertical displacements are great, which shows that plastic yield occurs at the interface, and restraint effect of geogrids on the deformation of expansive soil slope is reduced. Relative displacements between geogrids and the fills at slope surface are shown in Figure 6. It may be seen that relative displacement between geogrids and the fills for Schemes 1 and 3 is small, and their difference is small, which shows that the interface is at elastic state, restraint effect of geogrids on expansion is governed by tangential rigidity coefficient k, and is independent of strength parameter at the interface. Compared with Schemes 1, Scheme 2 obtains an obviously great relative displacement at the interface, this shows that when friction angle at the interface keeps at

Shear stress /kpa

16.0 12.0 8.0 scheme1 scheme4

4.0 0.0

0.8

2.4 4.0 5.6 7.2 8.8 Vertical distance from slope shoulder/m

b S he a r s tre s s a t the inte rfa ce a t the e nd of ge ogrid within s lope

Figure 7. Shear stress at the interface between geogrids and fills when different elastic modulus of geogrids are used.

10 degree, interface adhesion is reduced from 16 kPa to 10 kPa, the interface is changed from elastic state into plastic yield state, increasing the lateral deformation at the slope, decreasing the restraint effect of geogrids on the slope deformation. As for Schemes 1 and 4, Figure 7a and 7b show the shear stresses at the interface at the end of geogrids

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at slope surface and within the slope, respectively; horizontal displacements at slope surface points are shown in Figure 8, relative displacement at the interface at slope surface are shown in Figure 9. It may be seen that the greater the elastic modulus of geogrids, the larger the shear stress at the interface, the smaller the deformation at the slope surface, the more obviously the restraint effect of geogrids on slope deformation, the smaller the elongation of geogrids, the greater the relative displacement at the interface. 3.2

Restraint effect of geogrids on the shallow failure of expansive soil slope under long term weathering action

Affected by radiation and wetting/drying cycle under atmospheric influence, cohesion of clay is gradually reduced. The atmospheric influence zone is set to be as deep as 2.0 m, cohesion of compacted expansive soil under long term weathering action is reduced to 2.0 kPa, but friction angle is constant (Chandler et al. 1974). Calculation schemes and results are shown in

Horizontal displacement /m

0.16

Table 3. Safety factor and potential slip plane positions at different vertical space between geogrids. Horizontal Vertical length of space geogrids Safety Scheme (m) (m) factor Slip plane position 5 6 7 8 9 10

0.12

0.08

Table 3, elastic modulus of geogrids takes 1,200 MPa, tangential stiffness of interface takes 2.3 × 106 N/m3 , interface adhesion is 16 kN/m2 , and frictional angle at interface takes 10 degree. It may be seen from Table 3 that after geogrids are reinforced, safety factor of slope is more than 1.55, and the closer the geogrids arranged, the greater the safety factor; when horizontal embedded length of geogrid is 4.5 m, namely geogrids penetrate just through weathered layer, vertical space between geogrids will affect greatly the safety factor of slope. The potential slip plane of non-reinforced expansive soil slope under weathering action is showed in Figure 10, which is a typical shallow slip plane with a safety factor of 0.94; a potential slip plane of slope reinforced with geogrids (Scheme 9) is showed in Figure 11, with the safety factor increasing to 1.74. After geogrids are reinforced, potential slip plane moves toward the deep

scheme1 scheme4

0.4 0.6 0.8 0.4 0.6 0.8

4.5 4.5 4.5 6.7 6.7 6.7

1.68 1.64 1.57 1.75 1.74 1.73

geogrids’s end geogrids’s end geogrids’s end non-weathered soil non-weathered soil non-weathered soil

0.04 0.00 0.4

2.0

3.6

5.2

6.8

8.4

Vertical distance from slope shoulder /m

Figure 8. Horizontal displacement of slope surface points when different elastic modulus of geogrids are used.

Relative displacement /m

0.10

Figure 10. Typical slip plane of expansive soil slope without geogrids (safety factor = 0.94).

0.08 0.06 0.04

shceme1 scheme4

0.02 0.00 0.8 2.4 4.0 5.6 7.2 Vertical distance from slope shoulder /m

8.8

Figure 9. Relative displacement at the interface between geogrids and fills at slope surface when different modulus of geogrids are used.

Figure 11. Typical slip plane of expansive soil slope reinforced with geogrids (safety factor = 1.74).

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part of slope, and its safety factor is greatly increased, which shows that geogrids can effectively restrain the shallow failure of expansive soil slope. Calculation schemes and results obtained by considering effect of moisture content on strength parameters at interface and vertical space between geogrids are shown in Table 4, the horizontal embedded length is 6.7 m; potential slip planes obtained by Schemes 12, 13 and 10 are shown in Figures 12, 13 and 14, respectively. It can be seen from Table 3, Table 4 and Figure 12 that when vertical space between geogrids is 0.6 m, Table 4. Effect of strength parameters at the interface on stability of reinforced slope. Vertical space c  Safety Slip plane Scheme m kPa degree factor position 11 12 13

0.6 0.6 0.8

16 1 1

16 10 10

1.76 1.72 1.64

non-weathered soil non-weathered soil weathered soil

Figure 12. Potential slip plane obtained from Scheme 12 (safety factor = 1.72).

even if interface adhesion is reduced to 1 kPa and friction angle at the interface is 10 degree, it exerts little effect on the safety factor of slope and the position of potential slip plane, which shows that shear stress at the interface is less than its shear strength, so there is little relationship between stress state of slope and strength at the interface. In comparison with Schemes 10 and 13, if vertical space between geogrids is 0.8 m, when interface adhesion is reduced to 1 kPa, potential slip plane is located in shallow weathered layer; when interface adhesion is kept at 16 kN/m2 , potential slip plane moves into nonweathered layer, this shows that when strength at the interface is small, the interface is at plastic yield state, stress state in reinforced zone of slope is redistributed, so safety factor of slope is reduced. In comparison with Schemes 12 and 13, strength parameters at the interface are the same (interface adhesion is 1 kPa, friction angle is 10 degree), when vertical space between geogrids is 0.6 m, potential slip plane lies within non-weathered layer; when vertical space between geogrids is 0.8 m, potential slip plane is located at shallow weathered layer. At the same time it also shows that even if interface adhesion between geogrids and expansive soil is reduced to 1 kPa, keeping the interface at plastic yield state, safety factor of slope is still increased from 0.94 before reinforced to 1.64 after reinforced. When interface adhesion is maintained at 16 kN/m2 , vertical space between geogrids exerts little effect on the position of potential slip plane and the safety factor. It follows that reasonable space vertically arranged between geogrids is dependent on the strength parameters at the interface between geogrids and expansive soils. In addition, it may be affected by the horizontal embedded length of geogrids etc.

4

Figure 13. Potential slip plane obtained from Scheme 13 (safety factor = 1.64).

Figure 14. Potential slip plane obtained from Scheme 10 (safety factor = 1.73).

CONCLUSIONS

In this paper, the non-linear FDM is adopted to analyze behavior of expansive soil slope reinforced with geogrids under moisture increment of unsaturated expansive soils. The strength reduction method based on stress state is introduced to study the restraint effect of geogrids on the shallow failure of expansive soil slope under long term weathering effect. It comes to the following conclusions: 1. Under a given space between geogrids and embedded length condition, strength parameters at the interface plays a leading role on the reinforced effect; When the interface is at plastic yield state, the restraint effect of geogrids on the deformation of expansive soil slope will be reduced. 2. The larger the elastic modulus of geogrids, the smaller its elongation, the greater the relative

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3.

4.

5.

6.

displacement at the interface, the smaller the deformation at slope surface, the more obviously the restraint effect of geogrids on slope deformation under moisture increment of expansive soil. The shallow failure of expansive soil slope under long term weathering effect can be effectively restrained by reinforced geogrids; after reinforced, potential slip plane moves toward the deep zone, and safety factor of slope will be increased greatly. When strength parameters at the interface is lesser due to moisture increment, it may make the interface reach plastic yield state, redistribute the stress state in reinforced zone of slope, reduce the safety factor of slope; even if cohesion of the interface is reduced to 1 kPa, geogrids still can make the slope maintain at steady state. Reasonable space vertically arranged between geogrids is dependent on the strength parameters at the interface, and may be affected by the embedded length of geogrids etc. The strength reduction method based on stress state used to analyze the stability of reinforced slope can exactly reflect the action mechanism of reinforcement.

REFERENCES Chandler R.J. & Skempton, A.W. 1974. The design of permanent cutting slopes in stiff fissured clays. Geotechnique, 24 (1): 457–466. C.w.w. N.G., L.T. Zhan, C.G. Bao, et al. 2003. Performance of an unsaturated expansive soil subjected to artificial rainfall infiltration. Geotechnique, 53(2): 143–157. Hatami K., Richard J. Bathurst 2005. Development and verification of a numerical model for the analysis of geosynthetic-reinforced soil segmental walls under working stress conditions. Can. Geotech. J., 42: 1066–1085. Mitchell J.D. 1996. State of the art: Limit equilibrium and finite element analysis of slopes. Journal of geotechnical engineering, ASCE, 122(7): 577–596. Shen Z.J. 2004. Weathering resistant design: An important aspect of future development of geotechnical engineering design. Chinese Journal of Geotechnical Engineering, 26(6): 866–869 (in Chinese). Wang Z. & Wang X.Q. 2000. Some problems on foundations reinforced with geosynthetics. Chinese Journal of Geotechnical Engineering, 22(2): 503–505 (in Chinese). Zhu B.F. 2003. Thermal stresses and temperature control of mass concrete [M]. Beijing. China electric power press: 153–155 (in Chinese).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Geotechnical properties for a rainstorm-triggered landslide in Kisawa village, Tokushima Prefecture, Japan G. Wang & A. Suemine Research Centre on Landslides, Disaster Prevention Research Institute, Kyoto University, Japan

ABSTRACT: A typhoon (called Typhoon No. 10) attacked Shikoku Island and Tyugoku area of Japan in 2004. Due to this typhoon, rainstorm fell on Shikoku Island, giving a new daily precipitation record of 1,317 mm and triggering a huge number of landslides in Tokushima Prefecture. One of the catastrophic landslides was triggered on Shiraishi area of Kisawa village, which destroyed more than 10 houses and caused an unstable block on the source area. The unstable block kept moving after the event, showing accelerating/decelerating movement during/after the rainfall and reaching a displacement of several meters before the countermeasures were finished. To examine the mechanism for this landsliding characteristic, we took samples (weathered serpentinite) from the field, and their shear behaviours were examined based o ring shear tests. The test results revealed that the residual shear strength of the samples is positively dependent on the shear rate, which may provide an explanation for the continuous accelerating-decelerating process of the landsliding. 1

INTRODUCTION

Typhoon Namtheun (the 10th tropical storm in the western Pacific in 2004) originated west of Minamitorishima Island of Japan on July 25, 2004. It made landfall on Shikoku Island on July 31, then passed through the Seto Inland Sea and Hiroshima Prefecture, and moved toward the eastern part of the Korean Peninsula, losing energy to become a tropical depression. Accompanying this typhoon, heavy rain fell in Shikoku area of Japan (Figure 1), especially in the Nakagawa town (on the southwest part) of Tokushima Prefecture. The total precipitation from July 30 to August 2 was more than 2,000 mm (Figure 2). This is several times of the normal precipitation for the months of July and August in this area. Hourly precipitation reached more than 120 mm (Figure 2). A daily precipitation of 1,317 mm was recorded on 1 August; this value value gives a new Japanese daily precipitation record, which exceeds the previous one of 1,114 mm. Note that this old record was obtained in Kito village (about 16 km southwest of Kisawa village) on September 11, 1976 accompanying Typhoon Fran. The area where precipitation exceeded 1,500 mm for the storm was centered on Kisawa village and Kaminaka town, as a very narrow area of 5–6 km in the east-west direction, and 10–20 km in the south-north direction (Figure 1). In this area, many landslides were triggered, among which a catastrophic one occurring

on the source area of Furon valley of Shiraishi district destroyed more than 10 houses and caused an unstable block (about 20,000 m3 ) along the valley bed on the source area. Figure 3 shows the damaged area by the August event. Figure 4 shows the distribution of the unstable block on the source area of the gulley. This unstable block was a reactivated one, which kept moving with significant displacement after the event. If this unstable block fails and shifts to debris flow, a larger area on the down stream will be damaged (as shown in Figure 3). Because of this risk, the residents on the downslope area evacuated and countermeasures were undertaken immediately after the August

Figure 1. Rainfall distribution in the Shikoku area from July 30 to August 2 during the typhoon (after Nakagawa River Office, Shikoku Development Bureau, Ministry of Land, Infrastructure and Transport, Japan, 2004).

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A B C

Figure 2. Hyetograph of the heavy rainfall (data courtesy of Shikoku Electric Power Co., Inc.).

Figure 3. Traveling path of Shiraishi landslide as well as the predicted damage area if the unstable block becomes debris flow.

event. The monitored displacement revealed that the unstable block suffered accelerating and decelerating movement during and after rainfall, the accumulated displacement was several meters, but no rupture failure occurred. To clarify the movement mechanism, samples were taken from the unstable block area, and their residual shear strength characteristics were examined based on the ring shear test results. 2

LANDSLIDES ON SHIRAISHI AREA

Shiraishi area was designated as the landslide prevention area on 1962, based on the Japanese Landslide Preventive means, and the mountain stream was also designated as the debris-flow-risk-rich stream. Landslide on the source area of Furon valley was trigged around 20:00 hr on 1 August (hereinafter term the first landslide). The first landslide on the source area sized 70 m in length, 35 m in width, and 2–3 m in depth, and was originated on a slope of about 29 degrees

Figure 4. Plan of the source area of Shiraishi landslide, and the unstable block after the landslide event (area marked by dashed line) as well as the locations of extensometers and sampling.

(Hiura et al 2004). The scarp of this landslide was at the elevation of 540 m, and toe was at the elevation of 290 m reaching the national road, and the horizontal travel distance was about 550 m. The displaced landslide mass was approximately 5,000–7,000 m3 . The landslide transformed to debris flow when it came to the lower slope part. Many houses were destroyed or damaged by this event (Figure 3). Fortunately, the residents of this area noticed some strange phenomena, such as the ground water flowing out from the middle par of the mountain slope and great sound by moving rocks. They recognized these landslide precursors, and started to evacuate from 15:00, and then there was neither fatality nor injury. Nevertheless, after this event an unstable block (hereinafter term Shiraishi landslide, which is the target of this study) on the upper part of the source area formed and kept moving. Three extensometers were installed to monitor the movement (Figure 4). The monitored displacement and rain precipitation from September to December of 2004 are presented in Figure 5. Great displacements were observed. The landsliding was very sensitive to rainfall, and shoshowed accelerating/decelerating movement during/after each rainfall event. For S-1, S-2 and S-3, the observed displacement rates were 6.4–25, 2.5–17.5, 1.3–11.5 mm/h respectively during raining days; and 0.9–1.5, 0.6–0.9, and 0.3 mm/h during dry days. Due to the continuous movement and the great possible damage area, an evacuation order had been issued for months and, countermeasures were performed immediately. The displacement rate became smaller with progress of countermeasures. When the countermeasures were finished, no displacement was observed. Normally the displacement of landslide on crystalline schist area is not so great. However, Shiraishi landslide had an accumulated displacement greater than

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Figure 7.

Simulating the sliding surface in ring shear test.

Figure 5. Monitored rain precipitation and displacement.

Figure 6. Longitudinal section of the unstable block.

4 m. Clarifying the mechanism of this kind of landsliding will be of importance for disaster mitigation of similar landslides. The area where the landslides occurred is characterized by deep river valleys with steep slopes, and many of the mountain slopes have steep chutes. Most of the settlements are located on gentle slopes formed by old landslides, or on narrow streamside terraces. According to the subsurface geological map of Tokushima Prefecture (Tokushima Prefecture, 1983), this area is mainly underlain by Paleozoic greenstone, Paleozoic and Mesozoic pelite and greywacke, and serpentinite of the Mesozoic Kurosegawa terrane, as well as limestone and chert. Figure 6 shows a longitudinal section of the unstable block. The reactivated part is mainly composed of colluvium deposits, and can be divided into three sliding parts. The biggest block has its sliding surface mainly on weathered serpentinite layer. The thickness of the landslide mass is about 8–12 m, and the sliding surface sloped approximately 12 degrees.

3

RING SHEAR TESTS

Ring shear apparatus has been widely used for obtaining the residual shear resistance of soils at large shear displacement. In this work, the used ring shear

apparatus has a shear box sized 120 mm in inner diameter, 180 mm in outer diameter, and 115 mm in height, and a maximum shear velocity of 10 cm/s. The principle of simulating the sliding surface in ring-shear test is illustrated in Figure 7. The sample in the shear box is laterally confined between pairs of doughnut-shaped upper and lower confining rings. Landslide is simulated by loading normal and shear stresses that exist at the sliding surface in the field. Three samples (A, B and C) were taken from three parts of landslide. Sample A (weathered but does not include clay) was from the outside of the landslide, while Sample B and C were taken near the sliding surface of the landslide (see Figure 4). During tests, grains in the sample greater than 475 μm were sieved out due to the size limitation of the shear box. The specific gravities for sample A, B, and C are approximately 2.70, 2.70, and 2.73, respectively. For Sample B, the plastic index is 6.3, and clay content is about 18%; for Sample C, the plastic index is 32, and clay content is about 33%. Note that we did not measure the above mentioned index for Sample A, because it does not include clay. The specimen was made by dry-deposition method as introduced by Ishihara (1993): the oven-dried soil was poured into the shear box freely in several layers, and each layer was tamped differently to achieve different initial densities. All specimens were saturated by CO2 and de-aired water. After saturation, the specimen was consolidated under a given normal stress, and then was sheared to residual state using shearspeed-controlled method. No saturated process was performed for dry sample. From Figure 5, it is seen that the landsliding block was characterized by accelerating movement during rainfall, and decelerating movement after rainfall, i.e., the landsliding was at different rate at different time. Therefore it is reasonable to believe the displacement rate may affect the residual shear behavior of soils along the sliding surface. In the examination of the rate effect, two kinds of test methods have been normally used (Suzuki et al. 2004): (I) use different specimen for each test at different shear rate; (II) shear one specimen

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to residual state under a given normal stress and then change the shear rate without changing specimen. It was also reported that these two kinds of method do not show obvious difference in the results. Hence, in this study, we use method II to examine the shear rate effect.

4

Shear resistance (kPa)

100

Because Shiraishi landslide originated on a gentle slope of about 12 degrees, it is necessary to have a better understanding of the residual shear strength, although the possible friction angle of soils along the sliding surface had been evaluated through back analysis of slope stability during the planning stage for the countermeasures. Drained shear test was performed on sample A at first. After consolidation under a normal stress of 150 kPa, the specimen was sheared at a displacement rate of 0.11 mm/min. Figure 8 shows the obtained stress ratio (shear resistance/normal stress). In direct shear test, it is required to shear the specimen at a rate of 0.02 mm/min to obtain the residual shear strength. However, based on ring-shear test results, Yokoda et al (1995) found that when the shear rate is smaller than 1.01 mm/min, the rate does not affect the residual shear strength. Thus, the residual shear strength at the shear rate of 0.11 mm/min should be proper. It is seen from Figure 8 that the residual state had been reached when the shear displacement reached 40 mm. After this test was finished, the normal stress was changed to 110 and 70 kPa to measure the residual shear strength at different normal stresses. Figure 9 plots the results. The line connecting these three points projects to the original, giving an inclination angle of 30 degrees. Therefore, it is concluded that for sample A, the cohesion is zero and inter friction angle is about 30 degrees. For sample B, the residual shear strength obtained at different normal stress is shown in Figure 10. In this test, the specimen was consolidated under 274 kPa, and then sheared to residual state at a shear rate of

Figure 8. Drained shear test on saturated sample A.

30 0 0

TEST RESULTS AND DISCUSSION

Sample A Shear rate: 0.11 mm/min

50

Figure 9.

Figure 10.

50

100 Normal stress (kPa)

150

200

Residual shear strengths for sample A.

Residual shear strengths for sample B.

0.24 mm/min. Therefore, the specimen was kept shearing at this rate, while the normal stress was decreased very slowly. Through this kind of method, the residual failure line was obtained. From Figure 10, we can see that an inter friction angle of about 20 degrees is obtained. The test for sample C (Figure 11) was performed using the same method as that shown in Figure 9. Differing from samples A and B, sample C shows an inter friction angle of 16 degree with a cohesion of about 10 kPa. Many studies have been performed to examine the shear rate effect on the residual shear strength of clay (Skempton 1964; Kenney 1977; Yatabe et al 1997; Tika et al 1996; Suzuki et al 2004; among others). The possible effects of permeability and clay content as well as the structure of shear zone on the changing range of residual shear strength at different shear rates have been clarified. In this study, the possible rate effects for the above mentioned three samples have also been examined. After the above mentioned shear tests were finished, the specimens were sheared at different shear rates. The measured residual shear strengths are shown in Figs. 12–14. It is seen that the residual shear strength become greater with shear rate. To clarify the reason for this kind of increasing tendency, ring shear tests at different shear rates were also performed on dry sample A. The results are superimposed in Figure 12, where it can be noticed that the residual shear strength becomes greater with increasing of shear rate. Nevertheless, similar tests on

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Figure 11.

Drained shear test on saturated sample C.

Figure 14. Residual shear strength against shear displacement rate for sample C.

Figure 12. Residual shear strength against shear displacement rate for sample A.

Figure 15. Residual shear strength against shear displacement rate for sample from colluvial deposit.

Shear zone

Figure 13. Residual shear strength against shear displacement rate for sample B. Figure 16.

the sample collected from the colluvial deposit of the unstable block showed that the shear rate does not have, or at least has very small effect on its residual shear strength (Figure 15). To clarify this kind of increasing tendency of residual shear strength with shear rate for serpentinite, the shear zone formed during the shearing for each test and the orientation of soil grains were observed. After the shear test, the shear box was opened, and the specimen was observed. Shear zone was formed in each test, and shiny surfaces, i.e., slicken sides, had been

Shear zone formed on the test on sample B.

well developed. Figure 16 shows the shear zone as well as the specimen below the shear zone for the drained test on saturated sample B. The specimens above the shear zone were moved for the observation. The soil in the shear zone became very clay with different color. Similar phenomena had also been observed on the test on dry sample A. Using a laser microscope, the surface of the shear zone was observed. Figure 17 shows the observed surface. Through this method, the roughness of the

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Percent finer than by weight (%)

Surface shape

100 80 60 40 above shear zone

20

shear zone below shear zone

0 0.001

0.01

0.1 Grain size (mm)

1

10

Figure 19. Grain size distribution for the samples at different parts of the shear box.

Surface roughness measuring line Figure 17.

Observed shear surface of the shear zone.

Figure 18.

3D image of the shear surface.

surface can be measured. The roughness profile along the line shown in the lower part of this figure is presented on the upper part of the same figure. Figure 18 gives a 3D image of Figure 17. The soil grains show the structure like fish scale. It is desirable to clarify the relationship between the surface roughness and the mobilized shear strength from the microstructure of the shear surface. Detailed analysis is on going. From Figure 16, it is seen that the shear zone has become quit clay due to the particle breakage. To analyze the degree of particle breakage, samples were taken from, above and below the shear zone respectively, and grain size analysis was performed. Figure 19 shows the results. The soils from the shear zone became much finer than those from the upper and lower parts, indicating that very serious particle breakage occurred within the shear zone. The little difference between the upper and lower parts may be due to the settlement of fine particles from the upper part during the preparation of sample and shearing.

In the time prediction of landslide occurrence, the method based on the creep rupture theory has been widely used (Saito and Uezawa 1966; Fukuzono 1985; Hayashi 1988; among others). This method uses the displacement rate during the late half stage of secondary creep and the onset of tertiary creep to predict the rupture failure time. For Shiraishi landslide, although the sliding showed the tendency of late half stage of secondary creep and early stage of the tertiary creep almost at each rainfall event, there was no rupture failure, even when the accumulated displacement reached several meters. As for the main reason, besides the involvement of ground water during rainfall, the shear rate dependent residual shear strength may play a key role. During the rainfall ground water table increases with the infiltration of rainwater, and then reduce the effective normal stress and shear resistance. Then the landsliding will accelerate. However, increament in the sliding speed will increase the shear resistance of soil along the sliding surface, and then decelerate the sliding, such that the rupture failure can not be triggered. Nevertheless, this interpretation needs further checking, because the ring shear tests were performed at the shear rates greater than the monitored landsliding. Shear tests at the real sliding speed are planed in coming studies. 5

CONCLUSIONS

A series of ring shear tests was performed on the serpentinite samples taken from the Shiraishi landslide area. The basic shear behavior of these samples were examined by shearing them at different normal stresses and shear displacement rates. The results can be summarized as follow. 1. Samples with different weather degrees showed different shear behavior. Sample A from the outside of the landslide showed an inter friction angle of 30 degrees, while sample B from the landslide body and C near the sliding surface had the inter friction angles of 20 and 16 degrees, respectively.

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2. The residual shear strengths for all these three serpentinite samples increase significantly with increase of shear rate. However, the colluvial deposit of the landslide did not show this kind of rate dependency, indicating the special shear characteristics of serpentinite. 3. In each test, shear zone was formed after long displacement of shearing. Particle breakage occurred within the shear zone, and the soil within the shear zone became very clayey. 4. The increase of residual shear strength with shear rate may be a main reason for this unstable block that experience accelerating-decelerating movement repeatedly without rupture failure even when the accumulated displacement reached more than 4 meters. ACKNOWLEDGEMENTS This study was funded by a scientific research grant (No. 18380094) from the MEXT of Japan. The authors are grateful to the Nanbu General Bureau of Tokushima Prefecture, the previous public office of Kisawa Village, and Mr. Tamura in Yonden Consultants Inc. for their help in the field work and sampling. Thanks also go to Mr. Ochi in KEYENCE Corporation, Japan for his help in the observation of shear surface of the shear zone.

Hayashi, S., Park, B., Komamura, F., Yamamori, T. 1988. On the forecast of time to failure of slope (II)—Approximate forecast in the early period of the tertiary creep. Journal of Japan Landslide Society 23 (3): 11–16 (in Japanese). Hiura, H., Kaibori, M., Suemine, A., Satofuka, Y., & Tsutsumi D. 2004. Sediment-related disasters in KisawaVillage and Kaminaka-Town in Tokushima Prefecture, Japan, induced by the heavy rainfall of the Typhoon Namtheun in 2004 (prompt report). Journal of the Japan Society of Erosion Control Engineering 57 (4): 39–47. Ishihara, K. 1993. Liquefaction and flow failure during earthquakes. Géotechnique 43 (3), 351–415. Kenney T.C. 1977. Residual strength of mineral mixtures. Proc. of the 9th ICSMFE, 155–160. Saito M., Uezawa H. 1966. Forecasting the time of occurrence of a slope failure. Journal of Japan Landslide Society 2 (2): 7–12 (in Japanese). Skempton, A.W. 1964. Long-term stability of slopes. Géotechnique 14 (2): 75–101. Suzuki, M., Kobayashi, K., Yamamoto T., Matsubara T., Hukuda J. 2004. Influence of shear rate on residual strength of clay in ring shear test. Research Report, School of Engineering, Yamaguchi University 55 (2): 49–62, 2004. Tika, T.E., Vaughan, P.R., and Lemos, L. 1996. Fast shearing of pre-existing shear zone in soil. Géotechnique 46 (2): 197–233. Yatabe, R., Yokoda, K., Yagi, N., & Nochi, M. 1997. Considerationon the mechanism of landslides at Serpertine Belt. Journal of Japan Landslide Society 34 (1): 24–30. Yokoda, K., Yatabe, R., & Yagi, N. 1995. Strength characteristics of weathered serpentinite. Journal of the Japan Society of Civil Engineers 529 (III-33): 155–163.

REFERENCES Fukuzono, T. 1985. A new method for predicting the failure time a slope. Proc. IV Int.Cof. and Field Workshop in Landslide, Tokyo: 145–150.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Yigong rock avalanche-flow landslide event, Tibet, China Qiang Xu, Shi-Tian Wang, Hu-Jun Chai, Zhuo-Yuan Zhang & Simeng-Dong The state key laboratory of geoharzards prevention and geoenvironment protection, Chengdu University of Technology, Chengdu, Sichuan, China

ABSTRACT: The paper described a large rock avalanche and flow landslide in detail, which happened in Yigong, Tibet, China on April 9, 2000. According to the basic characteristics of the landslide, the whole event can be devided into three sections: rock avalanche, flow landslide, and deposition. The paper presented a detailed description of each section. Attention should be paid to some unique characteristics of the deposition section such as depositional cones, liquefaction holes, and remarkable wind damaged trees. Furthermore, in terms of some distinctive characteristics, the complex landslide can be subdivided into several subsections. The paper also provided a brief discussion on the mechanism of the large complex landslide to raise the research interests of the scholars both at home and abroad. 1

INTRODUCTION

A remarkable catastrophic rock avalanche-flow landslide occurred at 19:59 on April 9, 2000, (Ren et al. 2001) in Zhamulong gully of Yigong area, Bomi county, Tibet, China. The longitude and latitude of the rock avalanche-flow landslide (Hunger 2001) are 94◦ 55′ ∼95◦ E and 30◦ 10′ ∼30◦ 15’N, respectively. About 3×107 m3 rock fell down 1500 m from the top of the mountain, and entered into the loose debris filled Zhamulong Gully. During that time, the ice become melt under the debris, so it turn the debris into the debris-flow. It moved 8∼10 km along the gully and then accumulated at the outlet of Yigong lake. The rock avalanche-flow landslide travelled a horizontal distance of 8.5 km with a vertical elevation difference of 3000 m. The maximum velocity of the slide is more than 44 m/s. The immediate consequence which would later lead to disaster was that a natural dam was formed. The dam looked like a fan which was 4.6 km in length, 3 km in width with the height varying from 60 m to 110 m. The volume of the dam is about 3 × 108 m3 . The dam blocked the outlet of the lake which rasied the water level substantially. The rock avalanche-flow landslide is that it destroyed totally 8 km2 forest. Yingong tea plantation was flooded. About 4000 people were affected by the disaster. Fortunately, no one was killed.The directly economic losese were more than 20 million dollars. After the landslide, a man-made channel was dug on the natural dam in order to drain the water off from the barrier lake. However, the channel was washed away after 62 days later, at 7 p.m., on 10 June and the water bursted from the barrier lake. The velocity of the flood

was up to 9.5 m/s, and the discharge of the water was as much as 2940 m3 /s. The flood fiercely raised the water level of Yigong River, Palong River, and Yaluzangbu River. All of the bridges on the rivers were destroyed. A lot of roads and communicate equipments which were set along the road damaged. Furthermore, more than 30 secondary avalanches, slides, and debris-flows were triggered by huge flood. The object of paper is to present detailed description of the rock avalanche-flow landslide as well as to point out some distinctive features such as depositional cones, liquefaction holes, and twisted-off trees. 2 2.1

GEOLOGICAL SETTING Topography and geomorphology

The rock avalnache-flow landslide is located in the Yigong area, the northen part of Tibet. The gully presents a narrow and deep rocky gorge. The lowest elevation is 2188 m at the bottom of Yigong River. The highest elevation is 5520 m on the peak of the snow mountain above the Zhamulong Gully. The elevation difference is 3332 m from the peak to the bottom. The rock of the mountain has much potential energy due to the high steep cliffs and therefore avalanche prone to happen. Zhamulong Gully is shaped like a ‘‘takenup bag’’ with a wide up-slope entrance area and a very narrow down-slope out-let area. The height of the sides of the gully ranges from 150 to 200 m at its outlet. The bottom of the outlet of the gully only has a width of 50 meters. The top of the gully has a width of 300 meters. Moreover, there was a huge volume of colluvium deposited in the gully, with an elevation

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difference of 1100 m, cracks well developed in the gully. Such landforms provide a basic condition for a high-speed landslide. According to history records, a large-scale landslide, which was very similar to the one of April 9, 2000, had occurred in the year of 1900 in Zhamulong Gully. The debris blocked the outlet of Yigong River and built a channel type lake-—Yigong lake. After that a lot of small-scale landslides occurred in Zhamulong Gully. As a result, more and more debris accumulated at the outlet end of the gully and formed a natural fanshaped dam. It is calculated roughly that the volume of the dam was increasred to 6.95 × 108 m3 before the latest rock avalanche-flow landslide occurred. After the previous landslide which occurred 100 years ago, the type of the lake had changed from initial channel type into a braid type lake due to the gradually infilling. The catchment of the lake was about 9.8 km2 just before the rock avalanche-flow landslide occured. 2.2

Geological setting

Zhamulong Gully is chiefly formed in metamorphic rocks and granite. The metamorphic rocks belong to the Gangdisi group, and outcrop below 4100 m elevation. These rocks consist of marble, schist, gneiss and some mylonited hornstone. The metamorphic rocks generally dip to the SSW, and the dip angle is more than 30◦ . The granite outcrops above the 4100 m level and is part of the Jianggula granite botholith which was intruded into the metamorphic during the Himalayan orogeny. Overlying the bedrocks is a large volume of Quaternary side wash and colluvium in Zhamulong Gully and on the banks of Yigong River. In addition, there are some alluvial and terrace deposits in both banks of Yigong River. 2.3 Structural geology The landslide area (Zhamulong Gully) is located at the intersection of two faults. One fault named YigongPalong fault strikes northwest to southeast. The other named Yigong-Lulang fault strikes northeast to southwest. The water is identified as a strike-slip fault. There are four direction of joints in the area. The strikes are NE∼ENE, NNW, NNE and NWW. The NNW and NE∼NEE joints are much more developed than the other sets In fact, the rockf avalanche-flow landslide is mainly affected by the ENE joints directly. 3

Gully had occurred for one year before the rock avalanche-flow landslide. The temperature of the region had been relatively high for several days before the huge avalanche happened. It made much snow and ice thawed into water which increased the supply of water to the area of the landslide. Furthermore, it rained in the day on which the rock avalanche-flow landslide occurred. Thus, a large quantity of water seeped into the joints which softened the structure planes and rasied groundwater pressure. This is one of the main triggering factors of the rock avalanche-flow landslide. The rock avalanche-flow landslide occurred at 7:59 p.m. on April 9, 2000. At that time, people nearby the gully saw a huge stream of dust rising from the gully and heard a tremendous explosion sound. The landslide travelled very fast, dashed out from Zhamulong Gully, crossed over Yigong River, and stopped at the south bank of the river. A large stream of vapor and dust was seen to accompany the debris flow. Furthermore, some trees were damaged by the air blast was caused by the landslide. The total event of the rock avalanche-flow landslide lasted only about 3 minutes. 3.2 The key characteristics of the rock avalanche-flow landslide After studying the characteristics of the movement of the complex landslide, it is suggested to divided the Yigong rock avalanche-flow landslide into three zones: rock avalanche section, flow landslide section II and deposition section III-see Figure 1. The characteristics of each zone are presented below. 1. Rock Avalanche Section [I] The rock avalanche section is located at the fountainhead of Zhamulong Gully. The elevation of the section is between 4300 m and 5500 m. So the elevation difference is 1200 m. The gully has been formed mainly in granite which with lots of joint structures. Among them, two groups with N58◦ E strike and dip 32◦ to southeast, and N48◦ E strike and dip angle is 59◦ to northeast. The pattern of two groups of jointsis

IMPORTANT CHARACTERISTICS OF THE ROCK AVALANCHE-FLOW LANDSLIDE

3.1 Description of the avalanche and landslide According to the descriptions from the people who live near the area, small scale avalanches in Zhamulong

Figure 1. landslide.

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Cross section drawing of Yigong rockfall-

part of the slide-bed are lower. Overall, the slide-bed is shaped like a ‘‘spoon’’, which dips to the outlet of the gully. The inlet side of Zhamulong Gully is wide, while the outlet area is narrow. Thus, the gully provides favorable condition for the deposit of abundant rock debris. Also provides a topographic condition for high-speed landslide and debris-flow if impacted by large volume of debris from above.

Figure 2. Landform of avalanche area after the failure (the direction of the view is about NE45◦ ).

of ‘‘X’’ type shearing joints. The joints stretch in very long distance. The main surface of rupture (Varnes 1978) of the avalanche is mainly formed along the joint planes. The basement of the avalanche body is formed as a‘‘V’’ type hanging gully (Figure 2), because the body of the avalanche is generally wedge shaped. It is suggested that the volume of the avalanche body is about 3∼4 × 107 m3 . According to the investigation, after the rock detached from the mountain, it dashed against the northwest side firstly and then against the southeast side of the gully. The vibration caused by the avalanche was extremly powerful that the tea-plantion who live 10 km away from the gully could feel and heart. 2. The Flow Landslide Section (II) The section can be subdivided into instantaneous high-speed slide sub-section and high-speed debrisflow sub-section. 1. Instantaneous high-speed slide sub-section (II-1) This sub-section distributed in the bed and side of Zhamulong gully from NE to SW. There was some colluvial deposit in it. Based on the remaining debris in the gully, it can be estimated that the original deposit in the gully might have a length of 2500 m, a width of 1500 m, and an area of about 3000 m2 before the landslide. The thickness of the accumulations ranged from 50 m to 80 m. Thus the volume of the original material was about 3 × 108 m3 before the landslide occurred. The flow landslide section has an elevation of 3700 m at on the crown and an elevation of 2800 m at the toe (Varnes 1978). Due to the steepness of this section almost all of the deposited materials were moved out of the gully, and the bedrock can easily be found in the gully bottom after the landslide.The bedrock was the slide plane of the instantaneous landslide. The north, east and west sides of the slide-bed are relatively higher, but the south side and the central

2. High-speed debris-flow sub-section (II-2) The moving materials slide very fast along the flank of the gully, grabbing and entraining rock blocks from the gully walls. The channel of the gully is orientated NE-SW with a length of 600 m. The bottom and top widths of its south bank are 150 m and 300 m, respectively. The bottom and top widths of its north bank are 50 m and 150 m, respectively. Its height is 150 m. Thus, the channel is shaped like an inverted trapezoid. The slope angle of the channel is 45◦ ∼70◦ . The bottom of the channel is flat, the slope angle is only 3∼5◦ , and dips to SW. The channel has been carved out of the metamorphic rocks of the Gangdisi Group. There were some pine forests in this sub-section but during the slide, almost all of the pine trees were dug out and carried away. There are obvious scratches on the surface of both sides of the channel. From the stretches we can tell that the movement direction of the landslide is from north to south and with a parabola track . 3. Characteristics of deposition section (III) The deposition section of Huge Yigong rock avalanche-flow landslide presents some very distinctive appearances as follow 1. The features of sub-sections (III-1) Reference to Fig 3, we can divided the deposition section into boulder accumulating sub-section (III-1), fragment and fine grain accumulating sub-section (III-2), scraping-cutting and accumulating sub-section (III-3), and blast affecting sub-section (III-4). Boulder accumulating sub-section (III-I) is situated in the central of the deposition section. It has a length of 2600 m, an average width of 800 m, and an average thickness of 70 m. The deposited materials of this sub-section consist mainly of angular to sub-angular rock fragments (>90%), in addition, there are some sub-rounded boulders. Some of them occur as subround (so called ‘‘spherical stones’’). In the north part, 80% of the rock fragments have a diameter more than of 3 m. The biggest one is 42 m, with a volume of 30,000 m3 . The long-axis of the fragments is set with the same movement direction. Toward the southwest, the diameter of the rock fragments become smaller gradually. The diameter of the rock fragments in the southwest is generally from 0.2 to 0.3 m. Furthermore, the long-axis of the fragments has been changed from parallel to the direction of the slide movement

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N

Figure 3. Geological map of the sediment deposition section of Yigong landslide.

to perpendicular to it. Usually, there are some voids between the larger fragments infilled with matrix of smaller sized particle. The ratio of the granite fragments and the metamorphic rock fragments is 9:1. But the metamorphic rock fragments are found mainly at the southwest end of the area. The topography of this sub-section gently dips from northeast to southwest generally. The dip angle is about 8◦ . Its central part is slightly higher than the other parts. There are some small depressions in the deposited materials on both west and east sides. Some fine debris and mud are deposited on the surface of the depressions. The fragment and fine grain accumulating subsection III-III is situated in the flanks and front of the boulder accumulating sub-section. Clastic, sand and silt are the main compositions in this sub-section, and with small amount of large fragments (usually less than 20∼30%). Furthermore, the content of big fragments becomes less towards the edge of the sub-section. The debris consists mainly of granite and metamorphic rock. Its topography mainly occurs as small depression. But in the front part of this sub-section, small hillocks and small depressions distribute alternatively. The long axis direction of the hillocks and depressions

is orthogonal to the direction of landslide movement (Figure 4). There are some unusual phenomena in this subsection. They are ‘‘depositional cones’’, ‘‘liquefaction holes’’, and ‘‘blast-spattered mud’’. There are a lot of depositional cones in the front of the sub-section . The depositional cone is a kind of unique micromorphologic landscape in landslide area. Dipositional cones usually occur in the positive landform areas in the sub-section. Talus cones in the center of the subsection are usually larger than those in the edge of the sub-section. The central ones have diameters which vary from 3 m to 17 m at the base, and have heights ranging from 1.5 m to 15 m, and the top of them is commonly pointed. The diameters of the depositional cones on the edge ranging from 1 m to 8 m at the base with heights varying from 0.5 m to 3 m; and the top of them are commonly rounded (Figure 5). Most of the depositional cones appear double layers structure. The inner layer consists of relatively large debris fragments, but the outer layer consists of reatively small debris fragments. Furthermore, some of them have a thin coat of very fine sand and mud. Sand craters also occur where a liquefied mixture

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Figure 4. Alteranting hillocks and depressions in the fine grain accumulating sub-section.

Figure 6. sand crater in fragment and fine grain accumulating sub-section (hammer is for the scale).

Figure 5. Depositional cone with a round top and a coat of mud.

of water and sand has spouted out from the debris that occured (Figure 6). We call ‘‘liquefaction holes’’ (Zhang et.al 2002). These holess are very similar to the sand boils and ground cracks after an earthquake. The diameter of the craters range from a few centimeters to 30 cm, and the depth of them are no more than 10 cm. Some fine grain sands are found in the bottom of the holes. Scraping-cutting and accumulating overlap subsection (III-III) is situated out side of sub-section (III-II). The original topography is gully. The elevation is obviously higher than sub-section III-1. Some places of the area is only scraping and cutting without the deposition. In this area, the dip angles of the slopes ranges from 40◦ to 50◦ . The bedrock composed by the metamorphic rocks of Gangdisi Group. These were overlain by soil and eluvium. There was a dense forest on the sides also. After the landslide, the soil, eluvium and forest were scraped and cut totally only left a trimline. There are some large striations, plough marks and dash pits on the bedrock (Figure 7). In

Figure 7.

Striations on the eroded flanks.

some places, the scraping-cutting zones are overlain by patches of accumulaed materials. These places are located in the lower reaches of Zhamulong Gully and south bank of Yigong River. During the landslide, the forest was totally destroyed and about 0.5 to 5 meter thick of the soil and eluvium were scraped. Simultaneously, some new debris was accumulated in the channel. The eroded surfaces are from 6 m to 45 m higher elevation than the zones of accumulating. Blast affecting sub-section (III-IV) is situated outside of the accumulation area. There are only a few small debris with diameters less than 40 cm in the subsection, which was thrown to the surface by the blast. But 70% to 90% of the trees grown in this sub-section were pulled out from the roots or cut off in the middle. All of the damaged trees were laid down in the same direction as the blast movement (Figure 8). Some of

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The landslide scraped deposits within and on the flanks of the gully and destoryed a large number of trees. The landslide blocked the Yigong River and caused secondary disasters. 2. Some distinctive features of Yigong rock avalancheflow landslide include:

① After the materials moved out of the outlet of ②

Figure 8. Destroyed trees laid down in the same direction in blast affecting sub-section.

③ ④

them were thrown about 100 m by the blast. Furthermore, some trees were cut into pieces of wood. In the front of this sub-section, there is a thin layer of mud on the surface of both the damaged and undamaged trees. This kind of mud was called ‘‘blast-spattered mud’’ This indicats that the agent that was damaged the trees is not the debris of the slide, but the blast which was produced by the high-speed moving air current. It is indicated by the field investigation that when the landslide move into the gully, its energy was concentrate because of the narrow topography of the gully. However, after the debris rushed out from the gully and into the river, its energy was released quickly because the topography suddenly becomes much wider. This made the boulders and debris carried by the slide suddenly spread and deposited in front of the gully and formed a huge accumulation fan. The fan is very flat. Its slope angle is only 8◦ . Some alternating hillocks and depressions distribute alternatively in the front of this sub-section. The pattern of the topography is very like the that of a desert.

⑤

the gully, their moving track is shaped like a parabola. The accumulation distribution of the slide can be divided into several zones. The boulder accumulating sub-section is situated in the central part of the gully. It is surrounded by accumulations of pebble, sand, silt and clay. A wide air blasting and water spraying section surrounds the accumulations. The deposition of the material is characteristic of ‘‘stretching in level’’ along the movement direction and ‘‘spreading out’’ across the movement direction. There are some distinctive topographic features, such as depositional cones, liquefaction holes and trees damaged by blast wind.

3. The whole geohazard chain sequentially is mountain avalanche, landslide, mudflow, river blocked and establishment of the Barrier Lake, burst of the lake, flood, secondary avalanche and landslide. ACKNOWLEDGEMENTS The authors wish to acknowledge Mr. Baosheng Wang, Mr. Baoben Xia, Mr. Wei Liu, Mr. Xiangde Fan, Mr. Zhihua Wang, Mr. Ning Liu, and Mr. Jianjun Jiang for their valuable help. Thanks also to Mr. Jidong Sun, Mr. Chuanglu Xu, Mr. Jietang Lv, Mr. Jia Jialin, and Mr. Yaoping Yin for instructive conversation. REFERENCES

4

SUMMARY

The following summaries can be given on the study of the Yigong rock avalanche-flow landslide. 1. The process of Yigong huge rock avalanche can be described as: the stability of the rock mass was greatly reduced because of the large amount of rainwater and the water melt from snow which seeped into the rock. This caused about 3 × 108 m3 rock failure from 5000 m elevation and fell down 3000 m to the loose debris on the bottom of Zhamulong Gully. The rock avalanche mobilised a large quantity of colluvium in the gully, resulting in a fast-moving landslide along Zhamulong Gully.

Oldrich Hunger, Evans S. G., Bovis M.J. & Hutchinson. J.N. 2001. A Review of the Classification of Landslides of the Flow Type. Environmental & Engineering Geoscience, Vol.VII, No.3, 221–238. Ren Jinwei, Shan Xinjian & Shenjun. 2001. Topographic and dynamic study on rock avalanche- landslide—debirs flow in Yigong, Tibet [J]. Geology Comment, 47 (6) 642–647 (In Chinese). Zhang Zhuoyuan, Chen Shangming, & Tao Lianjin. 2002. 1983. Sale Mountain landslide, Gansu Province, China, in S.G. Evans & J.V. DeGraff, ed., Catastrophic Landslides: Effects, Occurrence, and Mechanisms: Geological Society of America, Reviews in Engineering Geology, v. XV, p.149–163.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Key issues of emergency measures and comprehensive remediation projects to control the Danba landslide, Sichuan province, China Qiang Xu, Xuan-mei Fan, Liang-wei Jiang & Peng Liu The State Key Laboratory of Geohazards Prevention and Geoenvironment Protection, Chengdu University of Technology, Chengdu, Sichuan, China

ABSTRACT: The Danba landslide is located behind Danba County in Sichuan Province, China. The total volume is about 2.2 × 106 m3 . It was triggered by the excavation of the foot of the landslide. The deformation become obvious in October 2004, and accelerated in February 2005, with the maximum value to reach 5 cm per day. A large arc main scarp appeared in the crown, which connects cracks on the flanks. If the landslide occurred as a whole, the entire Danba County and its 4600 people would be in extreme danger. The related agencies have taken emergency aid measures to reduce the deformation of the landslide as soon as possible. The paper studied and discussed some key problems of the technical measures such as monitoring, early warning, prediction, and emergency decision making).

1

INTRODUCTION

Danba County is located in the west of Sichuan province, China, and behind the county there exists a very large ancient landslide deposit which is named of Baijiashan. Being restricted by the topography condition, the county had to evacuate the foot of the deposit to build constructions. This behavior made the ancient landslide reactived. In February 2005, the movement of the landslide became very obviously and prone to slide down. According to the primary calculation, if the landslide occurs in total, the entire Danba County and its 4600 people will be in extremely danger. Furthermore, the landslide will block the river and be a threat to people who live in the upper and lower catchments of Dadu River. In order to ensure the safety of the people in the county and keep the stability of the society, the related governments organized a group of experts to implement the twenty four hours monitoring and early warning. In the meanwhile, it took emergency construction measures to stop the landslide sliding down. More than 7000 m3 sand bags were put in the toe area of the landslide and 6 rows of tensioned anchors were inset into the front part of the main body. With such efforts, the deformation speed of the landslide become under controlling. However, for the long time stable of the landslide, we need to execute the comprehensive remediation project based on the emergency construction works. The comprehensive remediation project completed in the early month of 2007, and then the landslide is stable which can be telling from the

monitoring data. The object of the paper is to discuss some key problems in the whole process, such as the implement the monitoring and early warning plan; the forecast of the landslide and decision making during such emergency condition.

2

THE DANBA LANDSLIDE [1][2]

The landslide resembled as a chair. It is 200 m to 250 m wide 290 m long 30 m thick and with the volume of about 2.2 × 106 m3 . It is classified as the large deposit landslide. The depletion zone is composed by loose ancient landslide deposit. The loose deposit is mainly composted by the sand and macadam. According to the important literature (Q. Xu et al. 2006), we divided the landslide into three sections with the consideration of its deformation process, mechanism and the sliding characteristics (Figure 1). 2.1

Section I

Section 1 can be further divided into two subsections. Subsection I-1 is the initial failed slid area, which is the real main zone of detachment. Its volume is 1.5 million m3 . The main moving direction is 353◦ (travels the valley). Deformation started since 2003, and till December 2004, the deformation in the toe area became obvious and showed up features which indicated the landslide entering into an acceleration deformation stage. To 20 February 2004, the tension

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3 3.1

Figure 1. The Danba landslide and Danba county.

cracks on the crown and on the flanks almost continued and closed, the toe moved ahead to the street surface. Ten more buildings were destroyed by such push. Early in March 2004, the one meter wide large tension crack on the crown developed. Therefore, the deformation continued expanding to the back part of the main body and formed the sub slid body in the section I-2. Because the flanks are steep slope in section I, lots of materials slide down during the move process which developed lots of cracks and become through gradually, so section II and section III established. Effected by the landform, the moving directions of section II and III are different from section I, both pointing to section I at an angle (Figure 1).

MONITORING EARLY WARNING AND EMERGENCY WORKS Summary

Deformation process is a reflection which is effecting by both the internal and external structural factors (Jin Xiaoguang et al. 2000). As to the landslide which is on the slid margin, implement the monitoring project is the key issue. Started form the 21 January, 46 professional monitors and 26 simple monitors been installed on the main body, and someone was send to observe the deformation behavior time to time. In the meanwhile, three deep displacement monitor holes were drilled on the centerline of the main body (ZK10, ZK12, ZK13 in Figure 2 and Figure 3). The surface displacement indicated that the landslide entered into an accelerated deformation stage in the middle of February 2005. Experts studied on the monitoring data before 20 February, by using the existed forecast module, then came to the conclusion that if no action is taken to stop the deformation, the landslide will slide down totally between 2 March and 15 March. The experts made evacuation plan and took some emergency measures immediately. The evacuation scale includes the direct dangerous area and the potential dangerous area and evacuated 4923 people in the direct dangerous area. People in the potential dangerous area received early warning cards in case

2.2 Section II According to the investigation results, the volume of section II was about 5 × 104 m3 , lots of tension cracks developed on the crown and because that the left flank is a steep slope (slope angle varies from 50 to 80 degree), some small scale collapses often happened on it. Two large collapses happened on March 9th and March 14th of which volumes reached to hundreds m3 . Thanks to the in time early warning, there is no one hurt in such collapse event.

2.3

Section III

Section III located on the right back of main body with the volume of 3.5 × 105 m3 . Referring to the monitoring data, its deformation scale is smaller than the section I and section II, and with the tension anchor project carrying out in subsection I-1, the deformation become slightly and stopped gradually.

Figure 2. The layout of the emergency aid and comprehensive remedial project.

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Figure 4.

The velocity of deformation in monitor 9.

Figure 5.

The accumulated displacement in monitor 9.

Figure 3. The layout of the comprehensive remedial project and the deep displacement monitors locations.

they were in danger. The emergency measures include: stacking the sand bags in the toe area in order to slow down the sliding speed and save time for the tensioned anchor project; constructing the tensioned anchors in the section I, from row A to F and 244 anchors; adding row G and H on the top area in section II, 40 anchors in total, and also some netting on some collapse area on section II (Figure 2 and Figure 3). The first anchor in section I was tensioned on March 15th and the whole project finished by the end of April. 3.2

Analyse on the monitoring data and emergency measures

(1) Displacement data The velocity of deformation curve (Figure 4) which is data from the monitor 9 on the back area of section I indicated that the acceleration deformation of the main body was start from February 6th, and reached to the maximum value of 30.3 mm on February 22nd, then the deformation slowed down. The Figure 5 shows the accumulated displacement on the back area in section I reached to 1.2 m. The big tension crack on the crown almost seperated the main body in subsection I-1 from the slope. In addition, the monitoring data shows the deformations on the crown were much larger than the ones on the toe. (2) Deep displacement data The bedrock of the slope was located in a very deep distance and the maximum depth reached to 45 m in the crown. During the emergency investigation, obvious slid plane was not found so the contact surface between the bedrock and the loose materials could be considered as the slid plane. However, in order to figure out the real location of the slid plane, three deep monitor holes ZK12, ZK10, ZK13, (Figure 2 and Figure 3) was installed along the centerline of subsection I-1 during the late period of the emergency aid. The curves which

show the relation between the depth of the hole and the displacement were given in Figure 6 to Figure 8. From the results of the deep displacement monitoring, the location of the real slid plane can be found. The location was a big difference as it was considered before. For instance, the monitoring data in ZK12 shows the real slid plane is 29 m underneath (Figure 6), but the depth of the contact surface is 36.32 m in the same place, the difference is 7.32 mm; the monitoring data in ZK10 shows the real slid plane is 28.5 m underneath (Figure 7), but the depth of the contact surface is 41.29 m in the same place, the difference is 12.79 m; The slid plane shows as a belt in ZK13 the depth is between 12 m and 25 m while the depth of the contact surface is 31.43 mm (Figure 8). Therefore, the Danba landslide was not reactive along the contact surface which is between the bedrock and the loose cover, most of the failures occured in the loose materials. (3) Emergency measures ① Utilize the monitoring data to analysis the comes out of the emergency aid. The monitoring data both from the surface and the deep place reflected the result of the emergency aid project. It can be seen clearly from Figure 4 and Figure 5 that the landslide experienced a start-stop process which was from the acceleration stage in the early February to the slowing down stage after 22 February then to the stable stage after middle of

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2005 - 4- 23 2005 - 4- 30 2005 - 5- 6 2005 - 5- 30 2005 - 6- 5 2005 - 6- 28 2005 - 7- 1 2005 - 7- 10 2005 - 7- 30 2005 - 8- 2 2005 - 8- 24 2005 - 9- 3 2005 - 9- 30 2005 - 10- 4 2005 - 10- 15 2005 - 10- 31 2005 - 11- 3 2005 - 11- 15 2005 - 11- 19 2005 - 11- 22 2005 - 11- 25 2005 - 11- 28 2005 - 12- 1 2005 - 12- 4 2005 - 12- 7 2005 - 12- 10 2005 - 12- 13 2005 - 12- 31 2006 - 1- 18 2006 - 2- 20 2006 - 4- 6 2006 - 4- 28 2006 - 5- 18 2006 - 6- 19

10

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2005 - 4- 17 2005 - 4- 30 2005 - 5- 3 2005 - 5- 15 2005 - 5- 30 2005 - 6- 2 2005 - 6- 12 2005 - 6- 28 2005 - 7- 1 2005 - 7- 10 2005 - 7- 13 2005 - 7- 30 2005 - 8- 2 2005 - 8- 15 2005 - 8- 31 2005 - 9- 3 2005 - 9- 15 2005 - 9- 30 2005 - 10 - 4 2005 - 10 - 15 2005 - 10 - 31 2005 - 11 - 3 2005 - 11 - 15 2005 - 11 - 19 2005 - 11 - 22 2005 - 11 - 25 2005 - 11 - 28 2005 - 12 - 1 2005 - 12 - 4 2005 - 12 - 7 2005 - 12 - 10 2005 - 12 - 13 2006 - 1- 18 2006 - 2- 20 2006 - 4- 6 2005 - 12 - 31 2006 - 4- 28 2006 - 5- 18 2006 - 6- 19

6

20 22 24 26 28 30 32 34 36 38 40 42 44 46 48 -20

0 20 40 60 80 100 120 140 Cumulative Displacement (mm)

Figure 7. Relation between the displacement and depth of the monitoring hole collected from inclinometer, ZK10.

110

Cumulative Displacement (mm)

Figure 6. Relation between the displacement and depth of the monitoring hole collected from inclinometer, ZK12.

April. During the stacking sand bags period from February 22nd to March 1st, the deformation velocity dropped down from 30.3 mm/d to 20 mm/d. This action made the landslide sliding slowly and saved enough time for the tensioned anchor installing project. The tensioned anchor row A to row F started to be installed in end of February. On March 15th, the first anchor becomes tensioned, but due to the limit amount of the anchors, the landslide deformation did not effect by such slight behavior. Till April 15th, the total 244 tensioned anchors in section I were being installed, and 166 anchors of them were pretensioned by different stress. After this, the results of the anchor project become obviously. Seeing from Figure 4, after April 15th, the velocity of the deformation in the front part of the main body slowed obviously and come to stop gradually, the displacement was 1 to 2 mm per day. The displacement in the middle of the main body dropped down to 5 mm per day. Figure 6 to Figure 8 show the velocity of the landslide deformation decreased from April 2005 to May 2006. All the

evidence show that the emergency aid project carried out successfully. ② Analyze the results of the emergency aid project by utilizing the numerical simulation Based on the monitoring data of the landslide deformation and used the back analysis method (Zhang Wuyu, 1999, Gao Dejun, 2006), also combined the investigation results, the parameters of the rock and soil on the landslide are determined. The numerical simulation by FLAC3D (Coetzee et al. 1993) also implement for the analysis of the results of the emergency aid (Figure 9 and Figure 10). By comparing Figure 10 with Figure 11 it can be seen clearly that after the aid project, the area of the plastic zone on the main body become reduce heavily. The plastic zone near the slide plane developed through to the shear strain increment zone. 3.3

Key issues of the monitoring and early warning as well as emergency aid

1. It is very necessary to reinforce the importance of the monitoring and timely data analyse is on the landslide which is on its slide margin. In the early time of the emergency aid, we pointed out

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Depth (meters)

three critical evidences from the analyzing on the characteristics of the Danba landslide: ① several small scale collapses occurred at the same time in the toe are of the landslide; ② the velocity of the deformation increased sharply; ③ according to the threshold time of the landslide to give early waning, time period should shorter than 24 hours, one of the three conditions turns out, we suggested the implement of evacuation of people both in the direct dangerous area and the potential dangerous area immediately. 2. Identifying the mechanism and deformation stages of the landslide correctly is the foundation of the early warning project. During the emergency aid, the experts come to a conclusion that it is possible to stop the landslide sliding down totally if we undertake some effective measures when the landslide just enters into its acceleration stage. As 0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 -20 -10 0 10 20 30 40 50 60 70 Cumulative Displacement (mm)

2005 - 5- 12 2005 - 5- 21 2005 - 5- 30 2005 - 6- 2 2005 - 6- 15 2005 - 6- 28 2005 - 7- 1 2005 - 7- 10 2005 - 7- 13 2005 - 7- 30 2005 - 8- 2 2005 - 8- 11 2005 - 8- 22 2005 - 8- 31 2005 - 9- 3 2005 - 9- 9 2005 - 9- 19 2005 - 9- 30 2005 - 10- 4 2005 - 10- 15 200 - 10- 31 2005 - 11- 3 2005 - 11- 9 2005 - 11- 15 2005 - 11- 22 2005 - 11- 28 2005 - 12- 1 2005 - 12- 4 2005 - 12- 7 2005 - 12- 10 2005 - 12- 13 2005 - 12- 31 2006 - 1- 18 2006 - 2- 20 2006 - 4- 6 2006 - 4- 28 2006 - 5- 18 2006 - 6- 19

we know the trigger factor of Danba landslide is the slope foot cutting, the most effective and simple remedial measure is to stack sand bags under the slope foot. Because of the fast right decision, we saved enough time for the later anchor project and avoided the landslide sliding down totally. 3. It is difficult to identify the slide plane of the Danba landslide by regular investigate methods, we decided to use the inclinometer to measure the deep displacement. In addition, we used the deformation curves to back analysis the parameters of the rock and soil in the landslide and built numerical simulation models to search for the slide plane.

Figure 9.

The calculated model in profile 2-2.

Figure 10. The shear strain plastic zone in profile 2-2 before the emergency project.

Figure 8. Relation between the displacement and depth of the monitoring hole collected from inclinometer, ZK13.

Figure 11. The shear strain plastic zone in profile 2-2 after the emergency project.

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When the inclinometer can not get the displacement data, the simulation method should be the best choice. The location of the slide plane is almost fitted from the FLAC3D numerical simulation and the deep displacement monitoring (Figure 12).

4

THE COMPREHENSIVE REMEDIAL PROJECT

4.1 Summary The entire emergency aid project was finished in June 2006. After the completion of the project, the velocity of the deformation reduced from 2∼3 cm/d to 1 mm/d, but the deformation haven’t stopped yet. The sand bags probably lost their effect after a long time period. Therefore, it is necessary to implement the comprehensive remedial project afterwards. It should include: installing 32 tensioned anchors- anti slid pills on the toe area of the landslide after clearing the sand bags; adding four rows of anchors and beam (L, K, J, I) from top to bottom in the middle and back areas of the section I, and enlarging the existing rows C and D to the east crack margin. there are 162 new ones in section I; adding four rows of anchors and beam (M, N, O, P) from top to bottom in the middle and back areas of the section III to ensure the long time stability (Figure 2 and Figure 3); building the ground surface water interception and drainage systems on the surface of main body.

experts thought the location should be between the rows A to F (Figure 3), which is to densify the anchors to reinforce the anchoring strength; the other experts thought the location should be on the middle and back areas of the man body (Figure 3), because the displacements there are always larger than the front area, it is more important to control the deformation on the middle and back areas of the main body. Two methods which are the deep displacement monitoring and the numerical simulation have been used to analyze which plans should we use in the remedial project. Because of these reasons:

The key issue of the comprehensive remedial project is which place is the best setting location of the new adding anchors rows (I, J, K, L) on the main body of the landslide. Two different opinions existed. Some

1. After carefully checking on the deep displacement data from ZK12, ZK10 and ZK13, it is easy to find there is a big difference of the deformation characteristics in the slide plane area, especially in under the slide plane. The data from ZK12 shows that the slide plane in the back area of the main body appeared very clearly, however, the data from ZK13 shows that the slid plane in the front area of the main body appeared as belts which distributed from the depth of 12 m to 25 m, and it is hard to tell where is the real location of the slide plane. The data from ZK10 also indicated the location of the slide plane, but the rock and soil under the slide plane have been disturbed. All the phenomena could be explained as below: The slide plane had already formed before the emergency aid and made the downslope movement of the landslide, fortunately the movement was stopped by the emergency actions, however, the actions did not stop it completely but just slowed it down. And after the remedial project, the results are different from the location to location. The back area almost did not effect by the project which mean the main body still sliding down along the slide plane. This is the reason for we can see the slide plane clearly in ZK12. The front area of the main body showed the results

Figure 12. The connection zone of the shear strain plastic zones and the sliding plane monitored by the deep displacement monitors displayed by FLAC3D .

Figure 13. The shear strain plastic zone in profile 3-3 (where the locations of the I∼L anchors are in the front areas of the landslide main body).

4.2 Analyze on the key issues of the comprehensive remedial project

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were coming out during the process, then come to the conclusion given as below:

Figure 14. The shear strain plastic zone in profile 3-3 (where the locations of the I∼L anchors are in the middle and back areas of the landslide main body).

of the project and turn to slow down gradually. At the same time, it is also pushed by the sliding body from the middle and back areas, so the data form ZK13 shows a pushed deformation belt formed in the deep part. From these analyses, we decided to set the anchor rows in the middle and back areas of the main body (Figure 3). 2. Comparison has been made between the outcomes of the two plans by using the FLAC3D numerical simulation (Figure 13 and Figure 14). If the new adding rows were setting on the front part of the main body, it will cause a large scale plastic zone in the middle and back areas which show in the cross section 3-3 (Figure 2), the partial deformation will deteriorated (Figure 13). But if we set the anchors in the middle and back areas, the plastic zone will get smaller which makes the landslide more stable (Figure 14). Therefore, it should come to the conclusion that the newly added anchor rows I to L be installed on the middle and back part of the main body of the landslide. In June 2006, the anchor project on rows I to L was almost finished. The monitoring data on June 18th 2006 (Figure 6 to Figure 8) indicated that the accumulated displacement of the ZK12, ZK10 and ZK13 on the slid plane and the drill hole mouth from may 2006 to June 2006 decreased obviously which indicated the good result of the new anchors. Install project. 5

1. We should reinforce the monitor and identification of the omen of the landslide’s marginal slid phase on such landslide which is prone to slid, and utilize the monitoring data to forecast the landslide as well as to make the comprehensive remedial project. This is considered to be the best effective method to avoid such kind of landslide to occur. 2. The effective decision making on the emergency aid and remedial project is relying on the deep analysis and utilizing the monitoring data. At the same time, the monitoring data is an effective evidence to prove the quality of the results of the emergency aid and the comprehensive project. The paper utilized the monitoring data to judge the results which come out from different phase and different practical measures. In addition, it suggests the widely used numerical simulation method in analyzing the results of the remedial project and helping on making decision on which remedial measure should be taken. However, the most important standard in judging whether the numerical simulation is fitting for the original geological module should be the checking on how it fits the monitoring data. To conclude, monitoring plays an important role in the whole process. REFERENCES Xu Qiang, Huang Runqiu, Fan Xuanmei, et al. 2006. Study of monitoring, prewarning and control engineering for the Danba landslide, China. IAEG2006 Engineering geology for tomorrow’s cities, 2006: 3–562. Jin Xiaoguang, Li Xiaohong, Wang Lansheng, et al. 2000. Characteristics of landslide deep displacement curve and stability discriminant. Journal of Mountain Science, 18(5): 440–444. Zhang Wuyu. 1999. Methods and applications of parameter back analysis. Journal of Qinghai University (Natural Science), 17(6): 6–10. Gao Dejun, XuWeiya & Guo Qida. 2006. The parameter back analysis applied in the big block landslide calculation in Three Gorges area. Journal of Hehai University (Natural Science), 34(1): 74–78 (in Chinese). Coetzee M. J. et al. 1993. FLAC basics. Itasca Consulting Group Inc.

CONCLUSIONS

The paper gave described details in the emergency aid and the comprehensive remedial project of the Danba landslide, and analysis on the key problems which

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Enhanced slope seepage resulting from localized torrential precipitation during a flood discharge event at the Nuozhadu hydroelectric station Mo Xu, Ying Ma & Xiaobing Kang State Key Laboratory of Geo-Hazard Prevention and Geo-Environment Protection, Chengdu University of Technology, Chengdu, China

Guoping Lu Earth Sciences Division, Lawrence Berkeley National Laboratory, Berkeley, California, USA

ABSTRACT: The Nuozhadu Hydroelectric Power Station, located on the Lancang River in Yunnan, China, represents a fairly typical high dam within a narrow, V-shaped canyon. Its flood control in rainy season requires concentrated release of surplus flood water, producing intensified misty rain over the energy dissipation area. This intensified misty rain would spread over a large area and fall on a steep slope (s), producing a higher infiltration condition lasting for days. This occurrence causes concern over the water table rise and saturation increase in the slope, especially near the foot of the slope where the stress is often focused. In this paper, we developed a 3-D water table flow model for a fractured granitic groundwater system that is recharged by rain precipitation. Results show that the precipitation from the flood release could elevate the water table by an amount greater than what would be a seasonal fluctuation. The modeling results provide the basis for further analysis of the effective rock stress in the slope caused by increased saturation, leading to the reduction of slope stability. This slope-flow study represents one of the first field study cases in this active research area of intensified infiltration during flood discharge events at large hydroelectric stations with high dams and narrow V-shaped canyons. Its methodology and general conclusions might be generally applicable to other slope sites within hydroelectric power stations.

1

INSTRUCTIONS

During the flood season at a typical hydroelectric station, the discharging (when necessary) of its excessive water could cause localized rain precipitation and mist and result in enhanced slope seepage in the site’s downstream slope (s). The precipitation resulting from such a discharge can be as intense as a torrential storm. For example, at the Nuozhadu Hydroelectric on the Lancang River in Yunnan, China, the localized rain precipitation caused by the discharge can last for as long as 5 days, and the core fraction of the rain precipitation could be as heavy as 200 to 300 mm per hour. These conditions could extend as far as 500 m along the river, 300 m across the river, and 200 m in altitude above the river. (Figure 1) Such heavy rain precipitation can cause alterations in slope flow conditions around the energy-dissipation area of the discharged floodwater, which in turn adversely affect the stability and safety of the engineering slope. Conventional practice treats the subsurface as saturated, neglecting the suction of the rock matrix associated with unsaturated conditions.

Figure 1. Plan view of the Nuozhadu dam site at Lancang River in Yunnan, China.

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Under saturated conditions, rock sheer strength has an additional contribution from matrix water-capillary pressure (suction). Hence, unsaturated rock has higher sheer strength than saturated rock. Thus, assuming a saturation condition when estimating what is in fact an unsaturated condition could lead to a severe underestimation of the effective sheer strength, what could reduce unnecessarily conservative and expensive engineering in protecting the slope. The Nuozhadu site has a tropical climate with distinct dry and rainy seasons, and an annual average air temperature of 21.7◦ C. Its long-term annual precipitation rate is 1,162 mm, with 84.5% of this rate contributed during the rainy season. Maximum precipitation intensities measured during the rainy season (mid-June to mid-September) from 2002 to 2004 are listed in Table 1. 2

ESTIMATION OF MISTY AREARESULTING FROM DISCHARGING EVENT

(Table 2), including Ertan and Laxiwa, which have dam heights and water-level drops similar to Nuozhadu Station. Definitions for rain precipitation and mist phenomena in the following discussion are in order. Mist occurrence is referred to as the rain precipitation and fog flow generated from flood water discharging down the spillway, forming a water tongue that diffuses into and mixes with air and splashes upon the water in the lunge pool. At the Nuozhadu site, mist forms as flood water discharges from the left bank of the river into the lunge pool, splashing into the air and then falling on the surrounding areas, including the right-side bank of the river (the study area for this paper). The estimated size of the mist area is listed in Table 3. Mist areas are also delineated in Figure 1, with Area 1 susceptible to moderate-to-strong mist rain storm (>16 mm per hour), Area 2 to dense and thin mist-rain (16∼2.5 mm per hour), and Area 3 to thin mist-rain (0.5–2.5 mm per hour)/light mist-rain (100.0 131.0 150.0 65.0 51.0 160.0 210.0

300.0

Engineering names

Max. Dam height (m)

Dense mist area

Wujiangdu Baishan Liujiaxia Fengtan Tuoxi Quanshui Hongyan Laxiwa Ertan

165.0 149.5 147.0 112.5 104.0 80.0 60.0 250.0 240.0

510.0 420.0 500.0 360.0 312.0 180.0 130 670 745.0

Thin and light mist area 900.0 900.0 700.0 800.0

1720

390 360

1690

Figure 2. Plan view of geological map at the 680-m Horizon.

3

SITE DISCRIPTION

The study slope site is located southeast of the Nuozhadu Station, directly across from the river, on the right side of the lunge pool (Figure 1). The slope runs N53 W, dipping NE with an undisturbed natural slope angle of approximately 45◦ . A field survey shows that the surface-covering layer is generally about 1∼2 m thick, and bedrock is of granitic origin. The right-bank’s slope (across from the lunge pool) exposes granitic rock of late Variscan-Indosinian period (γ43 ∼γ51 ), as well as Quaternary alluvium (Qal ) and a pedalfer layer (Qdl ). The plan view of the layering at the 680 m horizon is shown in Figure 2, and the cross sections marked in Figure 2 are plotted in Figure 3. The three profiles shown in Figure 3 are the geological sections of the slope, characterized by boreholes and horizontally dug adit. As mentioned above, layering of the granitic units is primarily based on degree of weathering in the rock, which is controlled by the trends of the land surface and correlated with the relative depth below the land surface. Structure surfaces marking the discontinuity of rock structure exist on different scales at the slope site. They are compressed in nature. The F1 fault and F2 fault are regional faults (Figures 1 and 2), with the F1 fault coinciding with the Huoshaozhai Valley and F2 running along the southeast edge of the site. In addition, structural surfaces (of compression nature) from 10 to 58 m long were found at the field site (Figures 2 and 3). Fractures in the granite generally close, short, and disconnected, are found in the adits. The layering of hydraulic properties at the field site (shown in Figure 3) is based on those geological layers identified by field surveys, by the six boreholes (80.08 to 120.18 m in depth), and by the eight adits.

Figure 3. Geological cross section along sectional profiles (in Figure 2).

4

HYDRAULIC CONDUCTIVITY OF FRACTURED ROCKS

The hydraulic conductivity tensor is determined by anisotropic properties of the fractured media’s structure. Further detailed discussion can be found in Snow (1965, 1969). The average measured values for hydrological layers are listed in Table 4. These values follow a normal distribution by geological layer. The hydraulic conductivities measured from the field tests are of composite and anisotropic nature arising from fracturing: their vertical values are approximately half its corresponding horizontal value.

5

CONCEPTUAL AND NUMERICAL MODELS

These field observations at the slope-study site determine the choice of an appropriate model. The rocks

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were treated as a single continuum, based (as mentioned before) on observations from the drilled cores and horizontally dug adits. Rock fractures are short, sparse, and disconnected. The model domain covers the slope on the right bank of the lunge pool. Groundwater flows from the southwest headland to the Lancang River bed. The X-axis starts from the Huoshaozhai Valley and extends to the F2 fault area, spanning about 1,300 m, whereas the Y-axis is approximately perpendicular to Lancang River, starting from a portion of the elevated flat in the southwest and extends to the left bank of the river for approximately 980 m. The vertical Z-axis extends from an elevation of 400 m msl, up to the maximum hill elevation (955 m). The domain is divided into 98 rows and 130 columns of equally spaced 10 m intervals in both directions, vertically discretized into 23 layers 0.2–30 m thick (to be detailed later in this section). In the regional flow direction, the model is bounded by noflow boundaries along the Huoshaozhai F1 fault on the northwest side and the F2 fault on the southeast side. The bottom was also treated effectively as a noflow boundary, with the bottom layer set to a location sufficiently below the water table. No clearly defined hydrogeologic boundary features exist along the southwest upstream boundary of the mesh. A head-dependent flux boundary was therefore imposed on this side of the mesh, to minimize the artificial boundary effect in the solution. The head was approximated from the trend of the land surface. Infiltration rates initially take 15% of the observed regional multiple-year annual precipitation rates as a first guess (later to be adjusted through model calibration). Modeling layers, constructed along the geological layers, Table 4. Measured hydraulic conductivity (K) at the Nuozhadu. K (m/s) Borehole

Qdl

γ43 ∼γ51 w

ZK441 ZK443 ZK445 ZK449 ZK551 ZK553

9.87E-07 8.00E-07 2.85E-07 5.92E-06 4.05E-07 7.67E-07

2.47E-074 7.37E-07 6.86E-07 2.07E-06 3.50E-07 4.74E-07

γ43 ∼γ51 f 7.05E-07

2.90E-07

were assigned the hydraulic conductivity values listed in Table 5. Variable discretization is imposed in the vertical direction to accommodate thickness changes in the layers. (A layer may vary in thickness, and layers may differ from one another in thickness.) Changes in the borehole water table data. This adjustment was justified because head data and recharge are sensitive parameters within the flow model (as discussed later). For the estimation of recharge rate, we define an objective function to be minimized as the root mean squared of the residual for calculated head vs. observed head at the boreholes. First, the hydraulic head at the upstream boundary was adjusted until an optimal residual was obtained, with the infiltration initially estimated to be 15% of the average annual precipitation. Second, the preliminary estimated upstream boundary head was used for the estimation of recharge rates. Using this upstream boundary head as the basis, the resultant model was then run to obtain the objective function values at varied recharge rates. The model was then modified for a number of sets of overall increase or decrease in the upstream boundary heads, and each set was run for different recharge rates. The optimal boundary heads were chosen based on the residual mean and thickness between adjacent layers were limited to a factor with a maximum of 2.0, to minimize numerical problems.

6

MODEL SIMULATIONS, CALIBRATION, RECHARGE ESTIMATION AND MODEL VALIDATION

The 3-D numerical simulation was originally performed with MODFLOW (McDonald and Harbaugh, 1988) for the water table flow field. Our 3-D groundwater table flow model was then calibrated using field-measured water table head data from boreholes (Figure 2). The head in the upstream boundary was adjusted by trial-and-error to match root mean square of the simulated and measured heads at the boreholes. This optimal boundary is marked as H, with the objective functions selectively listed at the optimal recharge rate (16%) in Table 6. The objective functions at selected recharge rates and upstream boundary head are listed in Table 6. Objective function values show that the resultant upstream boundary heads is at least close to optimal.

Table 5. The response of hydraulic head of the calibrated flow field to varying upstream heads. Head at upper boundary

H−30

H−20

H−10

H

H+10

H+20

H+30

H+40

H+50

Root mean squared

24.51

20.72

20.51

18.73

20.28

20.19

22.34

25.7

28.46

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Table 6. The response of hydraulic heads at boreholes of the calibrated flow field to varying recharge ratesa . Recharge

10%

12%

14%

15%

16%

17%

18%

20%

22%

30%

40%

Residual mean Root mean squared

−24.49 30.79

−20.42 27.24

−13.45 21.37

−6.52 18.73

−2.42 18.64

2.61 18.76

6.97 18.96

9.23 19.74

14.33 26.71

25.18 38.97

45.20 57.97

Note: a The recharge was given as the percentage of the average annual precipitation.

the overall layer-wise average hydraulic conductivity was used in our model, and that there are significant elevation change throughout the model domain. ZK443 Borehole name stands for . 764.96 Water table elevation (m)

Figure 4. Simulated groundwater table (in m) with water table measurement in boreholes. Table 7. Comparison of calibrated and measured water table data. Well

Calc. (m)

Obs. (m)

Residual (m)

ZK441 ZK443 ZK445 ZK551 ZK553 ZK831 ZK833 ZK837 ZK839 ZK843 ZK845 mean

679.55 758.07 665.59 684.06 754.00 603.10 605.73 638.23 651.05 653.54 661.74

703.58 764.96 682.84 670.88 775.55 604.06 605.12 651.04 680.75 652.64 655.01

−24.03 −6.89 −17.25 13.18 −21.55 −0.96 0.61 −12.81 −29.7 0.90 6.73 −4.43

In addition, the recharge rate is estimated to be in the neighborhood of 15–18% of the annual precipitation, correspondingly yielding a mean residual of −6.52 to +6.97 m (Table 6). In this paper, we use 16% annual precipitation as the calibrated recharge for our calibrated model. The calibrated water table of the flow model is shown in Figure 4. The comparison of calibrated heads with field-measured water table data, provided in Table 7, show reasonable agreement, considering that

The model was validated using seasonal variation of hydraulic head data. No directly observed data of seasonal head change existed for the modeling site. However, the annual water table head change at the modeling site is expected to be similar to that of the boreholes on the dam’s right shoulder, which have a seasonal fluctuation of about 2 to about 30 m. Using annual precipitation and with a portion (16%) of which assigned as infiltration, we first run the model to steady state, and then for 4-month rainy season from mid-June to mid-October, followed by an 8-month dry season for a few years. Simulated seasonal variations in groundwater levels are shown in Figures 5 and 6. 7

THE IMPACT OF A FLOOD DISCHARGE EVENT ON THE SLOPE SITE

The simulated response of a water table aquifer to intensified mist rain precipitation was based on the calibrated flow model. Areas 2 and 3 (marked in Figure 1) were assigned average precipitation rates of 9.25 and 1.25 mm per hour, respectively, based on the precipitation ranges specified in Table 3. The flood discharge event was simulated for 5 days, which in practice is a commonly assigned duration. Simulated results show the water table to be responsive to the discharge event (Figures 5, 6). Cross sections of the water table are shown in Figure 5 for the dry seasons, rainy seasons, and under the lasting torrential storms of a flood-discharge event. Results indicate that because of the torrential precipitation alone, the head would be somewhat elevated in the borehole near the Huoshaozhai Valley F1 (about 1 m at ZK833), and elevated to in a larger extent in the borehole near the middle part of the downstream location (about 40 m at ZK551). This finding indicates that intense precipitation during a flood discharge event would cause a significant water table rise. Saturation changes in the unsaturated zone were also simulated. To systematically understand changes

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612

head(m)

610 608 606 604 602 600 1

4

7

10 1

4

head1

7 10 1 4 month

head2

7 10

steady head

1

4

7

10

Observed

(a) ZK833 760

head(m)

740 720 700 680 660 640

1

4

7

10

1

4

7

10

1

4

7

10

1

4

7

10

month head1

head2

steady head

Observed

Figure 6. Simulated seasonal fluctuation at water table aquifer (head1 in diamond) and water table’s response to foggy rain precipitation (head2 in square): (a) Borehole ZK833 near the F1 boundary, and (b) Borehole ZK551 in the down-stream middle part of the domain. Figure 5. Predicted groundwater table responses along profiles in Figure 2. A marks the minimum water table elevation for dry season and B the maximum water table elevation for rainy season, and C for foggy rain precipitation from torrential flood discharging.

Dry

Elevation (m)

in saturation within the geological layers, we set up a column to represent a typical scenario in the unsaturated zone (Figure 7). The simulation was then run to steady state. We found that saturation in response to the flood discharge event was higher than that in the rainy season, changing from 35–41% during rainy season to 52.5–62.5%. The completely weatheredgranite has less saturation than the weakly weathered granite, which in turn is less saturated than the slightly weathered and fresh granite (Figure 7). Note that results show only a limited saturation increase in the unsaturated zone. The water table rise caused by intensified foggy rain precipitation causes the submerged rock to become fully saturated, with rock strength thus reduced to a minimum. However, rocks in the unsaturated zone are still far from fully saturated. The saturation is related to the strength of the rock. Generally speaking, the rock is significantly weakened with increased saturation (Xiang, 1986). The rocks that are about half saturated in the unsaturated zone would have a rock

Strongly Weathered Granite

800

Wet

Foggy Rain

750 Increased Saturation Weakened Strength

Weakly Weathered Granite Slightly Weathered Granite

700

650

0

0.25

0.5

0.75

1

Satutation Figure 7. Saturation under different recharges in dry seasons, wet seasons and under torrential storm.

strength significantly greater than that of fully saturated rock. From an engineering point of view, treating an undersaturated condition as if it were a fully saturated condition would significantly underestimate the strength of the rock at this site.

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Table 8. Sensitivity coefficients of the calibrated flow model of the Nuozhadu Site. K m/s

Parameters Sensitivity coefficients

8

All layers

Q

γ43 ∼γ51 w

γ43 ∼γ51 f

−0.0046

−0.00008

−0.0036

−0.0109

SENSITIVITY ANALYSIS

We also performed a sensitivity analysis on the calibrated model. Perturbation was generated on model parameters, and the resultant head variation was obtained. The ratio of relative head change to relative change of the perturbed individual parameter is defined as the sensitivity coefficient (Lu et al., 1999). Sensitivity coefficients for the calibrated flow model were computed and listed in Table 8. Results show that the most sensitive parameter is the overall hydraulic conductivity, followed by recharge, whereas the head at the upstream boundary is the least sensitive parameter. With respect to hydraulic conductivity, the slightly weathered to fresh granitic layer is more sensitive than the weakly weathered granitic layer. The strongly weathered layer Q is the least sensitive. 9

SUMMARY AND CONCLUSIONS

This paper characterizes how flow could affect slope stability when factoring in the intensified precipitation from the multiple-day discharge event during flood season at the Nuozhadu. First, a field-scale flow model was developed for the slope area affected by foggy rain precipitation. It was then calibrated against field-observed groundwater table data, producing a reasonably good match. In addition, the calibrated flow model was validated against the field-observed seasonal fluctuation of the water table in response to rainy seasons and dry seasons. We predict that any lasting torrential discharge events will add additional significant recharge to the rainy season’s groundwater level and cause significant water table rise. The water table rise would lead to a sudden, significant saturation increase in the highly

Head at upper stream boundary

Recharge

0.0002

0.0009

weathered and weakly weathered soil zone, but only to a limited saturation increase in the unsaturated zone. The saturation increase in the water fluctuation zone weakens rock strength, based on rock strength’s inverse relation to rock saturation. These findings provide the basis for evaluating the effect of an intensified precipitation event on the slope’s underground flow during a flood discharge. Future work will include a hydrological-mechanical coupling to analyze slope stability. REFERENCES CDUT (Chengdu University of Technology) and China Hydroelectric Consulting Group Kunming Hydroelectrical Investigation, Design and Research Institute (KHIDI). 2006. The Nuozhadu Hydroelectrical Power Station Site’s High Slope Engineering Features and Its Stability Study. Lu, G., T.P. Clement, C. Zheng & T.H. Wiedemeier. 1999. Natural attenuation of BTEX compounds, model development and field-scale application. Ground Water 37(5): 707–717. McDonald, M.G. & A.W. Harbaugh. 1988. A Modular Threedimensional Finite Difference Groundwater Flow Model. U.S. Geological Survey Techniques of Water Resources Investigations, Book 6, 586 p. Wu, Yu-Shu, G. Lu, K. Zhang & G. S. Bodvarsson. 2004. A Mountain-Scale Model for Characterizing Unsaturated Flow and Transport in Fractured Tuffs of Yucca Mountain. Vadose Zone Journal 3:796–805. Snow, D.T. 1965. A parallel-plate model of fractured permeable media, 331 pp., Ph.D. thesis, University of California, Berkeley. Snow, D.T. 1969. Anisotropic permeability of fractured media. Water Resources Research 5(6): 1273–1289. Xiang, Jiannan. 1986. The Effect of Water on the Mechanic Properties of Soft and Weak Layer, in ‘‘Hydrogeology and Engineering Geology Series 1’’, Geology Press.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

An issue in conventional approach for drainage design on slopes in mountainous regions Z.Q. Yue Department of Civil Engineering, The University of Hong Kong, Hong Kong, China

ABSTRACT: This paper uses a case example to show an issue in the conventional approach for provision of surface drainage system on slopes in mountainous regions. The issue is that the conventional approach has the potential to lead a large amount of unexpected surface runoff flowing down on man-made or natural slopes. The large amount of water flow can cause major slope instability and landslides.

1

INTRODUCTION

It has been well recognized that heavy rainfall is one of the most direct causes of landslides in mountainous regions (Lumb, 1975; Lee & Chen, 1997; Yue & Lee, 2002; Li et al. 2003). This recognition is due to the fact that a majority of landslides happened during heavy rainstorms. Rainfall interacts with the ground via infiltration and surface runoff. Infiltration is the movement of water from the ground surface into the soil or rock via the pores or interstices of the ground soils and rocks. Surface runoff is the flow of water on the ground surface from a higher catchment area to a lower place. Infiltration can substantially change the groundwater and the pore water pressures in the slope soils and affect the stability of slopes. Landslides due to infiltration and groundwater have been investigated extensively in the geotechnical communities around the world (eg., Li et al. 2005; Li et al. 2005a, b). The effect of surface runoff on slope stability is usually handled with the provision of surface drainage system on slopes. Design and construction of surface drainage system on natural terrains and urban areas are conventional topics and have been studied in hydrology for many years (DSD, 1995; GEO, 1994; Viessman et al. 1989; Zheng et al. 2006). According to GEO (1994), the main purpose of surface drainage is to improve slope stability by reducing infiltration caused by heavy rainstorms. The slope drainage should collect runoff from both the slope and the catchment area upslope and lead it to convenient point of discharge beyond the limits of the slope. Results of literature reviews, however, show that there seem a few papers available in the open literature on the improper provision of surface drainage

system on slopes in relationship with the slope stability in the geotechnical communities. One of the few case examples was given by GEO (1996, 2000) and Yue (1999, 2000). This case was about a major landslide that occurred on the hillside below Nam Long Shan Road in Hong Kong on August 13, 1995. Large amounts of unexpected water were observed flowing down on the road during rainstorms, which was considered one of the main reasons causing the major landslide. This case study has led to the provision of the regulations on the additional requirements about the surface drainage design on roads and highways and natural hillsides in Hong Kong (GEO, 2000; Leung & Leung, 2001). The purpose of the present paper is therefore to present an issue in the conventional approach for the provision of surface drainage systems on man-made slopes in mountainous regions. The issue is that the surface drainage system designed according to the conventional approaches can cause a large amount of unexpected runoff channeled to flow down slopes. The large amount of channeled water can cause major landslides with severe consequences. A case example is used to demonstrate the issue. It is hoped that the issue will draw attention of researchers and engineers in the geotechnical communities. 2

THE INCIDENT OF INSTABILITY

The incident in this case example is about the instability and deformation of a major engineered fill slope with a retaining wall during heavy rainfall in Shenzhen on August 19 and 20, 2005. Figure 1 is a satellite image of the entire site. As shown in Figure 1, the fill slope was part of a landfill site in a mountainous region. It was located above the access road to the landfill area.

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Figure 2.

A geological cross-section of the fill slope.

Figure 1. Location of the deformed fill slope above the access road to a landfill in a mountainous region.

The access road was above the waster water treatment plants of the landfill site. During the heavy rainstorms on August 19, 2005, many tensile cracks were observed and were being developed in the fill slope and the retaining wall. Besides, a large amount of groundwater was seeping out of the retaining wall from its weepholes and cracks. The retaining wall was made of masonry blocks, was of 10 m high and immediately above the access road. In realization of high consequences in the event of the slope failure, emergent protection measures were applied immediately. They included (a) covering the fill slope platforms with waterproof canvas, (b) backfilling the wall toe with sandbags and (c) temporarily closing the access road. After the emergent protection measures, fortunately, the fill slope and the retaining wall were not collapsed on the next day when it had heavier rainstorms. After the rainfall stopped on August 21, 2005, a landslide investigation was carried out. Details of the landslide investigation can be found in Li et al. (2006).

3

THE FILL SLOPE AND RETAINING WALL

The fill slope and retaining wall were constructed in 1993 during the construction of the landfill. The construction of the landfill was completed in 1997 and subsequently started its operation. As shown in Figure 2, the total height of the fill slope and the retaining wall was 45 m high. The retaining wall at the slope toe (Figure 3) was made of masonry blocks. The length of the retaining wall was about 200 m. The overall slope angle of the fill slope was about 26◦ . The fill slope has two wide platforms 1 & 2 (Figs. 4 and 5). The platforms 1 and 2 have their levels at 160

Figure 3. The masonry retaining wall as toe of the fill slope and above the access road.

Figure 4. A longitudinal crack sealed with cement on centre of platform 1.

and 140 mPD (meter above the datum, i.e., mean sea level), respectively. Several houses were constructed on the platforms 1 and 2 (Figs. 4 to 6). The platform 2 has its width varying from 10 to 40 m. The upper fill slope between the two platforms had a slope angle of 40–45◦ and its surface was protected with concrete grillages. The lower fill slope between the platform 2 and the wall had a gentle slope angle of 25 to 30◦ and its surface was heavily vegetated.

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4

Figure 5. Cracks in the platform 2 of the deformed fill slope.

THE SURFACE DRAINAGE SYSTEM

The fill slope had three U-channels on its surface. They were located in front of the retaining wall, the toe of the lower slope and behind the crest of the retaining wall, and the toe of the upper slope and above the platform 2 (Figure 6), respectively. Furthermore, there were three major drainage channels in the natural hillside above and adjacent the fill slope. As shown in Figure 7, the first major drainage channel was a major stepped channel on the natural hillside and led the surface runoff water from the point B (223 mPD) on the natural hillside to the point D (160 mPD) on the platform 1. The second was a major open drainage channel constructed on the inside extension of the cut platform 1 and intercepted and discharged surface runoff water from the point C (170 mPD) to the point D (160 mPD). This major open drainage channel was made of in-situ

Figure 6. Inclination of a U-channel wall due to bulging of the fill slope behind a building on the platform 2.

There is a steep cut slope above the platform 1. Natural hillside slopes were above the cut slope and densely covered with vegetations and trees. The natural hillside slope had an average slope angle of 26◦ . The in-situ ground underneath the fill at the slope site was highly decomposed sandstone. A majority of the fills forming the fill slope were highly decomposed sandstones excavated from the adjacent in-situ ground. The fills were found up to 17 m in thickness. The fill slope had high permeability and adequate effective shear strength. The instability of the fill slope can be observed in the site photographs in Figs. 4 to 6. The photographs were taken in the afternoon of August 28, 2005. Figs. 4 shows a longitudinal crack sealed with cement grout at the centre of the platform 1. Figure 5 shows a transversal crack in the platform 2 which was covered with the waterproof canvas. Figure 6 shows an inclination of the U-channel wall at toe of the upper slope and behind a house on the platform 2.

Figure 7. Surface drainage system and catchment area associated with the natural hillside above the fill slope.

Figure 8. An open drainage channel from the point C to the catchpit at the point D along the toe of the natural hillside above the cut and fill slopes of the landfill.

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weathered sandstone as shown in Figure 8. It was about 210 m in length, 2 to 5 m in width and 1 to 2 m in depth. The third was a major stepped channel and discharged water from the point D to the point E on the access road below the fill slope. As shown in Figure 9, this stepped channel was made of masonry blocks pointed with concrete grout. It was about 200 m in length, 2 to 3 m in width and 1 to 2 m in depth. The point D was of the most importance in the surface drainage system for the natural hillside above the fill slope because it was the junction position for redirecting and re-charging the surface runoff from the natural hillside to the toe of the fill slope. As shown in Figure 10, it was a deep vertical catchpit made of masonry blocks pointed with cement grout. It was about 3 m deep and had a square cross-section of 1.5 m in width. It was connected with a cross-road curvet beneath the platform 1 and led water into the stepped channel in Figure 9.

Figure 9. A large stepped channel directing water flow from the catchpit at the point D to the point E on the access road.

The site inspection on August 28, 2005 revealed that the catchpit was partially filled with debris. The debris was mainly cobbles and gravels of sandstone, as shown in Figure 10. Besides, open cracks were observed in the catchpits and stepped channels.

5

DISCUSSIONS AND FINDINGS

The natural hillside above the platform 1 of the fill slope had a large catchment area. Figure 7 gives the boundary of the catchment area on the natural hillside. The highest point of the catchment area is the point A of 310 mPD marked on Figure 7. From the point A, the height of the catchment area gradually decreases to 160 mPD at the catchpit point D. The natural hillside had five natural stream courses. The surface runoff water was led to the first and second drainage channels through the five natural stream courses. The platform 1 cut off the natural stream courses so that the runoff water was re-directed to the point E at the toe of the fill slope through the catchpit D and the third stepped channel, as shown in Figure 7. The catchment area was of 90,000 m2 on the horizontal plane, which was about ten times larger than the catchment area of the deformed fill slope. As in Hong Kong, the climate in Shenzhen is subtropical. Heavy rainstorms are common in Shenzhen in wet season of each year. The daily rainfalls recorded by Shenzhen Observatory Bureau in August 2005 are listed in Table 1. The rainfalls on August 19 and 20, 2005 were 243 and 303 mm, respectively, which are extremely high. Accordingly, the total rainfalls on the natural hillside catchment area were 21,870 and 27,270 m3 on August 19 and 20, 2005, respectively. The total rainfalls on the fill slope were 2,187 and 2,727 m3 on August 19 and 20, 2005, respectively. Due to the partial blockage of the catchpit D, a large amount of the runoff could be infiltrated into the fill slope through the cracks of the catchpit and the stepped channel Table 1.

Figure 10. The deep catchpit at the point D for directing water flowing down through a cross-road drainage culvert beneath the platform 1.

Daily rainfall (mm) in August 2005 (Li et al. 2006).

Date

Rainfall

Date

Rainfall

Date

Rainfall

01 02 03 04 05 06 07 08 09 10

15 0 0 0 0 0 0 0 48 26

11 12 13 14 15 16 17 18 19 20

12 9 83 2 14 49 51 39 243 303

21 22 23 24 25 26 27 28 29 30

17 0 0 16 0 2 0 2 9 21

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walls. Such large amount of water supplies could substantially arise the groundwater table in the fill slope and make the fill slope fully saturated. Besides, the discharging capacity of the catchpit D was substantially reduced due to the partial blockage of the debris. The runoff water led to the catchpit D from the natural hill slope during the extremely heavy rainfall could cause flooding on the crest of fill slope and supply water to flow on the slope surface and to leak into the cracks. As a result, the fill slope was fully saturated for a long period of time even though the seeping of groundwater from the retaining wall at the slope toe. The fill slope eventually became instable. Cracks and permanent deformation occurred in the fill slope. Such inferred infiltration and surface runoff of large amount of water into the fill slope through the catchpit D was evidenced by the fact that the large amount of water seeping out of the 10 m high retaining wall. Furthermore, the overall direction of the instability and deformation of the fill slope was found to adopt the catchpit D as its centre of crack propagation and extension (see the array lines in Figure 7). It was believed that the design of the fill slope and retaining wall as well as the surface drainage system was carried out suitable engineers and further checked by suitable engineers. They should be up to the current design standards because of the importance of the landfill to the city of Shenzhen. The landfill was the largest one in Shenzhen with more than 10 millions people. However, the design of the surface drainage system had a severe drawback that could cause the concentration of a large amount of water flow from the large hillside catchement to the catchpit D. Once the catchpit D was partially blocked and broken, the large amount of water could infiltrate into the fill slope and flow down on the fill slope, which eventually could cause the failure in the fill slope. Therefore, there was an issue in the conventional approaches in the provision of the surface drainage system to slopes in mountainous regions. 6

CONCLUSIONS

This paper has presented a case example to show the inadequacy in the approaches of the conventional surface drainage design on hillside slopes. The case example is the slope instability incident in the largest landfill site in Shenzhen in August 2005. The surface drainage design was carried out according to the conventional standards and had a drawback on the possibility of collection of large amount of water onto the fill slope below the upper natural hillsides. The large amount of water could infiltrate into the fill slope and flow down on the fill slope, which could cause severe slope instability.

The case example has shown a severe issue in the approaches on the provision of drainage design system that needs to collect surface runoff water and discharge the water to a safe location. Such collection of surface runoff water could be a driving force triggering major landslides if any part of the drainage system was blocked or partially blocked during a heavy rainstorm. The existing stream courses on the hillsides had evolved for thousands and millions years. The alteration of the matured topography and stream courses by engineering was extremely fast and could result in unsuitable and untested new drainage systems, which could inevitably cause large amounts of unexpected channeled water flowing down slopes. ACKNOWLEDGEMENTS The author acknowledges the financial support from Hong Kong’s Research Grants Council and thanks Dr. A.G. Li, Mr. G.X. Huang and Mr. A.J. Zhang for their assistance on the case study. REFERENCES DSD. 1995. Stormwater Drainage Manual. Planning, Design and Management (Second edition). Drainage Services Department (DSD), Hong Kong Government, HK. GEO. 1994. Geotechnical Manual for Slopes (Second Edition). Geotechnical Engineering Office (GEO), Civil Engineering Department, HK. GEO. 1996. Report on the Shum Wan Road Landslide of 13 August 1995, Volume 2: Findings of the Landslide Investigation; and Documents A to J. HK. GEO. 2000. Highway Slope Manual. Geotechnical Engineering Office (GEO), Civil Engineering Department, Government of HKSAR, HK. Lee, C.F. & Chen, H. 1997. Rainfall infiltration and landslides in Hong Kong. Hydrogeology and Engineering Geology, Vol.156, No.4, pp. 34–38. Leung, J.K.Y. & Leung, M.H. 2001. Upper Catchment Stormwater Diversion with Drainage Tunnels, Proceedings of the Hong Kong Institution of Engineers, Vol. 3, No. 1, July 2001. Li, A.G., Yue, Z.Q., Tham, L.G., Lee, C.F. & Law, K.T. 2005. Field monitored variation of soil moisture and matric suction in a saprolite slope, Canadian Geotechnical Journal, 42 (1): 13–26. Li, A.G., Xiong, J.N., Nan, L. & Qiu, J.J. 2006. The study of rainfall induced sliding of a fill slope in Shenzhen, China, Proceedings of the 10th International Congress of the International Association of Engineering Geology, Nottingham, United Kingdom, Sept. 6–10, 2006. paper number: 3–810, page 1–8 (CD RAM softcopy). Li, S., Yue, Z.Q., Tham, L.G., Lee, C.F., Yan, S.W. 2005a. Slope failure in under-consolidated soft soils during the development of a port in Tianjin, China. Part 1, field investigation, Canadian Geotechnical Journal. 42 (1): 147–165.

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Li, S., Yue, Z.Q., Tham, L.G., Lee, C.F., Yan, S.W. 2005b. Slope failure in under-consolidated soft soils during the development of a port in Tianjin, China. Part 2, analytical study, Canadian Geotechnical Journal. 42 (1): 166–183 Li, Y., Kang, Z.C., Yue, Z.Q., Tham, L.G., Lee, C.F., Law, K.T. 2003. Surge waves of debris flows in Jiangjia Gully, Kunming, China, Proceedings of International Conference on Fast Slope Movements: Prediction and Prevention for Risk Mitigation, Editor: Luciano Picarelli, Publisher: PÀTRON EDITORE, Bologna, Naples, Italy, May 11–13, 2003, Vol.1: 303–307 Lumb, P. 1975. Slope failures in Hong Kong. The Quarterly Journal of Engineering Geology, Vol. 8, pp. 31–65. Viessman, W. Jr., Lewis, G.L. & Knapp, J.W. 1989. Introduction to Hydrology, (3rd Edition), Harper & Row, Publishers, Inc., Singapore.

Yue, Z.Q. 1999. Surface Drainage Analysis of the 1995 Shum Wan Road Landslide. Expert witness report submitted to the Department of Justice, The Government of Hong Kong Special Administration Region. HK. Yue, Z.Q. 2000. Supplementary Report on Surface Drainage Analysis of the 1995 Shum Wan Road Landslide Site. Expert witness report submitted to the Department of Justice, The Government of Hong Kong Special Administration Region. HK. Yue, Z.Q. & Lee, C.F. 2002. A plane slide that occurred during construction of a national expressway in Chongqing, SW China, Quarterly Journal of Engineering Geology and Hydrogeology, 35: 309–316. Zheng, Y.R., Chen, Z.Y., Wang, G.X. & Ling, T.Q. 2006. Engineering Treatment of Slope & Landslide, China Communication Press, Beijing.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Analysis of geo-hazards caused by climate changes L.M. Zhang Department of Civil Engineering, The Hong Kong University of Science and Technology, Hong Kong, China

ABSTRACT: This paper analyzes the effect of climate on the generation of possible geohazards. The rainfall and evaporation changes in Hong Kong in the past four decades are first reviewed based on records from the Hong Kong Observatory. Then the effect of climates on the generation of emerging geohazrzds is analyzed through a series of transient infiltration analyses taking the climate conditions as initial conditions. Three climate conditions; namely, extreme drought condition, extreme wet condition, and steady-state condition, are studied. Extreme yearly weather variations are shown to be the key to the generation of interchanging extreme hazards such as landslides and floods. The analysis results demonstrate that, in a prior extreme drought condition, an intermediate rainfall process can result in large surface runoff and thus surprising floods. In addition, dissipation of suction only occurs in the shallow soils. Hence, storm water infiltration into a dry ground is likely to cause shallow-seated landslides or debris flows under the combined effect of shallow perched ground water and surface erosion from increased runoff. On the other hand, in extremely wet conditions, the ground water table can rise substantially and failure of some slopes that have been stable for a long time can be triggered even by a moderate rainfall event. 1

INTRODUCTION

In the past decades, climate phenomena became more and more abnormal. Such abnormal climate phenomena as extreme droughts and extreme storms can result in aggravated landslide-flood cycles. However, the mechanisms behind these emerging geo-hazards caused by abnormal climate conditions have not been well understood. In this paper, analyses on the geohazards caused by climate changes are studied through characterizing the initial conditions created by extreme drought and wet conditions and then studying the triggering of landslide or the generation of flood by a new rain event. Analysis of a normal rainfall condition is also undertaken for comparison purposes.

2 2.1

EMERGING GEO-HAZARDS FROM CLIMATE CHANGES Phenomena

Based on statistics of the Hong Kong Observatory (HKO) (Leung et al. 2004), there have been noticeable changes in several climate elements in Hong Kong since 1947: 1. The annual total evaporation, measured using evaporation pans with evaporation surface 0.18 m above ground, decreased by 40% from the 1960s to 2003, at a rate of 184 mm per decade, with the mean

value between 1964 and 2002 being 1405 mm (Figure 1a). 2. In the 56-year period between 1947 and 2002, the annual total rainfall at HKO increased from 2265 mm in the 1950s to 2518 mm in the 1990s. It represents an increasing trend of about 65 mm per decade, though not statistically significant at 5% level (Figure 1b). 3. The annual number of heavy rain days (i.e. the days with hourly rainfall greater than 30 mm, which is the criterion for issuing Amber Rainfall Warning) has been increasing from about 4.5 days a year in 1947 to about 7 days in 2002, though not statistically significant at 5% level (Figure 1c). While the HKO findings appear to suggest only a gradual minor change in average annual rainfall and evaporation, the yearly variations have become more extreme globally. Abnormal climate phenomena such as extreme droughts, storms, typhoons and tides occur more frequently than ever in the last decade. In the first half of 2004, Guangdong experienced five extreme climate events: one extreme drought, one 20-day cold front, one extreme heat wave, eight major storms, and seven major tidal events. This was unusual. In year 2005, there were severe droughts in parts of Africa, Western Europe, and Australia; record-breaking heavy rain in India; and an extreme active hurricane season in the north Atlantic. According to the HKO, in Hong Kong, year 2005 was the third wettest year on record,

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In extreme drought conditions, decrease in soil moisture content results in a substantial reduction in water permeability of the soil. After a long period of drought, little rainwater can infiltrate into the ground during a rainfall event due to the initially very small permeability of the soil. As a result, floods can be generated more easily than in the normal climate conditions. The flood in early July 2004 in Beijing and Shanghai indeed occurred under only about 40 mm of precipitation. The study of hydro-geological conditions at extreme drought conditions is essential to understand the formation of the ‘‘unexpected’’ floods. Storm water infiltration into a dry ground is likely to cause debris flow and shallow-seated landslides because of the combined effect of shallow perched ground water and surface erosion by increased runoff. The possible occurrence of such debris flow and landslides due to erratic pattern of extreme climate conditions needs to be addressed as they pose a threat to the safety of the public and the environment. On the other hand, in extremely wet conditions, the ground water table can rise substantially and a not-so-heavy rain event can trigger deep-seated slope failures. Failure of some slopes that have been stable for a long time can also be triggered.

(a) Annual total evaporation at King’s Park (1964–2002)

(b) Annual rainfall at HKO Headquarters (1947–2002)

(c) Number of days with hourly rainfall greater than 30 mm at HKO Headquarters (1947-2002)

3

ANALYSIS METHODOLOGY

In this paper, the effect of climate on emerging geohazards is considered in two aspects:

Figure 1. Changes in rainfall and evaporation over time in Hong Kong (After Leung et al. 2004).

mostly due to a very active southwest monsoon in June and August. The total rainfall of 3214 mm was 45.2% above normal years. June 2005 was the fourth wettest since 1884, and August 2005 the second wettest. The rainfall in these two months alone amounted to 1865 mm, about 84% of the normal annual rainfall. 2.2 Mechanisms of geohazards triggering Climate conditions affect engineering behavior of soils and the geological environment. With the occurrence of extreme droughts, storms and tides, many unprecedented ‘‘surprising’’ geohazards have been induced. These unusual geohazards are emerging challenges that have started to affect the environment and socioeconomic development of Hong Kong and the Pearl River Delta region.

1. The extreme weather creates extreme initial conditions in the ground, either extremely dry after a period of drought or extremely wet after a period of sustained rainfall. These conditions will affect the infiltration of water into the ground and the generation of surface runoff during a subsequent storm. 2. Depending on the initial conditions and the erodibility and shear strength of soils, a subsequent storm can cause various hazards such as slope instability, debris flow, and flooding. In this paper, the pore water pressures in a slope and the surface runoff on the slope will be analyzed at a benchmark condition (normal initial moisture content), after a drought, and after a very wet period. Figure 2 shows the profile of a soil slope considered. It is 30 m high, with a slope angle of 32 degrees. The slope consists of two soil layers; the lower layer is the natural soil and the upper layer is a loose fill. The loose fill is assumed to be a loose completely decomposed granite (a silty sand), with a porosity of 0.41 and a saturated permeability of 4.79×10−6 m/s. The lower layer is of less concern in this paper, with a porosity of 0.28 and a saturated permeability of 8.36×10−9 m/s. These parameters are similar to the mean values found

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The initial conditions at shallow depths of a slope vary throughout the year. McFarlane (1981) measured changes of suction with time in 1980 at five sites in Hong Kong. Figure 4 shows the variations of suction at two vegetated cut slopes in completely decomposed granite (CDG); one at King’s Park and another at Lung Cheung Road. The largest and most rapid changes in suction occurred at shallow depths. The suction level was high during the dry season, particularly in April, but low during the wet season. Particularly in July and September, the suction maintained was less than 10 kPa. In order to study the effect of climate conditions on rainfall infiltration and generation of runoff, three initial conditions are generated in this study:

45

A

40

35

El evation (m)

30

25

20

15

10

5

0 0

5

10

15

20

25

30

35

A

40

45

50

55

60

65

70

75

80

85

90

95

Distance (m)

Figure 2. Profile of the soil slope considered.

1. Steady state condition. The pore water pressures are obtained by subjecting the slope to a constant rain intensity equal to the annual average rainfall intensity, 2518 mm/year or 7.98 × 10–5 mm/s. 2. Extremely wet condition. The pore water pressures above the initial ground water table are assumed to be zero, representing the scenario when the suction in soil is completely destroyed due to sustained rainfall prior to a new rainfall event. 3. Extreme drought condition. The pore water pressures are obtained by subjecting the slope to a constant rate of evaporation/transpiration equal to one-half of the annual average rate of evaporation, 1405 mm/year or 4.46 × 10–5 mm/s. The initial pore water pressure distributions along section A-A (in Figure 2) are shown in Figure 5. After considering the various initial conditions created by the aforementioned climate conditions, three rainfall processes, with intensities of 70, 30, and 15 mm/hour are imposed onto the slope. Possible geohazards that can be generated by the rain events and their relation to the respective climate condition are then analysed. 4 Figure 3. The soil-water characteristic curves and permeability functions for the slope soils.

at the Sau Mau Ping slope (Hong Kong Government 1976). The soil-water characteristic curves and permeability functions for the two soils, shown in Figure 3, are generated based on the grain-size distributions and the porosity and saturated permeability values following the methods developed by Fredlund & Xing (1994) and Fredlund et al. (2004). The boundary conditions are shown in Figure 2. The bottom boundary is impervious but the ground surface is subject to either evaporation or rainfall infiltration. An initial ground water table is assumed.

RESULTS AND ANALYSIS

4.1 Influence of climate conditions on landslide triggering Rainfall infiltration is a well-known landslide trigger as it causes reduced soil suction (thus reduced shear strength of soil) and added hydrodynamic loading. The climate conditions prior to a particular rain event have a significant effect on the pore water pressures in the slope. Figure 6 shows the effect of prior-rain climate conditions on distributions of pore water pressure in the slope along section A-A (see Figure 2). The rainfall intensities in Figures 6a, b and c are 70, 30 and 15 mm/hour, respectively, but the instants the pore water pressures are plotted are so taken that the total rainfall amount is all 58.4 mm in the three cases.

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40

Suction (kPa)

35 30

Depth normal to slope (m)

Elevation (m)

Vegetated cut slope – King’s Park

25 20 15 10 5 0 -200

Hydrostatic condition Extreme evaporation Steady rain condition Extreme wet condition -160

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-80-

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0

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Initial pore pressure (kPa)

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Elevation (m)

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Suction (kPa)

Vegetated cut slope – Lung Cheung Road

25 20 15

Depth normal to slope (m)

10 5 Extreme evaporation 0 -40000

-30000

-20000

-10000

0

10000

Initial pore pressure (kPa)

Figure 5. Initial conditions along section A-A possibly caused by extreme climate conditions.

Figure 4. Variations of suction at two vegetated cut slopes in completely decomposed granite (After McFarlane 1981).

Given a rain event of a limited duration, the analysis results show that 1. If the condition prior to the rain event is very wet, say a new rain after a sustained heavy rain, the ground water table in the slope will be high and the pore water pressures in the slope will be positive. The high positive pore pressures will result in reduced shear strength of the slope soil and added seepage forces. All these will lead to reduced safety factor of the slope, and possibly deep-seated failure of the slope. Given the same rainfall amount, a smaller but longer rain process will cause larger pore water pressures in the slope and hence decrease the stability of the slope more significantly. Therefore, shortly after a sustained major rain event, a moderate new rain event may be able to trigger a landslide that has not occurred in a previous, much heavier rain event. 2. If the condition prior to the rain event is relatively wet, say a new rain after a sustained but light raining period, small suctions can be maintained in the slope even during the new rain event. In addition,

the changes of pore water pressure during the new rain event are limited; hence the safety of the slope is less affected by the new rain. Given the same rainfall amount, the rainfall intensity does not appear to affect the pore water pressures significantly. 3. After a period of extreme drought, the suctions in the ground become very high and may not be destroyed in a new rainfall event of limited duration. Deep-seated failure of the slope is unlikely caused by the new rain event although shallow seated failures may be triggered due to the loss of suction at shallow depths. The rainfall intensity plays a minor role in this special case because most of the rainwater does not infiltrate into the slope. This important point will be detailed in the next section. Although the rainfall characteristics (e.g., intensity, duration, and pattern) are known to be key to landslides triggering, the above analysis results show that the climate conditions play an equally important role. Note that the initial conditions may have a lessened effect in the case of a sustained new rain event. 4.2 Extreme drought-flood cycle caused by climate conditions Slope stability is only one special type of geohazard. A significant feature of recent disasters is the occurrence of repeated extreme droughts and floods within a particular year. This tendency can now be analyzed

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Figure 6. Effect of prior-rain climate conditions on distributions of pore water pressure in the slope after a rainfall amount of 58.4 mm.

through the effects of climate conditions on surface runoff generation. During a rain event, the amount of water infiltration can be obtained by de- fining a flux section along the ground surface and recording the flux across the section over time. The water infiltration rate obtained can be further used to calculate the surface runoff rate, defined as the ratio of surface runoff to the total rainfall. The runoff rate values for the three rain events under the aforementioned climate conditions are calculated and shown in Figure 7. Values of the cumulative average surface runoff per unit area are shown in Figure 8. Note the analysis does not consider any surface ponding and thus may overestimate the surface runoff.

Figure 7. Effect of prior-rain climate conditions on surface runoff rate.

Figures 7 and 8 clearly demonstrate that 1. The surface runoff rate strongly depends on the prerain climate conditions. It increases as the pre-rain ground condition becomes drier. This is reasonable since the permeability of soil becomes very small when the soil is desaturated (see Figure 3b). The effects of prior climate conditions are particularly significant when the rain intensity is relatively low (e.g. 15 mm/hour in Figure 7c). When the prior climate condition is very wet, the runoff rate from the new rain event is as low as 0.4. In contrary, after a period of extreme drought, the permeability of soil becomes so low that little rainwater infiltrates and the runoff rate reaches over 0.95.

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0.16

of precipitation. When the rainfall intensity is much larger than the saturated permeability of soil, the prior climate conditions have a minor effect on flood generation (Figure 8a). 3. This analysis reveals the mechanisms behind increasing, interchanging geohazards caused by climate changes; namely, extreme droughts followed by floods or debris flows, or extremely wet conditions followed by deep-seated landslides.

(a) Rainfall intensity = 70 mm/hour

Runoff per unit area (m3/m2)

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Hong Kong and the vicinity, as with other parts of the world, are subject to climate changes. Clime changes include the tendency of long-term changes and the tendency of more drastic yearly variations, with the latter tendency causing more geohazards based on the results of analysis in this paper. If the climate condition prior to a new rain event is very wet, the ground water in the slope concerned will be high and the pore water pressures will be positive. These will lead to reduced safety factor of the slope, or possibly deep-seated failure of the slope. Particularly, a smaller but longer new rain event will cause larger pore water pressures in the slope and hence decrease the stability of the slope more significantly. After a period of extreme drought, the suctions in the ground become very high and may not be destroyed in a new rainfall event of limited duration. Shallow-seated failures or debris flow may be triggered although deep-seated failure are unlikely. More importantly, the runoff amount generated by a moderate rain event after a long period of drought can be twice that generated after a sustained wet period, causing ‘‘surprising’’ flood disasters that are not likely under normal prior climate conditions.

Runoff per unit area (m3/m2)

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(c) Rainfall intensity = 15 mm/hour

Runoff per unit area (m3/m2)

0.14 0.12 0.10 0.08 0.06 0.04 Rainfall at a steady-state initial condition Rainfall after an extreme wet period Rainfall after an extreme drought period

0.02 0.00 0.0

2.0

4.0

6.0

8.0

10.0

12.0

14.0

CONCLUSIONS

16.0

Rain time (hour)

Figure 8. Effect of prior-rain climate conditions on surface runoff generation. A rainfall event after an extremely drought will generate a flood approximately twice as much as that in the normal condition, which explains why extreme geohazards (say extreme droughts followed by floods) become more often.

2. In response to the drastic changes in runoff rate due to prior climate conditions, given the same moderate rain event (Figure 8c), the runoff amount generated after a long period of drought can be twice that generated after a sustained wet period. This explains many cases of ‘‘surprising’’ flood disasters that were caused by moderate rains. This tendency may be aggravated by increasingly paved ground conditions in an urban area. For example, the floods in early July 2004 in Beijing and Shanghai indeed occurred under only about 40 mm

ACKNOWLEDGEMENTS This research is supported by the Emerging High Impact Areas (EHIA) Program 2004/05 of the HKUST (Project No. HIA04/05.EG02 ‘‘Emerging Geohazards in Hong Kong and Pearl River Delta due to Climate Changes’’).

REFERENCES Leung, Y.K., Yeung, K.H., Ginn, E.W.L. & Leung W.M. 2004. Climate Change in Hong Kong. Technical Note No. 107, Hong Kong Observatory, Hong Kong SAR. McFarlane, J. 1981. Soil Suction and Its Relation to Rainfall. GCO Report No. 13/81, Geotechnical Control Office, Hong Kong.

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Fredlund, D.G. & Xing, A.Q. 1994. Equations for the soil-water characteristic curve. Canadian Geotechnical Journal 31: 521–532. Fredlund, D.G., Xing, A. & Huang, S. 1994. Predicting the permeability function for unsaturated soils using the soil-water characteristic curve. Canadian Geotechnical Journal 31: 533–546.

Government of Hong Kong. 1976. Report on the Slope Failures at Sau Mau Ping 25th August 1976. Vols. 1–3. Hong Kong: Hong Kong Government Printer.

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Slope stabilization and protection

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Back experience of deep drainage for landslide stabilization through lines of siphon drains and electro-pneumatics drains: A French railway slope stabilization example S. Bomont Ecole Nationale des Travaux Publics de L’Etat, Lyon, France

ABSTRACT: The prime trigger for landslips is high standing groundwater levels within the slope. Drainage techniques which lower the standing groundwater level within the unstable soil mass are a proven method of stabilising landslips. The field of technique are important to lower and maintain the groundwater levels within slopes but techniques using one or more lines of vertical or inclined drainage wells could have advantage as siphon and electro-pneumatic often used in Europe. Back experience is presented for a case in France where the construction in 1995 for the high-speed railway line from Lyon to the Mediterranean included significant earthworks. In 2000 in the southern area of Lyon, a landslide 30 m in depth and 300 m width occurred. To solve the problem, 600,000 m3 of material was removed from the slope to reduce the movement at the upslope part of the railway. However on completion of these works slow ground movement continued over a 150 m length with groundwater levels being near ground surface. After further studies, the Railway Agency in 2006 have chosen to install about 47 electro-pneumatic drains wells to lower groundwater to 10 to 15 m depth. Details of the design, monitoring and geological context are described. 1

RAILWAYS TGV LANDSLIDE IN CHABRILLAN—FRANCE

1.1 Introduction—the problem The ‘‘Chabrillan’’ is a major cutting excavated in July 1998 as part of the TGV high speed train link to the Mediterranean coast of France. The cutting is located at the 530.300 km mark, approximately 20 km South of Valence. The cutting is approximately 1000 m in length and has a maximum depth of 35 m, one of the deepest cuttings on the TGV Mediterraneanline. Following construction, the SNCF (French National Railway Company) implemented a routine inspection program to regularly assess and maintain the condition of its infrastructure. In June 2000, minor deformation of the pavement to an access road approximately 200 m east of the Chabrillan cutting was reported during a routine inspection of the road by the local road authority. Later, in September 2000, larger scale movements resulted in a narrow fissure some centimeters in width appear within the slope to the east and above the road. Instrumentation including piezometers and inclinometers were installed within the slope to assess groundwater behaviours and ground movements. By November

2000 the fissure had developed into a major feature and extended for a length of 30 m. A preliminary assessment suggested the slope movements were limited in its extent and confined locally to the slope above the access road. However in January 2001 further discussion between SNCF and their consultants and examination of instrumentation data suggested that ground movements and slope failures were not only progressing but could affect the cutting and the high speed train line. Following the establishment of the ground movements, the TGV Mediterranean office proposed to excavate 600,000 m3 of material from behind the cutting slopes to reduce the risk of major failure affecting the railway line and slopes. An emergency program of further ground instrumentation was implemented including slope indicators, inclinometers and piezometers in the cutting slope and a precise total station survey network established to monitor key installations such as electric pylons, rail lines and signal gantries. In March 5th 2001, inclinometer data confirmed maximum lateral ground movement of 1 mm/day was recorded towards the high speed rail line and an area of 1,200,000 m3 was concerned by the problem with two failure surfaces identified.

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Borings and records of the area confirmed the geological sequence (from top to bottom) to comprise:

Initial crack Inspection road

Excavation Railway line

• Calcareous sandstone and sandy limestone of Miocene age with the presence of karstic features such as widened jointing. • Marly limestone of Oligocene age. • Variegated clayey plastic marls of Oligocene age.

Sandstone Marls Landslide failures line

Limestones

Inclinometer records confirmed the presence of two main levels of ground movement at 19 m and 30 m depths below the crest of the cutting generally coinciding with the main levels of groundwater seepage and marl layers dipping towards the cutting. Further ground investigation boreholes and instrumentation confirmed the ground movement affected the cutting confined to the two levels of movement for a length of up to 300 m.

Figure 1. Cross section Km 530.300.

1.3 Engineering intervention Preliminary estimates by SNCF indicated the total volume of the material in movement was of the order of 1,175,000 m3 . The preferred engineering solution was to excavate material from above the cutting slope (i.e. between the cutting and the access road) to the level of the upper failure surface to relieve the disturbing forces driving the slope instability. This option would provide a factor of safety in the region of 1.5. This solution limited the volume of the material to be excavated to 600,000 m3 .

Figure 2. Cross section Km 530.520.

Profil axis I32

I8 I3 PZ1

I2

PZ2

Area of excavation in 2001 PZ3 I24

PZ4

I4pz16

1.4 Additional deep drainage

PZ5

PZ4bis

I1 (7)

(3)

PZ I6 I6

(11)

Ib

Ia (13)

(5024)

Railway

PZsc2

530267-0

530287-1

PZ7

530334-1

530312-0

(4) (10)

PZsc3 I19 Ibb

530357-0

530373-1

I13a (21) pz13

Km 530+300

(20)

I22

530398-0

530398-5

530413-1

PZ9

530438-0

PZsc6

I14

530438-5

530458-1

530483-0

530508-1

530528-0

530552-1

530528-5 530483-5

PZ

PZ6 I20 I13

pz14

I7

(15)

(5)

I18

530357-5

530312-5

530267-5

Ic Icbis

(14) (5028)

(5027)

I5

pz15

(18)

I23 (19)

(16)

Km 530+500

530242-1

sc5

(5026)

Km 530+400

(5023)

After the excavation groundwater levels have been monitored in standpipe piezometers since 2001 to 2005 and shown that in the base of the excavation groundwater level reach level from 0 m to 8 m depth. In December 2002 movements were again recorded by two inclinometers installed during the 2002 ground investigations and by surface movement and datum

(2)

(6)

(17)

I21 (22)

Figure 3. Plan drawing. 280

Altitude mNGF

278

1.2 Geology The geology of the Chabrillan cutting comprises alpine molasses deposits principally formed by sandstone and fresh water limestone. Alpine tectonics have affected the region and the geological structure has been complicated by the presence of geological faults and low angle thrust planes inclined at 15◦ to 20◦ from east to west (i.e., towards the cutting). Groundwater was observed mainly at two levels within the slope generally at depths of 20 m and 33 m below the crest of the cutting.

276

Ground level

274 272 270 268

266

janv-01

janv-02

janv-03

PZ1 Figure 4.

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janv-04

PZ2

janv-05

janv-06

janv-07

PZ3

Water level on piezometer before construction.

point monitoring when the 4AER exceeded c.380 mm which was equated to a triggering groundwater level at the site. On completion of the first stage excavation a 10 m deep trench for a length of 150 m over the most affected area of ground movement. The trench drain would be excavated within the base of the excavation to control the groundwater seepages occurring within the lower landslide zone (−30 mbgl). However this solution suffered a number of limitations as follows: • A 6 m deep trench was difficult to construct and as a tension feature would increase the risk of shear failure over a section 150 m in length. Water level variation of more than 5 m in winter and summer would reduce the effective drawdown of the trench drainage, • Karstification of the aquifer would reduce the predictability of intercepting flows at higher levels. Other solutions have been design but without success as subhorizontal bores drains which involved problems of construction due to their important length and their efficiency in a context of low permeability and complex aquifer. To achieve the required level of drawdown a series of lines of deep pumped wells was considered the most appropriate approach and a number of pump options were considered including electrical submersible, electro-pneumatic and gravity fed siphon wells. Based on the thickness of the landslide that had to be penetrated by the wells, the level of drawdown required and a cost-benefit analysis of the whole life costing of each option, design based of deep electropneumatic or gravity fed siphon wells was designed. Effective cost and duration of construction without acknowledges permitted finally to choice the solution of electro pneumatic drains solutions for the following reasons: • Immerged pumps don’t have good efficiency in low permeability and for low flow—and require high energy in each well. • Limitations of the design of the siphon drainage system included a wide area to drain with a very shallow gradient of the finished ground surface that would have required the use lengths of siphon tubing in excess of 300 m. Previous site experience has shown the effectiveness of the siphon system performance is reduced where the length of tubing exceeds 150 m. This experience was confirmed following a trial carried out on site using an existing piezometer acting as a siphon well. To ensure the length of tubing did not exceed 150 ml, the design included the construction of a 3.5 m diameter vertical shaft to 12 m depth that acts as an outlet manhole for the siphon drainage system. All the automatic flushing hydraulic accumulators will be installed within the single outlet

manhole. This solution was desert for the cost of construction of the shaft. Detail of siphon drain principle is describes below because this solution could offers good alternative for landslide drainage. 2

SIPHON DRAIN AND ELECTRO-PNEUMATIC DRAINS PRINCIPLE

2.1 Siphon drain In the landslip, small diameter siphon drains are placed in vertical drilled drainage wells. These wells are generally spaced at between 3 to 6 metre centers and must be sufficiently deep to reach the layers to be drained. The wells are pumped using siphon tubes and the slope of the ground under the influence of gravity, by introducing the upstream ends of pipes of variable diameters all the way down to the bottom of each well and the down stream end towards an outlet manhole, situated along the slope (Figure 5). If the water level rises in the well, the siphon will flow and abstract water out of the well. The flow will continue until water levels in the well fall, provided that the flow rate in the siphon is sufficient to keep the siphon primed. As the water rises towards the top of the siphon the pressure falls, and may approach a perfect vacuum. In the upstream section the low pressure causes small bubbles to appear. These bubbles tend to coalesce into larger ones further downstream. Two forces act on the bubbles, firstly buoyancy and secondly hydraulic force due to the flow in the pipe. If buoyancy becomes the major force, the bubbles will collect at the summit of the pipe and combine into a single large bubble, which in time would break the siphon flow. This can be avoided by using a system that automatically flushes out bubbles by turbulent flow. The flushing system consists of an arrangement of PVC pipes at the downstream end of the siphon pipe which acts as a hydraulic accumulator. When the water level in the upstream end in the drainage well is nearly the same as the accumulator there is no flow in the

Figure 5.

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Cross section of a siphon drain network.

Figure 6. Siphon tubing and flushing system running.

Figure 8.

Figure 7. Typical outlet manhole and flushing system and typical manhole with siphon tubing and well.

siphon pipe. When the water in both the well and PVC accumulator rises and reaches a certain level, the stored water is quickly emptied by a simple flushing. The sudden lowering of the level of water in the accumulator causes flow in the siphon pipe which is sufficient to flush out any air bubbles within the siphon (Figure 3). It is important that both the siphon pipe and pipework in the accumulator is correctly sized to achieve a sufficient flow rate and duration to remove the air bubbles from the siphon tube. The flow continues until the water level in the well is lowered to the same level as in the accumulator. The water level will then again rise in both the well and accumulator flushing system up to the predetermined level and then the flushing cycle starts again. The siphon system is shown in Figures 2 and 3. 2.2 Electro-pneumatic drains The electro-pneumatic pump has been developed to stabilise landslides by intercepting groundwater at greater depths or lowering groundwater to lower levels than that capable using siphon drain techniques. The electro-pneumatic drains are designed similar to that for the siphon drains with a network of manholes and ducting for electrical cabling, the pipes for water discharge and the compressed air supply. This pumping system is simple, low cost and designed for long term use. The wells are equipped with slotted uPVC well casing of 110 mm internal diameter and centralisers

Electro-pneumatic drains principle.

and fine gravel filter to ensure its central location and filtering of incoming water. The compressor and the air tank can be located up to 3000 meters away from the control panel, if the configuration of the site needs it, and the air coming into the control panel is cleaned and dried by an air dryer and filters. Operating principle: The pump comprises a chamber with ball type non-return valves top and bottom and two electrode sensors. The pump body fills as the water level rises in the well through the lower non-return valve. When the water level reaches the upper electrode a solenoid valve is tripped in the compressor chamber and compressed air at a sufficient pressure is supplied to the pump. This air pressure immediately closes the lower non-return valve and displaces the water in the chamber out through the upper non-return valve to the discharge line. When the lower electrode is exposed the solenoid valve is tripped again and the air supply is stopped allowing the pump chamber to refill. It can be seen that the pump only operates when water is available for pumping minimising energy consumption. The pump operating sequence is given below: 1. Water enters the pump through the lower non-return ball valve 2. When full the upper electrode trips a relay 3. The relay opens a solenoid valve, allowing compressed air to fill the pump, and to push the water out through the outlet tube 4. Once lower electrode senses when the pump chamber is empty and the solenoid valve is tripped An electrical cable, which is connected to the water level detector, is linked through the duct to a control panel. The control panel contains a relay and

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1

2

3

Figure 11. system.

Compressor chamber scheme with the air

Figure 12. pictures.

Compressor chamber and site installation

4

Figure 9. Operating principle.

Figure 10.

Geological plan and well scheme.

solenoid that controls the operation of the compressed air-supply pumps. The solenoid switch controls the compressed air supply allowing it to pass from the air compressor and its air tank to the compressed air inlet tube to the pump.

3

SCHEME DETAILS

The design comprised 47 No. deep drainage wells installed to depths of up to 15 mbgl to 20 mbgl, equipped with gravity electro-pneumatic drains to achieve a groundwater lowering from 10 mbgl to 20 mbgl. The drainage wells were drilled at 200 mm diameter and a slotted uPVC casing of 103 mm internal diameter installed to full depth. The annulus was filled with a fine graded gravel filter of 2 mm to 4 mm size.

Each electro-pneumatic drain was installed within a well through the base of each manhole 1.5 m in depth. All air tubing and electrical cable are going from each well to a compressor chamber where are installed the compressors, and a control panel. The compressor station has been installed in a little house 4 m × 6 m and included: • Two compressors 30 kW to prevent the system from stopping in the event of one unit failing.

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PZ = Piezometer Dot DEi : Well with monitoring system + : Flowmeter

Figure 14. View of the actual excavation—drainage layout—and monitoring point.

Figure 13.

Pump and protective manhole.

• Two air reservoirs 900 Liter • Air Filters • A Condensate management system 3.1 Design and performance verification A comprehensive system of instrumentation and monitoring was established as part of the construction works. The purpose of the instrumentation and monitoring is to assess the effectiveness and performance of the slope drainage works against design assumptions and predictions and the performance specification and to verify long term performance of the scheme. The instrumentation comprised the following: • 5 Pressure transducers and dataloggers installed in observation wells to measure water levels between rows of electro-pneumatic pumping wells. • 9 Transducers piezometers installed in selected electro-pneumatic wells and linked to multi level dataloggers to monitor water level drawdown in the wells. • 3 Borehole inclinometers installed at various locations within the landslide to monitor lateral ground movements. • 1 open channel flow logger • An alarm system on GSM modem to alert in case water level rising, or air control problem,

Figure 15.

Internet access for monitoring control.

• Internet explorer access is available to check system operating—restricted access is control by password and login. All data could be update and unload. 3.2 Discussion and conclusions 3.2.1 Abstraction rates Flow conditions to the wells are variable reflecting the anticipated variation in ground conditions and permeability. Flow rates from electro-pneumatic pump wells and rates of abstraction are variable. The variation in flow rates measured during installation and follow up visits for the electro-pneumatic pump wells are generally in the range from 200 to 1000 l/hr respectively. Total amount of the flow rate recorded was from 0.5 m3 /h to 17 m3 /h. The actual normal flow is between only 0.1 m3 /h to 0.5 m3 /h.

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• The compressor is the essential part of the process. It is generally a screw compressor, which is more reliable, but requires maintenance every 6 months about. The technique has been used recently in European famous landslide problem like on these project:

Figure 16.

Pumps cycles counting monitored on each wells.

3.2.2 Groundwater drawdown In general, the minimum drawdown at the line of wells and the performance specification has been achieved and the ‘‘as constructed’’ drainage system is performing as predicted. However a number of wells give more water than other as recorded on the site as we can see below on the pump cycles monitoring. Piezometer confirmed too the drawdown required between 10 to 20 mbgl. One year of monitoring is actually in service to check the efficiency during one completed year. Actually during the last 6 months, the compressor was only at 10% of its running time in air production so an electrical consumption was about 5 kWatt for all the pumps to keep a lower water level on all the wells. All the pumps give variable flow as we can observed on the monitoring of the pumps cycle counter; the major flow was coming from the middle of the well system. 3.2.3 Electro-pneumatic drains discussions The technique of electro-pneumatics offers many advantages for landslide deep drainage, however it be used also as a temporary solution for dewatering during big excavation because it could be used with only line of wells up to 40 m depth, against well points system which are limited to 7 meters depth. The other advantages are: • Air fed by a small compressor on the surface • Ability to pump muddy or dirty water/low maintenance • Running only when there is water to pump • Drainage to great depths (tested up to 40 m) • Flows up to 30 L/min • Installation either in temporary work, or permanently • Easy installation • System adapted to low permeability grounds (equal or lower than 10−5 m/s). • Remote monitoring association possible The disadvantages of the system are

• Source of power supply required (progress to find to use more wind force or solar panel)

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Castlehaven, located on the southernmost point of the Isle of Wight (UK), has in recent years been under threat from coastal erosion and landslide reactivation. It forms part of the Undercliff landslide system which is one of the largest developed landslide systems in Western Europe approximately 12 km in length and extending up to 0.6 km inland from the coast. The site at Castlehaven is within this environmentally sensitive setting that a solution was sought to improve the stability of the developed landslide while mitigating the significant environmental impact. Extensive ground investigations and monitoring and the development of an accurate ground model confirmed landslide reactivation is being caused by recession of the sea cliffs due to marine erosion, high winter rainfall and groundwater levels and by susceptible geology. This resulted in a risk to both properties and infrastructure valued in excess of £18 m. To reduce the risk of landslide reactivation, but minimising the environmental impact, an innovative geotechnical solution was developed using deep and shallow slope drainage combined with coast protection measures. The groundwater control system developed in France and used for the first time in the UK, comprised 35 electro-pneumatic pumps® driven by compressed air and 121 gravity fed siphon wells® installed within 151 deep drainage wells to depths of up to 25 mbgl. The performance of the drainage system to achieve the specified drawdown was assessed using the Observation Method in conjunction with a comprehensive system of instrumentation, monitoring and reporting. Results of monitoring confirm that the minimum drawdown of groundwater levels has been achieved and the as-constructed drainage system is performing as predicted. To arrest marine erosion a 550 m length of revetment comprising 2 layers of 3–6 tonne rock armour was placed at the toe of the sea cliffs. The scheme won the national British Geotechnical Association Fleming Award in 2005. This paper describes the details of the scheme and the relationship between landslide movement and climate.

REFERENCES A.R. Clark, D.S. Fort, J.K. Holliday, A. Gillarduzzi & S. Bomont. 2007. Allowing for climate change; an innovative solution to landslide stabilisation in an environmentally sensitive area on the Isle of Wight— International Conference on Landslides and Climate Change—Challenges and SolutionsVentnor, Isle of Wight, UK. Bomont S., Fort D.S. & Holliday J.K. 2005. Two applications for deep drainage using siphon and electro pneumatic drains. Slope works for Castlehaven Coast Protection Scheme, Isle of Wight (UK) and slope stabilisation for the Railways Agency, France. In, Proceedings of the International Conference on Landslide Risk Management. 18th Annual Vancouver Geotechnical Society Symposium. Bomont S. 2004. Back experience from four landslides stabilized through lines of siphon drains® in Normandy (France)—9th International Symposium on Landslides 2004—ISL RIO.

Clark A.R., Storm C.V., Fort D.S. & McInnes R.G. 2002. The planning and development of a coast protection scheme in an environmentally sensitive area at Castlehaven, Isle of Wight. Procurement International Conference on Instability, Planning and Management. pub. London: Thomas Telford. pp. 509–518. Clark A..R., Storm C.V., Fort D.S. & McInnes R.G. 2002. The planning and development of a coast protection scheme in an environmentally sensitive area at Castlehaven, Isle of Wight. Proc. Int Conf on Instability, Planning & Management, Thomas Telford. Bomont S. 2002. Drainage with electro pneumatic drains® (Conference JNGG 2002 NANCY—FRANCE). Gress J.C. 2002. Two sliding zones stabilized through siphon drains—International conference on Landslide, slope stability of infrastructures. (SINGAPORE). Gress J.C. 1996. Dewatering a landslip through siphoning drain—Ten years experiences. Proc 7th International Symposium on Landslide (TRONDHEIM).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Experimental geo-synthetic-reinforced segmental wall as bridge abutment R.M. Faure Centre d’Etude des Tunnels, Bron, France

D. Rossi Edico, Nice, France

A. Nancey Tencate Geosynthetics Europe, Bezons, France

G. Auray Texinov, Saint Didier de la Tour, France

ABSTRACT: We present here the results of a full scale experiment on a geosynthetic-reinforced segmental wall used as a bridge abutment. This innovative use of geosynthetics gives the opportunity of mixing three oncoming technologies (woven fiberglass inside the geosynthetic for deformation measurements, segmental blocks, high modulus geosynthetic) and a new computational approach of safety. 1

INTRODUCTION

New technologies allow new building systems. Coming from segmental reinforced walls a technology including concrete knowledge, geosynthetic innovations, new measurement features and new stability evaluation approach is used for an experimental work. (Rossi et al, 2006) The Saint Saturnin abutment, in the west part of France, was chosen for a full scale experimentation including segmental walls, geosynthetic reinforced soil, up to date measurement of geosynthetic deformation by woven glass fibres and a new approach in stability computation. This paper describes these four technologies and presents the experimental abutment, now in use, supporting a new road with heavy traffic.

2

CONCRETE BLOCKS FOR AN EASY USE AND QUICK SETTING

Segmental walls reinforced by geosynthetics are more and more used in road construction. They have technical and economical advantages in comparison with other techniques. Technical improvements are: • Easy furnishment as concrete blocks and geosynthetics can be delivered rapidly.

Figure 1.

The Saint Saturnin abutment.

• Easy building works, as workers can be quickly operational and all blocks being man transportable, without any special system. The possibility of curved walls and architectural aspect are easy to set. • Strong links between two blocks allow taking in account the shear strength of the wall. The code described here uses this strength, limited as experiment show, by the overburden stress at the joint level. • Massive resistance of the wall to shocks when boulders in a river may be projected on the wall gives to the builder a better confidence in its work.

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Table 1.

Figure 2. A concrete block for high wall. This block may be cut in two or four parts for smaller wall.

Product

Material

Mass g/m2

PP150/50 GX07 GX05 GX01 PP450/50

Polypropylene Polyester Polyester Polyester Polypropylene

390 380 365 490 1050

Figure 4.

Tensile strength at break kN/m

Elongation at break %

100 150 100 200 450

20 11 11 11 20

Saint Saturnin abutment profile.

Mass per unit area is obtained following NF EN ISO9864, tensile strength and elongation following NF EN ISO 10319. At Saint Saturnin abutment PP150/50 was used.

Figure 3. Experimental wall to failure.

Figure 2 shows an extruded concrete block, cutable allowing wall more than 10 meters high. With very strong lugs this kind of blocks may be also used in seismic zones. Trials were made by Japanese (Mori et al, (1999)) on seismic table with very strong seismic solicitations, until 1 g horizontal. Figure 3 shows a French static experiment leading the wall to failure.

3

Characteristics of products.

ENLARGED GEOSYNTHETICS CHOICE FOR REINFORCEMENT

During these last years the choice in geosynthetics for civil engineering increases with knitting technology and specific use of new molecules. Warp knitting technology with weft and warp insertion offers high strengths. Other geo-synthetics may be used following the requirements, with other material and other characteristics as listed in Table 1.

4

WOVEN GLASS FIBRE INTEGRATED IN GEOSYNTHETIC FOR DISPLACEMENT EVALUATION

The Geodetect system® , a glass fibre woven inside the geosynthetic and scarred by Bragg nets (Fibre Bragg Gratings), allows the measure of the deformation inside the abutment. (Nancey, 2004) Classical topography gives the displacements on the face of the wall. Inside the abutment three layers of geosynthetics are equipped with Geodetect® and give deformation measures when the apparatus is connected. Measurement is done at every important step of the abutment life. (Wall erection, bridge construction, bridge testing with lorries loads). Figure 4 gives a profile of the abutment and the equipped layers position. Figure 5 gives the evolution in time of layer1 deformation, and Figure 6 the distribution of a layer deformation at a given time. These measures will be

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ma x résiduel

r

Figure 7.

Figure 5. Layer 1 deformation.

Soil behaviour representation.

account the displacements given as a field, from which local displacement is interpolated. (Faure et al, 1988). 16 dots are enough for a good determination of this maximal displacement and like in the CARTAGE code (Delmas et al. 1986), 10 steps of calculation are done. The operator have so, a good idea of the necessary displacement leading to an adequate equilibrium. (Faure et al. 2006) Behaviour law of the soil follows the usual representation of strength strain curves for soils. Figure 7 shows this representation obtained from simple shear tests. By the Coulomb law τmax is obtained and the two other parameters Rf and εr are given by the operator. Rf is defined as:

Figure 6. Deformation along the three layers with Geodetect.

used in the definition of the field displacement given to the computational code. 5

Rf = τrésiduel /τmax

(1)

τmax /τ = (c′ + σn′ tan ϕ′ )/(τred on the curve )

(2)

F1 = (τmax /τ) dl/ dl

(3)

ri = (τmax − τ)/Wi sin αi

(4)

F2 = ri / dl

(5)

For the geosynthetics two sets of fives points give their behaviour in term of shear strength and tensile strength. σn , at each point of the surface failure, is computed in a static manner using the perturbation method (Faure et al, 1996), (Faure, 1985). A local safety factor is computed at each point of the surface failure following: and allows the computation of a global safety factor as:

A NEW CODE FOR TAKING IN ACCOUNT SOIL DEFORMATION

A new computer code, taking in account the displacement at each point of the potential surface failure, quantify a global safety factor and also a safety margin that give to the engineer an accurate knowledge of safety. This code is an extension of Nixes-Mur, a code devoted to the calculation of blocks wall reinforced by geosynthetics and issued from the experimental norm XPG38064. One of its possibilities is to take in account non circular failure surfaces like the one that partially follow a layer of geosynthetics. (Faure et al, 1976), (Faure et al, 1988) This new code Nixes-Mur-Dep still uses elements from Nixes-Mur and adds the possibility to take in

When a geosynthetics crosses the failure surface its contribution is evaluated using the given displacement and is integrated in the equilibrium balance. One can also compute the ratio This ratio can be called ‘safety margin’ better than ‘safety factor’. This safety margin vanishes as the 10 displacement steps are considered. The code gives the summation of this expression following: that is a global safety margin, vanishing as the displacement increases.

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6

ABUTMENT SPECIFICATIONS AND RESULTS

In case of bridge abutment technical features must minimize the displacements. The authorities asked for a strong survey of displacements. The equipment of three layers by Geo-detect system gives a satisfactory answer. Although the authors were confident with displacement calculation, it was decided to double the number of geo-synthetics layers. It was also for creep reason, because it is still a research subject. What is the behaviour of geo-synthetics for the next 100 years (the predicted life of French works) must be evaluated. The abutment was loaded and measures done at each step. The analysis of the numerical results shows that the reinforcement by geo-synthetics, might be divided by two, giving in these conditions, an extra deformation of one centimetre. This interesting result is presented on Figure 8 and Figure 9 which show global factor of safety (increasing curves) and global safety margins (decreasing curves) in two cases, with 23 layers of geo-synthetics and with only 12 layers.

Figure 8. Global factor of safety (increasing curves) and global safety margins in case of 23 layers.

Figure 9. Global factor of safety (increasing curves) and global safety margins in case of 12 layers.

For a target value of global safety factor of 1, one can see that this value is reached at 30% of maximum displacement for 23 layers, and at 60% for 12 layers. That can be interpreted saying that using half of layers, for the same safety, the displacement should be twice. The Saint Saturnin abutment was elected by the French ministry of publics works as a reference work, giving it the IVOR label.

7

CONCLUSIONS

This abutment is an important step in the use of new technologies for civil works. If some questions (how taking in account more accurately creep behaviour of geo-synthetics) remains open, and ask for new laboratory experiments, full scale experiments are for the engineers the best way to integrate and check new ideas.

REFERENCES BSNR, géotextiles et produits apparentés, géomembranes. 2000. Avant projet de norme expérimentale XP G 38064. 43p. Delmas P., Berche J.C. & Gourc J.P. 1986. Le dimensionnement des ouvrages renforcés par géotextile. Bull. liaison Ponts et Chaussées, 142, pp33–44. Faure R.M., Magnan J.P., Moreau M. & Pilot G. 1976. Calcul sur ordinateur des ouvrages en terre. RGRA, 338, pp 25–38. Faure R.M., Rajot J.P. & Chan K.S. 1988. Prise en compte du déplacement pour l’évaluation de la stabilité d’une pente. Proc of 5th ISL Lausanne. Gouria F. 1998. Renforcement des sols par géotextiles. Thèse INSA Lyon, 204p. Faure R.M. & Auray G. 2006. Exploitations des mesures de déplacement faites sur la culée de Saint Saturnin RENCONTRES GEOSYNTHETIQUES 2006, Colloque Francophone, 12–14 Juin 2006, Montpellier. Rossi D., Faure R.M., Ducol J.P. & Nancey A. 2006. Culée de pont porteuse réalisée avec un mur fait de blocs, aspect pierre éclatée et renforcée par des géotextiles. RENCONTRES GEOSYNTHETIQUES 06, Colloque Francophone, 12–14 Juin 2006, Montpellier. Faure R.M., Jolly P., Pham M. & Robinson J. 1996. Stabilité des pentes en trois dimensions. Proc. of 7th ISL congress, Throndeim. Faure R.M. 1985. Analyse des contraintes dans un talus par la méthode des perturbations.; Revue Française de Géotechnique no 33 pp49–59. Nancey A., Rossi D. & Boons B. 2006. Survey of a bridge abutment reinforced by geo-synthetics, with optic sensors integrated in geotextile strips. Geosynthetics, Rotterdam, p 1071–1074. Mori S., Matsuyama T. & Ushiro T. 1999. Shaking table tests of concrete block retaining walls. Slope Stability Engineering Int. Conf., Matsuyama pp657–663.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Rock slope stability analysis for a slope in the vicinity of take-off yard of Karun-3 Dam Morteza Gharouni-Nik Iran University of Science & Technology, IUST

ABSTRACT: Karun-3 is a 205 m high concrete dam constructed at 28 km to the south of Izeh in Iran. The objectives of the construction of this Dam & Hydro-electric Power Plant are electric Power supply and flood control. Karun-3 power generators are connected to the National Power network as the Peak Power generation. With this power plant being operated, with the capacity of 2000 MW, and an average electric power generation of 4137 GW.h/y, a major portion of the electric power shortage in the country will be met. The underground powerhouse complex in this dam is one of the largest of its kind in the world, including powerhouse and transformer caverns, inlet valve gallery and related tunnels, two of which transfers electricity from the transformer cavern to the take-off yard at the surface. The take-off yard has been placed at the toe of a very dangerous slope named G2M, which seemed to be instable after the excavation of the yard. This paper is dealing with the stability analysis of rock slope placed above the take-off yard and proposes the method of stabilizing the slope without which the power plant would confront many difficulties.

1

INTRODUCTION

Karun3 hydropower development project, located in about 25 km north-East of Khoozestan, Iran, comprises a 205 m high arch concrete dam and a gigantic underground complex. This project located on the Karun River and in the Zagros mountain range, is the country’s largest hydro-power scheme. The first stage of the scheme will produce 2000 MW of electricity, which will be upped to 3000 MW in the second stage. The rock slope, called G2M, placed at the top of access road to the spillway of right bank of the dam, is one of dangerous places in dam site. This slope is formed of the layers of limestone, marly limestone and marl and these layers plus tension cracks make the slope as a slope with the potential of sliding. The history of formation of this slope shows that it has been formed by moving the layers and sinking a big wedge of earth materials inside the layers and therefore pushing front layers towards the valley in which access road and take-off yard have been built (Fig 1). It has been seen repeatedly that falling small and large rock blocks endangered the G2M road. Beside that, the surface power installations (Take off yard) placed at the toe of the slope may also expose to rock fall.

Figure 1. Deformation of layers and formation of the earth material wedge at G2M.

2

GEOLOGY OF G2M SLOPE

G2M slope is placed on the south-west shoulder of Keifmalek anticline. The core of anticline is formed of Pabdeh Formation consisting of marly limestone, shale and siltstone, while the shoulders are formed of Asmari Formation consisting of limestone, marly limestone and marl. The dip direction of the layers is 230◦ , their dips are between 35◦ and 80◦ and their strike is NW-SE. Due to placement of the hard layers of limestone among the loose and permeable layers of marl and

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layers. These cracks, which have been observed at the upper parts of the slope, are up to 100 centimeters wide especially after passing the foregoing access road. 3.3

Precipitation

Intensity of rainfall has an important role for initiation of sliding so that sinking a large amount of water inside tension cracks and bedding planes of G2M slope develops high water pressure, resulting in reduction of resistant and increasing destructive forces, which finally causes a reduction in shear strength and intensity of sliding. Figure 2. Type and angle of vertical layers of G2M and formation of toppling type of failure.

marly limestone in the region as well as G2M slope, high ridges of limestone with the heights from tens of centimeters to even ten meters are observed which shows erosion ability and wetherability of marl and marly limestone. At the top portion of the slope, there are various layers of limestone width from 35 cm to 11 m. The width of limestone decreases in lower part of the slope to the average of 90 cm and the amount of limestone has been superseded by marl and marly limestone with interlayer of shale. At the top parts, the dip of the layers varies from 35◦ to 50◦ to the south-west (Fig 2). 3

MAIN REASONS FOR INSTABILITY OF G2M

With regard to foregoing matters, different factors caused instability in G2M slope and one can say that failure of rock blocks and rock masses has been under the effect of combination of following factors: 3.1 Faults Fault at the upper parts of the slope, perhaps is of the primary reasons which separates G2M rock mass from north-east heights and causes slope instability and moving the rock mass downwards. 3.2 Human factor While natural factors are potentials for instability of G2M, human interference has aggravated it with construction of the road to the right bank which has eliminated the toe of the slope. Also the excavation of the take-off yard in front of G2M slope, even in lower level, has deteriorated the factor of safety for this slope. As the forming layers of the slope are mostly limestone, marly limestone and marl which has high potential for sliding, tension cracks has developed in the

4

PRIMARY INSTRUMENTATION OF G2M

After construction of the access road and observation of tension cracks at the top parts of G2M and locating exact place of the cracks, an instrumentation and monitoring program was established in order to find out the direction and rate of probable movements after excavation of the take-off yard. The readings were used in slope stability analysis to determine the critical limits of the slope. General methods in this regards consist of using crack meters for displacement control of tension cracks and surveying points for controlling rock mass movement as a whole. Thirteen 3D crack meters were installed on various points of the slope for observation of crack movements in different directions. These readings were checked with the results of surveying twelve points, conducted in parallel with recording of crack meters readings in dry and wet seasons in order to determine the total movement of the slope. It may be concluded from monitoring stage that in all stations on the slope, the movement of rock mass has been recorded specially in the southern part, which moves towards south west. The rate of movement is also varied and has been intensified for the upper parts of the slope. After monitoring the slope for a period of time in dry and wet seasons and considering the effect of various parameters on instability of the slope, the location of critical sections which have potential for instability were estimated. Three critical sections of this type were chosen for analysis of slope stability with regard to the geological map, condition of the layers, and data collected from boreholes drilled in the rock mass of the slope. 5

SLOPE STABILITY ANALYSIS OF G2M

The main reason for the activities mentioned in previous sections is analysis of stability of the rock slope. In order to perform this analysis and determine

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probable slip surfaces after excavation of the take-off yard, two methods, i.e., analytical and numerical methods were used. The limit equilibrium method presented by SARMA and UDEC software as a distinct element method was utilized for analytical and numerical methods, respectively. Both methods were used for analysis of the foregoing three chosen sections. The most important part for analyzing, particularly numerical method, is to determine input parameters and rock mass characteristics. 5.1

Modeling of G2M slope

In order to analyze the stability of the slope, there is a need for modeling of the problem which simplifies the complicated condition of the site. In this model, topographical condition of the slope, determination factors for construction control, the parameters for mechanical characteristics of rock materials, state of the stress and effective external parameters are to be defined. All the aforementioned factors were used in the models. Effective stresses were estimated from the height of overburden. Perfect elasto plastic behaviour and Mohr Coulomb failure criterion were chosen for rock materials. Detailed characteristics of present joints and layers in the slope were determined and are shown in Table 1. 5.2

Slope stability analysis with SARMA

As mentioned before, SARMA is a software prepared based on the limit equilibrium method for rock slope stability analysis. It may be utilized for:

Table 1. Characteristics of joint sets and layers of G2M slope. Joint sets and layers

Angle of dip

Angle of Persist. Spacing strike (m) (m) φ

C MPa

J1A J1B J2 J3 J4A J4B J5A J5B J6 Main layers Toppled layers Slope surface

83 86 49 39 82 85 39 45 27

233 027 299 039 326 153 322 155 144

2–10 2–10 2–5 2 5–30 5–30 2–10 2–10 1–100

4 6 2–10 5–20 18 18 3 3 10

30 30 40 40 35 35 45 45 35

0.05 0.05 0.05 0.05 0.05 0.05 0.05 0.05 0.05

60–80 230

2–10

4.5

30

0

80–90

2–10

4.5

–

–

30 0 – – – –

50

40–50 235

– non-circular slip surfaces, – rock masses with different characteristics in heterogeneous media, – non vertical slices for simulation of shape and geometry of the slope, – considering water level surface and uplift pressures, – solving the problem for static and dynamic stability state in the same time, and – possibility of computer simulation for solving the problem quicker and easier. As mentioned before, three critical sections were chosen for analyzing with SARMA. Beside the data presented in Table 1, mechanical properties of rock mass were used in this software, as listed in Table 2. With regard to the presented data and running the software for the three sections A-A, B-B and C-C, the following results may be obtained after simulating the excavation of take-off yard in the program: – In general, for most of performed analysis, particularly for sections A-A and B-B, minimum required factor of safety, i.e. 1.4, was not achieved. It is obvious that for worse conditions such as probable earthquake and saturation of the rock mass in wet seasons, this will be deteriorated and the safety factor underlies allowable limits. – Factor of safety decreased in sections A-A and B-B with increasing the level of assumed floor surface. This result supported the presence of a local unconsidered instability in the area between the two sections. – The conditions for section C-C showed a relatively stable situation. However, as the take-off yard is located precisely at the lower part of this section, after complete excavation of this area, factor of safety decreases dramatically along this section. 5.3 Slope stability analysis using numerical method As mentioned in previous sections, distinct element method is used for analyzing stress–strain relation in discontinuous media. In this method, discontinuities play a very important role for the behaviour of rock mass. Simulation of a jointed rock mass in this method is performed using discontinuities in a continuous media and therefore the media is divided to

Table 2. Type of rock

Mechanical properties of rock mass in G2M slope. Lime Marly Crashed stone Marl limestone rock

γ (kN/m3 ) 26 φ (deg) 38 C (MPa) 1.2

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26 35 0.54

26 32 0.9

24.5 25 0.055

Bedding surface – 30 0

Table 3. Input parameters of discontinuities for UDEC. Discontinuities

Dip direction

Angle of dip

φ Deg.

C MPa

Sh. Stiffness (MN/m)

N. Stiffness (MN/m)

Angle of dilat.

Bedding plane Old slip surface Tension crack J2 J5A J5B

229 230 180–250 229 322 155

80 60 90 49 39 45

30 25 0 40 45 45

0.05 0 0 0.05 0.05 0.05

640 320 0 320 320 320

6200 3200 0 3200 3200 3200

0 0 0 5 5 5

Table 4. Input parameters of rock mass for UDEC. Type of rock

γ (kN/m3 )

Bulk Modulus (GPa)

Shear Modulus (GPa)

φ Deg

C MPa

Angle of dilat.

Lime stone Marl Marly limestone Crashed rock

26 26 26 24.5

25 13 18 3

15 8 12 1.8

38 32 35 25

1.2 0.54 0.9 0.055

9 7 8 5

Table 5. The results of analysis for three sections. A-A

B-B

C-C

Section Variable condition

Before TOY excav.

After TOY excav.

Before TOY excav.

After TOY excav.

Before TOY excav.

After TOY excav.

Max. displace. (cm) Max. shear displace. (cm) Points on yield surface Yielded points Shear failure Max. displace. on road trench

125 Start shear 6 731 277

235 139 12 686 302

25 11 18 1689 103

53 19 28 1633 125

32 Start shear 16 2186 255

36 14 24 2120 283

111

238

16

48

some distinct blocks. These blocks may be rigid or deformable. Therefore, one of the general advantages of distinct element methods is the ability of kinematics analysis of rock blocks and probability of modeling of jointed rock masses. As G2M slope contains some discontinuities such as bedding plane and joint sets, the media may be considered as completely discontinuous. On the other hand, this slope has been separated from main anticline. Thus it has been released from tectonic stresses and in-situ stresses are merely of gravitational one. Therefore, structural control of the features in this slope is more important than the effect of stress

10

11

concentration. For these reasons, UDEC software which has prepared based on distinct element method was utilized in this regard. The most important part of working with numerical methods is input parameters so as the validity and accuracy of outputs directly depends on the accuracy of inputs some of which has been listed in Tables 3 and 4. With regard to the aforementioned data and running the UDEC software for the three sections A-A, B-B and C-C, the following results may be obtained after simulating the excavation of take-off yard in the program:

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• With a glance to this point that rock mass is a discontinuous media, present conditions and existence of tension cracks may be considered as the results of small shear displacements in the rock mass and the active forces in producing this condition are still in an instable situation and are directing rock mass to instability with passing of time. • Investigation of resulted block displacements showed that the effect of structural control parameters were more than the effect of state of the stresses in the region of the slope and dominant behaviour of slope rock mass was the behaviour of discontinuities, not the behaviour of the rock material. • In sections A-A and B-B, potentially instable surfaces located in a level upper than the access road and there is a possibility for local instability and block sliding on the surfaces resulted from combination of present joints. • Section C-C had an appropriate condition of stability and there were not any instable surfaces for present situation before excavation of take-off yard.

• Complete excavation of take-off yard had the most important influence in A-A and C-C sections so as for section A-A, beside intensifying the local instable situation, a total instability may be occurred, while for C-C section only after excavation of takeoff yard, potentially instable surfaces would be presented. 6

CONCLUDING REMARKS

Slope stability analysis was performed using limit equilibrium method (SARMA) and distinct element method (UDEC) for three sections. The comparative results obtained from these two analysis methods have been listed in Table 5. REFERENCE Manual of Universal Distinct Element Code (UDEC), 1993, Itasca consulting group Inc. Minnesota.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Stabilization of a large paleo-landslide reactivated because of the works to install a new ski lift in Formigal skiing resort J. González-Gallego, J. Moreno Robles, J.L. García de la Oliva & F. Pardo de Santayana Laboratorio de Geotecnia. CEDEX. M ◦ de Fomento, Spain

ABSTRACT: The paper presents the study and stabilization of a paleo-landslide reactivated because of the excavation works carried out at its foot to construct a new lift in Formigal skiing resort. Inclinometers were installed and surface movements were controlled by GPS techniques in order to analyse the sliding surface and to define the stabilization measures to undertake. The data obtained from the instrumentation was used to perform a back-analysis that allowed to determine the kinematics of the movement as well as to define the appropriate stabilization measures. Presently, the evolution of the movement is controlled by GPS. 1 1.1

GEOLOGICAL CHARACTERISTIC Geology of the area

The slope affected by the landslide is located in the so-called ‘‘Axial Zone’’ of the Central Pyrenees, in north-eastern Spain. The geology of the area is characterized by the presence of Paleozoic materials, mainly Silurian, Devonian and Carboniferous, affected by intense folding corresponding to the Hercynian orogenic processes, and reactivated later on by the Alpine orogeny. The materials present a low degree of metamorphism and a high intensity of fracturing. Colluvial soil deposits, originated from erosion, weathering and transport of Paleozoic materials, are found on stop of the bedrock. These colluvial materials are characterized by the presence of brown to dark grey clays, sometimes highly plastic, with shale, slate, sandstone and limestone pebbles. The presence of different paleo-slides is clearly visible at the hillsides on both banks of Gállego river. The slope studied here is affected by one of the most significant of these paleo-landslides. A deep valley was dug by the Gállego river after the uplift caused by the last phase of the Alpine orogeny. Due to the high alterability of the shales, a thick layer of weathered material was formed on the slopes, consisting of clays with limestone and slates pebbles. The paleo-slide that can be observed presently at the site was originated by the eroding action of the Gállego river at the foot of the hillside and was probable triggered by seismic shaking. As the paleo-slide invaded the river bed, a new channel was dug by the river through the slid mass, without reaching again the elevation of the underlying bedrock.

By means of the indirect methods used by García Ruiz et al. (2003), a period of about 20,000 years BP of significant ground movements, coinciding with the melting process of the Pyrenees glaciers, was determined. 1.2 Geomorphology of the paleo-slide The paleo-slide is 1,500 m long, the difference of elevation between crest and toe being of 400 m. It is bounded laterally by two small creeks oriented parallel to the line of maximum slope angle and separated a distance of 200 to 500 m. At the crest zone, slates dipping 30◦ –40◦ towards the Gállego river outcrop. Below, the slope angle of the hillside is between 8◦ and 12◦ . The surface of the slope is completely convex in the area occupied by the slid mass, and creeping phenomena are frequent there. At the toe of the slide, the Gállego river is in a process of excavating the slid mass tongue. This is the cause of the existence of small slides near the river course. In addition, some ponds, with water all the year around, can be found at mid-height of the slope, due to the local run-off and to the imperviousness of the slid material. Figure 1 shows a panoramic view of the paleo-slide, as well as the location of the main cracks and the pillars of the ski lift. 2

KINEMATIC OF THE SLIDE

2.1 General characteristics As shown in figure 1, the slide was reactivated at an area of accumulation of paleo-slide materials. This

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Figure 3. Evolution of the movement of a pillar located close to the toe of the slide. Figure 1. Panoramic view of the paleo-slide.

upwards, indicating a deep rotational-translational kind of mechanism for the slide.

2.2 Control of movements

Figure 2. Cracks at the crest of the slide.

new slide has a length of 550 m, an average width of 250 m and a depth around 35–40 m. The slide was detected in October 2004 and the cause of the reactivation was associated to the excavation works carried out at the foot of the hillside to construct a new ski lift. The most important cracks appeared at mid-height of the hillside, between pillars 7 and 8 of the ski lift. These cracks, with jumps up to 2 m (see figure 2), form a big circle that cuts all the width of the paleo-slide. Other secondary cracks were observed at a higher elevation, around pillars 9 and 10, as a consequence of the lack of support originated by the slide below. At the toe, the slide reaches the river. No surface cracks were visible at that area. However the movement of a pillar of a bridge over the river, the one located at the paleo-slide bank (see figure 3), shows clearly the occurrence of movements in depth at that zone. Visibly, the head of the slide exhibited a significant settlement, whereas the toe area moved slightly

A total of 93 surface control points by GPS were installed in order to analyse the kinematics of the slide. The points were located in the area presenting significant movements, as well as in zones presumably stable. In addition, 11 inclinometers were installed to detect the situation of the failure surface in depth. The cross-section corresponding to the alignment of the ski lift pillars, located at the center of the slide and along a line of maximum slope, was chosen for analysing the slope movement. The points located at this cross-section at the pillars near the crest of the slide exhibited downward vertical displacements, whereas those situated at the toe of the slope showed a slight upward vertical displacement (figure 4). A similar pattern of surface movements was observed over the entire slide area. Movements were measured from October 2004 to January 2007 and they are still being controlled. As to horizontal movements, figure 5 shows the horizontal projection of the displacement vectors determined for the different control points. The vectors are oriented towards the lower zone of the slope. Displacement rates of up to 7 m/year were registered. The readings of the inclinometers indicated significant rates of movement and showed clearly the position in depth of the failure surface. In general this surface was detected at a depth of about 15 m, near the head of the slide, and of 40 m at the center. It should be noticed that by the river, the failure surface was detected at a depth of 30 m, due to the fact that the river bed is still located in paleo-slide materials, far above from the bedrock elevation.

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0,2 0,1 0

Cracks

-0,1

-0,3 FOOT

-0,4

Vertical Displacements Z(m)

P7

-0,2

P5

P4

P6

P3 P2 I3

I2

I6

I5

I4

river

-0,5 HEAD

-0,6 -0,7

Figure 6. Possible failure circles determined from displacement vectors and inclinometer readings.

-0,8 Pillar 2

Pillar 3

Pillar 4

Pillar 5

Pillar 6

Pillar 7

Pillar 8

Pillar 9

Figure 4. Vertical displacements measured of ground surface at the ski lift pillars.

Table 1. Ground strength parameters obtained in backanalyses. Surface 1

Cohesion Cohesion Cohesion Cohesion

Figure 5. Displacement vectors measured at the surface marks by topographical control.

2.3

Obtaining the failure surfaces

By drawing the perpendiculars to the displacement vectors corresponding to the ski lift pillars, an array of straight lines was obtained converging very accurately on the same point, showing that the movement of the sliding mass could be assumed to be circular. So, by drawing circles passing through the failure points determined by the inclinometers, four different possible failure surfaces were obtained. Back-analysis

A stability analysis was performed considering the four possible failure surfaces, determined as men-

φ(◦ )

SF local

0.5 1.0 1.5 3.0

20.5 19.5 18.5 15.5

0.798 0.828 0.852 0.911

tioned above, in order to determine strength parameters of the materials affected by the landslide. Considering the movements that were going on, a global factor of safety equal to unity was assumed. Therefore, the strength parameters of the materials were estimated on the base of a factor of safety of 1, the actual geometry the slope and the four possible failure mechanisms that had been obtained. In order to calculate the strength parameters of the natural ground, a fixed value was considered for one of the parameters (cohesion or angle of friction), and then the other was obtained by imposing a factor of safety of the order of unity. According to this, four couples of strength parameters were considered for each failure surface geometry in the study. Values of 5, 10, 15 and 30 kN/m2 were adopted for cohesion, the corresponding values for the friction angle being those shown in table 1. In addition, as shown in table 1, local factors of safety were calculated which indicate the minimum factor of safety that take place in localized failures. 3 3.1

2.4

T/m2

STABILIZATION MEASURES Design of stabilization measures

As mentioned before, the origin of the landslide reactivation process was attributed to the excavation work realized at the toe to build a new ski lift.

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Head River

Head

Stabilizing fill Foot

River

Figure 7. Example of one of the failure surfaces analysed.

Figure 8. Cross-section of the landslide showing one of the failure surfaces and the fill at the toe area.

Stabilization measures had to be adopted following two main lines of action:

Table 2.

• Construction of a stabilizing fill in the previously excavated area, to provide a surcharge higher than the weight of the soils that existed before the excavation works. • Drainage of the ground surface to get a depressed water level position in the area of the stabilizing fill. It is important to point out that any stabilization measure to be adopted had to take into account the environmental legal restrictions in force in the area. One of them prohibited drainage operations at the upper part of the landslide, to protect some important ponds for protected species. To determine the stabilization measures, different situations and geometries were considered in the calculations: • Pre-existing situation, before the excavation works; • Situation after execution of drainage and fill; • Situation after construction of fill, but admitting failure of the drainage works. The strength parameters obtained in back-analysis were used in these analyses, performing additional calculations for each of the possible failure surfaces. The geometry adopted for the second case, (situation after execution of drainage and construction of fill) is shown in figure 8. In this case, the water level was considered to be located at the position of the natural ground under the fill. Table 2 presents a summary of the calculations carried out, as well as the resulting factors of safety. From the results obtained in these analyses, it was inferred that the hillside stability presented a factor of safety between 1.14 and 1.26 before the excavation works, situation that being not very satisfactory, was though sufficient. These low values were due to the fact of the slope being formed by a paleo-landslide.

c c c c

0,5 1.0 1.5 3.0

Summary and results of analyses carried out.

Intial state SF

Failure state Fill+drai. Fill without SF SF drai. SF

1.268 1.262 1.256 1.243

0.798 0.828 0.852 0.911

1.545 1.514 1.498 1.455

1.308 1.297 1.288 1.264

As to the second case, with the fill and the effect of the drainage, factors of safety increased significantly to values between 1.28 and 1.54, considered sufficient for the slope stability, taking into account the assumptions adopted in the analyses, in particular, the values adopted for the strength parameters, deduced from back-analyses. In the third case, factors of safety between 1.14 and 1.31 were obtained, clearly below the values calculated with drainage. 3.2

Description of the stabilization measures undertaken

The stabilization measures were carried out to fulfill to main objectives: • Increase of safety against sliding at the toe area by placing there a surcharge fill; • Lowering of water table at the bottom of the landslide area. The fill was constructed with the same material extracted from the excavation, but improving its properties by a proper placing procedure and compaction. Besides, additional weight was provided by increasing by 30% the height of the pre-existing material. In some areas, where the landslide material was softened or presented unsatisfactorily properties, it was substituted also by proper compacted material.

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1600.00

1580.00

Stabilizing Fill

1560.00

1540.00

Original topography

1520.00

Excavated topography 1500.00

River P.C. 1480.00

Figure 9. Detail of final cross-section after construction of the fill compared with the pre-existing geometry.

Figure 10.

Figure 11.

Placement and compaction of fill material.

Figure 12. fill.

Panoramic view of the location of the stabilizing

Construction of the drainage trenches.

At the lowest part of the slope, by the river, buttresses of granular material, 0.6 m wide and 3 m deep, oriented according to the maximum slope direction, were constructed to provide further reinforcement. To allow drainage of the surface prior to the placement of the fill, a system of draining trenches, also oriented in the direction of the maximum slope, was constructed. These trenches were 0.5 m wide and 0.5 m high, and were provided with a filter geotextil, granular material and slotted tube at the bottom of the trench. Distance between trenches was 10 m. Some other trenches were made following an oblique direction in order to facilitate interconnection between them and a better way out for the drainage water. Figure 10 shows the construction of the trenches, whereas figure 11 shows the works carried out to place and compact the fill. Finally, figure 12 shows a panoramic view of the slide after completion of the stabilization works. In the figure, the broken lines represent the main cracks at the head of the slide, whereas the location of the drainage works and stabilizing fill is drawn in continuous line.

4

RESULTS OF THE STABILIZATION

After carrying of the stabilization works, an intense control of movements was implemented to observe the behaviour of the landslide and to verify the efficiency of the solution adopted. This control was performed by GPS techniques using the same net of surface points as before the stabilization. Figure 13 shows how the displacement rates of the control points have decreased after the construction of the drainage system and stabilizing fill. Although the whole slid mass movement continues, as a consequence of its considerable big inertia, it is foreseeable that these displacements will diminish in time. The drainage trench system has proved to be very effective, as a considerable volume of water, up to

1735

the landslide. The fill was constructed using materials obtained from the slid mass, but improving their properties by adequate compaction. Present-day control with GPS techniques permits to carry out an adequate observation of the kinematics of the landslide and has shown that the stabilizing measures undertaken turned out to be effective.

8,00 Stabilizing Fill 7,00

6,00

V (cm/day)

5,00

4,00

3,00

ACKNOWLEDGMENTS

2,00

The authors of this paper would like to thank Javier Andrés (ARAMON, S.A.) and Jordi Castellana (Folia, S.A.) for their contribution to the success of this work.

1,00

0,00

20-10-05

7-8-06

16-8-06

30-8-06

18-9-06

27-9-06

9-10-06

26-10-06 15-11-06

17-1-07

Figure 13. Displacement rates at control points located at the ski lift pillars.

3 l/min in September 2006, has been observed in some of the trenches. 5

CONCLUSIONS

The reactivation of the paleo-landsilde was caused by the excavation works carried out at its toe zone. Control of surface movements by means of GPS techniques permitted to make a preliminary analysis of the kind of failure mechanism that was occurring, which was confirmed from the data obtained with the inclinometers that were installed in the slid mass. The construction of an adequate drainage system and a stabilizing fill has led to the deceleration of

REFERENCES García-Ruiz, J.M., Chueca, J. & Julián, A. 2004. Los movimientos en masa del Alto Gállego. Geografía Física de Aragón. Aspectos generales y temáticos. Universidad de Zaragoza e Institución Fernando el Católico (in Spanish). González-Gallego, J., Moreno, J., García de la Oliva, J.L. & Pardo de Santayana, F. 2006. Análisis de la ladera donde se encuentra el telesilla B-20 en Formigal (Huesca). Diseño de los trabajos necesarios para su estabilización. Technical Report for ARAMON. Laboratorio de Geotecnia. CEDEX. Ministerio de Fomento (in Spanish). Ríos Aragües, S. (Editor). El medio físico y su peligrosidad en un sector del Pirineo Central. Instituto Geológico y Minero de España, serie: Medio Ambiente, num. 1/2001. Ministerio de Ciencia y Tecnología (in Spanish).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

A case study on rainfall infiltration effect on the stability of two slopes M.W. Gui Civil Eng. Dept., National Taipei Univ. of Technology, Taipei, China

K.K. Han Geotechnical Consultant, Bandar Baru Nilai, Negeri Sembilan, Malaysia

ABSTRACT: Two landslides occurred in Malaysia after a heavy rainfall period in 1999. An investigation consisting of site reconnaissance, topography survey, subsurface investigation, laboratory testing and back analysis has been carried out. Rainfall records spanning over a period 42 days before the landslides incident till the occurrence of the landslides were compiled and used to examine the relationship between the landslides and the rainfall. The behavior of these slopes during the rainfall period was examined using a two-stage approach: (1) analyze the infiltration and the ground water flow to obtain the distribution of pore water pressure in the slopes; and (2) analyze the stability of the slopes with the pore water pressure profile obtained in (1). The analysis confirmed that the landslides were indeed related to the long period rainfall.

1

INTRODUCTION

There are many slope failures caused by long period of heavy and intense rainfall in the world, especially in subtropical area. In tropical regions, the hot and humid weather coupled with high annual rainfalls have resulted in the rapid weathering of rock formation and development of a deep overburden of residual soils. The weathering process has made much of tropical area desiccated with degrees of saturation being significantly less than 100% (Han 2001). Research in the area of slope stability has brought about the realization that most slope failures are caused by the infiltration of rainwater into a slope (Gasmo et al. 2000). The main purpose of this paper is to present a case study on the effect of rainfall infiltration on the stability of two residual soil slopes that failed in 1999 after a heavy rainfall period in Malaysia. The stability of these slopes during the rainfall period was examined by first modeling the infiltration and ground water flow in order to obtain the distribution of pore pressure in the slope, and then analyzed the stability of the slopes using the pore pressure obtained. 2 2.1

THEORETICAL BACKGROUND AND METHOD OF STUDY

shear stress τ, net normal stress (σ − ua ), and matric suction (ua − uw ) space (Fredlund & Rahardjo, 1993): τff = c′ + (σf − ua )f tan φ ′ + (ua − uw )f tan φ b (1) where τff is the shear strength on the failure plane; c′ is the intercept of the extended Mohr-Coulomb failure envelope on the shear stress axis where the net normal stress and the matric suction at failure are equal to zero; (σf − ua )f is the net normal stress state variable on the failure plane at failure; (ua − uw )f is the matric suction on the failure plane at failure; and φb is the angle indicating the rate of increase in shear strength relative to the matric suction. The existence of the pore-air and pore-water in the unsaturated soils resulting in the formation of surface tension (matric suction) among the soil particles, which pulls together the soil particles and increases the shear strength of the soils. During rainfall, the increasing water content will decrease the matric suction and, hence, reduce the shear strength of the soil. The shear strength for saturated soils may be expressed in mathematical term by letting ua approaches uw in Eq. 1, and Eq. 1 is now reduced to the renowned Mohr-Coulomb failure envelope: τff = c′ + (σf − uw )f tan φ ′

Shear strength of unsaturated soil

The shear strength of an unsaturated soil may be represented by the extended Mohr-Coulomb envelope in the

(2)

Because obtaining the shear strength of unsaturated soil requires the use of advanced and costly triaxial system, Vanapalli et al (1996) had proposed an empirical formulation that can be used to estimate the shear

1737

strength of unsaturated soil. His empirical formulation simply replaces the term tan φb in Eq. 1 with the following term:   θ − θr tan φ b = tan φ ′ (3) θs − θ r

where θs is the saturated volumetric water content; θr is the residual volumetric water content; and θ is the current volumetric water content. Therefore, by knowing the current volumetric water content of a soil, one can estimate its unsaturated shear strength parameter. 2.2 Total cohesion formulation Conventional slope stability analysis mostly employed the limit equilibrium methods of slices without giving any consideration to the displacement in the soil mass. For saturated slope, effective shear strength parameters c′ and φ′ are generally used in the limit equilibrium calculation. In this case, the shear strength contribution from the negative pore-water pressure above the ground water table is ignored. The shear force mobilized at the base of a slice Sm in Figure 1 can be represented using the following equation: Sm =

β ′ [c + (σn − ua ) tan φ ′ + (ua − uw ) tan φ b ] F (4)

where β is the length across the base of a slice; σn is the total stress normal to the base of a slice; F is the factor of safety which is defined as the factor by which the shear strength parameters must be reduced in order to bring the soil mass into a state of limiting equilibrium along the assumed slip surface (Fredlund & Rahardjo, 1993). The factor of safety for the cohesive parameter c′ and the frictional parameters tan φ′ and tan φb are assumed to be equal for all soil involved and for all slices (Fredlund & Rahardjo 1993). If we consider the

suction term as part of the cohesion of the soil, i.e. by defining the total cohesion c as c = c′ + (ua − uw )f tan φ b

as a result, the conventional factor of safety equation can be used without any modification since the mobilized shear force at the base of a slice now becomes: β Sm = [c + (σn − ua ) tan φ ′ ] (6) F which is identical to the one used in the conventional form and, hence, it is possible to use any slope stability computer program written for saturated soils to solve unsaturated soil problems (Fredlund & Rahardjo 1993). 2.3 Method of study The main purpose of this paper is to present a case study on the effect of rainfall infiltration on the stability of two residual soil slopes in Malaysia. The behavior of the slope during the rainfall period was examined using a two-stage approach: firstly, the infiltration and the ground water flow were analyzed in order to obtain the distribution of pore water pressure in the slope; secondly, the stability of the slope was analyzed using the pore water pressure profile obtained in the previous stage. The respective programs used were SEEP/W and SLOPE/W, developed by Geo-Slope International Ltd. The program SEEP/W is a finite element program that analyzes groundwater seepage, and excess porewater pressure dissipation problems (Khan 2004a). The program utilizes two functions to derive a solution for unsaturated flow: (1) hydraulic conductivity function; and (2) soil-water characteristic function (Khan, 2004a). The magnitude of the maximum negative pore-water pressure is dependent on the shape of the hydraulic conductivity function and, to a lesser extent, on the rate of infiltration. The capability of the soil to store water under changes in pore-water pressures is represented by the soil-water characteristic function. The program SLOPE/W was formulated in terms of moment and force equilibrium factor of safety equations. The method assumed slip surface was circular and two-dimensional plain strain conditions. 3

Figure 1. Forces acting on a slice through a sliding mass with a circular slip surface (after Fredlund & Rahardjo, 1993).

(5)

CASE STUDY

Within a period of thirteen hours between 4.30 pm of May 14 and 5.30 am of May 15, 1999, two massive landslides occurred in the state of Selangor, Malaysia. The cross-sectional views of the two landslides are shown in Figure 2(a) and (b). Both landslides were

1738

(a)

Figure 3.

study area. The daily rainfall intensity, from April 4 to May 15, 1999 is presented in Figure 3. Hence, the rainfall pattern and intensity at the study area could be postulated. There was an intense rainfall for a few hours prior to the first landslide. A total of 33 mm of rain was recorded one hour before the first landslide. The rainstorm on May 14 prior to the landslide was preceded by 81 mm of rain on May 12. A total of 308 mm of rain was recorded thirteen days before the landslide.

(b)

Figure 2. Sectional view of slopes: (a) 1 and (b) 2.

about 1000 m apart. Fortunately, there was no loss of life, except one injury, was reported for the two landslides. About 15,000 m3 of earth and granite boulders have fallen down from the second landslide on to the access road located at the toe of the second landslide. An investigation encompassed site reconnaissance, topography survey, geotechnical investigation, laboratory testing and geotechnical analysis has been carried out to determine the probable causes of the failures. Rainfall analysis was also carried out to improve our understanding in the effects of intense rainfall on the landslides. 3.1 Geomorphology The topography of the collapsed sites was characterized as rugged topography. It was made up by steep granite slopes with gradients ranges between 28◦ to 40◦ . The toe of the slopes was gentle with gradients of about 3◦ to 8◦ . Cutting and filling have been carried out on the slopes to facilitate nearby development projects. To channel the runoff, surface concrete drains had been installed on some part of the slopes but apparently they did not have sufficient capacity to cater the high surface runoff from the upper part of the slopes. 3.2

Daily rainfall between April and May of 1999.

Rainfall

Rainfall analysis was based on data collected at the rainfall station located some 1.5 km away from the

3.3 Soil properties Ground investigation was conducted after the landslide to determine the subsurface conditions of the study area. During the subsurface investigation, disturbed and undisturbed samples were retrieved for visual inspection and laboratory testing. The subsurface profile of the study site could be divided into fill material, residual soil and granite bedrock. Fill material of about 2 to 15 meters and 2 to 30 meters overlying the residual soil in the landslide area adjacent to slopes 1 and 2, respectively. The thickness of the fill material was thicker at the top of the slope and became relatively thinner at the toe for Slope-1, while the thickness of the fill material was thinner at the top of the slope and became thicker at the toe for Slope-2. The sandy clay to clayey sand residual soil was the result of the weathering of the medium to coarse-grained granite. 3.3.1 Shear strength parameters The effective shear strength parameters $(c′ , \φ ′ )$ for the saturated soils involved were obtained from the Consolidation Isotropic Undrained (CIU) triaxial tests on undisturbed samples taken from the failed sites. For Slope-1: the fill has unit weight γ = 17 kN/m3 , apparent cohesion c′ = 2 kPa and angle of friction φ = 32◦ ; the residual soil has properties of γ = 18 kN/m3 , c′ = 9 kPa and φ′ = 32◦ . For Slope2: Layer 1 fill has properties of γ of 17 kN/m3 , c of 1 kPa and φ′ of 30◦ ; Layer 2 fill has properties

1739

of γ = 17 kN/m3 , c′ = 2 kPa and φ′ = 30◦ ; the residual soil has properties of γ = 18 kN/m3 , c′ = 5 kPa and φ′ = 30◦ . 3.3.2 Soil-water characteristic function As mentioned earlier, one of the required input parameters for a transient analysis is the soil-water characteristic function (SWCC). Since it can sometimes be difficult or time-consuming to obtain the SWCC, it may be possible to estimate the SWCC using a closedform solution that requires user-specified curve-fitting parameters. This study used the van Genuchten (1980) four-parameter closed form equation for estimating the SWCC, in terms of volumetric water content, of the study soil: θs − θr θw = θr +    n  m ln 2.71828 + ψa

(7)

where θw is the volumetric water content, θr is the residual volumetric water content, θs is the saturated volumetric water content, ψ is the matric suction, and a, n, m are curve fitting parameters. By taking θs = 0.51, θr = 0.2, a = 70, n = 2, and m = 1, we obtained the SWCC for the study area. 3.3.3 Hydraulic conductivity function The difficult task of measuring unsaturated hydraulic conductivity function directly can be overcome by estimating the function from either a measured or predicted volumetric water content function. One of the methods for predicting the unsaturated hydraulic conductivity from SWCC has been proposed by Green and Corey (1971). They concluded that their method is sufficiently accurate for most field applications. Their equation is: k ( )l =

m ks 30T 2 ξ p  (2j + 1 − 2i) h−2 i ksc μgη n2 j=i

(8)

where k( )1 is the calculated hydraulic conductivity for a specified water content; ks /ksc is the ratio of the measured and calculated hydraulic conductivities; i is the last water content class on the wet end; hi is the negative pore-water pressure head; n is the total number of pore classes between i and m; is the volumetric water content; T is the surface tension of water; ξ is the water-saturated porosity; η is the viscosity of water; g is the gravitational constant; μ is the density of water; and ρ is a parameter that accounts for the interaction of pore classes (Khan 2004a). The term 30T 2 ξ p μgη n2

(9)

is a constant for a particular function and, for simplicity, it has been taken as 1.0 (Khan, 2004a) in the

Figure 4. Contour of pore-water pressures (kPa) on day-1 for slopes: (a) 1, and (b) 2.

determination of the shape of the hydraulic conductivity function. Hence, Eq. 9 and 10 were use to obtained the conductivity function to be used in the following analysis.

4

ANALYSIS

A complete analysis consisted of two stages: (1) seepage analysis and (2) slope stability analysis. The seepage analysis was performed to obtain the pore-water pressure distribution in the slope due to the daily variation of rainfall. Using this pore-water pressure result, the slope stability analysis was then performed to obtain the critical slip circle and factor of safety of the slope. 4.1 Seepage analysis They are two types of flow: steady state and transient flows. Steady-state flow refers to the condition where the temperature, pressure and flow rate of the fluid at any single point in the soil do not change over time; while transient flow refers to the condition where the temperature, pressure and flow rate of the fluid changes over time, as a result, the effective stresses in the soil also affected. By using the 42-day rainfall data (Figure 3) as input, we could then obtain the transient flow or more specifically the pore-water pressure changes over time for slopes 1 and 2 over a period of 42 days. The contour of the pore-water pressure on day 1 of slopes 1 and 2 are presented in Figure 4(a) and (b).

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Figure 5. Contour of pore-water pressures (kPa) on day-42 for slopes: (a) 1, and (b) 2.

It can be seen that there were some suction (negative pore-water pressure) near the surface of the slopes. The contour of the pore-water pressure on day-42 showed that the suction diminished as there were only positive pore-water pressure near the surface of the slopes (Figure 5(a) and (b)). The pore-water pressure profiles over time for section A-A and B-B of Slope-1 are presented in Figure 6(a) and (b), respectively. On day-1, there was a linearly distributed pore-water pressure profile in the slope with some suction value of −4 kPa at the surface of the slope (Figure 6(a)). At day-15, the profile drastically shifted to the right of the figure implicating there was a quick build-up in the pore-water pressure in the slope. At this point, the slope has already become fully saturated. The build-up ended on day-42. Figure 6(b) shows that the pore-water pressure build-up at section B-B of Slope-1 was negligible. The pore-water pressure profiles over time for section A-A and B-B of Slope-2 are presented in Figure 7(a) and (b), respectively. On day-1, there was a linearly distributed pore-water pressure profile in the slope with a suction value of -18 kPa at the surface of the slope (Figure 7(a)). At day-15, the profile shifted to the right of the figure implicating there was a build-up in the pore-water pressure in the slope. The build-up continued until day-42 where the slope became fully saturated. Figure 7(b) also shows a similar pore-water pressure build-up, albeit the build-up was only clearly observed after day-30.

Figure 6. Pore-water pressure profiles at (a) A-A, and (b) B-B of Slope-1.

Using the daily pore-water pressure profiles of each slope and the soil parameters mentioned earlier as the input data for SLOPE/W, it was possible to obtain the factor of safety of the slope throughout the period of interest. It can be seen from Figure 8 that, the factor of safety of the slope at day 1 (April 4) was at about 1.35 and 1.25 for slopes 1 and 2, respectively. On April 5, the factor of safety for Slope-2 started to increase to a value of approximately 1.4 and remained rather constantly until April 23. On the other hand, the factor of safety for Slope-1 started to decrease from 1.35 to 1.22 as a result of the 20 mm rain on April 6. It then increased to and remained at about 1.4, also until April 23, following the 19 days long dry period. 4.2

Slope stability analysis

The factor of safety of Slope-1 then reduces to 1.3 due to the 18 mm of rain on April 24, rebounds to 1.4 and remains constant until April 28 where the raining season began. The factor of safety of Slope-1 then reduces to 0.95 on May 15, at which point the slope

1741

Figure 9.

Slip surface at day-42 for slopes: (a) 1, and (b) 2.

profile. This could be due to the thicker soil layer it has compared to Slope-1. As a result, Slope-2 required a longer time than Slope-1 for the shear strength and mobilized shear stress to be altered.

Figure 7. Pore-water pressure profiles at (a) A-A, and (b) B-B of Slope-2.

5

CONCLUSIONS

The effect of intense rainfall on two landslides has been numerically study in this paper. Using a 42 days rainfall data, it has been possible to show that the stability of the two study slopes have been affected by this long rainfall period. The infiltration of the rainfall into the slopes has reduced the matric suction in the slopes and, hence, the shear strength of the slopes, which is vital in sustaining the stability of the slopes. This is further worsening by the fact that the infiltration of rainfall would also increase the self-weight and, thus, the mobilized shear stress of the slopes. ACKNOWLEDGEMENT

Figure 8. Factor of safety vs time.

failed. On the other hand, the factor of safety of Slope2 increased to 1.7 on April 26, decreased slightly to 1.6 on April 28 before it climbed to 1.9 on May 3. At this point, the factor of safety decreased gradually to 1.7 on May 11, after which it decreased drastically to 0.98 on May 15 where failure occurred. The slip surfaces are shown in Figure 9. The calculated factor of safety profile of Slope-2 did not seem to follow very well with the daily rainfall

The first author acknowledges the partial financial support received from NSC-94-2211-E-027-006 during the course of this work. Thanks are also due to Mr. P.C. Su who has helped in the above analysis. REFERENCES Abramson, L.W., Lee, T.S., Sherma, S. & Boyce, G.M. 2002. Slope Stability and Stabilization Methods. Second Edition, Wiley-Interscience Publication.

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Agus, S.S., Leong, E.C. & Rahardjo, H. 2001. Soil-water characteristic curves of Singapore residual soils. Journal of Geotechnical and Geological Engineering, 19, 285–309. Fredlund, D.G. & Rahardjo, H. 1993. Soil mechanics for unsaturated soils. John Wiley & Sons, Inc., New York. Gasmo, J.M., Rahardjo, H. & Leong, E.C. 2000. Infiltration effects on stability of a residual soil slope. Computers and Geotechnics, 26, 145–165. Van Genuchten, M.T. 1980. A closed-form equation for predicting the hydraulic conductivity of unsaturated soils. Soil Science Society of America Journal, 44, 892–898. Green, R.E. & Corey, J.C. 1971. Calculation of Hydraulic Conductivity: A Further Evaluation of Some Predictive Methods. Soil Science Society of America Proceedings, 35, 3–8.

Han, K.K. 2001. Modeling rainfall-induced landslides. Proc 10th Forum on Engineering Geology & Geotechnics of Slopes, Kuala Lumpur. Krahn, J. 2004a. Seepage Modeling with SEEP/W. GEOSLOPE International Ltd, first edition, 412p. Krahn, J. 2004b. Stability Modeling with SLOPE/W. GEOSLOPE International Ltd, first edition, 406p. Petterson, K.E. 1955. The early history of circular sliding surfaces. Geotechnique, 5, 275–296. Vanapalli, S.K., Fredlund, D.G., Pufahl, D.E. & Clifton, A.W. 1996. Model for Prediction of Shear Strength with respect to Soil Suction. Canadian Geotechnical Journal, 31, 379–392.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Consolidation mechanism of fully grouted anchor bolts Siming He, Yong Wu & Xinpo Li Key laboratory of Mountain Hazard and Surface Process, Chinese Academy of Science, Chengdu, China Chengdu Research Institute of Environment and Disaster in Mountainous Regions of Chinese Academy Science, Chengdu, China

ABSTRACT: Fully grouted anchor bolts have been widely used in slope reclamation and mining reinforcement. However, their consolidation mechanism has not been fully understood still. Problems such as the load-transfer character and the side resistance distribution rule also need further study. Taking the anchor bolt, the grouted materials, the surrounding rock and their interfaces as the research objects, based on the basic shear-lag models and given reasonable assumed conditions, the authors studied the mechanism of the axis force and the side resistance in the fully grouted bolt system. The results have shown that the resistance mainly exists in the head of the anchor bolt and its exponential distribution along the axes, so it is impossible to improve the capacity bearing of bolt effectively by increasing the length of bolt.

1

INTRODUCTION

Fully grouted anchor bolt, which is widely used in the underground engineering, mining engineering and rock slope engineering, plays an important role in controlling the rock deformation and strength distribution properties. Especially in fissured rock, the effect of reinforcement performs more perfectly. However, even been applied so widely, its mechanism especially about the load-transfer character and the side resistance distribution still has not been studied clearly so far. All current design specifications, such as ‘Technical Code for Building Slope Engineering’ (GB50330–2002) and ‘Code for Design on Retaining Structures of Railway Subgrade’ (TB10025–2001), assume the side resistance distribute uniformly along the bolt, but in fact results of many researches (Kilic et al. 2002, Farmer 1975, Jiang 2001, He et al. 2006 & He et al. 2004a) show that the side resistance does not follow the uniform distribution but exhibits a peak value on the head of the anchorage and bolt and increases gradually to zero on the end. Obviously, the current design theories cannot meet the actual projects’ needs. The Shear-lag theory (Cox, 1952) is mainly used to study the interface load-transfer character between the fiber and the body of mix-material. Since being proposed by Cox in 1952, it has been becoming an important tool to study the mechanism of the mix-material and obtains great development and improvement (Gao & Li 2005, Landis & McMeeking 1999, Wang & Hu 2005, Steen & Valles 1998).

In this text, taking the anchor bolt, the grouted materials, the surrounding rock and their interfaces as the research objects, based on the basic shear-lag models, the authors studied the mechanisms and deduced the computing formulas of the axis force and the side resistance in the fully grouted bolt system, studied the side resistance distribution rule and the load-transfer character. All these are benefit for understanding the mechanisms correctly and conducting the anchor bolt design. 2

MECHANISM OF THE FULLY GROUTED ANCHOR BOLT

Figure 1 shows the model of fully grouted anchor bolt system. Base on the Saint-Venant principle, it assumed that the pullout force p only has an influence within radius c to the surrounding rock. Obviously, the system can be divided into three parts—the anchor bolt, grouted material and surrounding rock—by two interfaces among them (see Figure 1). Under the condition of fully grouted, there is no stick-slide on the interface and the harmonious relationship between force and deformation is satisfied. Thus, base on the fundamental solution of shear-lag model and under the rational hypothesis above, the authors studied the mechanism of the parts and their interfaces respectively. 2.1 Load-transfer character of the anchor bolt Base on the fundamental equation of axisymmetric problem and the equilibrium condition of the bolt’s

1745

Integrate r in formula (3) from a to r:

P c b

wm (z, r) − wm (z, a) =

o

(4)

grouted material

Where, wm (z, a) is the axial deformation on the interface between grouted material and anchor bolt. According to the condition of fully grouted on interfaces, it is easy to conclude that:

surrounding rock

l

surrounding rock

grouted material

wm (z, a) = wf (z, a); Z

Figure 1. Schematic illustration of fully grouted anchor bolt.

physical force, some relationships are set up as below (Cai et al. 2004a, b, Zidi et al. 2000 & Zou 2004). dσf (z, a) 2 = − τ (z, a)i dz a

a τ (z, a) r

(2)

Load-transfer character of the fully grouted material

According to the formula (2), the shear stress at any positions in grouting body can be expressed as below: Em a τm (z, r) = τ (z, a) = r 2 (1 + vm ) ×

dwm (z, r) (a ≤ r ≤ b) dr

(5)

Where, wf (z, a) is the axial deformation of the anchor bolt; εm (z, a), εf (z, a) respectively represent the interface axial strain of grouted material and anchor bolt. Specially, the axial deformation on the interface between grouted material and anchor bolt is: wm (z, b) = wm (z, a) +

  2a (1 + νm ) b τ (z, a) ln Em a (6)

After differential, Eq. (4) can be expressed as:

where τ (z, r) is a shear stress of the anchor bolt on any place (include grouted material and surrounding rock). 2.2

εm (z, a) = εf (z, a)

(1)

where σf (z, a) is the axial stress of the anchor bolt at any depth, a is the radius of the anchor bolt, τ (z, a) is the shear stress which exists in the interface between the anchor bolt and grouted material. τ (z, r) =

r 2a(1 + νm ) τ (z, a) ln Em a

εm (z, r) − εf (z, a) =

2a (1 + νm ) dτ (z, a)  r  ln Em dz a (7)

According to the constitutive equation of grouted material and anchor bolt, formula (7) can be written as: σm (z, r) σf (z, a) 2a (1+νm ) dτ (z, a)  r  − = ln Em Ef Em dz a (8) where σm (z, r) and σf (z, a) represent axial stress at any depth which respectively exist in grouted material and anchor bolt; Em , Ef respectively represent elastic modulus of the grouted material and anchor bolt; the meanings of the other signs are as same as before. So, the axial force exists in the grouted material at any depth can be written as: σm (z, r) = 2a (1+νm )

dτ (z, a)  r  Em ln σf (z, a) + dz a Ef (9)

(3)

Where, Em is the grouted material’s elastic modulus; νm is the Poisson’s ratio of the grouted material; b is the drill radius of the anchor bolt; wm (z, r) is the grouted material’s axial deformation on any place.

2.3 Load-transfer character of the surrounding rock under the fully grouted conditions Using the formula (2), the shear stress of the surrounding rock on the any radial position can be expressed as follows:

1746

τr (z, r) =

a Er dwr (z, r) τ (z, a) = (b ≤ r ≤ c) r 2 (1 + vr ) dr (10)

where Er is the surrounding rock’s elastic modulus; νr is the surrounding rock’s Poisson’s ratio; c is the affect radius; wr (z, r) is the surrounding rock’s axial deformation on any place. Integrate r in the formula (10) from b to r:

Base on the equilibrium condition of the anchor bolt system’s axial force at any section, we can get: 0 = π a2 σf (z, a) + +

r 2a (1 + νr ) τ (z, a) ln (11) Er b

wr (z, r) − wr (z, b) =

where wr (z, b) is an axial deformation on the interface between grouted material and surrounding rock. According to the surrounding rock and grouted material’s fully grouted conditions: (12)

wm (z, b) = wr (z, b)

Using formula (6), the formula (11) can be written as: wr (z, r) − wf (z, a) =

2.4 Load-transfer analysis of the anchor bolt system

r 2a (1 + νr ) τ (z, a) ln Er b   b 2a (1 + νm ) τ (z, a) ln + Em a

dθ 0



b

dθ 0



b

a

σm (z, r)rdr

c

σr (z, r)rdr

0 = ξ σf (z, a) + ζ

dτ (z, a) dz

2a (1 + νr ) dτ (z, a) ln Er dz b

r

  b 2a (1 + νm ) dτ (z, a) ln + Em dz a (14) σr (z, r) σf (z, a) 2a (1 + νr ) dτ (z, a)  r  − = ln Er Ef Er dz b   2a (1 + νm ) dτ (z, a) b + ln Em dz a (15) The surrounding rock’s axial force on any place can be expressed as follows:

(17)

(18)

      where ξ = a2 + b2 − a2 EEmf + c2 − b2 EEfr

  b ζ = (1 + νm ) 2ab2 ln − ab2 + a3 a    c 2 + (1 + νr ) 2ac ln − a c2 − b2 b   Er b 2 c − b2 + 2a (1 + νm ) ln a Em Eq. (1) can turn to Eq. (19) by differential: d 2 σf (z, a) 2 dτ (z, a) =− dz 2 a dz

After differential, Eq. (13) can be expressed as:

Er σr (z, r) = σf (z, a) Ef

r + 2a (1 + νm ) + 2a (1 + νr ) ln b   Er b dτ (z, a) × ln Em a dz

2π

2π

Arrange formula (17):

(13)

εr (z, r) − εf (z, a) =





(19)

Using formula (18), the formula (19) can be written as below: d 2 σf (z, a) − k 2 σf (z, a) = 0 dz 2 Where, k 2 =

(20)

2ξ ζa ,

σf (z, a) = A sinh(kz) + B cosh(kz)

(21)

where A and B are undetermined parameters, they can be ascertained according to the boundary conditions. ⎧ ⎨σf (z, a) = p π a2 ⎩ σf (z, a) = 0

z=0

(22)

z=l

Where, l is the length of the anchor bolt.

(16)

⎧ p ⎪ ⎨A = − 2 coth (kl) πa p ⎪ ⎩B = π a2

1747

(23)

Table 1. Basic parameters of bolt and rock.

1800

a(mm)

b(mm)

c(m)

L(m)

Ef (GPa)

Value

10

16

20

4.0

210

Parameter

vf

Em (GPa)

vm

vr

Er (GPa)

Value

0.22

35

0.25

0.28

10

1600

Shear stress (kPa)

Parameter

1400 1200 1000

P=110KN

800 600 400 200

∗ The

physical meanings of the parameters in Table 1 is the same as before.

0

0.4

0.8

1.2

1.6

2

2.4

2.8

3.2

4

3.6

Length of Anchor bolt (m)

On the interface of the anchor bolt, the formula of the shear stress distribution can be written as follows: τ (z, a) =

p k [sinh(kz) − coth(kl) cosh(kz)] 2π a

Figure 2. Theoretical solution of the shear stress along a fully grouted rock bolt.

(24)

4000 3500

p τ (z, b) = k [sinh(kz) − coth(kl) cosh(kz)] 2π b

(25)

3000

Er/Em=0.079 Er/Em=0.142 Er/Em=0.714

2500 2000 1500 1000 500

Summarizing the formula (24) and (25), we can conclude that the distribution rule between anchor bolt and grouted material is the same as that between grouted material and surrounding rock. 3

Shear stress (kPa)

The formula of the shear stress distribution on the interface between grouted material and anchor bolt can be written as:

0

0.4

0.8

1.2

1.6

2

2.4

2.8

3.2

3.6

4

Length of Anchor bolt (m)

Figure 3. Theoretical solution of the shear stress along a fully grouted bolt under the different rigidity ratio between rock and grouted materials.

EXAMPLE

To prove the rationality of the theory, a rock slope consisted of The Jurassic Period sandstone is analyzed as an example. As shown in Table 1, the rock is perfect integrity and all its mechanical properties are good. The bolt is made of the 132 screw thread steel, which length is 4.0 m and the drill hole diameter is 90 mm. The grouting material is M30 cement mortar, which water cement ratio is 0.45. And the equipment used to pour mortar is the type 100/2.5 mortar pumps. According to the document (He et al. 2004b), the effects radius can be calculated out according to the formula: c = 2.5(1 − νm )l

(26)

More correlation parameters of the model are given in table 1. Figure 2 shows the side resistance distribution curve of anchor bolt which is predicted by theoretical model under pullout load, 110 KN. From the figure we can conclude: the side resistance reduced as the exponential rule along the axial and mainly exists in the head of anchor bolt. As is shown, the effective load which

is about 90% of the total loads is supplied by the front 2 meters; Other load which is less than 10% is shared by the rest of bolt, mainly the end of the anchor bolt about 2 meters. So it indicates that it isn’t economical and rational to increase the length of the anchor bolt. Figure 3 shows the relative rigidity variation curve of the anchor bolt’s shear stress. The results shows that the bigger the relatively rigidity of the rock layer, the side resistance is more concentrated on the head of the anchor bolt, the peak value of the shear stress is higher and the distribution of the shear stress along the anchor bolt is less uniform; to the contrary, the relatively rigidity of the rock layer is smaller or the rock layer is made up soft rock, the distribution of the shear stress is more uniform. The theories discussed above are equivalent to those which are deduced from the document (Xu et al. 2002 & You 2000), in other words, the theories discussed in the text are rational, they can be used to predict the distribution of anchor bolt’s side resistance and elaborate the anchor bolt’s load transfer character and mechanism correctly.

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4

CONCLUSION

Based on the basic shear-lag theory and given reasonable assumed conditions, the authors study the load-transfer characters of the fully grouted anchor bolt system and elaborate the system’s mechanisms, especially the distribution rule of the side resistance and axial force. Some conclusions are achieved as following: 1. The distribution of the side resistance on anchor bolt obeys the exponential rule. 2. The resistance mainly exists in the head of the anchor bolt. Contrarily it is very small at the end of the anchor bolt. So it is impossible to improve the bolt’s bearing capacity effectively by increasing its length. 3. The much bigger of the rigidity ratio between the rock and grouted materials, the more uneven of the shear stress distribution; To the contrary, the more softer the rock is, the more even the stress distribute. ACKNOWLEDGMENTS The authors would like to thank the Research Fund of NNSFC Projects for supporting this research. Grant Nos.40572158. REFERENCES Cai, Yue et al. 2004a. A rock bolt and rock mass interaction model, International Journal of Rock Mechanics & Mining Sciences 41(7): 1055–1067. Cai, Yue et al. 2004b. An analytical model to predict axial load in grouted rock bolt for soft rock tunneling, Tunnelling and Underground Space Technology 19(6): 607–618. Chad M. Landis & Robert M. McMeeking. 1999. A shearlag model for a broken fiber embedded in a composite with a ductile matrix. Composites Science and Technology 59(3): 447–457. Cox, H.L. 1952. The elasticity and strength of paper and other fiberous materials. Britain journal of applied physics 1952(3):72–79.

Farmer, I.W. 1975. Stress distribution along a resin grouted rock anchor. International Journal of Rock Mechanics and Mining Sciences 12(11): 347–351. Gao, X.L. & Li, K. 2005. A shear-lag model for carbon nanotube-reinforced polymer composites. International Journal of Solids and Structures 42(5–6): 1649–1667. Siming He et al. 2004a. Analysis on loading Deformation of pre-stressed cable based on damage theory. Chinese Journal of Rock Mechanics and Engineering 23(5): 786–792. Siming He et al. 2004b. Study on mechanism of pre-stressed anchoring cable based on modified shear lag model. Chinese Journal of Rock Mechanics and Engineering 23(15): 2562–2567. Siming He et al. 2006. Study on load transfer of bond prestressed anchor rope, Chinese Journal of Rock Mechanics and Engineering 25(1): 117–121. Jiang, Zhongxin 2001. A Gauss curve model on shear stress along anchoring section of anchoring rope of extensional force type. Chinese Journal of Geotechnical Engineering 23(6): 693–699. Kilic, A et al. 2002. Effect of grout properties on the pull-out load capacity of fully grouted rock bolt. Tunneling and Underground Space Technology 17(2002): 355–362. Steen, M & Valles, J.L. 1998. Determination of in-situ fiber, matrix and interface properties in a composite using tensile tests and an extended shear-lag model. Materials Science and Engineering, A 250(2):217–221. Xiaoling Wang & Gengkai Hu. 2005. Stress transfer for a SMA fiber pulled out from an elastic matrix and related bridging effect. Composites Part A: Applied Science and Manufacturing 36(8): 1142–1151. Nianfeng Xu et al. 2002. Exploration on working mechanism and calculating method of inner bonding section of prestressed cables within rock mass. Journal or Yangtze River Scientific Institute 19(3): 45–61. Chun’an You. 2000. Mechanical analysis on wholly grouted anchor. Chinese Journal of rock Mechanics and Engineering 19(3): 339–341. Zidi, M et al. 2000. Quantitative analysis of the microindentation behaviour of fibre-reinforced composites: development and validation of an analytical model. Composites Science and Technology 60(3): 429–437. Zou, D.H. Steve. 2004. Analysis of in situ rock bolt loading status, International Journal of Rock Mechanics & Mining Sciences 41:762–767.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Stability analysis for cut slopes reinforced by an earth retention system by considering the reinforcement stages W.P. Hong Chung-Ang University, Seoul, Korea

Y.S. Song Korea Institute of Geoscience & Mineral Resources (KIGAM), Daejeon, Korea

T.H. Kim Korea Maritime University, Busan, Korea

ABSTRACT: The purpose of this paper is to analyze the stability of cut slope reinforced by earth retention system according to reinforcement stages by use of a numerical program. The large-scale cut slope is located in the express highway construction site. During partial cutting slope for the construction of bridge foundation, tension cracks were occurred and slope movement was monitored. To improve the stability of failed slope, various earth retention systems such as stabilizing piles, anchors and nails were designed and installed in the upper and lower parts of slope through six stages. In order to ensure and check the stability of slope reinforced by earth retention systems, the slope stability analysis using a limit equilibrium method were performed according to six reinforcement stages. As the result of stability analysis, the safety factor of slope and the deflection of stabilizing piles were varied according to reinforcement stages. The most critical stage during slope reinforcement was modification and excavation stage for installation of nails. Also, the deflection of stabilizing piles was decreased after installation of nails and anchors due to increase resisting force of slope. This paper may be very useful to understand the stabilizing effect of earth retention system during reinforcement stages.

1

INTRODUCTION

During the last few decades, the frequency and distribution of landslides have increased in Korea due to human activities for development of hillside or mountainous areas. Also, the amount and scale of damage caused by landslides have increased every year. The danger of landslides may be variable according to the factors such as topography, soil type, vegetation, and climate etc. in the area. Other factors including non-homogeneity of soil layers, tension cracks, dynamic loading or earthquakes, and seepage flow also affect landslides (Coch, 1995; Keller and Blodgett, 2006). These factors are always acting together and increase the possibility of slope failure. Various analysis methods including finite element method (FEM), discrete element method (DEM), and limit equilibrium method (LEM) have been used for investigating and checking the slope stability based on the various theoretical backgrounds. Most of previous analyses of the reinforced slope stability have been conducted only after finishing slope reinforcement. However, slope stability in the

same location may be changed according to reinforcement stages. In this study, therefore, the slope stability according to the reinforcement stages is investigated analytically by one of LEMs. A large-scale cut slope with failure evidences such as tension cracks and slope movement was especially selected. To improve the stability of failed slope, various earth retention systems such as stabilizing piles, anchors and nails were designed and installed in the upper and lower parts of slope through six stages. Based on the analysis result, the safety factor of slope and the deflection of stabilizing piles are investigated according to reinforcement stages. 2

FIELD CONDITION

2.1 General description The large-scale cut slope is located in the express highway construction site in Donghae, Korea. Figure 1 shows the plan view of landslide outline. During cutting slope for the construction of bridge foundation, three tension cracks were generated at the

1751

The sliding surface of this slope may exist in the silty clay layer at depth of GL (−) 12.0 m–15.5 m. The N-value ranges from 13/30 to 50/26. The weathered rock originated from Phyllite has gray color and is placed at depth of GL (−) 15.5 m–23.0 m. Total Core Recovery (TCR) ranges from 14% to 80%, and Rock Quality Designation (RQD) is 0% on the average. The soft rock is composed Phyllite with gray color and partial limestone and is existed just below the weathered rock layer. TCR ranges from 42% to 100%, and RQD is 50% on the average. The properties of each soil and rock layers from field and laboratory tests are shown in Table 1. Meanwhile, groundwater level is existed at depth of GL (−) 11.7 m–23.7 m. As shown in Figure 2, the groundwater level is ranged from the soft rock at upper part of slope to the weathered soils at lower part.

Figure 1. Plan view of landslide outline.

3

SLOPE REINFORCEMENT

To improve stability of the failed slope, a row of stabilizing piles was constructed at the upper part of slope, and two rows of stabilizing piles with soil nailing were constructed at the lower part as shown in Figure 3. The two rows of piles at the lower part of slope were installed in zigzag type. The stabilizing piles were designed according to the design method presented by Ito et al. (1981, 1982). Table 1. LEM.

Figure 2. Typical profile and geometry of slope.

upper part of slope in the temporary road construction, as shown in Figure 1. The slope height is approximately up to 87 m and the slope width is up to 100 m in the collapsed area. Also, the slope angle ranges from 20◦ to 45◦ (Hong et al., 2004).

Properties of soil and rock layer for analysis of

Layer

Total unit Cohesion Internal fricweight (t/m3 ) (t/m2 ) tion angle (◦ )

Fill Colluvial soil Weathered soil Silty clay Weathered rock Soft rock

1.90 1.85 1.90 1.70 2.20 2.50

1.5 0.5 1.0 1.5 5.0 10.0

25 27 30 20 35 40

2.2 Ground condition Boring tests were performed to investigate the soil deposit profile of this area, and to collect the soils in order to do laboratory tests. Figure 2 shows one of typical profiles and geometries on this slope. The soil deposit from the ground surface is layered of colluvial soils, weathered soils including silty clay, weathered rock and soft rock at the boring locations. The colluvial soils having brown and gray color are composed of boulder, sand and clay at depth of GL (−) 0.0 m–6.5 m. The N-value in the standard penetration test ranges from 13/30 to 50/3. The weathered soils are composed of sand, silt and rock fragment including silty clay layer at depth of GL (−) 6.5 m–15.5 m.

Colluvial soils Weathered soils Silty clay stabilizing piles Weathered rock

508)

embankment

excavating and boring anchoring

Soft rock

stabiliziing piles

508)

soil nailing

Figure 3.

1752

Reinforcement stage to control the slope failure.

A stabilizing pile was constructed by use of steel pipe pile and H-pile. The steel pipe piles with 508 mm diameter were installed into the 1.5 m depth of the soft rock. Steel H-piles (H-250 × 255 × 14 × 14) were inserted into the steel pipe piles and the empty spaces around H-piles in the steel pipe piles were filled by cement grouting to prevent the corrosion of steel H-piles and to improve the stiffness of composition piles. The stabilizing piles installed in a row intervals of 1.5 m between piles. The pile heads were connected by wale and concrete capping. The soil nailing was placed in the front of lower part of the slope installed two rows of piles. The installation length of nails is 12 m, and the installation angle is 30◦ . Nails (HD29) are installed in vertical and horizontal intervals of 1.5 m, respectively. The horizontal drainage holes of 75 mm diameter were installed in order to prevent the rising of groundwater level and to induce the seepage water. As shown in Figure 3, the slope reinforcement was constructed through six stages as follow; 1. Install a row of stabilizing piles in the lower part of slope 2. Install two rows of stabilizing piles in the upper part of slope 3. Excavate the slope to install the soil nails 4. Install the soil nails 5. Install the anchors at the pile head 6. Road embankment in lower part of slope 4 4.1

SLOPE STABILITY ANALYSIS ACCORDING TO REINFORCEMENT STAGES Analysis program of LEM

Using the soil and rock parameters shown in Table 1, the slope stability analysis is performed by numerical method with SLOPILE (VER 3.0) program. SLOPILE (VER 3.0) program can analyze the stability of slope reinforced by earth retention system such as piles, nails and anchors. This program has been developed to analyze and design various slopes containing earth retention system on the basis of LEM (Hong & Song, 2006). Especially, this program can analyze the slope reinforced by piles using an analytical method proposed by Ito et al. (1981, 1982). SLOPILE (VER 3.0) can calculate the slope stability for both planar failure surfaces in the infinite slope and arc failure surfaces based on Fellenius and Bishop simplified methods. SLOPILE (VER 3.0) can be used to analyze and design piles installed in not only the fill slope but also the cut slope. Figure 3 shows a typical cross section of slope reinforced by earth retention system in this site for slope stability analysis. The slope stability analysis was performed partly in upper and lower parts. If the upper

part of slope moves, the lower part of slope will be moved or failed. That is, the stability in the lower part of slope is influenced by that in the upper part of slope. Also, the slope stability analysis was performed in dry and rainy seasons, respectively. The groundwater level in dry season was considered the measured groundwater level shown in Figure 2. The groundwater level in rainy season was used ground surface through the assumption of fully saturated ground. 4.2 Safety factor of slope Table 2 shows the change of safety factor of slope according to reinforcement stages in rainy season. The safety factor at the lower part of slope was increased about 0.50 by installing two rows of stabilizing piles. The safety factor at the upper part of slope also was increased about 0.30 by installing a row of stabilizing piles. When the lower part of slope was excavated for installing the soil nails, the safety factor at the lower part of slope was increased about 0.05 because of decreasing the soil weight, whereas the safety factor at the upper part of slope was oppositely decreased about 0.02 because of reducing resistance force. The safety factor at the lower part of slope was increased about 0.09 by installing soil nails, while the safety factor at the upper part of slope was maintained constantly. The safety factor at the upper part of slope was increased about 0.02 by installing anchors at the pile head. Also, the factor of safety at both the upper and lower part of slopes was increased together by road embankment. Consequently, the safety factor of slope gradually tends to be increased according to reinforcement stages. Especially, the most critical stage during slope reinforcement is the stage of excavating the slope to install the soil nails. Meanwhile, the result of safety factor of slope according to reinforcement stages during dry season was similar to the result of rainy season. 4.3 Behavior of stabilizing piles Figure 4 shows the comparison between behavior of piles measured by instrumentation and calculated by SLOPILE (VER 3.0) program. The deflection of piles measured by instrumentation was excerpted from a paper by Hong et al. (2004). Figure 4(a) shows the comparison result of a row of piles installed in the upper part of slope after road embankment. The measured deflection of piles is smaller than the analytical deflection. Also, the distribution of measured pile deflection is different in that of analytical pile deflection. This reason is the restrained effect of horizontal deflection caused by the jacking force at pile head during the anchor installation. Figure 4(b) shows the comparison result of two rows of piles installed in lower part of slope after road

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Table 2. Variation of the safety factor of slope according to reinforcement stages during rainy season. Safety factor of slope

Reinforcement stage Installing a row of piles in the lower part of slope Installing two rows of piles in the upper part of slope Excavating the slope to install the soil nails Installing the soil nails Installing the anchors at the pile head Road embankment in lower part of slope

Before reinforcement

After reinforcement

Lower part of slope

Upper part of slope

Lower Upper part of part of slope slope

0.582

1.046

1.044

1.046

0.582

1.046

1.044

1.353

0.601

1.017

1.090

1.336

0.601

1.017

1.180

1.336

0.601

1.017

1.180

1.353

1.086

1.031

1.420

1.361

(a) Deflection of pile installed in the upper part of slope

embankment. The measured deflection in the first row of piles is a little smaller than the analytical deflection. However, the distribution of measured pile deflection is similar to that of analytical pile deflection. The error of the two values between measured and analytical deflections ranges from 20 mm to 40 mm, and these errors are rather small. It can be concluded that the prediction of pile deflection using SLOPILE (VER 3.0) may properly simulate the safety factor of slope as well as the deflection of piles according to reinforcement stages.

5

CONCLUSIONS AND SUMMARY

A large-scale cut slope with failure evidences such as tension cracks and slope movement was selected to this study. To improve the stability of failed slope, various earth retention systems such as stabilizing piles, anchors and nails were designed and installed in the upper and lower parts of slope through six construction stages. The slope stability analyses using LEM were performed carefully according to reinforcement stages. Based on the analysis result, the safety factor of slope and the deflection of stabilizing piles were investigated according to reinforcement stages. From this study, the following conclusions could be drawn;

(b) Deflection of pile installed in the lower part of slope

Figure 4. Comparison between analytical and measured pile deflections after road embankment.

1. The safety factor of slope is changed according to reinforcement stages. Especially, the most critical stage during slope reinforcements is the stage of excavating the slope for installation of the soil nails. 2. The safety factor of slopes during rainy and dry seasons gradually tend to be increased according to reinforcement stages. 3. SLOPILE (VER3.0) program can analyze both the safety factor of slope reinforced by earth retention system and the pile behavior installed in slope. 4. The prediction of pile deflection using SLOPILE (VER 3.0) may properly simulate the real deflection of piles according to reinforcement stages.

ACKNOWLEDGMENTS This research was supported by a grant from the Safety Management NETWORK of Infrastructure Research Group, Ministry of Construction and Transportation.

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REFERENCES Coch, N.K. 1995. Geohazards, natural and human. Prentice Hall. 481. Keller, E.A. & Blodgett, R.H. 2006. Natural hazards, Pearson Prentice Hall. 395. Hong, W.P. Han, J.G. Song, Y.S. & Shin, D.S. 2004. Reinforcement effect of stabilizing piles in large-scale cut slope. Proc. 9th Inter. Symp. on Landslides. Rio de Janeiro. Brazil. 2: 1579–1583.

Ito, T. Matsui, T. & Hong, W.P. 1981. Design method for the stabilizing piles against landslide–one row of piles. Soils and Foundations 21 (1): 21–37. Ito, T. Matsui, T. & Hong, W.P. 1982. Extended design method for multi-row stabilizing piles against landslide. Soils and Foundations 22 (1): 1–13. Hong, W.P. & Song, Y.S. 2006. Development of a computer program to analyze stability of slopes reinforced by the earth retention system, Journal of Engineering Geology. 16 (1): 45–58 (in Korean).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Landslide stabilization for residential development I. Jworchan, A. O’Brien & E. Rizakalla Geotechnique Pty Ltd, Penrith, Sydney, Australia

ABSTRACT: A site in West Pennant Hills, Sydney, Australia, is to be subdivided for residential development. The site is located in an unstable hill slope. The Soil Conservation Service of New South Wales, Australia, indicates slope instability will impose major limitations on residential developments in the site. Therefore, geotechnical investigation and slope stability analyses were carried out to assess the risk of slope instability and ascertain if the site can be developed for residential purposes. Sub-surface profile across the site comprises a sequence of colluviums, residual soil and bedrock. The depth to bedrock varies from about 2.0 m–7.5 m from existing ground surface. Inclinometer readings indicate down slope movements of up to 5.0 mm in the upper 4.0 m of profile, confirming risk of slope instability. This paper presents results of geotechnical investigation and slope stability analyses and provides recommendations on appropriate slope stabilization works to make the site suitable for residential development.

1

2.2 Subsurface conditions

INTRODUCTION

A site at Fleur Close in West Pennant Hill, Sydney, is to be subdivided for residential development. The Urban Capability Study of the West Pennant Hills Area (The Soil Conservation Service of New South Wales, 1977) indicates that the risk of slope instability imposes limitations on residential development of the site. Therefore, geotechnical investigation and slope stability analysis were carried out to assess the risk of slope instability. This paper presents results of geotechnical investigation and slope stability assessment and provides recommendations on appropriate slope stabilization works to make the site suitable for residential subdivision. 2 2.1

SITE DESCRIPTION

The site is underlain by Middle Triassic Age Bringelly shale of Wianamatta Group, comprising shale, carbonaceous claystone, laminite, fine to medium grained lithic sandstone and rare coal (Herbert, 1980). The sub-surface profile across the site comprises a sequence of topsoil, colluviums, residual soil and bedrock shale (Geotechnique Pty Ltd, 2006). Fill is present in some portions of the site. Depth to bedrock varies from about 2.0 m–7.5 m, shallower in the northern portions of the site. Bedrock is extremely or distinctly weathered up to depth of about 10.0 m and then becomes slightly weathered or fresh. Possible slip surfaces were evident in three test pits excavated adjacent to boreholes BH10, BH11 and BH12 (Figure 1) at depths of about 4.0 m–6.5 m from existing ground surface, indicating slip surfaces are within residual/colluvial soils and/or at interface between soil and bedrock.

Site features

The site measures approximately 2.1 hectares in area and is located between Fluer Close and Castle Hill Road, West Pennant Hill, Sydney (Figure 1). Northern portion of the site is almost flat and ground surface in the remaining portions dips from the north towards the south. In the mid northern and south western portions, ground surface is dipping at about 20.0–25.0 degrees, with a gently sloping (5.0–10.0 degrees) platform between two steeply dipping portions. A steeply sloping embankment, with drainage depression at the toe, forms the south western site boundary.

2.3 Groundwater conditions Groundwater levels in two standpipes BH10 and BH13 (Figure 1) and rainfall in North Parramatta (nearest meteorological station from the site) as obtained from the Bureau of Meteorology web site (2006) are presented below in Table 1. It should be noted that fluctuations in the level of groundwater occur due to variations in rainfall and/or other factors. Table 1 indicates that the groundwater level during entire monitoring period remained above the slip

1757

Figure 1. Site plan.

surfaces. This shows groundwater conditions could adversely affect the stability of the site.

Table 1. Groundwater levels and rainfalls. Depth to groundwater (m) Measurement Weekly rainfall∗ date (mm) BH10 BH13 01/03/06 16/06/06 30/06/06 18/06/06 ∗ Cumulative

51.4 17.6 9.0 54.0

3.1 4.0 3.8 2.2

rainfall in preceding one week.

2.5 3.8 3.8 3.4

3

INCLINOMETER READING

Inclinometers installed in boreholes BH11 and BH12 (Figure 1) were monitored first on 7 April 2006 and subsequently on 19 June and 18 September 2006 (Geotechnique Pty Ltd, 2006). Results are summarized below:

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BH11 indicates that the sub-surface profile to depth of about 6.0 m is undergoing slope movements. In upper 4.0 m of profile, both down slope and cross slope movements of up to about 2.0 mm were noted. Below depth of 4.0 m, the slope movement was less than 1.0 mm. BH12 indicates that the sub-surface profile to depth of about 7.0 m is undergoing slope movements. In upper 4.0 m of profile, down slope and cross slope movements of up to 5.0 mm were recorded. Below depth of 4.0 m, the slope movement was less than about 2.0 mm. Movements recorded in BH11 are small and do not positively confirm continuing slope movements. Small erratic movements may be related to clay reactivity. However, movements in BH12 are significant to positively confirm some ground movements in upper 4.0 m of profile. Slope movements were observed to increase following rainfall events and rising of ground water levels. It should be noted that significant movements were not recorded in all the identified slip surfaces. 4

LABORATORY TESTS

Large Direct Shear tests were conducted on natural (colluvial and residual soil) and lime stabilized (5% by weight) soil samples compacted to dry density ratio of 95% to 98% standard, at about optimum moisture content, to determine peak and residual strength parameters. Samples were submerged and consolidated under normal stresses between 50 kPa and 200 kPa and then sheared at rate of 0.02 mm/minute. These samples were collected from test pits excavated in vicinity of the site where subsurface profile was similar to that within the site. Considering likely influences from test methods, indicative effective peak strength parameters of compacted natural soil are estimated by cohesion of 15–40 kPa and friction angle of 25.0–30.0 degrees. The corresponding residual strength parameters are 0–5 kPa and 10.0–15.0 degrees respectively (Geotechnique Pty Ltd, 2006). The strength parameters for lime stabilized soil samples are only marginally higher than those for natural soils. 5 5.1

Figure 1 and site conditions in each domain are briefly described below: GD1 is located along the south western portion of the site and extends beyond the western site boundary. This domain is steeply sloping (25.0–30.0 degrees) embankment slope, with a drainage depression at the toe. Some tension cracks are evident at the crest of the embankment. GD2 is located in the south eastern portion of the site, has hummocky ground surface and general slope of 5.0–10.0 degrees. GD3 is located in the mid northern portion of the site, which has been leveled by fill placement. Fill batter is steeply sloping at about 20.0–25.0 degrees. Trees at the toe of embankment are bent. GD4 is located along the northern site boundary and is almost flat with ground slope of less than 5.0 degrees. Based on ‘‘Landslide Risk Management Concepts and Guidelines’’ (Australian Geomechanics Society, 2000), the assessed risk of slope instability for four domains are presented in Table 2. Qualitative assessments indicate domains GD2 and GD4 have very low to moderate risk of instability. But domains GD1 and GD3 have moderate to very high risk of instability. 5.2 Back analysis Slope movements are taking place and likely to continue in some portions of the site, especially in domains GD1 and GD3. Therefore, back analyses were carried out to assess probable strength parameters of shear zone materials. Back analyses involve determining strength parameters for shear zone to achieve a factor of safety of 1.0, for anticipated groundwater level and realistic strength parameters for soils and rock above and below shear zone. Assumed realistic strength parameters for soils and bedrock are presented in Table 3. Back analyses were carried out for a typical section AA (Figures 1 and 2). This section shows two Table 2.

Qualitative risk of slope instability.

Qualitative measure

SLOPE STABILITY ANALYSIS

Likelihood

Risk of slope instability

The site is divided into four geotechnical domains, arbitrarily designated as GD1, GD2, GD3 and GD4, in terms of landslide risk. Approximate boundaries between geotechnical domains are shown in

Domain DG1

Almost certain Probability ≈10−1 Consequence Major Risk level Very high

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Domain GD2

Domain GD3

Domain GD4

Unlikely

Possible

Rare

≈10−4 ≈10−3 ≈10−5 Medium Medium Minor Low to Moderate Very moderate low

Table 3. Assumed strength parameters for natural materials. Soil/rock description

Unit weight (kN/m3 )

Cohesion (kPa)

Friction Angle (◦ )

Colluvial soil Shaley clay Bedrock shale

17.0 18.0 20.0

0.0–5.0 5.0–10.0 15.0–20.0

20.0 18.0–20.0 30

steeply sloping areas, domain GD1 and GD3, with a gently sloping portion, domain GD2, between them. Critical slip surfaces for back analysis were determined using the auto search facility and fully specified slip surfaces in the computer program (Slope/W). For assumed strength parameters presented in Table 3 and varying groundwater levels, estimated strength parameters for shear zone in sections AA, to achieve a factor of safety of 1.0, representing limit equilibrium state, are presented in Table 4. Factor of safety for lower slope (GD1) are critical. Table 4 indicates that slope movements are likely to be initiated when groundwater level is within 3.0 m–4.0 m of existing ground surface, when the strength of shear zone is represented by effective cohesion of 0.0–5.0 kPa and effective friction angle of 13.0–15.0 degrees. Therefore, significant slope movements are likely as groundwater level is expected to reach close to existing ground surface during adverse climatic conditions. Inclinometer measurements also indicate slope movements are likely when groundwater level is within 3.0 m–4.0 m from existing ground surface.

Table 4. Strength parameters for shear zone in Section AA to achieve factor of safety of 1.0. Depth to groundwater (m)

Cohesion (kPa)

Friction angle (◦ )

1.0 2.0 3.0 4.0 5.0 6.0

5.0 5.0 5.0 5.0 5.0 5.0

18.0 16.5 15.0 13.5 11.5 11.0

As slope movements have already occurred, the above cohesion and friction angles correspond to residual strength parameters. Analyses also indicate that the factor of safety for residual strength parameters is in the order of 1.3–1.4, when groundwater level is lowered to depth of 6.0 m from existing ground surface. This level of risk is assessed to be unacceptable for residential development. Therefore, we recommend implementation of appropriate site remediation and/or slope stabilization works to reduce the risk to acceptable level.

6

SLOPE STABILIZATION

Ground movements within domains GD2 and GD4 might have been ceased and/or insignificant. Therefore, domains GD2 and GD4 may be suitable for residential development after minor site preparation works. But significant slope stabilization works will

Figure 2. Cross section.

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be required in domains GD1 and GD3 to reduce risk of slope instability so that these domains are suitable for residential development. Appropriate slope stabilization works for four domains are detailed below. 6.1

Domain GD1

Risk of slope instability in this domain is very high. This domain will be influenced by instability in steep embankment along the site boundary. Appropriate slope stabilization works include: • Installation of surface and sub-soil drainage • Removal of unstable materials and replacing with controlled fill and/or • Installation of engineered retaining structures. If disturbance beyond the site boundary is not permissible during slope stabilization works and if a level platform is required within this domain, slope stabilization works should include installation of retaining structures. 6.2

Domain GD2

This domain will be influenced by instability in domain GD1. Therefore, appropriate stabilization works within GD2 will depend on works carried out in domain GD1 and include: • Installation of adequate surface and sub-soil drainage and • Removal of unstable soil mass and replacing with controlled fill. If retaining structures, if any, within domain GD1 can withstand lateral load imposed by unstable materials within GD2, these materials may be left as they are and future structures supported by pier founded on stable bedrock. If retaining structures within domain GD1 is not capable to withstand lateral load imposed by unstable materials within GD2, these materials should be removed and replaced with controlled fill. If the embankment slope in domain GD1 is battered for long term stability, existing soil mass within GD2 may be left as it is and future structures supported by pier founded on stable bedrock. 6.3

Domain GD3

There are evidences of previous ground movements within this domain. Therefore, appropriate slope stabilization works in this domain include: • Installation of adequate surface and sub-soil drainage and

• Removal of sliding material and replacing with controlled fill and/or • Installation of engineered retaining structures. Retaining structures should be installed if a leveled building platform is required within this domain. 6.4 Domain GD4 This domain covers almost flat ground with ground surface slope of less than 5.0 degrees. The risk of slope instability in this domain is very low and hence this portion of the site is suitable for residential development if the site is managed by normal slope maintenance procedures. Although no evidence of previous ground movement was observed within this domain, care should be taken to ensure landslides are not induced by construction activities. 6.5 Subsurface drainage Sub-surface drainage should be installed to lower groundwater level, and may comprise single graded aggregate of 20 mm–40 mm size, wrapped in nonwoven geofabric, or equivalent materials, installed in trenches excavated to depth of at least 0.5 m into stable bedrock. Sub-surface drainage should be designed for 1 in 100 year rainfall events. Surface water from domains GD3 and GD4 should be diverted to storm water disposal system before reaching domains GD1 and GD2 down slope. Surface regrading during site preparation should allow for no potential ponding of rainwater on ground surfaces. If depth to groundwater level during slope stabilization works is below bedrock level, the drainage system can be installed during removal and replacement of sliding materials. However, if groundwater is encountered above sliding surface, the drainage system should be installed first to ensure that the dry subgrade is prepared before placement of controlled fill and construction of retaining structures. 6.6 Retaining structures It is almost certain that landslide will occur in domain GD1 and this may trigger instability in GD2, which in turn could affect stability in GD3. Therefore, depending on preferred method of slope stabilization works, engineered retaining structures may have to be installed in domains GD1 and GD3. Appropriate retaining structures will comprise anchored contiguous piles or soldier piles, founded, socketed and anchored into shale bedrock. Design of anchored retaining structures (in terms of pile diameter, length, spacing, socket into bedrock and anchor

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• Place controlled fill in horizontal layers not exceeding 250 mm loose thickness and compacted to a minimum dry density ratio of not less than 95% standard and t moisture content within 2% of optimum moisture content. Controlled fill may comprise natural soils and shale obtained from excavations within the site after removal of unsuitable materials and treatment with 2% of hydrated lime. This treatment is desirable to improve shrink swell characteristics and improve strength. • Filled slope should be battered at 1 vertical to 3 horizontal, or flatter for long term stability.

Table 5. Recommended earth pressure parameters. Material description

Unit weight (kN/m3 )

Ka

Ko

Controlled fill Natural soils Weathered shale

18.0 18.0 20.0

0.50 0.55 0.20

0.60 0.65 0.30

Ka = Active earth pressure coefficient. Ko = At rest earth pressure coefficient.

strength, location and length) will depend on anticipated laterals load and acceptable deflection of the piles. Lateral loads may be estimated using earth pressure parameters presented in Table 5.

It will be preferable that removal and replacement works is carried out in conjunction with grading of drainage depression to the western side of the site.

6.7 Removal and replacement

7

Installation of sub-surface drainage and retaining structures, if used, will reduce the risk of instability. In fact, subsoil drainage might be sufficient to alleviate instability problems within the site, especially in domains GD2 and GD4. However, continual monitoring will be required to confirm the success of drainage system. Furthermore, it should be noted that the toe of the landslide in GD1 is located down slope beyond the site boundary. As such, it is unknown whether drainage installation will effectively relieve groundwater pressures within the neighboring site. Therefore, it is suggested that the drainage works are supplemented by installing retaining structures and/or removal and replacement of unstable material. Removal and replacement should commence from GD1 and continue upslope to GD2, GD3 and then GD4, as detailed below. • Excavate sliding/unstable materials to expose stable natural soil or bedrock. The excavation, in each domain, should be started in the higher side and continued down the slope. It is noted that this operation may extend into adjacent domains and beyond site boundary while working in GD1. • The excavation faces may be battered at about 1 vertical to 1 horizontal for short term stability or 1 vertical to 3 horizontal for long term stability. • If the excavation face, after removal of sliding materials, extends to the adjacent domain in higher side by more than about 3.0 times the height of the excavation face, we recommend the following procedures: • The excavation face should be benched so that the bench widths are at least 4.0 m and batter slope between the benches are 1 vertical to 1 horizontal, or flatter.

CONCLUSIONS

At the completion of recommended remediation/ stabilization works, the factors of safety against sliding are 1.3–1.5 and more than 1.5 when groundwater level is within or in excess of 3.0 m of existing ground surface respectively. Therefore the risk of slope instability across the site will be reduced to ‘‘low’’, which is considered to be acceptable for residential development provided drainage system is maintained properly so that groundwater level does not rise within 3.0 m from existing ground surface. Therefore, subsoil drainage might be sufficient to alleviate instability problems. However, continual monitoring will be required to confirm the success of drainage system. Furthermore, it should be noted that the toe of the landslide in GD1 is extends beyond the site boundary. Therefore, it is recommended that the drainage work is supplemented by installing retaining structures and/or removal and replacement of unstable material. REFERENCES Australian Geomechanics Society, Sub-committee on Landslide Risk Management. 2000. ‘‘Landslide Risk Management Concepts and Guidelines’’, 63p. Bureau of Meteorology. 2006. web site www.bom.gov.au. Herbert, C. (Editor), 1983. Geology of the Sydney, Geological Survey of NSW. Department of Mineral and Energy. Map Sheet 9130, 225p. Geotechnique Pty Ltd. 2006. Final Stability Assessment Report, 13 to 19 Castle Hill Road & 12 Fleur Close, West Pennant Hills, Report No 10208/1-AF, 161p. Krahn, J. 2004. Stability Modeling with Slope/w—An Engineering Methodology, Geo-Slope/W International Ltd, 408p. The Soil Conservation Service of NSW. 1973. The Urban Capability Study of the West Pennant Hills Area, Sydney, Australia, 63p.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Influence of load transfer on anchored slope stability S.K. Kim, N.K. Kim & Y.S. Joo Dept. of Civil, Architectural and Environmental System Engineering, Sungkyunkwan University, Suwon, South Korea

J.S. Park Geothech. Engrg. Research Dept., Korea Institute of Construction Technology , Goyang, South Korea

T.H. Kim & K.S. Cha Daewoo Institute of Construction Technology, Daewoo E&C Co, Ltd, Seoul, South Korea

ABSTRACT: This paper presents how the load transfer mechanism of the ground anchor affects on the stability analysis of anchored slope. The finite element analysis and the conventional limit equilibrium analysis on the anchored slope were performed and compared. The limit equilibrium analysis of the anchored slope is quite open used in design practice due to the easiness of the analysis. However, the load transfer mechanism is not considered properly for the analysis. When the failure surface passes through the bonded length of an anchor, the anchor load is disregarded and the factor of safety for the anchored slope is smaller than it should be. In this study, the load transfer distribution was incorporated into the limit equilibrium stability analysis of the anchored slope and the results were compared with the results of finite element analysis. 1

INTRODUCTION

Slope stability analysis is performed either by limit equilibrium methods or by finite element methods. The limit equilibrium methods are widely used in design practice and have proven to be a successful method for the assessment of the stability of the slope. In the limit equilibrium methods, the number of available equations is smaller than the number of the unknown. Therefore, all limit equilibrium methods of the slope stability analysis need to employ the same definition of the safety factor, employing certain assumptions with regard to equilibrium. However, the main disadvantage of the limit equilibrium methods is that the deformation analysis is not possible and the strain distribution along the slip surface is not available. A finite element analysis utilizes a stress versus strain model for the soils involved to calculate the stresses in the soil mass and produces an overall factor of safety that is an expression of the stability of the slope based on the calculated stresses within the slope. Also the finite element method has been used to evaluate the stability of slope reinforced with piles, anchors, or nails. Numerical comparisons have shown that the finite element method is a reliable method for assessing the safety factor of slope and corresponding critical slip surface. This method becomes more popular than before in design practice, but the users should

understand the basics of the method and review the results carefully. The limit equilibrium analysis of the anchored slope is quite open used in design practice due to the easiness of the analysis. However, the load transfer mechanism is not considered properly for the analysis. When the failure surface passes through the bonded length of an anchor, the anchor load is disregarded and the factor of safety for the anchored slope is too conservative. In this study, the load transfer distribution was incorporated into the limit equilibrium stability analysis of anchored slopes. The limit equilibrium analysis considering the load transfer distribution was performed and finite element method was used to evaluate the results.

2

STABILITY ANALYSIS OF ANCHORED SLOPES CONSIDERING LOAD TRANSFER DISTRIBUTION

2.1 Limit equilibrium method of slope stability analysis In the geotechnical field, the limit equilibrium method of analysis has proven to be successful for the assessment of the stability of slopes. In the slop stability analysis, a factor of safety along the assumed failure

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surface must satisfy the minimum requirement. The safety factor is defined as:

T=0 T = T2

FS =

shear strength of soil shear stress required for equilibrium

No . 1

No. 2 T1

(1)

T = T1 + T 2

T2

No .3 T1

In this study, method of slices using simplified Bishop’s rules was selected for the analysis. The anchor loads applied on the slope are treated as distributed radial stresses only along the failure surface as shown in Fig. 1. The radial stress on the midpoint of a slice is calculated using Flamant’s Formula (Tenier and Morlier, 1982): 2 × T × cos θT σr = π ×D

(2)

Where σr is radial stress, T is Equivalent anchor line load, θT is angle between the line of action of the anchor and the line between the point of application of the anchor on the ground surface and midpoint of a slice, and D is distance between the point of application of the anchor on the ground surface and the midpoint of a slice. The radial stresses are summed up into the calculation of normal stresses acting on the bottom of slices. The factor of safety can be defined as the following equation. Fs =

n=p n=1

1 (c′ bn + Wn tan φ ′ ) mα(n) , n=p n=1 Wn sin αn

mα(n) = cos αn + 2.2

T2 unbonded zone

failure surface bonded zone

T1 T2

T : Equivalent anchor line load ( =T1 + T2)

Figure 2.

Intersection of anchor with failure surface.

anchor and the soil. The load transfer rate at failure is approximately uniform along the anchor length (Ostermyer and Sheele, 1977). The effect of the anchor length should be considered for the stability of the anchored slope. Load transfer distribution between grout and soil was measured and predicted by Kim (Kim, 2003) in weathered soils. In this study, the load transfer distributions on anchors (Kim, 2003) were assumed to be uniform into the analysis. For example, the load of anchor No. 2 in Fig. 2 was determined to be T2 by assuming the uniform load transfer along the bonded zone outside of failure surface. For anchors which locate the bonded zone outside the failure surface, the anchor load T (= T1 + T2 ) was considered in the analysis. 2.3 Finite element method of slope stability

tan φ ′ sin αn Fs

(3)

Load transfer mechanism in the ground anchors

In a conventional limit equilibrium analysis of anchored slope dose not consider the load transfer of anchors so that it disregards the anchor load when the failure surface passes through the bonded zone of anchors. The loads of anchors No. 1 and No. 2 in Fig. 2 are disregarded and do not affect the factor of safety. The anchor load is transferred to the soil by the shear stresses developing at the interface between the

The finite element methods can be used with complex slope configurations and soil deposits in two or three dimensions to model virtually all types of mechanisms. General soil material models that include Mohr-Coulomb, Drucker-Prager and numerous others can be employed. Generally, there are two approaches to analyze slope stability using the finite element methods. The one approach is to increase the gravity load until slope becomes unstable and the other approach is to decrease the strength parameters of the slope until slope becomes unstable. In accordance to the shear failure, the factor of safety against slope failure is simply defined as:

bn FAILURE SURFACE

Wn

FS =

τ τf

(4)

T

T

PN n

P D

PT

Figure 1. Transfer of concentrated load to failure surface.

Where τ and τf are the shear strength of the slope material and the shear stress on the sliding surface respectively. The factored shear strength parameters cf and φf are as the following. cf =

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c FS

(5)

2.3H

0 .7

2H

1:

H

H

Figure 3. Finite element mesh.

φf = tan−1



tan φ FS



(6)

This method has been referred to as the ‘shear strength reduction method’ (e.g. Matsui & San, 1992). There are several possible definitions of failure, e.g. some test of bulging of the slope profile (Snitbhan & Chen, 1976), limiting of the shear stresses on the potential failure surface (Duncan & Dunlop, 1969), or non-convergence of the solution (Zienkiewicz & Taylor, 1989). In this study, the non-convergence option is taken as being a suitable indicator of failure. This actually means that no stress distribution can be achieved to satisfy both the Mohr-Coulomb failure criterion and global equilibrium. Slope failure and numerical non-convergence occur at the same time, and are accompanied by an increase in the displacements. In this study, the commercial program (ABAQUS, 2004) was used for the finite element analyses of an unreinforced slope and an anchored slope. A refined mesh was adopted to minimize the effect of mesh efficiency on the finite element analyses, as illustrated in Fig. 3. The mesh consisted of 5365 nodes and 1736 elements for the unreinforced slope and consisted of 5479 nodes and 1775 elements for the anchored slope, and the boundary conditions are given as vertical rollers on the left and right boundary and full fixity at the base. Details of the finite element modeling are presented in this paper. 2.3.1 Soil element model The finite element analysis was two-dimensional plain-strain analysis of elastic-perfectly plastic soils with a Mohr-Coulomb failure criterion utilizing eightnode quadrilateral elements. The Mohr-Coulomb constitutive model was used to describe the soil (or rock) behavior. The Mohr-Coulomb criterion relates the shear strength of the material to the cohesion, normal stress and angle of internal friction of the material. For Mohr-Coulomb material model, six material properties are required. These properties are the friction angle φ, cohesion c, dilation angle ψ, Young’s modulus E, Poisson’s ratio ν and unit weight of soil γ . The dilation angle, ψ affects the volume change during soil yielding. If ψ  = φ, the plasticity flow

rule is considered as ‘‘non-associated’’, and if ψ = φ, the plasticity flow rule is known as ‘‘associated’’. The parameters c and φ refer to the effective cohesion and friction angle of the soil. The effective cohesion and friction angle for the weathered soil were assumed to be 20 kN /m2 , 30◦ respectively. The elastic parameters E and ν have a profound influence on the computed deformations prior to slope failure, but they have little influence on the predicted factor of safety in slope stability analysis (Griffiths, 1999). In this study, slope material assumed to be weathered soils (e.g. E = 105 kN /m2 and ν = 0.3) which is widespread in South Korea. Slope stability analysis is relatively unconfined, so the choice of dilation angle is less important (Griffiths, 1999). Therefore non-associated flow rule with zero volume change (ψ = 0) during yield was used in this study. Soil material parameters used in the finite element analysis are summarized in Table 1. 2.3.2 Anchor and block element model In the simulation of the anchored slope, the anchor installation is simulated by applying a constant load at the location of the block on the slope surface and the same constant load, but in opposite direction, at the bonded zone with 2D eight-noded plain strain truss element of the anchor simultaneously as shown in Fig. 4. In the bonded zone, material properties of anchors are as follows: the cross sectional area of the bonded zone was 12, 668 mm2 and equivalent elastic modulus of the bonded zone was 2.1 × 107 kN /m2 . Then a spring was installed between at the block and at the beginning of the bonded zone to simulate load change in the tie rod (or unbonded zone of anchors). The load changes in the tie rod are governed by the tendon stiffness. The spring behavior can be assumed to be linear so that a constant stiffness value was used throughout the analyses in this study. The tendon stiffness of spring element, K, can be calculated as follow: F = Kδ =



 AE δ Lu

(7)

where F is the anchor lock-off load, Lu is the unbonded length of the anchor, A is the cross section area of the steel tendon, E is the Young’s modulus of the steel tendon, and K is the stiffness of the steel tendon. Table 1. Soil material parameters examined in the finite element analyses. Material

γ (kN/m3 ) ν

Weathered soil 21.0

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Es φ ψ c (kPa) (degree) (degree) (kPa)

0.3 105

30

0

20

3m

Table 3. Results of factor of safety in unreinforced and anchored slope.

Block (Beam Elements) Unbonded Zone ( Spring Elements)

F

Slope

3m 12m

4m

ABAQUS PCSTABL Spreadsheet Remarks

Unreinforced 1.186 Anchored 1.565

R=F

1.216 1.392

– 1.495

3m 4m

3m

= tz = Ko depth

Bonded Zone (Truss Elements)

Initial stresses

z

Figure 4. Schematics of slope reinforced with anchors.

Step 1. specify initial geostatic stress

Table 2. Anchor parameters examined in the finite element analysis. F (kN) 240

Lu (m)

A (m2 )

E (kN/m2 )

K (kN/m)

4.0

2.96 × 10−4

2.0 × 10−8

3.2 × 103

g

Step 2. apply gravity load

Anchor parameters used in the finite element analysis are summarized in Table 2. The block with vertical length of 2 m, horizontal length of 2 m and thickness of 0.3 m on the slope surface was simulated with 2D eight-noded plain strain beam element. This model was treated as a linear elastic material. The Young’s modulus and Poisson’s ratio of the block is 2 × 107 kN /m2 and 0.2 respectively. An interface element was not used between anchors and soil in this study. Therefore, no slippage is expected to occur between the anchors and the soil.

Step 3. install bonded zone of the anchors

2.3.3 Simulation of anchored slope In finite element analyses of slope, two dimensional plane strain conditions are typically assumed. The modeling of the anchored slope consists of five phases as follows: Step 1 – generate the mesh and set up a proper initial geostatic stress distribution in the mesh with all boundary condition fixed, bottom, side and surface; Step 2 – apply gravity load to the mesh and install anchors (truss elements) and blocks (beam elements); Step 3 – apply a constant load at the bonded zone of anchor and at the block; Step 4 – install unbonded zone (spring elements) between at the block and at the beginning of the bonded zone; Step 5 – remove fixed boundary condition of slope surface.

Step 4 apply lock-off load and install steel tendon

Step 5 remove boundary condition of slope surface

Figure 5. Finite element modeling sequence of anchored 3 RESULTS AND DISCUSSION slope.

In this study, unreinforced and anchored slopes were analyzed by using the conventional limit equilibrium

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method, the finite element method and the limit equilibrium method with proposed modeling of anchors. The limit equilibrium analyses considering load transfer mechanism of anchors were performed in spreadsheet by using simplified Bishop approach. The results of finite element method are compared with those of conventional limit equilibrium method and proposed spreadsheet program. In the analyses of unreinforced slopes, the factor of safety in the finite element method and the limit equilibrium method appears to be very similar and the location of the failure surface of developed procedure from the the finite element method is slightly deeper than that of the limit equilibrium method, as shown in Table 3, Figs. 5 and 6. In the analyses of anchored slope, the factors of safety from the finite element method, the conventional limit equilibrium analysis, and the proposed modeling are 1.55, 1.40, and 1.50 respectively. The location of the failure surface from the finite element method is similar to that from spreadsheet but the location of the failure surface in the limit equilibrium method differs from that in the finite element method. The factor of safety using conventional limit equilibrium analyses is calculated to be smaller than that using the finite element method and proposed method. The reason is due to the disregard of the anchor load when the failure surface passes through the bonded zone of anchors.

R=16.2 m

Figure 7.

4

Results of factor of safety in Case 1 and 2.

CONCLUSIONS

In this study, the load transfer distribution was incorporated into the limit equilibrium stability analysis of anchored slopes and the results were compared with the results of finite element analysis. The results of the proposed modeling were evaluated by comparing with the results of the finite element analysis. On the basis of the results of this study, the following conclusions can be drawn:

PCSTABL

a) unreinforced slope

R=10.8m

Spreadsheet

PCSTABL

b) anchored slope

Figure 6. Comparison of failure surface between ABAQUS and PCSTABL and spreadsheet. a) unreinforced slope b) anchored slope.

1. In the unreinforced slope stability analysis, the factor of safety from the finite element method appears to be very similar to that from the limit equilibrium analysis and the location of the failure surface of developed procedure from the finite element method is slightly deeper than that limit equilibrium method. 2. In the analyses of anchored slope, the factors of safety from the conventional limit equilibrium analysis was 1.4, 10 percent smaller than finite element method and the proposed method gave a factor of safety of 1.5, 8 percent larger than the conventional limit equilibrium analysis. The location of the failure surface from the finite element method is similar to that from the proposed limit equilibrium analysis but the location of the failure surface in the conventional analysis differs from that in the finite element analysis.

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ACKNOWLEDGMENTS

Griffiths, D.V., 1985, ‘‘Numerical modeling of interfaces using conventional finite element’’, In: Proc. 5th Int. Conf. Num. Meth. Geomech., Nagoya, pp. 837–844. This research was supported by a grant (C106A1000012Griffiths, D.V. & Lane, P.A., 1999, ‘‘Slope stability analy06A060001211) from Construction Core Technology sis by finite elements’’, Geotechnique, Vol. 49, No. 3, Program funded by Ministry of Construction & Transpp. 387–403. portation of Korea government. Kim, N.K., 2003, ‘‘ Performance of tension and compression anchors in weathered soil’’, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Vol. 129, No. 2, REFERENCES pp. 1138–1150. Kondner, R.L., 1963, ‘‘Hyperbolic stress-strain response: AASHTO, 1990, ‘‘Permanent ground anchor specification cohesive soils’’, Journal of the Soil Mechanics and In-situ soil improvement technique’’, AASHTO-AGCFoundations Division, ASCE, Vol. 89, SM1, pp 115–143. ARTBA TF27 Rep., AASHTO, Washington, D.C. Matsui, T. & San, K.C., 1992, ‘‘Finite element slope stabilABAQUS, Version 6.4, Hibbit, Karlson & Sorensen Inc., ity analysis by shear strength reduction technique’’, Soils Pawtucket, R.I., 2004. Found. Vol. 32, No. 1, pp. 59–70. Briaud, J.L. & Lim, Y., 1999, ‘‘Tieback walls in sand: Ostermayer, H. & Sheele, F., 1977, ‘‘Research on Ground numerical simulation and design implications’’, Journal of Anchors in Non Cohesive Soils’’, Sepcial session 4, IXth Geotechnical and Geoenvironmental Engineering, ASCE, ICSMFE, Tokyo. Vol. 125, No. 2, pp 101–110. PCSTABL5 Manual, Federal Highway Administration U.S. BSI, 1989, ‘‘British standard code of practice for ground Department of Transportation. anchors’’, London, England. Snitbhan, N. & Chen, W.F., 1976, ‘‘Elastic-plastic large Davis, E.H., 1968, ‘‘Theories of plasticity and the failure deformation analysis of soil slopes’’, Comput. Struct. 9, of soil masses’’, In: Lee, I.K., editor, Soil mechanics: 567–577. selected topics, London UK : Butterworth, pp 341–380. Tenier, P. & Molier, P., 1982, ‘‘Influence of ConcenDuncan, J.M., 1996, ‘‘State of the art: limit equilibrium and trated Loads on Slope Stability’’, Canadian Geotechnical finite-element analysis of slopes’’, J. Geotech. Engng, Journal, Vol. 19, pp. 396–400, Feb. ASCE, Vol. 122, No. 7, pp. 577–596. Weerasinghe, R.B. & Littlejohn, G.S., 1977, ‘‘Load transDuncan, J.M. & Dunlop, P., 1969, ‘‘Slopes in stiff fissured fer and failure of anchorages in weak mudstone’’, In: clays and soils’’, J. Soil Mech. Found. Div., ASCE, Vol. 95, Proc., Conference on Ground Anchorages and Anchored SM5, pp. 467–492. Structures, Institution of Civil Engineers, London. Fredlund, D.G. & Krahn, J., 1977, ‘‘Comparison of slope Zienkiewicz, O.C. & Taylor, R.L., 1989, ‘‘The finite element stability method of analysis’’, Canadian Geotechnical method’’, Vol. 1, 4th edn. London, New York: McGrawJournal, Vol. 13, No. 3, pp. 429–439. Hill. Fredlund, D.G., Krahn, J. & Pufahl, D.E., 1981, ‘‘The methods’’, Proceedings of Tenth International Conference on Soil Mechanics and Foundations Engineering, Stockholm, Sweden, 3, pp. 409–416.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Review of slope surface drainage with reference to landslide studies and current practice in Hong Kong T.M.F. Lau, H.W. Sun, T.H.H. Hui & K.K.S. Ho Geotechnical Engineering Office, Civil Engineering and Development Department, Hong Kong SAR, China

ABSTRACT: Landslide investigations undertaken by the Geotechnical Engineering Office have shown that inadequate surface drainage provisions can be a key contributory factor in causing landslides and washout failures in Hong Kong. This paper presents the findings of a review of the landslide hazard scenarios associated with potential surface water drainage problems, together with the design and detailing aspects of surface drainage provisions on slopes. The key issues that warrant particular attention are discussed.

1

INTRODUCTION

Hong Kong has very high annual rainfall of up to about 3300 mm with a mean value of 2200 mm. About 80 percent of the rain typically falls between May and September. One-hour and 24-hour rainfall intensities of more than 50 mm and 250 mm respectively are not uncommon. Landslide studies undertaken by the Geotechnical Engineering Office (GEO) of the Hong Kong SAR Government have shown that inadequate surface drainage provisions can be a major contributory factor in landslides and washout failures on slopes. This paper presents the findings of a review of the landslide hazards associated with the surface drainage problems, together with the design and detailing aspects of surface drainage provisions on slopes. Some areas with respect to surface drainage provisions on slopes that deserve attention and improvement are highlighted.

2

Temporary drainage is generally designed on the basis of a 10-year return period storm except that where the construction period extends over several wet seasons, the design parameters should be greater (GCO, 1984). The surface drainage system for slopes in Hong Kong typically comprises U-shaped concrete channels along slope berms, down-slope stepped channels or cascades, and catchpits at junctions of channels or any change in channel alignment to minimise excessive splashing and turbulence (Figure 1). The key considerations in respect of the design and detailing of surface drainage provisions on slopes are presented below. 2.1 Design guidance Detailed guidance on slope surface drainage design is given in GCO (1984), which emphasized that the promulgated design methods are largely empirical and

SURFACE DRAINAGE DESIGN PRACTICE IN HONG KONG

Guidance relating to slope surface drainage design and construction in Hong Kong is promulgated in various documents (e.g. GCO, 1984; GEO, 2000; DSD, 2000). In routine practice, the Rational Method is commonly adopted for estimating the maximum discharge in catchment areas for slope drainage assessment. The design of the permanent surface drainage system on steep slopes is based on a 200-year return period rainstorm, using the Intensity-Duration-Frequency (IDF) relationship derived from rainfall data from the principal raingauge at the Hong Kong Observatory (HKO).

Figure 1. Typical berm channel and catchpit (left) and stepped channel (right) on man-made slopes in Hong Kong.

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should be used with care, and that knowledge of local precedents for similar slopes is an invaluable aid.

3

HAZARDS ASSOCIATED WITH SURFACE DRAINAGE PROBLEMS

The effects of road drainage and road geometry, together with the role of adverse environmental factors, in causing roadside slope failures had been highlighted by Au & Suen (2001a, b). The importance of road drainage and regular maintenance was emphasized by the 1995 Shum Wan Road landslide, which resulted in two fatalities (GEO, 1996, HyD, 1999). The problems of uncontrolled overland flow in triggering landslides or major washouts have been demonstrated by a number of significant landslide incidents (e.g. MGSL, 2006). The main landslide hazard scenarios associated with potential surface water drainage problems in Hong Kong can be broadly summarized as follows: a. inadequate surface drainage provisions/detailing at slopes and stormwater drain inlets; b. unplanned/uncontrolled discharge of surface water from roads and paved areas, due to inadequate drainage provision upstream, blockage of drains, diversion of surface water flow by landslide debris and unfavourable changes in environmental setting, which has not been catered for in the surface drainage design for the downstream area; c. overflow from catchwater channel; and d. overflow from drainage lines due to blockage by landslide debris from hillside failures or blockage by unauthorised structures. Given the setting of a dense urban development on steep hilly terrain, the potential landslide hazards associated with inadequate surface drainage provisions may result in major social impact and economic losses. The following case studies illustrate some of the hazard scenarios described above.

4

CASE STUDIES

4.1 Washout failures of a newly upgraded fill slope below south lantau road Two washout failures with a total volume of about 150 m3 occurred on a newly upgraded fill slope below South Lantau Road (Figure 2). During heavy rainfall in May 2003, the road above the slope crest was flooded due to partial blockage of the road drainage system. Concentrated surface water flow overtopped the road onto vegetated fill slope at two low points on the road. One of the low points coincided with the

Figure 2. The 5 May 2003 washout failures on a fill slope below South Lantau Road.

location of a previous concrete apron, which originally served as an overflow spillway for the overland flow from the road prior to the slope upgrading works but the apron was removed during the works. The subsequent landslide investigation revealed that the washouts were mainly caused by inadequate design and poor detailing of the surface drainage provisions. This incident highlights the importance of giving due regard to the environmental setting of the site in the design and detailing of slope drainage provisions, in particular the proper detailing of the connections between cross-road culverts and the slope drainage provisions. A comprehensive assessment of all the potential hazards affecting the performance of the slope surface drainage provisions should be undertaken, especially for vulnerable sites with an adverse environmental setting and a history of problems associated with uncontrolled surface water flow. 4.2 Washout failure of a fill slope at Shek O Road Repeated minor washout failures of a newly upgraded fill slope below Shek O Road were observed since the completion of slope upgrading works in November 2003. Site inspections were made during intense rainfall and erosion was observed to have been caused by overspilling from a stepped channel that runs obliquely down the slope face at about 30◦ (Figure 3). The principal cause of the overflow was due to an abrupt change in the water flow directions between the upstream cross-road culvert and the downstream stepped channel across a catchpit. Inadequate depth of the stepped channel on the downslope side and lack of precautionary measures at the vulnerable location (such as provision of elevated sidewalls to the channel on the downslope side of the catchpit) may have contributed to the uncontrolled overspilling. This incident serves to highlight the importance of paying due attention to the proper detailing of slope

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drainage provisions and the appropriate amendment of the detailing during slope works to suit the actual site conditions. 4.3

The 2 July 1997 landslides at the Tao Fung Shan Christian Cemetery

Two major landslides (with a total volume of about 1300 m3 ) affected two man-made slopes within a cemetery (Figure 4). Several abandoned buildings

were damaged as a result of the landslides. A squatter structure was demolished by landslide debris and casualty was narrowly avoided in this ‘near-miss’ incident. The follow-up study (HAP, 1998) revealed that the hilltop development above the source areas of the landslides was paved shortly before the failures. Illegal earthworks involving the construction of a substandard retaining wall, together with inadequate surface drainage provisions within the hilltop development leading to uncontrolled and concentrated surface runoff onto the downhill slopes, were the key contributory factors to the landslides. These incidents highlight, inter alia, the importance of paying due attention to proper surface drainage provisions to avoid uncontrolled discharge of runoff onto vulnerable downhill slopes, particularly for sites with changes in the ground surface cover that could significantly alter the surface runoff characteristics.

5

Figure 3. Overspilling from stepped channel at a fill slope below Shek O Road.

EXAMPLES OF INADEQUATE DETAILING OF SLOPE SURFACE DRAINAGE PROVISIONS

Observations have been made on the detailing of slope drainage provisions in landslide inspections, slope maintenance inspections as well as site visits during heavy rainstorms. These have highlighted some examples of inadequate detailing or construction of surface drainage provisions. The following are some examples: a. presence of sharp bends in drainage channels with no catchpits or baffle walls provided to control potential splashing (Figure 5),

Sharp bend in channel

Figure 4. One of the 2 July 1997 landslides on a cut slope below Tao Fung Shan Christian Cemetery.

Figure 5.

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Example of a sharp bend in drainage channel.

b. inadequate capacity of the downstream drainage provisions to cater for the discharge from the slope resulting in overflow, c. drainage channels constructed with the fall in a direction that is opposite to that of the design, d. presence of obstructions in drainage channels (e.g. downpipes, tree roots, etc) leading to reduction in drainage capacity, e. poor workmanship in the construction of drainage channels, e.g. the top of the sidewalls being above the adjacent ground level, hence leading to erosion along the side of the channel, f. lack of upstands at the downhill side of roads to minimise the chance of uncontrolled discharge of runoff to the downhill slope at low points or vulnerable locations, g. lack of intersecting drains along a long sloping road (which may act as a conduit for surface runoff during heavy rainfall) to reduce accumulated discharge at certain points down the road, which could cause surface erosion or flooding, h. trees in close proximity to drainage channels, hence posing a risk of damage to the channels due to tree root growth (Figure 6), i. undersized drainage channels that can lead to splashing, hence erosion of the vegetated slope surface alongside the drainage channels, j. absence of trash grill or silt traps at inlets to main culverts/drainage channels, making them vulnerable to blockage, especially where the site setting involves major surface runoff during heavy rainfall

Figure 7. Water flow along stairway instead of along adjacent drainage channel.

k.

l. m. n.

o. p.

6

Figure 6. Surface channel damaged by tree roots.

leading to scouring and washout debris from the upstream/uphill area, inadequate protection of the headwalls at inlets to cross-road culverts against water ingress into the road embankment leading to potential subsurface erosion and ground movement, a sufficient number of drainage discharge points not provided, undersized catchpits that are prone to blockage, presence of a concrete stairway adjoining drainage channel that is liable to act as an interceptor and prevent surface runoff from getting into the channel (Figure 7), absence of crest channels or inadequate sizing of crest channels for slopes with a sizeable surface catchment, and poor detailing at the connection between the existing road/stormwater drainage provisions and the slope drainage systems. SOME OBSERVATIONS ON LOCAL DESIGN PRACTICE FOR SURFACE DRAINAGE CHANNELS ON SLOPES

With regard to the design of slope drainage provisions, one of the key uncertainties includes the assessment of rainfall intensity for the design return period. The current guidance as given in GCO (1984) provides an IDF chart based on data from the principal raingauge at the HKO. Evans & Yu (2001) noted that there is much spatial variability of rainfall in Hong Kong. The design rainfall intensity for a given return period could be under-estimated by up to about 50% between the data from the principal raingauge and a local raingauge. Designers are advised to adopt the higher rainfall intensity based on reference to

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both the HKO raingauge data and that of the local raingauges. The design approach for stepped channels has recently been improved by the GEO (GEO, 2007) to adopt the design methodology proposed by Chanson (1994) and DSD (2003), in which the design flow is assumed to be in a skimming flow regime. The improved methodology is to ensure that the design stepped channel would be deep enough to retain the flow in the uniform aerated flow region, where the flow depth is the largest. Irrespective of the design approach adopted, it would be prudent to make some suitable allowance in the sizing of stepped channels and drainage provisions, in order to cater for potential partial blockage of the channels, particularly for vegetated slopes, given the empirical nature of the recommended design approaches and the simplifying assumptions made. Overall, it would seem advisable for designers to take due cognisance of the relevant site-specific environmental factors in making a considered assessment of the necessary allowance for partial blockage of the uphill drainage provisions in order to enhance the redundancy and robustness of the drainage system. 7

AREAS REQUIRING ATTENTION

Extensive guidance has been promulgated in the past (GCO, 1984; GEO, 2000), with the broad principles highlighted to assist slope designers, maintenance agents, etc. Notwithstanding this, inadequate implementation of the recommended good practice has led to inadequate practice in some instances. Poor design, detailing or construction of surface drainage provisions could cause landslides as well as promote breeding of mosquitoes and growth of unplanned vegetation. Key issues that warrant special attention include the following: a. a holistic assessment should be made of the environmental conditions and the overall land drainage characteristics of a site, b. sufficient redundancy in the surface drainage provisions should be allowed for to cater for uncertainties, c. adequate attention should be given to the proper detailing of slope surface drainage provisions, and d. adequate site control by suitably experienced supervisory staff and proper construction of the surface drainage provisions, including timely and suitable amendments made to the detailing and layout to cater for the actual site conditions, are of the essence. The sizing of drainage channels should not be dictated only by hydraulic considerations. Due

allowance should be made for possible blockage and the relevant site-specific environmental factors so as to enhance the redundancy of the system. Surface runoff over vegetated or bare slopes in the uphill area is liable to bring along washout debris, fallen tree branches, leaves, etc. which would increase the chance of blockage of the surface drainage in the downstream areas. This emphasizes the importance of regular maintenance as well as the need for the provision of suitable precautionary measures (e.g. trash grills or debris screens at inlets of major culverts). Trash grills or debris screens, where provided, will require regular maintenance to clear blockages, as otherwise the blockage could back up the water flow and result in uncontrolled overspilling. An alternative approach based on a methodology similar to that in hazard and operability studies may be considered for the design and assessment of drainage systems, as suggested by some practitioners. Under this approach, designers would pose questions on what could happen to cause water to spill over various sections or points of the drainage system, assess the volume and possible flow paths of the spilled water, examine the consequences and assess what needs to be done if the consequences are not acceptable. In principle, designers should endeavour to take due account of the site-specific details and exercise judicious judgement in the design and detailing of slope drainage provisions.

8

CONCLUSIONS

Inadequate surface drainage provisions have been found to be contributory factors in some of the landslides and washout incidents. A review of slope surface drainage provisions based on the findings from landslide studies, together with field observations, has identified areas that deserve attention. The key issues that warrant attention are highlighted in this paper, with suggestions made on how to further enhance the practice in respect of the design and detailing of slope surface drainage provisions. ACKNOWLEDGEMENTS This paper is published with the permission of the Head of the Geotechnical Engineering Office and the Director of Civil Engineering and Development, Government of the Hong Kong Special Administrative Region. REFERENCES Au, S.W.C. & Suen, R.Y.C. 2001a. The effect of road drainage and geometry in causing roadside slope failures.

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The Practice of Geotechnical Engineering in Hong Kong, pp 73–79. Au, S.W.C. & Suen, R.Y.C. 2001b. Environmental factors in triggering slope failures. The Practice of Geotechnical Engineering in Hong Kong, pp 89–97. Chanson, H. 1994. Hydraulic Design of Stepped Cascades, Channels, Weirs and Spillways. Pergamon, Oxford, England, 261 p. Drainage Services Department. 2000. Stormwater Drainage Manual—Planning, Design and Management (Third Edition). Drainage Services Department, Hong Kong, 162 p. Drainage Services Department. 2003. Design of Stormwater Inlets (DSD Practice Note No. 1/2003 Version 2). Drainage Services Department, Hong Kong, 26 p. Evans, N.C. & Yu, Y.F. 2001. Regional Variation in Extreme Rainfall Values. GEO Report No. 115, Geotechnical Engineering Office, Hong Kong, 81 p. Fugro Scott Wilison Joint Venture. 2005. Detailed Study of Distress at Slope No. 15 NE-B/FR31 Shek O Road near Lan Nai Wan. Landslide Study Report No. LSR 4/2005, Geotechnical Engineering Office, Hong Kong, 92 p. Geotechnical Control Office. 1984. Geotechnical Manual for Slopes (Second Edition). Geotechnical Control Office, Hong Kong, 295 p. Geotechnical Engineering Office. 1996. Report on the Shum Wan Road Landslide of 13 August 1995. Volume 2: Findings of the Landslide Investigation. Geotechnical Engineering Office, Hong Kong, 51 p.

Geotechnical Engineering Office. 2000. Highway Slope Manual. Geotechnical Engineering Office, Hong Kong, 114 p. Geotechnical Engineering Office. 2003. Guide to Slope Maintenance (Geoguide 5) (Third Edition). Geotechnical Engineering Office, Hong Kong, 132 p. Geotechnical Engineering Office. 2007. Hydraulic Design of Stepped Channels on Slopes. GEO Technical Guidance Note No. 27. Geotechnical Engineering Office, Hong Kong, 16 p. Halcrow Asia Partnership Limited. 1998. Detailed Study of the Landslides at Tao Fung Shan Christian Cemetery on 2 July 1997. Landslide Study Report No. LSR 2/98, Geotechnical Engineering Office, Hong Kong, 46 p. Highways Department. 1999. Final Report on Discharge of Water along Roads at Crest of Slopes. Technical Report No. RD/TR/032, October 1999. Highways Department, Hong Kong. 20 p. Maunsell Geotechnical Services Limited. 2006. Review of the 20 August 2005 Debris Flood, Lo Wai. Tsuen Wan. Landslide Study Report No. LSR 5/2006, Geotechnical Engineering Office, Hong Kong, 100 p.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Analysis of dynamic stability about prestressed anchor retaining structure Hui Li, Xiaohong Yang & Hongfei Liu School of Civil Engineering, Xi’an Univ. of Arch. & Tech. Xi’an, China

Lili Du The 5th Metallurgical Construction Corp. Engineering Branch, Chengdu, China

ABSTRACT: The paper fully considers the influence of the design parameters on slope safety and stability factors, which include water, anchor parameters (such as, anchor length, spacing and layout, angle, surface layer thickness) and soil properties. The primary factor indexes of the retaining structure stability are expressed in trapezoid fuzzy numbers. The fuzzy utility value, function and the importance of each possible affecting factor on the slope stability are determined by using FMADM Buckley. The safety margin of retaining structure is taken as the objective function. The dynamic stability analysis models of retaining structure are established. The safety margin of prestressed anchor flexible retaining structure is calculated in all possible conditions. The result will serve as the comprehensive, reasonable and scientific decision-making basis for engineering design, construction and safety management. 1

INRODUCTION

The prestressed anchor retaining structure is a common retaining method in deep foundation pit and slope, with the characteristics of early and quick loadbearing, and the stresses in the surrounding soil are fully developed up to the bearing capacity. The soil layers in engineering design for prestressed anchor retaining structure basically belong to non-saturated soil. The influencing factors are not in the same degree and have certain constraints within each other because bolting prestressed structure slopes have many complex factors. Based on the fuzzy multiple criteria decision theory, there may be two different forms of incident uncertainties: randomness and fuzziness. Randomness is a causal external uncertainty, only involving the amount of information. Fuzziness is the uncertainty of the inherent structure related to the meaning of information, which has more profound uncertainty than randomness. In actual stability, the objective function and constraint conditions are hard to perform in classical decision model and take on the fuzziness in most cases. As there exist contradictory and balances among multi-attribute indexes, there is no optimal solution under the usual significance when the study of the given proposal evaluation is made. By using the FMADM (fuzzy multi attribute decision making) theory, the effective solution, the satisfactory solution, optimal solution and the ideal solution can be obtained, which have specific guidance.

The fuzzy multi-attribute decision-making method is used in this paper to sort all factors which affect prestressed anchor retaining structure and the careful analysis of the existing projects will be made. Genetic algorithm is a random search calculation, simulating natural selection and natural genetic mechanisms from living nature. It searches the optimal solution through the simulation of natural evolution process, instead of relying on the gradient information. It uses some kind of coding technology, acts on digital strings named chromosome, and simulates community’s evolution process which is composed by these strings. Genetic algorithm organizes and randomly exchanges information to recombine more adaptable strings and produce the new community strings. The analysis of the dynamic stability of the prestressed structure bolting is performed by using MATLAB software package to calculate the minimum safety of the slope and find the most dangerous sliding surface location. 2

THE DETERMINATION OF THE MAIN FACTORS

There are many factors which affect the slope safety stability (Zhu et al. 2003), including water, prestressed anchor size and losses, anchor various parameters (such as anchor length, spacing and layout, angle, layer thickness) and soil property. The main factors are shown in Table 1.

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Table 1. The primary factors that affect the prestressed anchor retaining structure stability. Factors

Main function

Anchor length A1

The building foundation maximum horizon displacement and the surface subsidence reduce along with the increase of the anchor length, and the reduction is slower when anchor rod is longer, and the change slightly when the anchor rod increases much longer. Pit deformation and anchor length are not linear, and anchor length needn’t be too long. The change of some row of anchor length will affect the internal force of this row and adjacent row. From the perspective of the deformation control, it is appropriate to make each row of anchors with equal lengths or make the crown anchors longer in isotropic soil. Pit maximum horizontal displacement and the maximum axial force increase along with the increase of anchor bolt spacing, and increase more slowly when bolt spacing is enlarged. Considering the deformation of building foundation pit and the project construction cost, it is better for bolt spacing within 2∼2.5 m. In the calculation of the vertical excavation slope, the best bolt angle is within 5◦∼10◦ , within which the effect of anchor bolt bound to the surrounding soil has been fully exploited, and the pit deformation is smaller. The surface layer of prestressed bolting is flexible, increasing the thickness of surface layer can enhance surface stiffness, and decrease pit lateral displacement. It is more appropriate for the surface layer thickness to be taken within 150∼200 mm. Rainfall increases soil moisture content, and reduces cohesion, internal friction angle and the suction, thus reducing the shear strength of soil. Rainfall will increase the soil unit weight, and infiltration down along the original soil fissures will reduce the skid-resisting capacity of the potential sliding surface. The increase in water will lower friction of anchor and the surrounding soil.

Anchor pitch A2 Bolt angle A3 Layer thickness A4 Water influence A5

For the detailed analysis, slope stability properties are divided into four levels: stable, basically stable, less stable and unstable. The preliminary design of prestressed bolting for a deep pit in Xi’an is as follows: slope height is 15 m; slope angle is 60◦ , the overloading q = 15 kN/m2 , γ = 18.9 kN/m3 , ϕ = 22.6◦ , c = 62.5 kN/m2 .

2.1

Determining the fuzzy weighing and the fuzzy utility function of every scheme

Regarding attribute C1:

⎛ ⎞1/5 5  a1 = ⎝ a1j  = (1 × 2 × 2 × 1/3 × 1/3)1/5 j=1

= 0.8503 ⎛ ⎞1/5 5  a2 = ⎝ a2j  = (1/4 × 1 × 1 × 2 × 2)1/5 j=1

Fuzzy weighing and the fuzzy utility function of every scheme are calculated by using FMADM (Fuzzy Multiple Attribute Decision Making) Buckley to determine fuzzy utility value and the function (Li et al. 2002, Lei et al. 2005). The comparison and ordering between fuzzy utility values are made to determine the distinct factors on the slope stability. Initially the quantitative analysis is made of the five factors above mentioned in three of the selected slope (attribute). As it is difficult to determine the actual quantitative data of different factors in different attributes, and it is not accurate to draw the relative level of importance in attributes, experts obtain the following four fuzzy inverse matrixes by the comparisons between every two factors. The first three matrices are drawn from various factors compared in the slope, which are used to determine Fuzzy utility function of every scheme. The last one is used to determine fuzzy weighing of attributes shown in Table 2 and Table 3.

= 1.0000 ⎛ ⎞1/5 5  a3 = ⎝ a3j = (1/4 × 1/4 × 1 × 1/4 × 1/4)1/5 j=1

= 0.3299 ⎛ ⎞1/5 5  a4 = ⎝ a4j  = (1 × 1/4 × 1 × 1 × 1/4)1/5 j=1

= 0.5743 ⎛ ⎞1/5 5  a5 = ⎝ a5j  = (1 × 1/4 × 1 × 1 × 1)1/5

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j=1

= 0.7579

Table 2. Fuzzy mutual matrix of factors. C1

A1

A2

A3

A4

A5

A1 A2 A3 A4 A5

(1, 1, 1, 1) (1/4, 1/3, 1/2, 1/2) (1/4, 1/3, 1/3, 1/2) (1, 1, 2, 3) (1, 2, 2, 3)

(2, 2, 3, 4) (1, 1, 1, 1) (1/4, 1/2, 1/2, 1) (1/4, 1/3, 1/2, 1) (1/4, 1/3, 1/3, 1)

(2, 3, 3, 4) (1, 2, 2, 4) (1, 1, 1, 1) (1, 2, 3, 4) (1, 3, 3, 4)

(1/3, 1/2, 1, 1) (2, 2, 3, 4) (1/4, 1/3, 1/2, 1) (1, 1, 1, 1) (1, 1, 2, 4)

(1/3, 1/2, 1/2, 1) (2, 3, 3, 4) (1/4, 1/3, 1/3, 1) (1/4, 1/2, 1, 1) (1, 1, 1, 1)

C2

A1

A2

A3

A4

A5

A1 A2 A3 A4 A5

(1, 1, 1, 1) (1/4, 1/3, 1/3, 1) (1/4, 1/4, 1/2, 1) (1/3, 1/2, 1, 1) (1/4, 1/2, 1/2, 1)

(2, 3, 3, 4) (1, 1, 1, 1) (1/4, 1/3, 1/3) (1/4, 1/2, 1/2) (1, 2, 2, 3)

(2, 2, 4, 4) (1, 3, 3, 4) (1, 1, 1, 1) (1, 1, 2, 3) (1, 2, 3, 4)

(1/3, 1/2, 1, 1) (1, 2, 2, 4) (1, 1, 2, 3) (1, 1, 1, 1) (1/4, 1/3, 1/3, 1)

(1, 2, 2, 4) (1/3, 1/2, 1/2) (1/4, 1/3, 1/2) (2, 2, 3, 4) (1, 1, 1, 1)

C3

A1

A2

A3

A4

A5

A1 A2 A3 A4 A5

(1, 1, 1, 1) (1/4, 1/2, 1/2, 1) (1, 1, 2, 2) (1/4, 1/3, 1/2, 1/2) (1/5, 1/3, 1/2, 1/2)

(1, 2, 2, 4) (1, 1, 1, 1) (1/3, 1/2, 1/2, 1) (1/3, 1/2, 1, 1) (1/3, 1, 1, 1)

(1/2, 1/2, 1, 1) (1, 2, 2, 3) (1, 1, 1, 1) (1/3, 1/3, 1/2, 1) (1/4, 1/2, 1/2, 1)

(2, 2, 3, 4) (1, 1, 2, 3) (1, 2, 2, 3) (1, 1, 1, 1) (1/3, 1/2, 1, 1)

(2, 2, 3, 5) (1, 1, 1, 3) (1, 1, 2, 4) (1, 1, 2, 3) (1, 1, 1, 1)

Table 3. Fuzzy mutual matrix of fuzzy weighing.

x˜ 31 =

W

C1

C2

C3

C1 C2 C3

(1, 1, 1, 1) (2, 3, 4, 5) (1, 2, 2, 3)

(1/5, 1/4, 1/3, 1/2) (1, 1, 1, 1) (2, 2, 3, 4)

(1/3, 1/2, 1/2, 1) (1/4, 1/3, 1/2, 1/2) (1, 1, 1, 1)

Therefore: a =

5

i=1

ai = 3.5124

x˜ 41

x˜ 51

Similarly, we may figure out: b1 = 1.0845, b2 = 1.3195, b3 = 0.4503, b4 = 0.8027, b5 = 1.0487, b = 4.08057; c1 = 1.3510, c2 = 1.5581, c3 = 0.4884, c4 = 1.2457, c5 = 0.3195, c = 5.9564 d1 = 1.0000, d2 = 2.0000, d3 = 0.8706, d4 = 1.4310, d5 = 1.8882, d = 7.1897 ˜ 1 we Therefore, from the fuzzy mutual matrix M may obtain fuzzily utility function x˜ i1 (i = 1, 2, 3)   a1 b1 c1 d1 , , , x˜ 11 = d c b a

x˜ 21

= (0.1186, 0.1821, 0.2811, 0.2847)   a2 b2 c2 d2 = , , , d c b a



a3 b3 c3 d3 , , , d c b a



= (0.0459, 0.0756, 0.3230, 0.5694)   a4 b4 c4 d4 , , , = d c b a = (0.0799, 0.1348, 0.2592, 0.4074)   a5 b5 c5 d5 , , , = d c b a = (0.1069, 0.1929, 0.2746, 0.5376)

˜ 3 and W ˜ ˜2 M From the fuzzy mutual matrix M we may obtain fuzzily utility function x˜ i2 x˜ i3 and fuzzy weighing w. ˜ Fuzzily utility function is shown in table 4. Fuzzy weighing : w˜ j = (w˜ 1 , w˜ 2 , w˜ 3 ) = ([0.0913, 0.1378, 0.1782, 0.3228], [0.1787, 0.2756, 0.4081, 0.5519], [0.2837, 0.4376, 0.5886, 0.9310]) 2.2

Determining fuzzy utility value

Fuzzy utility value of factor A1 is drawn from fuzzily utility function and fuzzy weighing: U˜ 1 = (w˜ 1 × x˜ 11 ) + (w˜ 2 × x˜ 12 ) + (w˜ 3 × x˜ 13 )

= 0.1144[0.02505, 0.08656], 0.1861, 0.4374, 1.0907 [0.13993, −0.079328]

= (0.1391, 0.2215, 0.3230, 0.5694)

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Table 4. Fuzzy utility function-˜xij of every scheme. Ai

x˜ i1 (c1 )

x˜ i2 (c2 )

x˜ i3 (c3 )

A1 A2 A3 A4 A5

(0.1186, 0.1821, 0.2811, 0.2847) (0.1391, 0.2215, 0.3230, 0.5694) (0.0459, 0.0756, 0.1016, 0.2479) (0.0799, 0.1348, 0.2592, 0.4074) (0.1069, 0.1929, 0.2746, 0.5376)

(0.1329, 0.2285, 0.2946, 0.6688) (0.0763, 0.1597, 0.2090, 0.4412) (0.0546, 0.0780, 0.1460, 0.3157) (0.0951, 0.1507, 0.2603, 0.4785) (0.0720, 0.1472, 0.2990, 0.4166)

(0.1405, 0.2239, 0.3844, 0.6764) (0.0927, 0.1697, 0.2477, 0.5443) (0.0981, 0.1949, 0.3086, 0.5316) (0.0597, 0.0952, 0.1877, 0.3053) (0.0433, 0.1032, 0.1423, 0.2451)

surface location, and analyze the slope stability in the various possible combinations of the significant factors.

and the function: ⎧ 0 x ≤ 0.1144 ⎪ ⎪ ⎪ ⎪ 0.1144 ≤ x ≤ 0.1861 ⎨α μUi = 1 0.1861 ≤ x ≤ 0.4374 ⎪ ⎪ ⎪ β 0.4374 ≤ x ≤ 1.0907 ⎪ ⎩ 0 x ≥ 1.0907  0.02505α 2 + 0.08656α + 0.1144 x= 0.13993α 2 − 0.79328α + 1.0907

3.1 x ∈ [0.1144, 0.1861] x ∈ [0.4374, 1.0907]

Similarly, Fuzzy utility value of factor A1 , A2 , A3 , A4 , A5 : U˜ 2 = (0.0807 [0.02373, 0.07240] , 0.1488, 0.2887, 0.9340 [0.17058, −0.81591]) U˜ 3 = (0.0642 [0.01855, 0.05548] , 0.1172, 0.2593,

Identifying the most dangerous slide surface

Assuming that the most dangerous slide surface is arc, and the force among blocks has little impact on the overall slope stability, so the force among blocks is ignored, Draw a slope section on a certain scale, choose optional radius of the potential sliding surface and divide slide surface soil into n vertical soil blocks, then analyze the unit width block i, only consider bolt rally, ignore its shear and flexural. Decompose each block weight γi hi bi and ground loads qi bi to normal force Ni and circumferential force Ti . From equilibrium of moments, the expression of safety factor K is as follows: Mn K= = MT

0.7491 [0.12191, −0.61176]) U˜ 4 = (0.0633, [0.0134, 0.04714] , 0.1018, 0.2625, 0.6798 [0.09328, −0.51056]) U˜ 5 = (0.0534 [0.02051, 0.05693] , 0.1124, 0.2547, 0.6316 [0.09014, −0.46707]) Obviously factor A1 has the strongest impact on slope stability, factors A2 and A3 stronger, and factor A4 and A5 factor relatively weak. It shows that the bolt length has greater impact on the slope stability than the bolt spacing and bolt angle. Bolt length, bolt spacing and bolt angle are the important factors in the slope stability.

n i=1 (Ni tgϕi + ci li ) n i=1 Wi sin θi

(1)

In the formula, Mn is the sliding moment of force and MT is the skid-resisting moment of force. The coordinate system of danger sliding in the most dangerous slope is expressed in chart 1. Some calculated results are shown in Table 5. In the most dangerous slide in which the safety factor is the minimum, the center coordinate is O3 (−14.2 m, −13.1 m) Slide arc radiu is R = 45.1 m; the minimum safety factor is 0.987, and the Slope is instable. It’s necessary to establish bolting structural measures to improve the safety of the structure in the most dangerous sliding surface. 3.2 The model for stability

3

THE ANALYSIS OF THE OVERALL STABILITY IN RETAINING STRUCTURE

The safety level of retaining structure serves as the objective function, and the MATLAB software (Mi et al. 2003) is used to calculate the minimum safety of the slope, search out the most dangerous sliding

Based on the Rankin earth pressure theory (Zeng et al. 2000), the soil can be divided into several layers. Calculate active pressure in each layer and segment the drawing maps into half. Transform the area of Irregular polygon or trapezoid into rectangular area equally, and then allocate the adjacent soil pressure to the corresponding bolt. Assuming the bolt length

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Table 5. Sliding surface locations of different safety levels.

Center coordinates Slide arc radius Safety factor

O1

O2

O3

O4

(−19.7, −25.9) 57.5 1.078

(−7, −25.3) 55.2 1.102

(−14.2, 13.1) 45.1 0.987

(−24.9, 18.2) 53.7 1.005

Sliding moment of force:

0 q

MT = T1 r + T2 r + · · · · · · + Tn r = r

i

n

Wi sin θi

i=1

(3)

Wi = γi hi bi + qbi

R

h

d

x

Under the conditions of slip equilibrium: 

Ri Ti = ci li + ci li + Ni + sin (αi + θi ) tgφi cos αi +

y q b

T

R

i

i

w

N

i

i

i

Figure 1. The model of the prestressed anchor retaining structure.

is L; the layout inclination is α; standard spacing is Sh; ; vertical spacing is SV . The active pressure borne by each bolt is: Ri = Ni · Sh · SV In the radial equilibrium conditions: Ni = Wi cos θi

Mn = K′ = MT

(2)

n

i=1



 ci li + Ni +

(4)

Skid-resisting moment of force: 

R1 Mn = r N1 + sin (α1 + θ1 ) tgϕ1 cos α1 

R2 sin (α2 + θ2 ) tgϕ2 + · · · · · · + r N2 + cos α2 

Rn sin (αn + θn ) tgϕn + r Nn + cos αn

i

i

t

Ri cos (αi + θi ) cos αi

Ri cos αi

 sin (αi + θi ) tgϕi + n i=1 Wi sin θi

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+ r (c1 l1 + c2 l2 + · · · · · · + cn ln )

R2 R1 cos (α1 +θ1 )+ cos (α2 +θ2 ) +r cos α1 cos α2  Rn + ··· + cos (αn + θn ) cos αn 

n 

Ri =r ci li + Ni + sin (αi + θi ) tgϕi cos αi i=1  Ri (5) cos (αi + θi ) + cos αi Thus the overall safety factor for stability:

Ri cos αi

cos (αi + θi )



(6)

Table 6. Calculated results under the conditions of different combinations.

Plan 1 Plan 2 Plan 3

Anchorage length L/m

Anchor pitch Sh × Sv /m × m

Angle α/◦

Safety factor

14 12 10

2 × 2.5 2×2 2 × 1.5

0 5 10

1.590 1.583 1.597

In the formula, r = sliding arc radius; θi = the angle between normal force Ni and vertical line Ni ; li = soil sliding arc length of section i; D = bolt hole diameter; α = angle between bolt and the horizon; τ = cohesive force between cement (sand) pulp and soil, which is based on the experimental results at the site. The bolt anchoring force depends primarily on cohesive force between cement (sand) pulp and soil, which is used to determine the parameters as security reserves when it is active soil pressure 1.5 times. That is 1.5 Ri = 1.5 Ni · Sh · Sv = L2 · π Dτ cos α, and substitute it into (6): Mn = K ′′ = MT

n

i=1



ci li +

2L2 π Dτ 3Sh ·Sv

[cos αi + sin (αi + θi )] tgφi + n i=1 Wi sin θi

L2 = anchorage length; L = L1 +L2 is the bolt length; L1 = anchor free length, which is transformed from the sliding scale based on the measurements of the most dangerous slide section. The optimal program is selected by using different anchoring lengths, anchor pitches and angles in different programs. The calculated results under the conditions of some combination are shown in Table 6. The slop must meet the requirements of the medium use and the safety factor should be in 1.5∼1.8. These three programs all meet the requirements. Program 2 is more economical and its construction is not complicated. It is more economical and reasonable for the optimal program than the other two programs.

4

2. According to the analysis of the prestressed anchor retaining structure slop based on the FMADM Buckley, we can know that factor A1 obviously has the strongest impact on slope stability, factors A2 and A3 become less in turn. 3. To establish the safety function with significant factors A1 , A2 , A3 as variables, to analyze the dynamic stability by using MATLAB, and search out the minimum value of the objective function as the most dangerous slide surface and establish optimization model for the retaining structure. 4. According to the security of prestressed anchor retaining structure in various possible conditions,

CONCLUSION

1. The analysis of the factors which affect the prestressed anchor retaining structure is made by using the theory of fuzzy decision. The significant impact factors are sorted through calculating fuzzy utility value and the function of each program, and then their effect on safety may be taken into consideration. In the decision-making process, the balance and constraints among various factors need considering and special survey should be made extensively so as to improve the reliability of the results.

2L2 πDτ 3Sh ·Sv

cos (αi + θi )



(7)

the parameters of retaining structure may be revised in order to raise safety, economy of retaining structure and provide a comprehensive, reasonable and scientific decision-making basis for engineering design, construction and safety management. REFERENCES Lei yingjie, Zhang shanwen, Li xuwu & Zhou chuangming. 2005. MATLAB Toolbox and Application of Genetic Algorithms. Xi’an: Xi’an University of Electronic Science and Technology Press. Li Rongjun. 2002. Fuzzy Multi-criteria Decision-making and Application. Beijing: Science Press. Mi Hongliang & Chen Zuyu. 2003. Genetic Algorithms in Determining Minimum Safety Factor of Slope Stability. Geotechnical Journal 25 (3); 671–674. Zeng Xianming, Huang Jiusong, & Wang Zuoming. 2000. Soil Nailing Design and Construction Manual. Beijing: China’s Construction Industry Press. Zhu Yalin & Kong Xianjing. 2003. Structural Parametric Study of Prestressed Anchor Flexible Retaining. Journal of Anhui Institute of Architecture & Industry, 11 (4); 30–36.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Safety analysis of high engineering slopes along the west approach road of ZheGu mountain tunnel T.B. Li, Y. Du & X.B. Wang State Key Lab of Geohazard Prevention and geoenvironment Protection, Chengdu University of Technology, Chengdu, Sichuan, China

ABSTRACT: The long-term safety of highway slopes in mountain areas with complex geological conditions has a direct relation not only to the safe operation of the highway itself, but also to the life and wealth security of the people who live along it. In this paper some typical high slopes along the west approach road of the ZheGu mountain tunnel, the long-term safety of the slopes were systematically studied through the in-situ tracking investigation, the deformation monitoring of slopes and stress of anchors, the FEM numerical simulation of interaction between slope body and anchoring structure, and the fuzzy comprehensive assessment method. As a result,, these slopes were classified in terms of safety. 1

INTRODUCTION

Along with the west development and the persistent basic construction in West China, there are more and more high engineering slope need to be treated and reinforced. Due to the complexity of geological condition and the limitation of slope reinforcement design theory, many project accidents of slopes occurred, which led to great loss. Lots of scholars have paid great attention to the reinforcement structure’s work state and the long-term safety of high engineering slope. However, there is not a set of perfect system and a material standard on engineering slope safety evaluation presently. In this paper, a typical high engineering slope in West Approach Road of Zhegu Mountain Tunnel on Sichuan-Tibet Highway is selected as an example to discuss the way and the method of high engineering slope long-term safety analysis and evaluation by numerical simulation, in-situ monitoring, and Fuzzy Comprehensive Assessment, in hope of promoting the construction of the safety evaluation system on high engineering slopes.

2

typical. Its natural slope angle reaches to 70◦ –80◦ . The slope is made of lamina slate, phyllite and carbonaceous phyllite of Triassic system top series, and colluvium and slope wash whose thickness is less than 1 m. The high slope is reverse slope or oblique slope with medium to low slope angle (Figure 1). The slope rock mass is very bad and medium weathered totally. A little of its top is strong weathered.

SURVEY OF TYPICAL HIGH ENGINEERING SLOPE

Zhegu Mountain Tunnel is an important part of the rebuilding project of Sichuan-Tibet Highway. The geological condition of the West Approach Road is very complex, where the high engineering slopes’ stability problem is very outstanding. Thereinto, the high and abrupt slope lying in GK7+709∼+999 is most

Figure 1. Typical GK7+930.

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geological

condition

in

profile

Unloading effect of the slope is not strong enough. The rock mass is cataclastic structure and part inlaid structure.Technical staff and experts debated intensely on choosing whether roadbed or bridge to pass this area during construction. Finally, the former was adopted after the long-playing argumentation and technique economy comparison. By optimization design, the slope gradient ratio was designed mostly between 1:0.3 and 1:0.5, the height of the highest slope was near 80 meters, and bolt and shotcrete was the main reinforce measures. The slope optimization design economized RMB 12,000,000 Yuan than the former reinforcement design. Because the rock mass of the slope is weakness and fragmented, and the slope is very high and steep, and the reinforce measure carried on very great optimization, some technical personnels and experts queried the long-term safety of the slope. So information-monitoring and safety evaluation were carried out during and after the construction. In this paper, the highest one of the slope lying in GK7+900∼+950 is taken as an example to discuss the way and method of slope safety analysis. After the first excavation was finished, the crushing tensile zone of deformation and relaxation came up into being and its depth was 5 to 6 meters in the slope of GK7+900∼+950. Several tensile fissures occured in the top of cutting plain at 5 to 9 meters from the groove. In optimization design, gradient ratio of 1:0.3 to 1:0.75 was adopted to clear away the unstable rock mass, three slope steps and two platform steps were set. The way that combined soil nail and anchoring and shotcreting together was adopted to support the slope.

8 meters long archor in the top and the base of slope, and 5 meters in the middle part. The concrete class was C20, and the concrete thickness was 15 cm. The pressure grouting anchors were adopted in the bottom to reinforce the carbonaceous phyllite (Figure 2). 3

NUMERICAL SIMULATION OF THE REINFORCEMENT EFFECT

In order to research the reinforcement effect of the support measures, such as bolt, shotcrete layer, etc., Finite Element Method was adopted to simulate the interaction between the slope rockmasses and reinforcement structures at GK7+900 and GK7+930. The results are showed as follows: 1. Due to the anchoring effect, the deformation of the slope was controlled effectively and its stability was improved. These bolts are especially useful in controlling the deformation of carbon phyllite in the bottom of the slope. 2. Before anchoring, there was broken area about 4–7 meters in depth at the slope toe, and the broken area had extended to breakthrough in the range

broken area

Figure 3. profile.

Broken area before anchoring in GK7+930

broken area

Figure 2. Layout of the reinforcement measures and monitoring instruments in profile GK7+930.

Figure 4.

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Broken area after anchoring in GK7+930 profile.

forced, the displacement to the direction of free face of Gk7+900∼+950 high slope was small, and the value was between 5 mm to 9 mm. The top of the slope has the greatest horizontal displacement, i.e., 12 mm and the displacement tended to be stable in the late. The displacement of the outer edge of the road was relatively stable and within 5 mm. So the deformation was small and the slope was totally safe.

of 22 meters high. Obviously, the shallow and surface layer of the slope might slide down (Figure 3). The calculation results after anchoring showed that the broken area at the slope toe diminished evidently, and the integral broken depth was less than 1 meter, which exert few influence on the slope stability (Figure 4). Under the control of bolt, the plastic zone became smaller in the inner part of the slope. 3. The fragmentized and soft rockmasses at the slope toe was reinforced by means of pressure grouting, which made its stability coefficient improve from 1.11 to 1.15.

4

4.2 Results of deep displacement monitoring The inclinometer holes were located at the platforms of the top, middle and lower part of the slope (Figure 2). The evolution of the displacement of the surface and deep part of the slope showed as below (Figure 5, Figure 6, Figure 7):

IN-SITU MONITORING AND ITS RESULTS

1. In rainy season the deformation rate increased and the total displacement raised fast. When entered November of each year, this rate slowed down quickly and reached its minimum value in January. 2. The deep displacement curves of top slope showed no evident signs of sudden change or inflexion. Its accumulating displacement was within the range of 55 mm (Figure 5), which reflected the strengthening effect and constraint to deformation of reinforcement measures.

Seven ground displacement monitoring points, three deep displacement monitoring boreholes and eight bolt force monitoring points are located in section GK7+900∼+950. 4.1

Result of ground displacement monitoring

The ground displacement monitoring, which lasted 18 stages in 558 days, showed that after being rein-

Displacement/mm

60 50

12-13-04

03-16-04

05-10-05

10-30-05

04-29-06

02-02-07

09-27-05

40 30 20 10 0

1

2.5

4

5.5

7

8.5

10 11.5

13 14.5

16 17.5

19 20.5

Depth/m

22 23.5

25 26.5

28 29.5

31 32.5

34 35.5 36.8

Figure 5. Displacement—Depth curves of inclinometer hole CX1.

120 12-5-04 3-16-05 7-2-05 10-30-05

80 60

12-13-04 5-10-05 9-27-05 7-20-06

40

Figure 6. Displacement—Depth curves of inclinometer hole CX2.

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20.2

19.5

18.5

17.5

16.5

15.5

14.5

13.5

Depth/m

12.5

11.5

10.5

9.5

8.5

7.5

6.5

5.5

4.5

3.5

0.5

0

2.5

20 1.5

Displacement/mm

100

Displacement/mm

18 16 14 12 10 8 6 4 2 0

9-27-05 3-18-06 2-2-07

0.5

1.5

2.5

3.5

4.5

5.5

6.5

7.5

8.5

9.5

Depth/m

10-30-05 4-29-06

10.5

11.5

12.5

1-14-06 6-21-06

13.5

14.5

15.5

16.5

17.5

Figure 7. Displacement—Depth curves of inclinometer hole CX3.

Table 1. Safety factor of the bolts’ force. Monitoring point R6 R7 R8 R9 R10 R11 R18 R19

Location Top Second platform First platform Slope toe

GK7+940 GK7+930 GK7+940 GK7+930 GK7+940 GK7+930 GK7+940 GK7+930

3. The displacement of the middle part of the slope was the greatest among the three parts, whose accumulating surface displacement was in the range of 100 mm, and its tendency showed the property of seasonality, which was the same as the top part of the slope. Although in the depth of 10.5∼12.5 m, evident inflexion appeared on the displacement— depth curves (Figure 6), this phenomenon didn’t appear among the top monitoring holes at the same location, which meant that there was no potential rapture plane that joined up two holes in the slope. This kind of sudden change in displacement might have relationship with the deformation of unloading fissures in the shallow and surface part of the slope. In the followed monitoring stages, the deformation of this part was small and showed no signs of persistent deformation. 4. The accumulating displacement of the lower slope, which was changing within the range of 16 mm, was very small (Figure 7). The deformation of the slope tended to be stable after March, 2006.

4.3 Result of bolt stress monitoring Bolt force monitoring points were located at the slope toe, the first platform, the second platform and

Max value (KN)

Design value (KN)

Safety factor

23.7 12.3 10.9 5.8 13.4 1.2 11.9 20.9

70 70 70 70 70 70 70 70

2.95 5.69 6.42 12.06 5.22 58.33 5.88 3.35

the peak in profile GK7+930 and GK7+940. The monitoring results showed as below: 1. All the data of the bolts’ force were distributed between 5 KN to 24 KN, which were far below the designed force (70 KN). Relatively speaking, the force of the bolts located at the top and bottom part of the slope was a bit higher than that of the middle part. For most of the bolts, their safety factor was between 3 and 6 (Table 1). So, the bolts’ safety margin was high. 2. In view of the bolts’ force of the whole slope was developed with time, it developed fast at first while its amount was relatively low. As time went on, the force increased gradually. By now, most of the bolts’ force had tended to be steady (The curves were convergent), except point R7 and R9 that were located at the upper to middle part of GK7+930 profile. These bolt’s force were still developing gradually, but the values were below 13 KN and the safety margin was still high. In summary, the bolt force was comparatively small and tended to be stable. All of these indicated that the bolts had a good effect on supporting the slope and the structure was safe.

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5 5.1

SYNTHETIC ANALYSIS AND EVALUATION OF SLOPE SAFETY General synthetic analysis

Since the end of construction, no great instantaneous deformation has emerged in the high slope of GK7+900∼950 and previous cracks have not accelerated any more, in addition, few new cracks exist in the trailing edge of slope. The surface of bolt and shotcrete support is intact, and the slope’s surface is well-drained so that no accumulated water exists. It can be justified that the bolt and shotcrete support has great effect on the stability of this high slope via monitoring data of surface deformation, inclinometer borehole and anchor force. Therefore, the slope is stable and secure. 5.2

Fuzzy synthetic evaluation

5.2.1 Selection of evaluative factors and criterion of classification The following 6 kinds of features are considered as the judgments factors for the safety fuzzy evaluation on the slopes. 1. Crack extent in the slope mass after reinforcement. 2. The area of plastic zone, indicated by the percentage rate (divided by the whole area of the section in FEM analysis). 3. Anchoring force. 4. Deformation on the ground surface. 5. Accumulated displacement velocity at the orifice of inclinometer monitoring hole, using the peak value of accumulated displacement velocity at the orifice of inclinometer monitoring hole in latest 6 months. 6. Weather the curves of deep displacement have sudden change. According to the safety acquirement of engineering slope, the safety criterion classify as 3 classes: safe,

almost safe and potentially unsafe, and the three are signified as I, II, III respectively. Based on concrete condition of high slopes in west approach road of the Zhegu Mountain tunnel after reinforcement, combined with the characteristics of various factors and in term of above-mentioned 3 safe classes, we confirm the classes criterion of each factor as table 2. 5.2.2 Membership function and weight In the synthetic evaluation of slope safety, according to basic principle of creating membership function, evaluative factors are separated into two parts to create membership function: if evaluative factors are qualitative, membership function consist of eigenfunction as they are discrete; if evaluative factors are quantitative, membership function are established by ‘‘falling semi-trapezoid distribution’’. Each evaluative factors’ weighting was confirmed by analytical hierarchy process, which result in the weighting of U1, U2, . . . , U6 are W (Ui) = (0.28, 0.05, 0.19, 0.28, 0.10, 0.10). 5.2.3 Synthetic safety evaluation of slope Results of Synthetic safety analysis and evaluation of slope could get by combination of weight matrix W and fuzzy relation matrix R. To establish fuzzy relation matrix R(A) by using the membership grade of evaluative factors’ eigenvalue (table 3) to safe classes calculated via membership function:

R(A)

⎡ 1 0 0⎤ ⎢ 1 0 0⎥ ⎥ ⎢ ⎢ 1 0 0⎥ ⎥ ⎢ = R(Ui ) = ⎢ ⎥ ⎢0.51 0.49 0⎥ ⎥ ⎢ ⎣0.83 0.17 0⎦ 0 0 1

Put the weight matrix W (Ui) and fuzzy relation matrix R (A) into synthetic evaluative formula,

Table 2. Grades of fuzzy evaluation factors for slope safety. Safety grade Factors

I

II

III

Extent of crack (U1 ) Area percentage of plastic zone (U2 ) Anchor force (U3 ) Surface displacement (U4 ) Displacement velocity of inclinometer monitoring (U5 ) Displacement curve of inclinometer monitoring (U6 )

No fractures 20% 35 KN 30 mm

Gentle fractures 20%∼40% 35∼70 KN∗ 30∼100 mm

Multiple fractures  40% 70 KN 100 mm

0.05 mm/d

0.05∼0.2 mm/d

0.2 mm/d

No mutation

Almost no mutation

Obvious mutation

∗ 70

KN is the designed lode of bolt which diameter is 22 mm.

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Table 3. Eigenvalue of evaluative factor. Factors

Characteristic or eigenvalue

Extent of crack (U1 ) Area percentage of plastic zone (U2 ) Anchor force (U3 ) Peak value of surface displacement (U4 ) Displacement velocity of inclinometer monitoring (U5 ) Displacement curve of inclinometer monitoring (U6 )

No fractures 20% 24 KN 12 mm 0.063 mm/d Obvious mutation partly

result in: B(Ui ) = W (Ui ) × R(Ui ) = (0.75, 0.15, 0.10) According to the maximum membership grade pinciple we can get that the safety class of the high slope located at GK7+900∼950 section in westapproach road of the Zhegu Mountain Tunnel is persistent safe. In the light of each membership grade value, the grade of potential unsafe class is 0.1 which indicate that parts of the slope may unsecured. Among all evaluative factors, the part unsafe phenomenon indicated as the visible mutation on the deep displacement curve of the inclinometor hole in the middle of the slope. Consequently, the unsecured parts in the middle of the slope should be monitored unceasingly. 6

CONCLUSION

1. This paper selects the high engineering slope in West Approach Road of Zhegu Mountain Tunnel located on Sichuan-Tibet Highway as an example to discuss the way and method for slope safety analysis and evaluation by numerical simulation, on site monitoring and fuzzy comprehensive assessment, which offers a good foundation for building the safety evaluation system of slope.

2. After the high slope had been reinforced in GK7+900∼950 section of West Approach Road of Zhegu Mountain Tunnel, the deformation of slope became slight and no new crack occurred in the slope’s top. The surface of bolt and shotcrete support is intact. It reveals that the bolt and shotcrete support has fairly good effect on the deformation and stability of this high slope through the monitoring datum analysis of ground and deep displacement and anchor force. The slope is totally stable and secure. 3. Fuzzy synthetic evaluation can consider each factor which affects on slope’s safety and it could get the slope safety grade by quantitive analysis. It is a effective method for the synthetic safety assessment of slopes. This paper determines weighting of every factor by analytical hierarchy process. It is concluded that the safety grade of high engineering slope in GK7+900∼950 section is persistently safe. 4. Deep displacement monitoring of GK7+930 section shows that the deformation curves have distinct inflexion at the depth of 10.5∼12.5 m in the middle of the slope (CX2 monitoring hole). At present, although the deformation do not aggravate, it should be monitored further to confirm the safety of this part of the slope. REFERENCES Li, T.B. 2003. Systematical research on the stability of high rock engineering slope and its control. Chinese Journal of Rock Mechanics and Engineering, Vol.22, No.2. Li, T.B. et al. 2007. Informational monitoring report on engineering high slope at GK7+709∼+999 section of D contract section in west approach road of ZheGu Mountain tunnel. Song, X.G., Zhang, S.F. & Li, Y.Y. 2005. In-situ dynamic monitoring and analysis of high slope stability. Rock and Soil Mechanics, Vol.26, No.7, 1153–1156. Xia, Y.Y. & Li, M. 2002. Evaluation method research of slope stability and its developing trend. Chinese Journal of Rock Mechanics and Engineering, Vol.21, No.7, 1087–1091. Zhang, Y., Zou, S.P. & Su, F. 1992. Fuzzy mathematic method and application. China Coal Industry Publishing House.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Landslide stabilizing piles: A design based on the results of slope failure back analysis M.E. Popescu Illinois Institute of Technology, Chicago, Illinois, USA

V.R. Schaefer Iowa State University, Ames, Iowa, USA

ABSTRACT: It is generally accepted that shear strength parameters obtained by back analysis of slope failures ensure more reliability than those obtained by laboratory or in-situ testing when used to design remedial measures. In many cases, back analysis is an effective tool, and sometimes the only tool, for investigating the strength features of a soil deposit. Procedures to determine the magnitude of both shear strength parameters (c′ and φ′ ) or the relationship between them by considering the position of the actual slip surface within the failed slope are discussed. Using the concept of limit equilibrium the effect of any remedial measure (drainage, modification of slope geometry, restraining structures) can easily be evaluated by considering the intercepts of the c′ − tan φ′ lines for the failed slope (c′0 , tan φ′0 ) and for the same slope after installing some remedial works (c′nec , tan φ′nec ), respectively. The above outlined procedure is illustrated to design piles to stabilize landslides taking into account both driving and resisting force acting on each pile in a row as a function of the non-dimensional pile interval ratio B/D. The accurate estimation of the lateral force on pile is an important parameter for the stability analysis because its effects on both the pile-and slope stability are conflicting. That is, safe assumptions for the stability of slope are unsafe assumptions for the pile stability, and vice-versa.

1

INTRODUCTION

Correction of an existing landslide or the prevention of a pending landslide is a function of a reduction in the driving forces or an increase in the available resisting forces. Any remedial measure used must involve one or both of the above parameters. The IUGS Working Group on Landslides (Popescu, 2001) has prepared a short checklist of landslide remedial measures arranged in four practical groups, namely: modification of slope geometry, drainage, retaining structures and internal slope reinforcement, as shown in Table 1. As many of the geological features, such as sheared discontinuities, are not well known in advance, it is better to put remedial measures in hand on a ‘‘design as you go basis’’. That is the design has to be flexible enough for changes during or subsequent construction of remedial works. Although slope instability processes are generally seen to be ‘‘engineering problems’’ requiring ‘‘engineering solutions’’ involving correction by the use of structural techniques, non-structural solutions including classical methods such as drainage and

modification of slope geometry, as well as some novel methods such as lime/cement stabilization, grouting or soil nailing, are increasingly being used (Popescu, 1996). The cost of non-structural remedial measures is considerably lower when compared with the cost of structural solutions. Terzaghi (1950) stated that, ‘‘if a slope has started to move, the means for stopping movement must be adapted to the processes which started the slide’’. For example, if erosion is a causal process of the slide, an efficient remediation technique would involve armoring the slope against erosion, or removing the source of erosion. An erosive spring can be made non-erosive by either blanketing with filter materials or drying up the spring with horizontal drains, etc. Morgenstern (1992) followed this theme when he noted that post-failure analyses can be used to provide a consistent explanation for landslide causal events. The back-analyses can then be used as a basis for design of the stabilizing measures if engineering works are required. This approach has the added appeal that the remedial design is normalized in terms of the post-failure analytical model.

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Table 1. A brief list of landslide remedial measures. 1. MODIFICATION OF SLOPE GEOMETRY 1.1.

1.2. 1.3.

Removing material from the area driving the landslide (with possible substitution by lightweight fill) Adding material to the area maintaining stability (counterweight berm or fill) Reducing general slope angle

2. DRAINAGE 2.1. 2.2.

2.3. 2.4. 2.5. 2.6. 2.7. 2.8. 2.9. 2.10. 2.11.

Surface drains to divert water from flowing onto the slide area (collecting ditches and pipes) Shallow or deep trench drains filled with freedraining geomaterials (coarse granular fills and geosynthetics) Buttress counterforts of coarse-grained materials (hydrological effect) Vertical (small diameter) boreholes with pumping or self draining Vertical (large diameter) wells with gravity draining Subhorizontal or subvertical boreholes Drainage tunnels, galleries or adits Vacuum dewatering Drainage by siphoning Electroosmotic dewatering Vegetation planting (hydrological effect)

3. RETAINING STRUCTURES 3.1. 3.2. 3.3. 3.4. 3.5. 3.6. 3.7. 3.8. 3.9. 3.10.

Gravity retaining walls Crib-block walls Gabion walls Passive piles, piers and caissons Cast-in situ reinforced concrete walls Reinforced earth retaining structures with strip/sheet polymer/metallic reinforcement elements Buttress counterforts of coarse-grained material (mechanical effect) Retention nets for rock slope faces Rockfall attenuation or stopping systems (rocktrap ditches, benches, fences and walls) Protective rock/concrete blocks against erosion

4. INTERNAL SLOPE REINFORCEMENT 4.1. 4.2. 4.3. 4.4. 4.5. 4.6. 4.7. 4.8. 4.9. 4.10.

Rock bolts Micropiles Soil nailing Anchors (prestressed or not) Grouting Stone or lime/cement columns Heat treatment Freezing Electroosmotic anchors Vegetation planting (root strength mechanical effect)

Most landslides must usually be dealt with sooner or later. How they are handled depends on the processes that prepared and precipitated the movement, the landslide type, the kinds of materials involved, the size and location of the landslide, the place or components affected by or the situation created as a result of the landslide, available resources, etc. The technical solution must be in harmony with the natural system, otherwise the remedial work will be either short lived or excessively expensive. In fact, landslides are so varied in type and size, and in most instances, so dependent upon special local circumstances, that for a given landslide problem there is more than one method of prevention or correction that can be successfully applied. The success of each measure depends, to a large extent, on the degree to which the specific soil and groundwater conditions are prudently recognized in an investigation and incorporated in design. In this paper a methodology involving back analysis of the slope and the use of piles to remediate the landslide are presented. 2 2.1

BACK ANALYSIS OF FAILED SLOPES TO DESIGN REMEDIAL MEASURES Failure envelope parameters

A slope failure can reasonably be considered as a full scale shear test capable to give a measure of the strength mobilized at failure along the slip surface. The back calculated shear strength parameters, which are intended to be closely matched with the observed real-life performance of the slope, can then be used in further limit equilibrium analyses to design remedial works. The limit equilibrium methods forming the framework of slope stability/instability analysis generally accept the Mohr-Coulomb failure criterion: τf = c′ + σ′ tan φ′

(1)

τf = A(σ′ )b

(2)

′

where τf and σ are the shear stress and effective normal stress respectively on the failure surface and c′ and φ′ are parameters assumed approximately constant for a particular soil. A significant limitation in the use of this criterion is that the constant of proportionality is not really a constant when wide range of stress is under consideration. There is now considerable experimental evidence to show that the Mohr failure envelope exhibits significant curvature for many different types of soil and compacted rockfill. Therefore, if the assumption of a linear failure envelope is adopted, it is important to know what range of stress is appropriate to a particular slope instability problem. To avoid this difficulty a curved failure envelope can be approximated by the following power law equation:

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which was initially suggested by De Mello (1977) for compacted rockfills and subsequently found appropriate for soils (Atkinson and Farrar, 1985). 2.2

Procedures for back analysis of slope failures

Shear strength parameters obtained by back analysis ensure more reliability than those obtained by laboratory or in-situ testing when used to design remedial measures. In many cases, back analysis is an effective tool, and sometimes the only tool, for investigating the strength features of a soil deposit (Duncan, 1999). However one has to be aware of the many pitfalls of the back analysis approach that involves a number of basic assumptions regarding soil homogeneity, slope and slip surface geometry and pore pressure conditions along the failure surface (e.g. Leroueil & Tavenas 1981). A position of total confidence in all these assumptions is rarely if ever achieved. While the topographical profile can generally be determined with enough accuracy, the slip surface is almost always known in only few points and interpolations with a considerable degree of subjectivity are necessary. Errors in the position of the slip surface result in errors in back calculated shear strength parameters. If the slip surface used in back analysis is deeper than the actual one, c′ is overestimated and φ′ is underestimated and vice-versa. The data concerning the pore pressure on the slip surface are generally few and imprecise. More exactly, the pore pressure at failure is almost always unknown. If the assumed pore pressures are higher than the actual ones, the shear strength is overestimated. As a consequence, a conservative assessment of the shear strength is obtainable only by underestimating the pore pressures. Procedures to determine the magnitude of both shear strength parameters or the relationship between them by considering the position of the actual slip surface within a slope are discussed by Popescu and Yamagami (1994). The two unknowns – i.e. the shear strength parameters c′ and φ′ – can be simultaneously determined from the following two requirements: a. F = 1 for the given failure surface. That means the back calculated strength parameters have to satisfy the c′ − tan φ′ limit equilibrium relationship; b. F = minimum for the given failure surface and the slope under consideration. That means the factors of safety for slip surfaces slightly inside and slightly outside the actual slip surface should be greater than one (Fig.1a). Based on the above mentioned requirements, Saito (1980) developed a semi-graphical procedure using trial and error to determine unique values of c′ and tan φ′ by back analysis (Fig.1b). An envelope of the limit equilibrium lines c′ − tan φ′ , corresponding to

Figure 1.

Shear strength back analysis methods.

different trial sliding surfaces, is drawn and the unique values c′ and tan φ′ are found as the coordinates of the contact point held in common by the envelope and the limit equilibrium line corresponding to the actual failure surface. A more systematic procedure to find the very narrow range of back calculated shear

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strength parameters based on the same requirements is illustrated in Fig.1c. The procedures discussed above to back calculate the linear strength envelope parameters, c′ and φ′ in equation (1) can be equally applied to back calculate the nonlinear strength envelope parameters, A and b in equation (2) (Popescu et al., 1995). The fundamental problem involved is always one of data quality and consequently the back analysis approach must be applied with care and the results interpreted with caution. Back analysis is of use only if the soil conditions at failure are unaffected by the failure. For example back calculated parameters for a first-time slide in a stiff overconsolidated clay could not be used to predict subsequent stability of the sliding mass, since the shear strength parameters will have been reduced to their residual values by the failure. In such cases an assumption of c′ = 0 and the use of a residual friction angle, φr is warranted (Bromhead 1992). If the three-dimensional geometrical effects are important for the failed slope under consideration and a two-dimensional back analysis is performed, the back calculated shear strength will be too high and thus unsafe.

2.3 Design of remedial measures based on back analysis results In order to avoid the questionable problem of the representativeness of the back calculated unique set of shear strength parameters a method for designing remedial works based on the limit equilibrium relationship c′ − φ′ rather than a unique set of shear strength parameters can be used (Popescu, 1991). The method principle is shown in Fig. 2. It is considered that a slope failure provides a single piece of information which results in a linear limit equilibrium relationship between shear strength parameters. That piece of information is that the factor of safety is equal to unity (F = 1) or the horizontal force at the slope toe is equal to zero (E = 0) for the conditions prevailing at failure. Each of the two conditions (F = 1 or E = 0) results in the same relationship c′ − tan φ′ which for any practical purpose might be considered linear. The linear relationship c′ − tan φ′ can be obtained using standard computer software for slope stability limit equilibrium analysis by manipulations of trial values of c′ and tan φ′ and corresponding factor of safety value. It is simple to show that in an analysis using arbitrary φ′ alone (c′ = 0) to yield a non-unity factor of safety, Fφ ∗ , the intercept of the c′ −tan φ′ line (corresponding to F = 1) on the tan φ′ axis results as: tan φ0 ′ = tan φ′ /Fφ ∗

(3)

Similarly the intercept of the c′ − tan φ′ line (corresponding to F = 1) on the c′ axis can be found

Figure 2. Limit equilibrium relationship and design of slope remedial measures.

assuming φ′ = 0 and an arbitrary c′ value which yield to a non-unity factor of safety, Fc ∗ : c0 ′ = c′ /Fc ∗

(4)

Using the concept of limit equilibrium linear relationship c′ − tan φ′ , the effect of any remedial measure

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shown in Figure 3. FD increases with the pile interval while FR decreases with the same interval. The intersection point of the two curves which represent the two forces gives the pile interval ratio satisfying the equality between driving and resisting force. The accurate estimation of the lateral force on pile is an important parameter for the stability analysis because its effects on both the pile-and slope stability are conflicting. That is, safe assumptions for the stability of slope are unsafe assumptions for the pile stability, and vice-versa. Consequently in order to obtain an economic and safe design it is necessary to avoid excessive safety factors. The problem is clearly three-dimensional and some simplification must be accepted in order to develop a two-dimensional analysis method based on the principles outlined above. However the only simplicity to be accepted and trusted is the simplicity that lies beyond the problem complexity and makes all details and difficulties simple by a sound and profound understanding.

3

APPLICATION

3.1

(drainage, modification of slope geometry, restraining structures) can easily be evaluated by considering the intercepts of the c′ − tan φ′ lines for the failed slope (c0 ′ , tan φ0 ′ ) and for the same slope after installing some remedial works (c′ nec , tan φ′ nec ), respectively (Figure 2). The safety factor of the stabilized slope is:   c0 ′ tan φ0 ′ F = min Fc = ′ , Fφ = (5) c nec tan φ ′ nec Errors included in back calculation of a given slope failure will be offset by applying the same results, in the form of c′ − tan φ′ relationship, to the design of remedial measures. The above outlined procedure was used to design piles to stabilize landslides (Popescu, 1991) taking into account both driving and resisting force. The principle of the proposed approach is illustrated in Figure 3 which gives the driving and resisting force acting on each pile in a row as a function of the nondimensional pile interval ratio B/D. The driving force, FD , is the total horizontal force exerted by the sliding mass corresponding to a prescribed increase in the safety factor along the given failure surface. The resisting force, FR , is the lateral force corresponding to soil yield, adjacent to piles, in the hatched area

620 600

New Fill

580

Existing Fill

560

Height (ft)

Figure 3. Driving vs. resisting force for stabilizing piles.

Site conditions

The described methodology is applied to a landslide in Ohio in the United States. The site is located along the Ohio River in south-central Ohio. A replacement bridge was proposed at the site and site preparations reactivated an ancient slide. The cross section is shown in Figure 4. The slope consists of shale bedrock overlain by shale weathered to a residual clay. Overlying the residual clay is alluvial silts and clays. Construction activities at the site led to the reactivation of an ancient slide. The slip plane discerned from surface scarps and inclinometer data is shown in Figure 4. It can be seen that the failure surface is planar in nature and occurs just above the shale bedrock in the weathered residual clay.

540

Alluvial Silts & Clays

Shale

Ohio River

520

Failure Surface

500

Residual Clay

480 460 0

100

200

300

Length (ft)

Figure 4.

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Cross section of slope.

400

500

600

6000

15 Force, Driving & Resisting, kips

Friction Angle, deg

5000

12

9

No Pile

4000 Resisting Force 3000 Driving Force 2000

1000

6

B/D = 0.75

0 0.30

0.40

Figure 6. ratio.

B/D = 0.50 0 0

0.5

Cohesion, ksf

0.60

0.70

0.80

Driving and resisting forces as a function of B/D

1

Figure 5. Back analyzed relationship between friction angle and cohesion.

To accommodate the new bridge a fill was proposed on the existing slope, which was now moving and would have exacerbated the instability. Hence the use of piles to stabilize the slope was proposed. 3.2

0.50 B/D Ratio

3

Back analysis

Back analyses were conducted of the slope failure using limit equilibrium techniques as described previously. The back analyzed friction angle and cohesion for the residual clay and the failure surface are shown in Figure 5. The resulting strength parameters vary depending upon the water level in the Ohio River. The relationship between the friction angle and the cohesion are shown in Figure 5 for cases of no pile and pile B/D ratios of 0.75 and 0.5. The back analyzed friction angle of about 13◦ for the no pile case compares favorably with residual shear test results. The results for the B/D ratios in Figure 5 were obtained using the methodology proposed by Liang (2002) and coded into an Excel spreadsheet. 3.3 Driving and resisting forces The driving forces were determined using limit equilibrium analyses utilizing the program XSTABL (Interactive Software Designs, Inc. 1994) and spreadsheet analyses. The driving forces are shown in Figure 6 for various B/D ratios. The resisting forces were determined using the Ito and Matsui (199x) method as outlined by Popescu (1995). The resisting forces are shown in Figure 6 for various B/D ratios. From the results in Figure 6 it can be seen that the resisting force and driving force cross at a B/D

ratio slightly larger than 0.5 with a required resisting force of about 1800 kips. A shear force of this magnitude could be obtained using six-foot diameter shafts; however, eight-foot diameter shafts were selected to provide a margin of safety for the drilled shafts.

4

CONCLUSIONS

This paper has outlined an approach to back analyzing the strength parameters in a slope failure and determining the force required to stabilize a slope using piles considering the back analysis results. The use of the technique has been demonstrated through application to a case history.

REFERENCES Atkinson, J.H. & Farrar, D.M. 1985. Stress path tests to measure soils strength parameters for shallow landslips. Proc.11th Int. Conf. Soil Mech. Foundation Eng., San Francisco, 2:983–986. Bromhead, E.N. 1992. Slope Stability. 2nd Edition, Blackie Academic & Professional, London, 411 pp. De Mello, V.F.B. (1977). Reflections on design decisions of practical significance to embankment dams. Géotechnique, 27 (3):281–354. Duncan, J.M. 1999. The use of back analysis to reduce slope failure risk. J. Boston Soc. Civil Eng., 14:1:75–91. Interactive Software Designs, Inc. 1994. XSTABL An Integrated Slope Stability Analysis Program for Personal Computers. Reference Manual, Version 5, Moscow, ID. Ito, T. & Matsui, T. 1975. Methods to estimate landslide forces acting on stabilizing piles. Soils and Foundations, 15:4:43–59. Liang, R.Y. 2002. Drilled shaft foundations for noise barrier walls and slope stabilization. University of Akron, Akron, Ohio, Report prepared for the Ohio DOT and FHWA. Leroueil, S. & Tavenas, F. 1981. Pitfalls of back-analyses. Proc. 10th Int. Conf. Soil Mech. Found. Eng., 1:185–190.

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Popescu, M.E. 1991. Landslide control by means of a row of piles. Keynote paper. Proc. Int. Conf. on Slope Stability Engineering, Isle of Wight, Thomas Telford, 389–394. Popescu, M.E. 1995. Keynote Lecture: Back analysis of slope failures to design stabilizing piles. 2nd Turkish Symposium on Landslides, 15–26 October, Adapazari, Turkey. Popescu, M.E. 1996. From Landslide Causes to Landslide Remediation, Special Lecture. Proc. 7th Int. Symp. on Landslides, Trondheim, 1:75–96. Popescu, M.E. 2001. A Suggested Method for Reporting Landslide Remedial Measures. IAEG Bulletin, 60, 1:69–74. Popescu, M.E. & Yamagami, T. 1994. Back analysis of slope failures—a possibility or a challenge? Proc. 7th Congress Int. Assoc. Eng. Geology, Lisbon, (6), p. 4737–4744.

Popescu, M.E., Yamagami, T. & Stefanescu, S. 1995. Nonlinear strength envelope parameters from slope failures. Proc.11th ECSMFE, Copenhagen, (1), p. 211–216. Morgenstern, N.R. 1992. Keynote Paper: The role of analysis in the evaluation of slope stability. Proc. 6th International Symposium on Landslides, Christchurch, 3:1615–1629. Saito, M. 1980. Reverse calculation method to obtain c and φ on a slip surface. Proc. Int. Symp. Landslides, New Delhi, 1:281–284. Terzaghi, K. 1950. Mechanisms of Landslides, Geological Society of America, Berkley, 83–123.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Landslides on the left abutment and engineering measures for Manwan Hydropower Project Xianliang Tang & Qiang Gao Kunming Investigation, Design and Research Institute, CHECC, Kunming, China

ABSTRACT: A large slope failure with a volume of 10.6 × 104 m3 occurred on the left abutment of Manwan hydropower Project was presented in this paper. Geological condition, failure mechanics and simulation, and remedial words, lessons learned, are also discussed in this paper. 1

SKETCH OF THE PROJECT

The Manwan project is located on the Lancang river between Yunxian county and Jingdong county of Yunnan province. The project is composed of concrete gravity dam, spillway tunnel and powerhouse. The dam has the maximum height of 132 m. The slope of left bank strikes at N40◦ W with three sides facing the river, see Figure 1. The angles of natural slopes range from 45◦ at left abutment to 35◦ at the end of downstream plunge pool. The elevation of the top of hill near the dam axis is 1040 m and strip like hill ridge has the ratio of width to height is 2:1. The excavation of dam foundation and

plunge pool cuts off the medium dipping structural surfaces. The slope elevation drops from 920 m to 911 m and produces large slide with total volume of 10.6 × 104 m3 . 2

GEOLOGY

The fresh rock of left bank slope is compacted without primary soft layers. The structural surfaces in rock mass are abundant, which can be classified to three categories: (1). NNW—NW, SW∠35◦ ∼45◦ ; (2). NNW—NW, SW—NE∠60◦ ∼90◦ ; (3) EW. The joint is distributed with interval of 20 m∼30 m (refer to Fig. 2). Figure 3 gives the geological cross section of 0 + 50. Figure 4 shows the statistical maps of fractures using upper hemisphere projection method. 3

DESCRIPTION OF LANDSLIDE

A catastrophic landslide happened during the left abutment excavation of Manwan concrete dam on January 7, 1989 when the elevation went down from 1017 m to 911 m. This case caused delay of the on-line

Figure 1. Geologic Plan view of left bank of Manwan dam site.

Figure 2.

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Joints along the slope of left bank.

Figure 3. 0+50 geological cross section.

Figure 4. Statistical maps of fractures in left bank rock mass (upper hemisphere projection) a) strong weathered zone b) strong weathered zone of outlets of three holes c) weak weathered zone d) weak weathered zone.

Figure 5. The landslide happened during the exaction of the left abutment of the Manwan concrete dam (a) Before the failure (b) After the failure.

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schedule of a 1500 MW power plant for one year. Figure 5 shows the pictures before and after the failure. Figure 6 shows the initial part of landslide passing through fault f393 . The steels and piles after failure are shown in Figure 7 and 8 respectively. 4

REINFORCEMENT MEASURES

Figure 6. The initial part of slide passing through fault f393 .

Several engineering measures such as cutting slope, unloading, anti-sliding piles, pre-stressed cables, shallow and deep drainage have been adopted to guarantee the stability after the landslide failures. An overview of left bank slope after the removal of failure deposits can be seen in Figure 9. Figure 10 shows the layout of cables and anchoring holes.

Figure 7. Failure of steels due to tensor.

Figure 8.

Failure description of piles.

Figure 9. An overview of left bank slope after the removal of failure deposits.

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Figure 10.

Layout of cables.

The monitoring results and the ten years of practice suggest that the engineering measures are valid to retain the slope stability. 5

EXPERIENCES

1. It is an important step to ensure the stability of slope by strengthening construction management, strictly following the construction steps and controlling blasting operation. 2. The excavation of the slope of rhyolite with abundant fractures leads to a landslide failure.

The happened landslides and the corresponding engineering measures indicate that:

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Factors resulting in the instability of a 57.5 m high cut slope J.J. Wang Research & Design Institute of Chongqing Communication, Chongqing, China College of River & Ocean Engineering, Chongqing Jiaotong University, Chongqing, China

H.J. Chai & H.P. Li Research & Design Institute of Chongqing Communication, Chongqing, China

J.G. Zhu Research Institute of Geotechnical Engineering, Hohai University, Nanjing, China

ABSTRACT: The stability of cut slopes is still an evolving art in building highways in mountain areas. The factors resulting in the instability of a cut slope with 57.5 m in height is investigated in present paper. Its shape features, engineering geological, hydrological geological and tectonic geological conditions, deformation behaviors are investigated using several prospecting methods such as boring, welling and drawing. The factors resulting in the instability of the cut slope may be summarized as five aspects as follows: (1) cracked rock masses by a fault and lots of joints, (2) soft discontinuous surfaces in the rock masses, (3) steep slope angle, (4) unreasonable cutting method, and (5) infiltration of raining water. 1

INTRODUCTION

More than 30 000 km highways are/will be constructed in mountain areas of China. During the construction, lots of high and steep slopes are/will be cut. The engineers and constructors should do their best to ensure the stability of the cut slopes. But the stability of the cut slopes, especially that of the high and steep slopes cut in soil mass and/or cracking rock mass, is still an evolving art. Many failure instances were reported by some researchers or engineers such as Goodman & Kieffer (2000), Pugh & Puell (2006), and Fell et al (2007). In present paper, the instability mechanisms of a highway cut slope with 57.5 m in height were investigated. The slope was cut and then instable during the re-construction of a second-class highway (Trade Standard of P.R. China, JTG-D30, 2004) in mountain area in the Northwest of China. The general view of the cut slope is shown in Figure 1. It is clear that the right part of the slope is instable. The mean length of the unstable part of the cut slope in its sliding direction or in the direction normal to the highway is about 62.5 m, and the mean width in the direction parallel to the highway is about 67.5 m. The mean thickness of the unstable part of the slope is about 13.4 m (see Figure 2), such that the area is about 4219 m2 , and the volume is about 56531 m3 .

Figure 1.

2

General view of the cut slope.

GEOLOGY CONDITIONS

According to the survey standards for engineering geology to building highways in P.R. China (National Standard of P.R. China, GB50021, 2001; Trade Standard of P.R. China, JTJ-064, 1998), the engineering geological, hydrological geological and tectonic geological conditions of the cut slope were investigated by several prospecting methods such as boring, welling

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Altitude (m) 570

S49˚E

560

Slope surface Secondary slip surface

550

dl

Q S1

540 530

Main slip surface S1

Highway

520 510

O3

Figure 2. Structure sketch of the cut slope. Figure 3. surface.

Argilliferous clay and scratch on main slip

and drawing, and they were described in present section. 2.1 Engineering geological conditions Prospecting results indicated that the geological structure of the cut slope can be expressed as the sketch shown in Figure 2. The figure shows that there are a Quaternary soil layer with 2.4∼7.9 m in thickness containing lots of rock blocks (Qdl in Figure 2) under the surface of the cut slope, heavy weathered Siluric shale mass (S1 in Figure 2) under the soil layer, and then Ordovician limestone (O3 in Figure 2). It is clear from the sketch shown in Figure 2 that there are two slip surfaces in the cut slope, i.e., main and secondary slip surfaces, respectively. The secondary slip surface is also the bottom boundary surface of the Quaternary soil mass, such that its depth is about 2.4∼7.9 m. Its dip angle is changing from 60∼77◦ at its upper part to 15∼30◦ at its middle part. The main slip surface, with about 10.8∼16.6 m in depth, is a discontinuous surface through the heavy weathered Siluric shale mass. Its dip angle is also changing from 68∼77◦ at its upper part to 10∼45◦ at its middle part. It may also be seen from the sketch shown in Figure 2 that the altitude of the after-edge of the instable cut slope, which is the intersection line between the main slip surface and the slope surface at its upper part, is about 569.8 m, and that of the leading-edge, which is also the intersection line but at its lower part, is about 525.2 m. The height difference between the after-edge and leading-edge of the instable cut slope along the main slip surface is therefore about 44.6 m. Along the two slip surfaces shown in Figure 2, some traces reflecting the sliding deformation or displacement of the cut slope were found in prospecting boreholes and wells. Especially on the main slip surface, a thin layer of argilliferous clay with many

Figure 4.

Discontinuous deformation of shale mass.

scratches indicating the instable sliding and the sliding direction of the cut slope was found (see Figure 3). From the scratches on the thin clay layer, it is known that the main sliding direction of the unstable cut slope is about S41◦ E as shown in Figure 2. Another trace indicating the position of the main slip surface and the sliding deformation of the cut slope is the discontinuous deformation or displacement of the shale mass at the front-end of the main slip surface, namely the intersection region between the main slip surface and the slope surface. (see Figure 4) 2.2

Hydrological geological conditions

The hydrological geological conditions of the cut slope should include two aspects, i.e., surface water and ground water, respectively. During dry seasons, no surface water can be found on or near the cut slope. But on rainy weather, only little surface water collected by raining water may be found on the slope

1800

surface because most of raining water has infiltrated into the soil/rock mass of the cut slope. The infiltrating water is an important factor inducing the instability of the cut slope because it can weaken the strength of the soil and/or rock mass of the slope, and increase the self-weight of the unstable soil/rock mass at the same time. On fine days, no ground water was found from the boreholes drilled in the cut slope, and also no spring was found on or near the cut slope. But after rainy weather, little ground water or spring was found seeping out along the two lines which are respectively the intersection lines between the slope surface and the secondary slip surface, and between the slope surface and the main slip surface. This indicated that the ground water did completely come from raining water, such that the instability of the cut slope may be induced during rainy weather. 2.3

Tectonic geological conditions

The tectonic geological conditions of the cut slope include two parts, i.e., fault and joint. The instability of the cut slope may be related to the fault at the middle of the whole cut slope (see Figure 1). The fault is the right boundary of the instable part of the cut slope, and it is shown in Figure 5. It is clear from the photo shown in Figure 5 that the rock masses besides the fault are different. The rock mass upper the fault, namely the hanging side of the fault, is the heavy weathered Siluric shale mass. But the rock masses under the fault, namely the heading side of the fault, are the cracking Siluric shale mass and the fracture Ordovician limestone. The position of the Ordovician limestone at the heading side of the fault was elevated, such that the fault is a normal fault. It may also be seen from the figure that the dip direction, dip angle and strike direction of the fault face are 268◦ , 63◦ and 178◦ , respectively. It is also clear from the photo in Figure 5 that the rock mass at the heading side of the fault is more cracking than that at the hanging side. On rainy weather, more raining water may infiltrate into the soil/rock mass at the heading side. The prospecting boreholes explored some soft structure surfaces such as thin clay, solution phenomena such as solution crack and/or cavity, which are related to the long-term interaction between ground water and rocks. The cracking rock mass at the heading side of the fault was proved to be related to the fault, even to be the results of the fault, such that the instability of the cut slope should be related to the fault. The other tectonic geological condition of the instable cut slope is the joints or fissures in the rock masses. There are four group fissures in the Siluric shale mass, whose dip directions and dip angles are respectively (1) 57∼61◦ and 27∼86◦ , (2) 100∼138◦ and 50∼76◦ ,

Figure 5.

General view of the fault.

Figure 6.

Shale rock mass cut by fissures.

(3) 152∼184◦ and 45∼75◦ , and (4) 192∼245◦ and 53∼85◦ . The cracking shale mass is resulting from these fissures. The shale rock mass cut by the fissures is shown in Figure 6. Two group joints were also found in the Ordovician limestone, and their dip directions and dip angles are respectively (1) 100∼125◦ and 62∼81◦ , and (2) 181∼200◦ and 73∼88◦ .

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3

DEFORMATION BEHAVIORS

3.1 Deformation on slope surface In order to cut the slope as quickly as possible during the construction, some special excavation methods were used to cut the slope, such as blasting methods. During cutting the slope, large deformation or displacement of the slope surface were observed by the constructors and field engineers, but the building work wasn’t stopped, even the excavation methods weren’t changed because of the observed overlarge deformation. Soon after the cutting work, larger and larger deformation was observed, such that the cut slope should be regarded as in unstable state. Figure 7 shows the fissures on/near the forehead of the cut slope surface. It is clear from the picture shown in Figure 7 that the rock/soil mass at/near the surface of the slope forehead are cracked by many open fissures. The open fissures should be induced by the deformation of the cut slope because their surfaces

are fresh and no plant is growing on their surfaces. The widths of the open fissures are about 0.05∼1.0 m, their depths are at least 0.1∼2.0 m, and their lengths are 4.2∼26.8 m. The relative displacement in vertical directions of the two faces of the fissures is about 0.05∼2.7 m, and that in horizontal directions is about 0.1∼2.4 m. Lots of cracks were also found at the middle, side-edge and after-edge of the unstable slope surface, respectively. Their widths, depths and lengths are respectively about 0.05∼0.5 m, 0.3∼4.0 m and 1.5∼45.5 m. Figure 8 shows the open crack with 0.3∼0.4 m in width, 45.5 m in length, and 0.5∼3.5 m in visible depth. It is clear that the crack should be the after-edge of the unstable cut slope. The relative displacement in vertical directions of the rock masses at two sides of the crack is about 0.5∼1.0 m.

Figure 9.

Prospecting well No.TJ01 (April 19, 2006).

Figure 10.

Prospecting well No.TJ01 (May 3, 2006).

Figure 7. Fissures on/near forehead surface of the slope.

Figure 8. Crack at the after-edge of the unstable slope.

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3.2

Deformation in exploratory excavation

During the investigation of the cut slope, some prospecting wells were excavated in order to explore its geological conditions and monitor its deformation or displacement. Figure 9 shows a new prospecting well near the after-edge of the unstable cut slope. It is clear that no large or open fissure can be observed on the wall and bottom of the new well. But after about 14 days, the wall and bottom of the well were cracked by the open fissure as shown in Figure 10. It should be stressed that there was no rain during the 14 days. It is therefore evident that the displacement of the cut slope is very obvious even on fine days. 4

INSTABILITY MECHANISMS

Based on the investigation for the geological conditions and deformation behaviors of the cut slope, its instability mechanisms can be summarized as follows: (1) Cracking rock mass. During inducing the normal fault at the middle of the whole cut slope (see Figs. 1 and 5) in tectonic process, lots of fissures in the rock masses especially in the shale mass and limestone near the fault surface in its heading side were also induced. The structure fractures cut the rock masses, the strength of the rock masses was therefore weakened; (2) Thin clay layers or soft discontinuous surfaces in the rock masses of the unstable cut slope. Under long-term weathering action and water-rock interaction, the fissures resulted from tectonic movement may be filled by soils such as clay, silty or sand, and some new fissures may also be induced. The strength of the rock masses was further weakened, and the stability of the slope even before cutting should be weak. (3) Steep slope angle. The natural slope angle before cutting is about 10∼30◦ , but cutting slope angle during its design and construction is about 55∼65◦ . The cutting slope angle about 55∼65◦ may be adequate or reasonable for the rock mass with few cracks and high strength values, but for the breaking shale mass and loose soil mass of the cut slope, it may be too steep. (4) Unreasonable cutting method. In order to shorten the time span used to building the highway, the blasting techniques usually for hard rock mass were used to cut the slope which was made of soil mass and soft shale mass. The rock/soil mass in the cut slope may further be loosened by the vibrating forces, such that the stability of the cut slope may be reduced. Another feature of the cutting method for

the slope is that the slope wasn’t cut step by step, but the whole slope was cut almost at the same time. It is well known that the method, in which the whole slope is cut at the same time rather than step by step, isn’t reasonable or scientific for most soil slopes and rock slopes with steep cutting angle or high height. (5) Infiltrating of raining water Almost every engineer understands the influence of the water (including both the surface and ground water) on the stability of the slope. During and after the cutting of the slope, the engineers and constructors at the field found that the deformation or displacement of the cut slope was much larger on rainy weather than that on fine days. This indicated that the infiltration of raining water into the soil and rock masses of the cut slope may weaken the stability of the slope, and accelerate the process of the instability of the slope. The influence of the infiltrating water on the stability of the slope at least came from two aspects, i.e., the increase of the self-weight of the soil/rock mass and the decrease of the shearing strength of the rock/soil mass along the main and secondary slip surfaces by the water.

5

CONCLUSIONS

The stability of cut slopes is very important to the construction and management of the highways in mountain areas. In order to improve the building quality of the highways and reduce the slope hazards, engineers and constructors should improve the stability of the cut slope. But the stability of the cut slope during the construction of the highways in mountain areas is still an evolving art. Some suggestions may be obtained from the instable cut slope investigated in present paper as follows: 1. Understanding mechanisms of the instability of the cut slopes is very useful to reduce the cut slope hazard. Therefore the investigation on the mechanisms of the instability of the cut slopes such as the slope described previously is a valuable tool to improve the highways in mountain areas. 2. Exploring geological conditions, such as engineering geological, hydrological geological and tectonic geological conditions related to the cut slope, are very important to design the slope structures such as its height and slope angle, and to choose reasonable excavation method cutting the slope. 3. Scientific cutting method should be given large attention in order to improve the stability of the cut slope, because the instability of many slopes were reported as results from unreasonable cutting methods.

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4. Water, especially ground water, is a very important bad factor affecting the stability of most cut slopes including soil and/or rock slopes. The unfavorable influence of the water on the stability of the cut slopes has been paid large attentions by more and more researchers and engineers. ACKNOWLEDGMENTS The authors gratefully acknowledge the financial supports from the Chongqing Municipal Education Commission of China under Grant No. KJ070416, the Natural Science Foundation Project of Chongqing Science & Technology Commission of China under Grant No. CSTC2007BB7422, the Science Foundation Project of Postdoctors in China under Grant No. 20070410206, and the Ministry of Communications of China under Grant No. 2005319740090, respectively.

REFERENCES Fell, R. Glastonbury, J. & Hunter, G. 2007. The eight Glossop lecture—Rapid landslide: the importance of understanding mechanisms and rupture surface mechanics. Quartely Journal of Engineering Geology and Hydrogeology 40: 9–27. Goodman, R.E. & Kieffer, D.S. 2000. Behavior of rock in slopes. Journal of Geotechnical and Geoenvironmental Engineering, ASCE 126 (8): 675–684. National Standard of P.R. China, GB50021. 2001. Survey standard for geotechnical engineering. China Architecture and Building Press, Beijing, P.R. China (in Chinese). Pugh, R.C. & Puell, F. 2006. First time landslide in mudstones in Aviles, Asturias, Spain. Quartely Journal of Engineering Geology and Hydrogeology 39: 241–247. Trade Standard of P.R. of China. JTG-D30. 2004.Design standard for subgrade of highways. Beijing: China Communications Press (in Chinese). Trade Standard of P.R. China, JTJ-064. 1998.Survey standard for engineering geology in highways. Beijing: China Communications Press (in Chinese).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Orthogonal analysis and applications on anchorage parameters of rock slopes E-chuan Yan, Hong-gang Li, Mei-jun Lv & Dong-li Li Engineering Faculty, China University of Geosciences, Wuhan, China

ABSTRACT: Anchorage engineering is the most effective means of protecting rock slopes, while till now the problem how to select anchorage parameters i.e. length, spacing and prestress value etc. is not well solved yet theoretically and practically. This thesis, taking Shi-zi-bao rock slope as an example, analyzes grouping approach of anchorage parameters via orthogonal experimental design method; adopts numerical stimulation method to study stability alteration when changing anchorage parameters separately. The study shows in terms of effect (from stronger to weaker) influencing factors over slope stability line in this way, prestress value, spacing, length. Based on the study, anchorage parameter and its designing pattern is decided, and definite difference method FlAC3D is employed for numerical analysis to verify anchorage effect. The engineering practices show that orthogonal design method is practical optical design of anchorage engineering. 1

INTRODUCTION

2

With the rapid economic development, massive rock slope problem appears in the course of resource exploitation and basic facilities construction. And due to enlarging engineering scope the stability problem becomes increasing prominent with growing slope height. So reducing rock excavation to minimum and safety prospects of slope are the main concern of both civil engineers and geological engineer. Under the circumstances of steep slope, serious safety problem or obvious influence over project expense (Cheng, 1996 & Cao, 2005) by slope designing, optical design for the rock slope protection become extremely important. In China, anchorage engineering is the most effective means of protecting rock slopes, while till now the problem how to select anchorage parameters i.e. length, spacing and prestress value etc. is not well solved yet. Most designs, just based on experience, can not achieve optical effect. Therefore how to determine the optimum scheme for anchorage is especially important (Sun 1998). This thesis, taking Shi-zi-bao rock slope as an example, firstly analyzes grouping approach of anchorage parameters via orthogonal experimental design method; secondly to study the slope stability condition after protecting, definite difference method FlAC3D is employed for numerical analysis and stability evaluation. The study provides important basis for optimizing the anchorage designing.

THE ENGINEERING GEOLOGICAL CONDITION OF SLOPE

The Shi-zi-bao rock slope is located beside road (from the Badong new town to Xi-xiang-po) along the Yangtse River. The slope starts from Tongpenxi bridge head in east, ends at road corner of Zhao-shuling slope in west with the length of 368 m and height of 10∼50 m. Most parts of the slope are nearly erect, leaning to north (about 350◦ ). The geological condition of each stratum inner the slope describe as follows (from ground surface to under): 1 the arable stratum: the substance is silty clay in snuff color and the thickness is 0.3∼0.6 meter; 2 the residual soils stratum: the substance is silty clay mixed up with macadam whose component is the weathering residues of the limestone and the muddy limestone. This stratum is in snuff color and thickness is 0.5∼5 meters. 3 Badong Formation of Middle Tertiary : the substance mainly is the muddy limestone, limestone and thin-bedded marlite which is medium bedded , thick-bedded or massive and are in color caesious, yellow-gray and fawn. The slope is mainly composed by this rock stratum whose thickness is greater than 50 meters (Tan 2004). 3

ORTHOGONAL ANALYSIS ON ANCHORAGE

As one scientific method for treating with the multifactors experiment, the orthogonal design method

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can chooses the combination conditions which are representative and less frequency by making few tests. Then, the better condition or scheme can be educed by simple calculating. By analysis of test results, the optimum experimental plan and the factors effect also can be obtained. In recently years, the orthogonal testing method is used widespread application on geotechnical engineering, and also obtains many significant conclusions for engineering.

Table 1.

Factor and level of parameters.

Level

Length meter

Spacing meter

Prestress ton

1 2 3 4

6 10 15 20

2.5 3.0 3.5 4.0

0 20 40 60

anchor arm

3.1 The stability calculation model of slope

3.2 Test arrangement and result Every factor is supposed no interactive in this paper. According to preceding text, the model with five factors and four level parameters had been established. Then the test can be made by selecting orthogonal table L16 (45 ). Successfully, the test only need do 16 times but needn’t do the whole which is 43 = 64 times. At the same time, limiting equilibrium analysis of slope is carried through the 4 levels parameters that are in Table 1. The test arrangement and the calculate result can be seen in Table 3.

anchor rope

1

lattice beam

2 3

anchor arm

4 5

22m

In this paper the 2–2’ section is chosen for calculating the stability of Shi-zi-bao rock slope. Firstly the slope is excavated and reinforced with anchor rope and anchor arm. As the slope will reload after excavating, we must take a series of reinforcement to control the relaxation displacement, improve the stress condition and the stability of slope, etc. which all are the important indicators to appraises the anchorage measure. Therefore, in order to reflect the reinforcement effect of the anchorage measure, the stability coefficients after slope protected is taken as the evaluating indicator in this paper. There are many factors to influence the anchorage effect, which not only have geometry parameters but also have physical mechanics parameters. So, it’s very difficult to comprehensive analyze all the factors. According to the importance and the special condition, the length, the spacing and the prestress which are three primary factors were researched in this paper. Four kinds of representational situations are selected for each factor (four levels of parameters) (Jin 1988). The factor and level of parameters are selected as fallows (Table 1). According to the different parameter combination, every calculation is taken as a test. Because the paper length limited, we only list one model for example (as in Figure 1). In the first test, the mechanical parameter of rock material is listed as Table 2. Besides, the Young’s modulus of anchor rod takes 200 GPa and the diameter is 25 mm.

6

Figure 1. Table 2.

Calculating model of experiment 1. Mechanical parameter of rock material (1).

Petrofabric Density ρ (g/cm3 ) Deformation modulus Em (GPa) Passion’s ratio µ Shearing strength cm (MPa) Shearing strength m (degree)

Argillaceous limestone

Marlite Limestone

2.66

2.26

2.69

8.6 0.32

2.2 0.37

11.2 0.37

0.09

0.017

0.2

28

20

42

3.2.1 Rang analysis The comparison during the test result can be done after performing the 16 experiments. However, it would be impossible to pairwise compare directly because every test are different from other. In other words they don’t have the foundation to compare. Nevertheless, if the test data was combined we will discover the comparability between them (Zhang 2001). For example, the first level (A1 ) of the length appears in the test 1, 2, 3, 4 of Table 4, and the average stability coefficient of the slope from the four tests can be depicted as A′1 . According to the same principle, other average

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Table 5.

Table 3. Mechanical parameter of sliding surface (2).

Rang analysis.

Sliding surface type

Shearing strength

Type

cj (MPa) φj (◦ )

Levels

A′i

Bi′

Ci′

0.18 0.26 0.14 0.35 0.28 0.21

1 2 3 4 Rang Order

1.492 1.497 1.513 1.519 0.0264 3

1.548 1.513 1.489 1.47 0.0774 2

1.389 1.472 1.544 1.615 0.2263 1

Rock character

Bedding plane

Limestone Argillaceous limestone Marlite Extension joint Limestone of parallel Argillaceous limestone slope surface Marlite

30 25 20 35 28 20

Stability coefficient

Table 4. Result of orthogonal experiment. B(s)

C(P)

—

—

Appraisal index

Test number

1

2

3

4

5

Stability coefficient

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16

1(6) 1 1 1 2(10) 2 2 2 3(15) 3 3 3 4(20) 4 4 4

1(2.5) 2(3.0) 3(3.5) 4(4.0) 1 2 3 4 1 2 3 4 1 2 3 4

1(0) 2(20) 3(40) 4(60) 2 1 4 3 3 4 1 2 4 3 2 1

1 2 3 4 3 4 1 2 4 3 2 1 2 1 4 3

1 2 3 4 4 3 2 1 2 1 4 3 3 4 1 2

1.406 1.479 1.525 1.559 1.505 1.392 1.589 1.503 1.595 1.629 1.382 1.444 1.685 1.554 1.460 1.375

stability coefficients

A(L)

1.63 1.58 1.53 1.48 1.43 1.38 6

10

15

Length (m)

20

2.5

3.0

3.5

4.0 0

Spacing (m)

20

40

60

Pre-stress (t)

Figure 2. Relations between stability coefficients and influence factors.

stability coefficient of the second, third, fourth level is depicted as A′2 , A′3 , A′4 . During the four tests in condition A1 , other factors not only are taken over all the level but also have the same used times in every level. Therefore the A′1 has the peculiarities of comparison because the difference between the A′i , where i = 1, 2, 3, 4, reflect the difference of four levels. In the same way we also can calculate the Bi′ and Ci′ , where i = 1, 2, 3, 4. After detailed calculation, the average value that influenced by the various factors are summarized in Table 5. And from it, the range of A is defined as the difference between the maximum and the minimum of the A′i , which is the important index that can appraise the factor how influence the index. According to Table 5, we can take the compounding of the maximum stability coefficient A4 B1 C4 from A′i , Bi′ , Ci′ as the optimum parameter scheme.

The concrete scheme was that the length of anchor rod is 20 m, the spacing is 2.5 m and the prestress value is 60 t. Obviously, the combination scheme isn’t included in the 16 experiments that have been carried. On the other hand, we can get a conclusion that the result educed by the orthogonal test design is comprehensive. According to Table 5, rang A is 0.0264, rang B is 0.0774, rang C is 0.2263. The bigger range indicates the bigger influence on the slope stability coefficient. So the prestress has the most effect on the stability of slope, followed by the spacing and the length thirdly. The relations between stability coefficients and influence factors are as follows Figure 2. 3.2.2 Variance analysis As the range method has the advantage of small quantity calculation and easy to understand, so it is widely used in intuitively analyses. However, the fluctuating data which is caused by the condition changing or the experimental error aren’t distinguished severely in the test. At the same time, the range methods also don’t put a standard to estimate if the factor effect has significantly influence. In order to make up the deficiency, the variance analysis is used to analyze the influence degree of each factor to achieve indirect evaluate whether the optimum method is proper or not. The specific methods are as follows: Firstly, we suppose the factor i that has remarkable influence on the

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result as long as it satisfies the following formula: Fi =

Si /fi > F1−a ( fi , fe ) Se /fe

(1)

Where: r is the number of level test; n is the number of whole test; Si , fi are the sum of departure and the freedom of the factor i, and fi = r −1; ST is the square sum of total departure; Se , fe are the sum of departure and  the freedom of the error, and fe = fT − m i=1 fi ; fT is the freedom of ST . In generally engineering, if Fi > F0.99 , the influence is special remarkable and labeled with‘‘∗∗ ’’; if F0.99 > Fi > F0.95 , the influence is remarkable and labeled with‘‘(∗ )’’; lastly, if Fi < F0.90 , the influence is not remarkable and isn’t labeled. The significance level is taken as 0.01, 0.05, 0.l in the F examination respectively. Then we have obtained F0.99 (3, 6) = 9.78; F0.95 (3, 6) = 4.76; F0.90 (3, 6) = 3.29. The variance analysis of test result are shown as Table 6, Table 7. The calculation results in Table 7 show that the influence sequence (from large to small) of the parameters is: pre-stress value, spacing and length. In other words, the prestress value has most influence on the slope stability, the spacing is moderate and the length is list. 3.2.3 Selection of optimal scheme The optimal scheme of anchorage parameter is educed by former orthogonal design in intuitively analyses. However, the scheme might not be the most economical and reasonable. According to Figure 2, the stability Table 6. Variance analysis.

Ki1 Ki2 Ki3 Ki4 Ki Si

A length (L)

B spacing (s)

C prestress (P)

5.968 5.989 6.05 6.074 36.2438 0.0018

6.190 6.054 5.956 5.881 36.255 0.0130

5.556 5.888 6.176 6.461 36.35 0.1080

∗

Where: Kij is the sum of test result in ith factor and jth level, and j = 1, . . . , r; yi is the result of the ith test.

coefficients has increased greatly when the anchor rod length changed from 10 m to 15 m, but the variation amplitude decreases when the anchor rod length increase from 15 m to 20 m. It shows that the anchor rod will can’t play its function if the length excessively. After a comprehensive though, the anchor rod length may choose 15 m instead of the optimal 20 m for economy. Besides, the anchor rod spacing was taken 3.0 m as the optimized result. Finally, we can take the compounding of A3B2C4 as the optimum scheme for bolting. The concrete scheme is that the length of anchor rod is 15 m, the spacing is 3.0 m and the prestress value is 60 t.

4

NUMERICAL SIMULATION OF ANCHORAGE IN ROCK SLOPE

4.1 Modeling According to the engineering geological investigation report and the filed condition, the FLAC calculation model was established as Figure 3, where the displacement boundary condition is adopted in the model. In addition, the plan of anchorage measure is shown in Figure 4.

4.2 Analysis of the calculate result The failure zone of slope after and before comprehensive treatment is shown as Figure 5 and Figure 6. As shown in Figure 5, the plastic zone is so wide so that the slope foot has occurred shear failure and the crack almost runs through the slide surface. Therefore, the slope must take a series of treatment to keep stability. The optimum scheme is that the length of anchor rod is 15 m, the spacing is 3.0 m and the prestress value is 60 t. After taking the anchorage measure as shown in Figure 6, the plastic zone and the shear failure zone are decreased significantly. Besides, there isn’t tensile stress zone appeared in the slope. Although the plastic zone still exist inner the shape of slope, the effect of anchorage measures is obvious.

Table 7. Result of variance analysis. Resources of variance Square sum S Freedom f Mean square V Value of F Fe Significance

A (L)

B (s)

C (P)

Error e

Sum

0.0018 3 0.0006 0.57 F0.99 (3, 6) = 9.78

0.0130 3 0.0043 4.11 F0.95 (3, 6) = 4.76 (∗ )

0.1080 3 0.0360 34.18

0.0063 6 0.0011

0.1291 15

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∗∗

F0.90 (3, 6) = 3.29

F L A C 3 D

3 .0 0

Step 19099 Model Perspective 20:19:54 Wed Jan 10 2007 Center: Rotation: X: 7.888e+001 X: 360.000 Y: -5.000e+000 Y: 0.000 Z: 7.133e+001 Z: 0.000 Dist: 4.363e+002 Mag.: 0.8 Ang.: 22.500

Block State None shear-p shear-p tension-p tension-n tension-p tension-p

Figure 6.

Figure 3. Calculation model.

F L A C 3 D

3 .0 0

Step 21224 Model Perspective 22:29:19 Thu Jan 11 2007 Center: X: 7.800e+001 Y: -5.000e+000 Z: 7.500e+001 Dist: 4.363e+002

Rotation: X: 0.000 Y: 0.000 Z: 0.000 Mag.: 1 Ang.: 22.500

Block Model: Mechanical mohr

SEL Geometry Magfac = 1.000e+000

Figure 4. Schematic plan of anchorage measure.

F L A C 3 D

3 .0 0

Step 7095 Model Perspective 22:27:03 Tue Jan 09 2007 Center: X: 7.800e+001 Y: -5.000e+000 Z: 7.500e+001 Dist: 4.363e+002

Rotation: X: 360.000 Y: 0.000 Z: 0.000 Mag.: 1 Ang.: 22.500

Block State None shear-n shear-p shear-n shear-p tension-p shear-p shear-p tension-p tension-p

Plastic zone of supporting.

Through the above analysis, the conclusions that the anchorage measure can obvious improve the stability of the slope and also have the control action on the deformation and failure of slope can be drown. 5

CONCLUSIONS

1. The anchor rod parameter is essential to rod‘s performance and anchorage effect. The existing research shows that the functioning efficiency of anchor rod would be less if the length is oversize. So when selecting length of shank, we should abide by the principle: full use of the anchor rod and achieve economic and reasonable anchorage effect. 2. Optical design on anchorage parameters of Shizi-bao rock slope with orthogonal design method indicates that in terms of effect (from stronger to weaker) influencing factors over slope stability line in this way, prestress value, spacing, length, slope stability coefficient being evaluation index. As a result, the optical parameter design for Shi-zi-bao slope protection is concluded, anchor rod length 15 m, spacing 3.0 m, prestress value 60 t. 3. With numerical simulation analysis of fore-andaft condition of slope via FLAC3D software, we learned that anchorage can effectively improve the mechanics parameter of situ rock mass, enhance the strength of rock mass in anchorage region, and alter stress distribution in the rock around to narrow plastic zone as well. Therefore in this way slope stability conditions can be obviously improved.

ACKNOWLEDGMENTS

Figure 5. Plastic zone without supporting.

This study was support by ‘‘The plan to support the New Century Excellence Talent of the Ministry of Education’’ (No:NCET-07-0775).

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REFERENCES Cao Xingsong & Zhou Depei. 2005. Design of Anti-slidepile with Prestressed Anchor Rope for High Slope of Weak Rockmass. Journal of Mountai Research 23 (4): 447–452. Cheng Liangkui. 1996. The using and progress of the Chinese ground anchorage technology [A]. Beijing: International Academic Publishers. Jin Liangchao. 1988. Optimize test. Beijing: National Defence industry press.

Sun Yuke. 1998. Stability analysis of the Slope Rockmass. Beijing: science press. Tan Juhong & YAN E’chuan. 2004. Influence and Implication of Water and Fissure on Slope Stability. Journal of Mountain Research 22 (3): 373–377. Zhang Lewen et al. 2001. The using and test research of grouted bolts for the slope reinforcement engineering. Chinese Journal of Rock Mechanics and Engineering 20 (supple1): 1209–1212.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Waste rock dump slope stability for a gold mine in California Yang Hong, Gregory C. Rollins & Minsoo Kim MWH Americas, Inc., California, USA

ABSTRACT: The development of the final closure plan for a gold mine in California, USA involves evaluation of slope stability of the Waste Rock Dumps (WRDs) of more than 100 m in height. Based on the results of field investigation and laboratory shear strength tests, slope stability analyses were performed considering both static and seismic conditions due to potential earthquake. The seismic slope stability analyses were conducted using pseudostatic method with various horizontal ground accelerations of up to 0.20 g to investigate the effect of the earthquake. Potential slip surfaces of different depths were also considered to evaluate the potential instability of mass of different volumes. The results of the slope analyses show that the WRD slopes have inadequate factors of safety against slope failure under both static and seismic conditions, and relatively deep-seated failure may occur. Based on the results, slope remediation measures would be required as part of the considerations for the mine closure.

1

INTRODUCTION

Waste rocks are products of mining activity and normally placed nearby the mining area; thus, WRDs are formed. WRD slopes are important engineered slopes and their geotechnical stability has received increasing attention due to their relatively large sizes and the potential impact of failure, as reported by many case histories in literature (e.g. Robertson 1982; Stormont and Farfan 2005). The gold mine in this study is located in the central Sierra Mountain foothills in California. The site includes a 259,000-m2 open pit, three heap leach pads referred to as Waste Management Units (WMUs) and five WRDs; and has been operated as a gold mine since 1850. The WMUs were closed by 1992 in accordance with a state-approved closure plan that involved flushing the heap leach pads of residual cyanide solutions. In addition, reclamation and closure work was completed at the site in 1995, including surface regrading and placement of vegetated soil cover. In recent years, however, post-closure monitoring of on-site wells identified groundwater impacts down gradient of the WMUs and the nearby reservoir. Therefore, development and implementation of a corrective action program and an effective, final closure of the mine were required to be implemented. Currently, various mine closure options have been considered which require the evaluation of the geotechnical stability and consequences of potential failure of the WRD slopes. A preliminary stability evaluation of the WRD slopes has indicated that three of the five WRDs in the site, namely WRDs-2 through -4 as shown in Figure 1,

Figure 1.

Site plan showing the WMUs and WRDs.

would require further evaluation. Therefore, the work was concentrated on these three WRDs and reported in this paper.

2

GEOLOGIC SETTING AND SEISMICITY

Geologic materials in the site vicinity are generally composed of a stratified sequence of the Paleozoic era,

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Pennsylvanian system, Upper Paleozoic series sedimentary deposits. There is a limited thickness of alluvium on site, mostly in the ancestral drainages, now buried beneath waste rock. The maximum reported soil depth is nine feet. The alluvium is underlain by bedrock that is comprised of slate, schist, phyllite, and andesite. Sulfide mineralization includes copper and iron sulfides such as pyrite, chalcopyrite and tetrahedrite. Documentation of the construction practices and conformance testing of the WRDs is largely absent from the site files. However, it was found that waste rock embankments for WMU-1 and WMU-2 may generally be underlain by one to four feet of soil overlying fractured bedrock, although the construction specifications required founding the waste rock on competent bedrock. Depth to groundwater across the site varies from approximately 20 to 200 feet below ground surface and generally mimics topography. The gold mine site is in a region characterized by low historical seismicity in terms of both the total number of earthquakes and the size, or magnitude of the earthquakes. The historical earthquakes are sparsely distributed in the area of the site. The majority of the recorded events that occurred with close proximity to the site (within 25 km) were under magnitude 3.8. The low historical seismicity and the low rate of tectonic activity result in the relatively low site peak ground accelerations (PGAs) that are derived from the probabilistic seismic hazard analysis, even for long return periods. In contrast, the site PGAs derived from deterministic seismic hazard analysis are quite high because of the maximum credible earthquake of magnitude 6.7 on a nearby (0.6–1.6 km distant) fault zone. Given that the majority of the waste rock dumps are above a high quality reservoir and the consequences of slope failure are potentially severe, a minimum of 2% probability of exceedence in a 50-year period design event (2,475 year return period) is considered. The 2,475-year return period PGA from the U.S. Geological Survey is about 0.15 g (USGS 2002). In addition, the 10,000-year return period of PGA is about 0.27 g (0.5% probability of exceedence in 50 years). 3

WM U-2

WRD-2 Erosion gull y

WM U-3 Sli de and t ension cracks WRD-3

Sli des

Figure 2. Aerial photo showing WRD-2, WRD-3, erosion gully and slides. Slide and tension cracks (on the bench) Erosion gully

Figure 3. Photography showing the WRD-2 slope and erosion gully.

SURFACE AND SUBSURFACE CONDITIONS

A site reconnaissance of the WRD slopes was conducted following the examination of the topographic maps (Figure 1) and aerial photos (Figure 2) of the site. A number of slope stability issues were observed and identified, which include 1) A large erosion gully developed at the western end of the intermediate flat deck of WRD-2 (Figures 2 and 3). The gully was more than 3 m deep in locations with side-slopes locally steeper than 1H:1V. The slopes of the gully were

Figure 4. Photography showing the slide and tension cracks on top of the WRD-2 slope (as indicated in Figures 2 and 3).

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4

observed to be actively sloughing during heavy precipitation. A large amount of sediment was observed extending downstream of the fill slope toward the nearby reservoir. 2) A shallow (estimated less than 2 m deep) failure was developing on the west side of the large erosion gully as evidenced by the tension cracks at the slope crest (Figure 4). This failure appeared to be related to the development of the gully and the resulting locally over-steepening of the slope. 3) Two significant slope failures (slides) were observed on the waste rock slopes located between WRDs- 2 and 3 as shown in Figures 1 and 2. This area is comprised of relatively thin layer of waste rock placed over a preexisting ridge. Although the specifications required founding the waste rock on competent bedrock, the failures could be the result of founding the waste rock on weathered rock/colluvium. A field exploration program comprising eight borings and four test pits were completed to characterize the subsurface geotechnical conditions of the WRDs. The waste rocks observed in the drilling program were serpentinite and schist with some andesite and lesser quantities of quartzite and phyllite. A significant amount of fines (silt and clay) were observed throughout the borings. In addition, locally saturated zones were also observed in some of the borings. Based on historical records of observations made during construction, the waste rock reportedly broke down during placement. These observations in conjunction with the field observations clearly indicated that the waste rock dumps are not fully free-draining with localized areas that perch water. WMU leakage is likely contributing to the source of water. However, naturally occurring seeps and springs in the pre-existing valley could also be contributing to source of water.

SLOPE STABILITY ANALYSES: INPUTS, RESULTS AND DISCUSSIONS

The WRD slope profiles were developed based on preconstruction and current topography in conjunction with the results of the drilling program. In general, the WRD materials of various thicknesses overlie the bedrock sandwiching a relatively weak layer of foundation soil throughout majority of the WRD slopes. The foundation soil is either highly weathered, incompetent bedrock or alluvium/colluvium due to the inadequate preconstruction striping. Field investigation indicates that the WRD materials are highly heterogeneous with a wide range of grain size distribution. The WRD material can be generalized with a classification of sandy, silty gravel with cobbles and boulders. The WRD material has exhibited a relatively wide range of shear strength values based on previous and recent laboratory shear strength testing. The effective friction angles range from 24 to 42 degrees, with majority of the friction angles ranging from 30 to 36 degrees. An average shear strength comprising a friction angle of 34 degrees and cohesion of 5 kPa was considered appropriate for the WRD materials. The shear strength values are comparable to those of mine waste materials that are similar to the WRD materials. For example, Stormont and Farfan (2005) investigated a mine waste pile in Colorado using largescale field and laboratory testing and found an average shear strength value of friction angle of 37 degrees and cohesion of 5 kPa for the materials with a silty gravel (GM) classification. In addition, for the foundation soil of the site, a shear strength value corresponding to a friction angle of 32 degrees and cohesion of 10 kPa was used based on laboratory shear strength testing. Total density for the WRD material and foundation

Elevation (feet) (+1200 feet)

600

WRD-2 Slope

A: Shallow slip surface B: Intermediate slip surface C: Deep-seated slip surface D: Foundation slip surface

500 Waste Management Unit

Emba nkme nt dam

Upper slope

400

WRD material D C B Lower slope

Bedrock

A

300 Foundation soil

Bedrock

200

100 0.0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1.0

1.1

1.2

1.3

Distance (feet) (x 1000) Figure 5. Slope cross-section and potential failure surfaces for lower WRD-2 slope (1 foot = 0.305 m).

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1.4

1.5

soil was estimated at 19.6 kN/m3 and 18.8 kN/m3 , respectively. To account for the seismic loading conditions in the limit equilibrium slope stability analysis, pseudostatic analyses were performed by assigning a horizontal ground acceleration (HGA). HGA values for pseudostatic analysis generally range from 0.05 g to 0.20 g depending on seismic characteristics and site conditions (Abramson et al. 2001). For this study, although an average HGA value of 0.10 g was considered appropriate based on PGA, HGA values ranging from 0 to 0.20 g were also used in the pseudostatic slope stability analyses. An HGA value of zero represents the static condition. To evaluate the stability of WRD slopes with respect to different potential slip surfaces, stability calculations were also completed on critical cross-sections of the slopes assuming a few potential slip surfaces of

different depths, namely, shallow, intermediate, deepseated and foundation slip surface. These slip surfaces are circular in shape, which resulted in lower factors of safety as compared with non-circular slip surfaces of similar depth. Although only the shallow slip surface is the most critical slip surface (i.e. resulting in the lowest factory of safety) and selections of other slip surfaces are arbitrary, all these slip surfaces correspond to potential failure masses of different volumes or extents (thus different level of risk in case of a slope failure). The analyses indicate that the lower portion is more critical than the upper portion for WRD-2 slope; and WRD-3 and WRD-4 slopes have similar results. Therefore, only slope profiles (with different slip surfaces) and the results of slope stability analyses for the lower WRD-2 slope and WRD-3 slope are presented, as shown in Figures 5 through 8. In general, the factors of safety decrease, as expected, with increasing HGA values and decreasing depth of slip surfaces (i.e. shallower slip surfaces correspond to lower factors of safety). The results also indicate that the WRD slopes, which were constructed at or near angle-of-repose, exhibit relatively low static factors of safety ranging from approximately 1.04 (lower WRD-2 slope) to 1.30 (WRD-3 slope) for shallow slip surface. The low factor of safety for lower WRD-2 slope is consistent with the shallow slide that has already occurred (next to the erosion gully) as indicated in Figures 3 and 4. For the lower WRD-2 slope (Figures 5 and 6), using the most appropriate HGA value of 0.10 g, the pseudo-static factors of safety are less than 1.0 for all the potential slip surfaces indicating that the slope is generally unstable and a deep-seated slide may occur. For WRD3 slope (Figures 7 and 8), the pseudo-static factors of

1.4 Lower WRD-2 Slope

Shallow slip surface Intermediate slip surface Deep-seated slip surface

Factor of Safety

1.2

Foundation slip surface

1.0

0.8

0.6 0.00

0.05

0.10

0.15

0.20

Horizontal Ground Accelaration (g)

Elevation(feet) (+1200 feet)

Figure 6. Factors of safety versus horizontal ground accelerations for lower WRD-2 slope.

600

D

500

400

C

Waste Management Unit Compacted soil

B

WRD-3 Slope A: Shallow slip surface B: Intermediate slip surface C: Deep-seated slip surface D: Foundation slip surface (cut into WMU)

A

WRD material 300

Foudation soil

200

Bedrock 100 0.0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1.0

1.1

Distance(feet) (x 1000) Figure 7. Slope cross-section and potential failure surfaces for WRD-3 slope (1 foot = 0.305 m).

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1.2

1.3

1.4

ering both static and seismic conditions. The seismic slope stability analyses were conducted using pseudostatic method with various horizontal ground accelerations of up to 0.20 g to account for the effect of earthquake. Potential slip surfaces of different depths were also considered to evaluate the potential instability of mass of different volumes. The results of the slope analyses show that the WRD slopes have inadequate factors of safety against slope failure under both static and seismic conditions, and relatively deepseated failure may occur. The results suggest that slope remediation measures should be implemented as part of the considerations for the mine closure.

1.6 WRD-3 Slope

Shallow slip surface Intermediate slip surface Deep-seated slip surface

Factor of Safety

1.4

Foundation slip surface (cut into WMU)

1.2

1.0

0.8 0.00

0.05

0.10

0.15

0.20

Horizontal Ground Accelaration (g)

Figure 8. Factors of safety versus horizontal ground accelerations for WRD-3 slope.

safety using an HGA of 0.10 g are between 1.05 (for shallow slip surface) and 1.15 (for foundation slip surface) indicating the slope is marginally stable under seismic conditions. However, the slope will be unstable with potential foundation failure when the HGA is larger than 0.175 g (Figure 8). Such foundation failure surface will encroach into the WMU-3 and impact the containment systems (Figures 2 and 7). Based on the results of the slope stability analyses, it was recommended that the WRD slopes be flattened to a maximum 2H:1V intermediate slopes in order to achieve acceptable factors of safety under static and seismic conditions. This can be accomplished by both regarding and buttressing the slopes. 5

SUMMARY AND CONCLUSIONS

ACKNOWLEDGEMENTS The findings and opinions presented in the paper are those of the authors of the paper only, and do not necessarily reflect those of the client or administration, nor coincide with those in the project report. REFERENCES Abramson, L., Lee, T.S., Sharma, S. and Boyce, G.M. 2001. Slope Stability and Stabilization Methods. John Wiley. Robertson, A.M. 1982. Deformation and Monitoring of Waste Dump Slopes. Proceedings 4th Canadian Symposium on Mining Surveying and Deformation Measurements, Banff, Alberta. Stormont, J.C. and Farfan, E. 2005. Stability Evaluation of a Mine Waste Pile. Environmental and Engineering Geoscience. Vol. 11 (1): 43–52. USGS 2002. Interactive National Seismic Hazard Maps, U.S. Geological Survey. Website: http://gldims.cr.usgs.gov.

Slope investigation and stability analyses were performed for the WRD slopes at the gold mine consid-

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Properties of the high rock slope of Hongjiadu hydropower project and its engineering treatment measures Zeyan Yang China Hydropower Engineering Consulting Group Corporation (CHECC), China

Wanchun Xiao & Dayong Cai Guiyang Hydropower Investigation Design & Research Institute, CHECC, China

ABSTRACT: The dam site of Hongjiadu CFRD is located in the area with high mountains, narrow valley and Karstic formation. The river bank is very steep and developed with soft intercalations. The height of the left abutment rock slope is 310 m and the height of the large scale bedding slope of intakes is more than 360 m. The scale and the height of those slopes are rarely finding both at home and abroad. It is the main difficulty for the construction of the project. By using different methods to determine the strength parameters and applying multiple slope stability analysis methods, several engineering measures were accepted, which include the timely anchoring during excavation, continuous narrow plinth with constant length, huge anti-slide piles plus anchor cables, etc. The measures have successfully solved the high slope problems of the project. Now, the slope has been safely operated for 3–6 years and still in good condition.

1 1.1

1.2 Introduction of the high rock slope

INTRODUCTION Introduction of the project

Hongjiadu Hydropower Project is located at Qianxi County, Guizhou Province in China. It is the first reservoir of the cascade hydropower development of the main stream of Wujiang River. The normal storage water level of the reservoir is 1140 m and the total storage capacity is 49.47 × 106 m3 . The reservoir is a carryover storage reservoir. The installed capacity of the hydropower station is 600 MW. The main structures of the project include a concrete face rockfill dam with the height of 180 m, spillway tunnel, flood discharge tunnel, power tunnel and surface powerhouse. Except the main dam, all the water carry structures are arranged in the mountain of the left bank. The project commenced in November, 2001 and completed in 2005. The main characteristics of the project are shown in following aspects: high rock slope, high dam in narrow valley, developed Karstic formation, multiple tunnels, large range of reservoir water level change.

Hongjiadu Hydropower Project is located in a deep narrow river valley in Karst region. With the requirement of structure arrangement, several high rock slopes were formed at the left and right abutment, water intake and outlet, outlet of flood discharge structures, rockfill quarries. For the left abutment slope, due to the arrangement of plinth, the height of slope reach to 310 m and the width reach to 350 m. As all the water carry structures are arranged in the left abutment slopes, a large scale bedding slope with the height of 360 m and the width of 540 m was formed at the intake of the tunnels. The steep high slope of the left abutment and the huge scale of the bedding slope of the intake are seldom encountered in previous projects. The safety of the high rock slopes are directly related to the safety of the concrete face rockfill dam with the height of 180 m. It is one of the most important technical issues of the Hongjiadu Project. During the construction of the project, several research works were conducted and the research results were applied in the engineering measures. The paper will present

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the main research results on the left abutment slope and the bedding slope of the intake. 2

LEFT ABUTMENT HIGH ROCK SLOPE AND ITS ENGINEERING TREATMENT

2.1 Basic geological characteristics The nature slope of the abutment is 30◦ ∼90◦ , the maximum height of the slope is 300 m. The angle between the dip of rock stratum and the direction of the slope is about 130◦ . The slope is stable in nature condition. The middle and upper part of the slope are thick or medium thickness and thin layer limestone. Soft intercalations are developed in the layers. The lower part of the slope is argillaceous shale. The rock stratum is monocline, with the dip angle of 25◦ ∼55◦ . In the limestone, medium and small size faults are developed. There are two group fissures with steep dip. With the preliminary analysis and design optimization, the excavation of the left abutment is set for two zones. The area near the dam axis is zone I and the area in upstream is zone II. The average slope of the excavation is 1:0.25. It is excavated vertically in step. The height of the step is 15 m and the width is 3.75. The total excavated slope height is 310 m. The plinth on the slope is a kind of narrow plinth. 2.2

Stability analysis and assessment

According to the ‘‘Standard on grade determination and design safety of hydropower project of China (DL5180–2003)’’ and the working conditions of the slope, the design criteria of the slope are shown in Table 1. The stability of the slope is mainly affected by the adverse combination of the geological structure planes. In the design, the rigid body limit equilibrium method was employed to analysis the possible sliding blocks determined from stereographic projection, the limit equilibrium method by using upper bound solution (energy method) was applied to analysis the overall stability of the slope, the nonlinear finite element method was used to analysis and Table 1.

simulate the whole procedures of the slope excavation and the reinforcement. (1) Stability analysis of the stereographic projection blocks by rigid body limit equilibrium method. By using the stereographic projection, the possible unstable rock blocks and its sliding mode were analyzed. There are 13 combinations for the potential sliding rock blocks and no direct sliding blocks. With the calculation of the wedge stability, most of the sliding body has the FOC larger than 4. Only few blocks at the downstream side of dam axis have relatively low FOC (F = 1.21), which should be reinforced. (2) Slope stability analysis by upper bound limit equilibrium method The slope stability analysis is conducted by using the ‘‘software system for stability analysis of rock slope’’ developed by China Institute of Water Resources and Hydropower Research. The main results are:

① Large scale statistic of the joints and fissures of rock mass were conducted in the investigation adit and the outcrop. 8000 structure plane were measured to determine the attitude of the representative structure planes. ② From the analysis of the sliding mode, the bedding plane (structure plane 1) oblique crossing with the excavation plane. It cannot cause bedding slide or wedge failure. The tendency of structure plane 2 goes all the way with the excavation plane. It may lead to unloading cracks on the top of the slope an also the bedding slide (Figure 1). ③ The shear strength parameters of the rock mass were determined by in-situ and laboratory rock test,

Safety criteria of the left abutment high rock slope.

Grade of the structure affected by Grade of Design Name the slope the slope cases

Load

Left abutment slope

Dead weight 1.30

Dam, grade 1

1

operation

Allowed FOC

Plane 1(bedding): strike 334 , dip 30 ; Plane 2(fissures ): strike 218˚, dip 71 ; Plane 3(fissures ): strike 141˚, dip 76 ; Slope: strike 228 , dip 73 , friction angle 33 Figure 1.

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Instability mode of the left abutment slope.

Hoek-Brown formula and the network simulation method based on Monte Carlo principle. ④ By applying the upper bound approach, the stability of the slope is analyzed by using the parameters determined from three different methods. The calculation results are shown in Figure 2 and Table 2, which demonstrate that the FOC are larger than 1.80 for all the cases. It can be considered that the slope is stable. (3) Simulation analysis by non-linear FEM The non-linear finite element method was applied to conduct the simulation analysis of the process of slope excavation and reinforcement. The computation results show that the interfering energy of the elements is larger than zero, which means the slope is in stable condition. Compare with the results of no reinforcement during excavation, the method of in time reinforcement with the excavation can reduce the horizontal displacement of rock mass up to 2.2∼3.2 mm, and the area of plastic zone is reduced significantly. This has demonstrated that the method of in time reinforcement during the excavation is necessary. The computation results show that: the unloading displacement towards outside of the slope is relatively small. Normally, the value is 4 mm. Near the area of excavation step, the excavation unloading will cause the upward displacement. Normally, the value is 7 mm.

2.3 Engineering measures From the analysis, the left abutment will be in stable status after the excavation. For avoiding the instability of the steep high slope due to the blasting excavation and secondary unloading, reinforcing of the slope is necessary. The general principle of the reinforcing is: the main reinforcing measure is the pre-stressed anchoring in the upper part and the lock bolt support with shotcret along the slope will be the secondary measures. Besides, the drainage system is arranged in the slope. (1) Pre-stressed anchor for locking the top part of the slope Under the second excavation step, 3∼4 rows anchor cable were arranged (as shown in Figure 3). The length of the anchor cable will across the plastic zone predicted by FEM analysis and the potential sliding surface. The inner anchorage section will be put in the integrated rock mass. The orientation of the anchor cable will be arranged to across as much bedding planes and main joint fissures as possible. The inclination angle of the cable should larger than 10◦ and the angle with the excavated slope should larger than 60◦ . There are 69 pre-stressed anchor cables with the stress of 1500 kN in zone I and 45 such cable in zone II. The length of the cables is 36 m or 44 m. The cables are crossed arranged, with the horizontal interval of 5.0 m and height difference 7.5 m. (2) System lock bolt support with shotcret Normally, the lock bolt anchor bar has a diameter of φ 25, interval of 3 × 3 m. The length of anchor bar inside rock is 4 m and the length of outside part is 0.1 m. The anchor bars are arranged in the form of

Figure 2. Sliding surface.

Table 2. Results of the stability analysis. Position

By rock test

By Hoek-Brown principle

By network simulation

Left abutment slope

2.246

1.830

2.176

Figure 3.

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Engineering treatment of left abutment.

cinquefoil. The orientation of anchor bar is determined by excavation plane and is designed to cross as much bedding plane and joint fissures as possible. For the area at the downstream part of dam axis and between elevations 1147.5 m∼1278 m, where the FOC of stability is relatively low, the length of anchor bar will be increased to 15 m. The specification of mesh reinforcement is φ 6@200 × 200 mm. The thickness of shotcret is 150 mm and the concrete is C20. For the small solution opening disclosed by excavation, it will be filled by concrete. For large solution cavern, the treatment will include the lining of the entrance and backfill inside the cavern. (3) Drainage Drainage holes were arranged on the surface of the slope. The diameter is 100 mm, interval is 3.0 m, inclined angle is 10◦ , the depth of drainage hole is 3∼5 m. At the outer side of the excavation steps, a water block bank (30 cm × 30 cm) is arranged. It could form a drainage ditch. Outside of the slope, a cutoff ditch was set for blocking surface water. (4) Control of the excavation height difference During the construction, the slope surface is excavated by presplit blasting. Anchoring was conducted at the same time with the undercut excavation. Normally, the anchorage support should not be lagged behind the excavation by two steps, viz. the height difference of 30 m. The instrumentation should also be conducted simultaneously with the excavation.

3

LARGE SCALE INTAKE BEDDING SLOPE

3.1

Basic geological characteristics

The slope is 360 m in height, with the strike of N45◦ E and the gradient of 25◦ ∼30◦ . The rock of outcrop is the same rock layer with the left abutment slope. The rock is limestone and dolomite limestone. Soft intercalations are developed in the layers, with the thickness of 0.5 cm∼3.7 cm. The soft intercalations are filled with clay and rock scraps. Some sections were eroded by water, where it is filled with mud and rock scraps.

Among those intercalations, eight of them are continuous (J1∼J8). The stratums are monocline and the strikes are almost parallel to the slope surface. The dip angles are 26◦ ∼34◦ . The tendency of stratum is crossed with the slope in a small angle, which form the bedding slope. The slope has 2 big faults, where the reverse fault F13 developed at the top of the slope, fault F6 located at the middle part of the slope. There are 4 group fissures with steep dip angle. Its connectivity is 15.8%∼59.2%. The nature ground water table is relatively deep. Dozens of intakes of different structures are arranged in the slope, which include 3 power tunnels, flood discharge tunnel, spillway tunnel, 2 diversion tunnels and several construction adits. After reservoir impoundment, the ground water level will move up. The original ground water status in the mountain will be changed. It will produce a significant impact on the stability of the slope. 3.2 Stability analysis and assessment According to the ‘‘Standard on grade determination and design safety of hydropower project of China (DL5180–2003)’’ and the working conditions of the slope, the design criteria of the slope are shown in Table 3. The slope is a kind of bedding slope. The leading edge of 4 shallow soft intercalations J1∼J4 is exposed on the slope surface, which form the bottom sliding plane. Fault F6 is the trailing edge cutting plane. The two group fissures are the upstream side cutting plane. The air face of the outcrop of J1∼J4 form the downstream cutting plane and leading edge air face. The analysis of the instability of the slope shows that: in 2D condition, the slope will slide along the bedding plane, in 3D condition, the slope will slide as a wedge which is formed by the bottom sliding plane and the upstream side cutting plane. The shear strength parameters determined by network simulation based on Monte Carlo principle and the laboratory tests of the undisturbed sample are shown in Table 4. From 2D analysis, in the nature condition and after slope excavation, the factors of safety of all sections are 0.8∼1.0, which is much less than the requirement.

Table 3. Safety criteria of the intake bedding slope.

Name

Grade of the structure affected by the slope

Intake bedding slope

Intake of water discharge structures, grade 1

Grade of the slope

Design cases

Load

Allowed FOC

Construction Operation

Construction: dead weight Operation: Deadweight + pore pressure

Construction:1.05 Operation:1.25

1

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Table 4. Shear strength parameters of the intake bedding slope. Rock

γ (kN/m3

c (MPa)

φ(◦ )

Soft intercaltion Jointed rock mass Fault plane Upstream side cutting plane

20.0 26.6 –

0.03 0.20 0.05

20 35 22

–

0.25

0

If the slope is to be treated to meet the required FOC of 1.25, the engineering quantity will be very huge. Actually, there were no slide happened in the construction of intake. This has proved that the 2D condition is not the real situation. It should be a typical 3D problem. By conducting the fitting calculation of the 3D geometric configuration of the soft intercalation J1∼J4, the 3D limit analysis method based on upper bound theorem of the mechanics of plasticity was applied. The FOC getting from 3D analysis has a significant increase compared with 2D results. From the above computation results, under the condition of rapid drawdown, the stability of slope cannot reach the requirement of the design stand. The drawdown of reservoir has a significant impact on the intake bedding slope. The sliding body with J2 as its bottom sliding plane presents the lowest FOC. It is the critical sliding plane. 3.3

(a) general layout

(b) typical section Figure 4. Engineering treatment of the large scale intake bedding slope. (a) general layout (b) typical section

Engineering measures

(1) Slope treatment design The local slopes of the intakes have already been reinforced. From the calculation, for keeping the stability of the whole intake bedding slope, 189,000 t sliding resistance force should be provided. The anchorage construction could not affect the construction of the lining of spillway, flood discharge tunnel, power tunnel and the impoundment of reservoir. The construction period is only 4∼6 months. So the anchor cable should cross intercalation J1∼J4 and not go through the intakes and the water carry tunnels. For speed up the construction, the existing investigation and construction adit should be utilized. According to the construction condition at site, the scheme of using huge anti-slide pile plus anchor cable was accepted by comparing several options. The anchor cables were arranged at the low elevation and the huge anti-slide piles were put in the interval of the tunnels (Figure 4). It can provide 218,500 t sliding resistance force, higher than the design requirement. The intakes of the water carry tunnels were arranged to distract from each other for avoiding the continuous cutting of slope base. In the area where tunnel and shaft

pass through the intercalation, the lining reinforcement was strengthened. This measure could provide the safety margin of the slide resistance. (2) Huge anti-slide pile The total number of the pile is 10. The dimension of the pile is: 20 m∼24 m in width, 5 m in thickness, 40 m∼80 m in height. The material of the pile is C30 concrete. The reinforcement is arranged by a 3 × 5 m unit. Each unit pile is connected by hoop reinforcement to make the pile take an integrate action. Each pile will provide 1.34∼1.59 t sliding resistance force and the total sliding resistance force will be 150,400 t. At the top of the pile, there is a small opening, with the area of 3 × 5 m. Vertical anchor bar with diameter of φ25, length of 9 m, interval of 2 m were arranged. (3) Pre-stressed anchor cable 227 anchor cables with the anchorage force of 3000 kN were arranged in zone I, II, III. In addition, there are 23 random anchor cables. 23 anchor cables were arranged in zone I, distributed in the area above EL. 1147.5 m. 51 anchor cables were arranged in zone II, distributed in the area of EL. 1147.5 m∼1056 m.

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Table 5. Results of the stability analysis of the intake bedding slope. FOC of reservoir rapid drawdown

Cases

FOC of nature slope

Upstream side sliding plane

J1

J2

J3

J4

J1

J2

J3

J4

J1

J2

J3

J4

1.440

1.323

1.417

1.401

1.384

1.308

1.375

1.371

1.195

1.134

1.207

1.200

c = 0.25 MPa

FOC during construction

Note: Upstream side sliding plane: φ = 0◦ , c = 0.25 MPa, 0.5 MPa. The coefficient of pore pressure is 0.2 when considering the case of rapid drawdown.

153 anchor cables were arranged in zone III, distributed in the area below EL. 1056 m, which is the elevation of the base plate of power tunnel. The function of the anchor cables in zone III is to lock the slope base. (4) Drainage of the slope By taking the reference of the experience of other project, the drainage will utilize the construction adits at different elevation. The U shape drainage hollow in the middle part of the lining has the size of 2 × 2.55 m. 3 drainage holes of φ50 were arranged in the radial direction. The hole is 5 m in the rock and the interval is 3 m. At the same time, all the investigation adit will be used as drainage holes if possible. (5) Stability analysis During the implementation, the 23 random anchor cables were not used due to the limitation of site condition. The grouting of 13 anchor cables hasn’t reached the design requirement. They are considered as failed anchor cable. The checking calculation shows that the minimum FOC of slope is 1.259 by considering the cancellation of random anchor cable and the failure of 13 anchor cables. (6) Technology for using reverse pilot shaft to release broken bits In the construction of the pile, several construction adit were arranged in the middle and lower part of the pile. It is to solve the problem of broken bit release and concrete pouring for the construction of the middle and lower part of the pile. In the construction, a pilot hole was drilled in advance. Then the hole was excavated to a vertical shaft with diameter of 2 m by using nacelle. The excavation of the pile starts from top to bottom. After the excavation, all the construction adits were backfilled by reinforced concrete to increase the shear strength and safety margin. 4

OPERATION OF THE SLOPES

The hongjiadu hydropower station started impoundment in April, 2004. The slope treatment has experienced the flood retaining in 2004 and the normal

operation in 2005, 2006, 2007. The maximum reservoir water level is 1134.90 m (2007) and the low water level is 1076 m (2006). The left abutment slope has already safely operated for 6 years. From the observation, the maximum relative horizontal inner displacement is 2.88 mm, which means the small movement between rock mass. There is no stress increase in anchor cable. The variation of the values of anchor cable gauge is small. The maximum attenuation is 75 kN. The opening of cracks is unchanged. All these indicate that the slope is in stable condition. It is considered that the existence of many Karst caverns makes the in-situ stress in the mountain cannot be increased too much. This could be one of the important reasons for the safely construction and operation of such a vertical high slope. The intake bedding slope has already been safely operated for 3 years. The observed movements are very small and the variation of the values of anchor cable gauge is small. The slope is in stable status.

5

CONCLUSIONS

The engineering treatment of the left abutment slope and the intake bedding slope of Hongjiadu Hydropower Project has experienced several flood seasons. The operation status is good. From the experience of the treatment, following conclusions can be drawn: 1. The scale, height and the gradient of the left abutment slope and the intake bedding slope of Hongjiadu Hydropower Project are rarely encountered both at home and abroad. To recognize the difficulties and to conduct related research works will provide a sound foundation for engineering treatment. 2. The left abutment slope is the inclined layered high slope. By using different analysis method and applying different parameters, the slope is stable in natural condition and after excavation. The applying of narrow plinth design, excavate

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the slope in step and the pre-stressed anchor cable have reduced excavation and slope treatment quantities significantly. It has guaranteed the safely operation of the 180 m concrete face rockfill dam. 3. The intake slope is a high bedding slope. The application of 3D analysis method has solved the problem of spatial effect that 2D analysis cannot take into consideration. It makes the slope stability analysis and reinforcing treatment of the slope in a reasonable and reliable base.

4. The key issue for solving the problems of engineering slopes include following aspect: comprehensive investigation, computation and stability analysis method, accurate stability analysis. The treatment of slope should not only take the reference of the existing engineering experience, but also apply innovation techniques. The treatment measure should consider the requirement of construction implementation, and also to install necessary observing instruments.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Typical harbor bank slopes in the Three Gorges reservoir: Landslide and collapse and their stability control Aijun Yao College of Architecture and Civil Engineering, Beijing University of Technology, Beijing, China

Chaoyang Heng Institute of Foundation Engineering, China Academy of Building Research, Beijing, China

Zaiming Zhang & Ruide Xiang College of Architecture and Civil Engineering, Beijing University of Technology, Beijing, China

ABSTRACT: The construction of the Three Gorges Project and the water impounding of the reservoir will widely and continuously affect the geological environment around the Three Gorges Area, and cause a large number of geological problems. With the rise of water level of the reservoir, there will be 340 public dock berths which belong to the 11 borough or county harbors, 80 town docks and even more exclusive use docks of enterprises to be submerged. So, lots of ports have to be constructed in geologically problematic area because of site constraints. Furthermore, the wide-range variation of water level and the storage loading may cause potential geological problems. The paper, taking Baima harbor district as a typical example, discussed these problems that affect the stability and deformation failure of harbor slopes. On the basis of physical modeling experiments and numerical analyses, stability control measures for bank slope were proposed. The numerical simulations indicate that the deformation failure mechanism of bank slope under loading and groundwater seepage is creep-shear failure with circular sliding plane in unconsolidated debris. The deformation failure process may be divided into two phases: creep deformation and shear failure. By back analyses, it is considered that the bank slope of Baima harbor district is in a limit equilibrium state under the action of groundwater seepage and variable loadings. With regard to measures taken to control the bank slope stability, the water level variation in the reservoir and the variable load in the harbor were focused on. 1

INTRODUCTION

The Three Gorges Project is a world renamed irrigation and electricity project. Its construction and impounding action of reservoir will seriously and continuously affect the geological environment around the Three Gorges Area. The construction or reconstruction of many harbors and docks which belong to 11 borough or county harbors will be affected. A large number of ports inevitably need to be constructed in geological problematic areas. Furthermore, groundwater effect due to wide-range variation of water level and loading action of storage may cause potential geological problems during the construction and operation of the harbor (Yueping Yin 2004, Faquan Wu 2002). Bank slope of Baima harbor district is typical bank slope in the three gorges reservoir. It has a 1000 tonnage goods berth, and design trafficability of 150,000 t/y. The designed slope on the dock is a slope with an angle of 1:2.3. Combining the construction of

ramp road and formation of construction site in harbor area, slope protection and slope stabilizing piles have been used to strengthen the bank slopes. The stability of these slopes is affected by many factors, seepage force induced by variation of water level and harbor variable loading are important factors. Because of the circulation adjustment and control of water-level and regional rainfall, the bank slopes suffer frequent, severe and large-scale variation of water level. Rock and soil masses are always under variable and strong seepage forces. It is also an important feature that the bank slope is different from others. Variable loadings include storage load, mechanical mobile load, ship load and nature load. The paper mainly considers uniformly distributed storage loads that act on the harbor surface and analyses the effects on stability of the bank slope. The research on the slope stability also takes the effect of variable water-levels. The work has important meaning on preventing landslide and collapse disasters.

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2

ENGINEERING GEOLOGICAL CONDITION OF TYPICAL HARBOR DISTRICT

Baima cargo port is located at Guijing village in the front district of new Fengjie county of the north bank of Changjiang river. Site topographical features are tectonic denudation and low mountainous ground. The level is high on the north and low on the south. Landform includes two gullies from south to north, two platforms and a gentle slope from east to west and a steep slope as shown in Figure 1. The geological structure of the site is complex. It is located at northwest part of Sichuan-Guizhou apophysis foldbelt, near the east edge of Sichuan foldbelt. The upper strata of the site are mainly composed of soil and blocky stone layers which include Quaternary al+pl alluvial-proluvial (Q 4 ), eluvial (Qel+dl3 ), clay stratiml fication and fill (Q 4 ). The underlying bedrock is limestone and argillaceous limestone (T2 b3 ) belonging to trias. The bedrock is highly jointed with a low-angle dip in harbor district. Interlayer foldings were extensively found in the bedrock layers. Outcrops at harbor district and vicinity have no large scale faults or joints. But the rock mass is highly jointed. Rock stratum has many folds and the dip of the rock stratum varies in a large range. Groundwater mainly supplied by atmospheric precipitation

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MECHANISM OF DEFORMATION FAILURE AT BANK SLOPES

Many factors affect the character of deformation and failure of the bank slopes, such as landform, geological formations and rock mass quality, fluctuation of groundwater and reservoir water level, human activity and so on. The slope of Baima harbor is an uncompacted debris slope and its stratum feature affecting the slope stability is mainly huge thickness eluvium (Qedl ). As shown from drilled cores, the eluvium can be divided into two layers. The upper layer is clay with corestone, whose average thickness is about 10 m and local thickness of 36 m; the lower layer is mainly blocky stone, whose thickness is nearly up to 40 m. The main engineering geological conditions which affect the slope stability are compactness of debris, shear strength, seepage force induced by fluctuation of reservoir water level and variable loadings (Transport Planning and Research Institute; Ministry of Communications, 2005).

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drains to the Changjiang river. Groundwater exists in joints of limestone and argillaceous limestone. The groundwater level changes with the fluctuation of the reservoir water-level (Changjiang Research Institue of Geo-Technique and Survey, 2000). Based on the figure of Chinese earthquake intensity distribution published in 1990, the earthquake intensity of the harbor district is VI. After reservoir impounding, the bank slope composed of uncompacted debris and bedrock layers may have to face reconstruction problem.

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Based on engineering geology conditions of the harbor, the paper considers variable loadings and seepages, using FLAC3D software to simulate the deformation and failure mechanism of the slope. As the results show, the top and toe of the slope which are located at uncompacted debris would incur large settlements and horizontal displacements (Figure 2). It indicates that under the action of load in the slope surface and groundwater seepage, the deformation and failure of the bank slope is creep with shear failure in uncompacted debris. The process of deformation and failure is divided into two phases: creep phase and shear failure phase, as shown in Figure 3. 3.2

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Figure 1. Plan figure of Baima harbor.

Siphon drainage test research

With the fluctuation of reservoir water level in the Three Gorges, the groundwater level and stability of the bank slope in Baima harbor will be changed. In

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Figure 2. Z-Displacement of the Baima harbor slope.

Figure 4.

Siphon drainage test device.

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Figure 3. The sketch map of deformation and failure in uncompacted debris bank slope.

order to obtain their response regularity and understand control technique of groundwater, a siphon drainage test in the bank slope was conducted, as shown in Figure 4. Experimental results indicate that the water level in the slope is very different from that outside the bank slope, and the surface of the bank slope slumps badly when siphon drainage is not used, as shown in Figure 5(a). When siphon drainage is used, the difference of water level between slope and outside bank slope reduces obviously, and groundwater

level inside the slope appeared double funnels around absorption wells (Figure 5(b)). Under the siphon condition, the surface of the bank slope slumps slightly. In Figure.5, number 0 is the water-yielding orifice, 1–4 are absorption wells, 5–13 are water gages. Comprehensive analyses indicate that following the siphon drainage in the bank slope, the appearance of double funnels around absorption wells obviously decreases the hydraulic gradient of groundwater in the slope and partly change its seepage direction. Siphon drainage helps to lower the groundwater level, hereby increases the stability of the bank slope.

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STABILITY CONTROL OF THE BANK SLOPE

Based on the above analyses, the bank slope stability of Baima harbor is affected by many factors. The falling speed of the reservoir water level and variable loading in the bank slope are the most important controlling factors. The sliding and collapse disaster can be efficiently prevented if those two factors are adjusted reasonably. For the bank slope of Baima harbor, the water level adjusting of the reservoir from 175.0−145.0 m at a duration 15 d, 20 d, 30 d and 40 d respectively. Geoslope software was used to analyze the relationship between dewatering speed and stability of the bank slope, as shown in Figure 6. From Figure 6 the following conclusions could be drawn: ① If the duration of change reservoir water level is longer, namely the water level change speed is

smaller, the safety factor of the bank slope is larger; ② if the duration is 15∼20 d when the water level drops from 175.0−145.0 m, the relationship between bank slope stability and reservoir level change speed is relatively large; if the duration is 30∼40 d, the result is opposite; ③ Only when the duration is longer than 30 d, namely the water level change speed is smaller than 1 m/d, the minimum safety factor of the slope is bigger than 1. The goods berth constructed on Baima harbor is a slope dock and the top surface load is 20 kPa. In 1.05 20kPa Safety Factor of bank slope

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(b)section 2-2 Figure 7. The relationship between stability of Baima harbor and distance of top surface loading.

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order to obtain the effects of variable loadings, the safety factors of the slope in section 1-1 and 2-2 are calculated while the distance from pack loading site to slope crest is 0 m, 10 m, 20 m respectively and dewatering speed is 1 m/d during normal impounding area at a running period (175.0−145.0 m). The results are shown in Table 1 and plotted in Figure 7. From Figure 7, the following conclusions could be drawn: ① With the increase of surface loading, the safety factor of the slope decreases; ② With the increase of the distance from surface loading to the slope crest, the safety factor of the slope increase; ③ The top surface load of Baima harbor should be no more than 20 kPa. To sum up, most bank slope in Baima harbor is at a state of limit equilibrium under seepage force and variable loading. In order to ensure the stability of the bank slope, the average dewatering speed of reservoir water level should be controlled less than 1 m/d. The proposed duration of reservoir operation level from 175 to 145 m before flooding should be 40 d, and the average velocity is 0.75 m/d. After fill of reservoir with floodwater, the water level drop should be at the same speed. Through the above study, it is known that the extent of variable loading at the dock could affect stability of the bank slope. The greater the distance from top surface load to the crest, the larger will the safety factor increase, the more top surface loading, the more decreases in safety factor. Hence, for the goods platform of Baima dock at the worst situation that the water level lowers from 175 m to 145 m or from 156 m to 135 m, the position of the goods yard on the harbor should be adjusted in time to the back area. The pileup value at slope crest should be kept at half of the largest pileup value. The loading in the top control area should be adjusted accordingly in time to avoid any dangerous situation. 5

CONCLUSIONS

The bank slope of Baima harbor is mainly composed of uncompacted debris. The main factors that

affect the slope stability are compactness of the debris, shear strength, seepage force induced by fluctuation of the reservoir water-level and variable loading. Bank slope deformation and failure mechanism is creep and secondly shear failure in uncompacted debris. The process of deformation and failure is divided into two phases, namely creep phase and shear failure phase. Siphon drainage technique is an effective type of drainage and pressure relief technique. It can reduce the hydraulic gradient in the slope at the lower part, or partly change the seepage direction. Through arbitrarily controlling the dewatering speed of the reservoir water level and variable loading at the top of bank slope, the slope stability can be efficiently controlled. Landslide and collapse disaster can be reduced and avoided.

ACKNOWLEDGEMENTS This paper was made possible through financial support of the technical traffic construction project for western part of China ‘‘Research on the technology of the prevention and control engineering for harbor geological hazards in the Three-Gorge Reservoirs’’. (Project No. 2003 328 220 47). REFERENCES Faquan, Wu. 2002. Landslide Disaster Treatment Project and many technique problems of 135 m water level slope brought by the Three Gorges Project. Geotechnical engineering world, 5(6):15–16. Yueping, Yin. 2004. Research on the grave geological casualty of migration-setting in the Three Gorges Reservoir. Beijing: Geological Publishing Company. The Changjiang Research Institue of Geo-technique and Survey. 2000. The engineering geological prospecting report of Shanmashan passenger station and Baima port belong to Wanzhou port authority Fengjie port station. Transport Planning and Research Institute; Ministry of Communications. 2005. The research report about prevention and cure technique of geological casualty in the Three Gorges Reservoir Port.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Weighting predisposing factors for shallow slides susceptibility assessment on the regional scale J.L. Zêzere, S.C. Oliveira, R.A.C. Garcia & E. Reis Geographical Research Centre, University of Lisbon, Portugal

ABSTRACT: The quality of landslide prediction models at the regional scale does not automatically increase with the increasing number of layers assumed to be landslide predisposing factors. Additionally, the significance of such layers as predisposing factors is frequently not evaluated. In this paper we apply a sensitivity analysis to a statistically-based landslide susceptibility model, performed for shallow translational slides occurred in a test site located north of Lisbon (Portugal). The model is applied individually to each layer (e.g. slope, aspect, transverse slope profile, geomorphology, lithology, superficial deposits and land use) and to different combinations of overlapped layers. The computation of success-rates and prediction-rates for such models allows concluding that the relationship between number of variables within the prediction model and the quality of predicted results is not linear, and it is possible to obtain an accurate landslide susceptibility map using a limited number of instability predisposing factors in the prediction model. 1

INTRODUCTION

The assessment of susceptibility associated to mass movements reveal in recent years significant improvements in indirect statistically-based methods (Guzzetti et al. 1999, 2005, Zêzere et al. 2004). Current Spatial Data Analysis (SDA) techniques allow the independent validation of results in post-processing operations, for prediction models based on both bivariate and multivariate statistical methods (Fabbri et al. 2002, Chung & Fabbri 2005). Therefore, validation is not anymore exclusively dependent on the occurrence of new instability events. Assessment of landslide susceptibility is always based on the assumption that future mass movements are more probable to occur in areas with conditions similar to those that originate slope instability in the past (Carrara et al. 1999). In this context, recent developments in Geographical Information Systems allow the development of models resulting from the spatial relationships between landslides and an increasing number of landslide predisposing factors. The quality of the landslide inventory and of the landslide predisposing factors database are of crucial importance for the quality of prediction results, independently on the statistical tools used for the modelling procedure. Usually, it is not easy to obtain systematic and detailed cartographic data that reflects directly the physical parameters involved in slope instability (e.g., shearing forces, soil shear strength, and spatial and temporal variation of pore water pressure). Therefore, it is

common to make recourse to the available cartography that may correlate with landslide distribution (e.g., terrain morphology, geology, land use). However, the significance of such themes as landslide predisposing factors is frequently not evaluated. The present study aims at evaluating quantitatively the relevance of different predisposing layers for shallow translational slides susceptibility assessment, using a spatial data set from a test site located north of Lisbon (Portugal). The main objectives of this study are: (i) to evaluate the relation between the number of variables within a statistic/probabilistic landslide susceptibility model and the quality of predicted results; and (ii) to assess the weight of each individual landslide predisposing factor by applying a sensitivity analysis, and to define the best variable combination by computing the corresponding success and prediction rates. 2

STUDY AREA

The sensitivity evaluation of landslide susceptibility models to the type and number of landslide predisposing factors was performed in a test site of 20 km2 located northward of Lisbon. The test site of Fanh˜oes-Tranc˜ao (Fig. 1) is characterized by the monocline geological structure, and the layers dip from 5◦ to 25◦ towards south and southeast. From the lithological point of view, the outcropping rocks are very heterogeneous and include conglomerate,

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composed of claystone, marl or volcanic tuff (Zêzere et al. 2005, 2007). 3

Figure 1. Study area location and distribution of shallow translational slides.

sandstone, claystone, marl, marly limestone, limestone, compacted basalt and volcanic tuff, dated from the Cretaceous to the Palaeogene. The monocline setting and the diversity of geological formations sustain a cuesta landscape, and the Fanhões-Trancão test site is located in the dip slope of the cuesta, i.e., a substructural slope defined by a general coincidence between the topographical surface and the dip of the strata. The geological setting also controls the fluvial system, and the most important rivers run in the same direction of the dip of strata. This is the case of the Fanhões river and of the Trancão river, located in the west and east side of the test site, respectively (Fig. 1). The Fanhões and Trancão valleys are the most relevant geomorphologic features within the study area because of their strong deep and the general steep slopes, although the altitude of the area does not exceed 335 m. The detailed geomorphologic mapping (scale 1:2000) of the study area allowed the identification and characterization of 100 shallow translational slides, resulting in 143,000 m2 of unstable area and corresponding to 0.7% of the total study area. Shallow translational slides within the test site affect almost exclusively colluvium deposits and have minor dimension (mean area, 1422 m2 ; mean volume, 357 m3 ). These landslides occur along planar rupture zones located usually from 0.5 to 1.5 m below the topographic surface. Shallow translational slides occurred in the study area during the last 4 decades have been triggered by intense rainfall periods ranging from 1 to 15 days (Zêzere & Rodrigues 2002, Zêzere et al. 2005). Intense rainfall is responsible by the rapid growth of pore pressure and by the loss of the apparent cohesion of thin soils, resulting in failure within the soil material or at the contact with the underlying impermeable bedrock

LANDSLIDE SUSCEPTIBILITY ASSESSMENT AND SENSITIVITY ANALYSIS

The susceptibility assessment to shallow translational slides occurrence is based on the favorability concept (Chung & Fabbri 1993, Fabbri et al. 2002). Within this concept, we assume that future probability of landslide occurrence can be quantitatively evaluated by bi-variate statistical relationships between the spatial distribution of past landslides and several types of independent spatial data sets that are understood as landslide predisposing factors. The landslide predisposing factors used in this study are the following: slope angle, slope aspect, transversal slope profile, lithology, superficial deposits, geomorphological units and land use (Table 1). More details about the data collection and database structure can be found in Reis et al. (2003) and Zêzere et al. (2004, 2007). Figure 2 summarizes the methodological procedures for the landslide susceptibility assessment and validation, as well as for the sensitivity analysis. Table 1. Predisposing factors used for shallow translational slides susceptibility assessment. Id

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Figure 2. Methodology for landslide susceptibility assessment and validation.

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The calculation of a priori and conditional probabilities was the first step in the cartographic data integration. These probabilities were estimated by overlapping the landslide map to those maps representing each landslide predisposing factor, and using the following equations: i. a priori probability of landslide occurrence Pps =



Sarea Tarea



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where Sarea = landslide area within the test site; Tarea = total area of test site. ii. a priori probability of occurrence of a class x belonging to the predisposing factor T Ppc =



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where Xarea = area of class x from the predisposing factor T . iii. conditional probability of landslide occurrence in the class x from the predisposing factor T  Cp = 1 − 1 −

1 Xarea

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where Sx = landslide area within class x from the predisposing factor T . Scores obtained by the application of equation (3) for each class of each considered landslide predisposing factor are interpreted as favorability values, or landslide susceptibility indicators. The probability of landslide occurrence given n landslide predisposing maps is obtained using the conditional probability integration rule through the next expression (Chung & Fabbri 1999, Zêzere et al. 2004): P =



The prediction model performance was assessed through the computation of success rate curves (Fabbri et al. 2002). These curves were constructed by crossing the distribution of the total set of landslides used to generate the susceptibility model with the prediction results, after sorting in descending order the susceptibility values corresponding to each pixel. Additionally, we compute the ‘Area Under the Curve’ (AUC) for each success rate curve, in order to quantify the model performance, and to allow the objective comparison among different success rate curves. The AUC values range from 0 to 1, and the quality of the prediction model increases with the increase of the AUC value. The landslide susceptibility model was applied, in a first step, to each landslide predisposing factor considered individually, and seven prediction rate curves were constructed. The AUC values corresponding to these curves were used to rank variables. In the next step, the landslide susceptibility model was performed using groups of 2, 3, 4, 5, 6 and 7 predisposing factors that were selected according to the above mentioned AUC-based ranking. Lastly, the models corresponding to the best variable combinations are tested by the computation of prediction rate curves (Chung & Fabbri 2005), based on the temporal partition of the original landslide data base in two parts: prediction set (used to develop the landslide prediction model) and validation set (used for the independent validation of the predicted results).

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RESULTS AND DISCUSSION

As previously referred, one of the main goals of the sensitivity analysis was to assess the weight of different landslide predisposing factors within a statisticallybased landslide susceptibility model. Therefore, Figure 3 and Table 2 illustrate, respectively, the success rate curves and the corresponding AUC obtained by

  PpT 1 ∗ PpT 2 ∗ · · · ∗ PpTn CpT 1 ∗ CpT 2 ∗ · · · ∗ CpTn PpslideTn−1 ∗ (T 1 ∗ T 2 ∗ · · · ∗ Tn)

(4)

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where T 1, T 2, . . . , Tn = set of landslide predisposing factors; Pp = apriori probability of occurrence of a class x from the predisposing factor T ; Ppslide = apriori probability of landslide occurrence; Cp = conditional probability of occurrence of a landslide in the class x from the predisposing factor T . The equation (4) was applied on a 5 m grid cell structure that is reasonably conform to the detail and resolution of the cartographic database (Zêzere et al. 2007). The obtained results (a score for each pixel of the study area) range between 0 and 1 and measure the susceptibility (or spatial probability) of occurrence of future shallow translational slides.

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Figure 3. Success rate curves corresponding to individual landslide predisposing factors.

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Table 2. Hierarchy of predisposing factors for shallow translational slides occurrence, according to success rate curves and AUC (Area Under the Curve).

Table 3. Area under the curve (AUC) of success rate curves corresponding to landslide susceptibility models obtained using from 2 to 7 predisposing factors.

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Figure 4. Success rate curves corresponding to landslide susceptibility models obtained using from 2 to 7 predisposing factors (2 variables = variable Id: A + F; 3 variables = variable Id: A + F + B; 4 variables = variable Id: A + F + B + E; 5 variables = variable Id: A + F + B + E + D; 6 variables = variable Id: A + F + B + E + D + C; 7 variables = total variable set).

applying the predictive model using separately each one of the slope instability predisposing factors. The obtained results demonstrate that the considered independent variables do not correlate in the same way with landslide distribution (AUC ranging from 0.67 to 0.80). Moreover, according to AUC records, ‘slope angle’ and ‘geomorphological units’ are the variables more able to predict the future occurrence of shallow translational slides. The variable ranking summarized in Table 2 was used to define the conjugation of landslide predisposing factors that support the next landslide susceptibility models (i.e. models running with 2 variables, 3 variables, 4 variables, 5 variables, 6 variables and 7 variables). Figure 4 illustrates the success rate curves of these landslide prediction models, and Table 3 summarizes the corresponding ‘Area Under the Curve’. Additionally, Figure 5 illustrates the variation on models prediction capability according to the number of variables within the model, for some standard areas of

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Figure 5. Variation on the predictive power of landslide susceptibility models according to the number of variables in the model (higher susceptibility scores corresponding to 5%, 10%, 20%, 30% and 40% of the total area were selected for comparison).

maximum landslide susceptibility (corresponding to 5%, 10%, 20%, 30% and 40% of the total study area). The analysis of Figures 4 and 5 and Table 3 allow concluding that: i. the quality of landslide prediction models demonstrates a slight tendency to improve with the increment on the number of variables within the model, as it is shown by the AUC values (Table 3). This is particularly true when we consider the top 5% and 10% of the total area classified as more susceptible to slope instability (Fig. 5); ii. if we consider the 30% and 40% of the total area classified as more susceptible, the predicted results tend to stabilize (Fig. 5), with maximum variations of 4% on success rate curves. These features demonstrate the low sensitivity of landslide prediction models to the increasing number of landslide predisposing factors; iii. the introduction of more variables in the landslide prediction model, does not generate necessarily better results in success rates. For instance, the model produced using 4 variables generate better prediction results, when compared with those

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obtained using 5 and 6 variables, as it is confirmed by the corresponding AUC values (Table 3). iv. it is possible to predict the future spatial occurrence of shallow translational slides in the study area with very satisfactory results, based on a restricted number of landslide predisposing factors. For instance, the susceptibility model produced with 4 variables (slope angle, geomorphological units, slope aspect and superficial deposits) shows results very similar to those obtained using the total set of variables, as it is confirmed by the shape of success rate curves (Fig. 4) and the corresponding AUC values (Table 3). Moreover, results obtained with the above mentioned 4 landslide predisposing factors are even better than those obtained with the complete set of variables when we isolate for analysis the 20% and 30% of area defined as more susceptible (Fig. 5). Figure 6 and Figure 7 show the susceptibility maps to shallow translational slides occurrence in the study area, based on 4 predisposing factors (slope angle, geomorphological units, slope aspect and superficial deposits) and on the total set of predisposing factors, respectively. In order to allow map comparison, we define 6 landslide susceptibility classes, which were generated in the same way for both maps as % of the total area, after sorting in descending order

Figure 7. Shallow translational slides susceptibility map based on the complete (7) set of predisposing factors. Susceptibility classes were defined as % of the total area, after sorting in descending order the susceptibility values corresponding to each pixel. STS = shallow translational slides. Table 4. Estimated probability (%) for landslide susceptibility classes represented in Figures 6 and 7, per 1% of total area. Top 5 5–10 10–20 20–30 30–40 40–100 Figure 6 5.9 Figure 7 7.0

Figure 6. Shallow translational slides susceptibility map based on 4 predisposing factors (slope angle, geomorphological units, slope aspect and superficial deposits). Susceptibility classes were defined as % of the total area, after sorting in descending order the susceptibility values corresponding to each pixel. STS = shallow translational slides.

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the susceptibility values corresponding to each pixel. Table 4 summarizes the spatial probabilities computed for each 1% of the total study area, for landslide susceptibility classes represented in Figures 6 and 7. The obtained results are very similar, despite the higher probability corresponding to the first susceptibility class in the map based on the complete set of predisposing factors (Table 4). The comparison between Figure 6 and Figure 7 allows concluding that the obtained spatial patterns of landslide susceptibility are very similar. This fact is more evident for the high susceptibility classes, which is in accordance with the success rate data. The main difference found between landslide susceptibility maps is the homogeneity level of representation of susceptibility classes, which is higher in the prediction model reported in Figure 6. This

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landslide predisposing factors are very similar (Fig. 8) allowing to conclude that prediction performance is equivalent for both models. This fact is also confirmed by the corresponding AUC: 0.809 for the model based on 4 variables; 0.808 for the model based on 7 variables.

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CONCLUSION

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100

Area classified as susceptible (descending order %)

Figure 8. Prediction rate curves of landslide susceptibility models obtained using 4 predisposing factors (slope angle, geomorphological units, slope aspect and superficial deposits) and 7 predisposing factors (corresponding success rate curves are also plotted for comparison).

difference is explained by the different number of ‘unique conditions’ (particular terrain combinations) presented within landslide prediction models. These ‘unique conditions’ result from the crossing among the set of predisposing maps considered in each model. Therefore, the prediction model based on 7 variables generates 15,636 unique conditions that contribute to a less regular pattern of susceptibility distribution (Fig. 7). On the other hand, the prediction model supported by 4 variables generates only 234 unique conditions, which justify the higher spatial homogeneity of susceptibility classes (Fig. 6). Finally, the prediction capability of landslide susceptibility models based on 4 and 7 landslide predisposing factors was assessed through the construction of prediction rate curves. These curves are shown in Figure 8 and were constructed by dividing the original landslide data base in two groups using a temporal criterion: the landslide prediction group (46 shallow translational slides occurred prior to 1980 that were used to develop a new landslide prediction model); and the landslide validation group (54 shallow translational slides occurred after 1980 that were used for the independent validation of the predicted results). As it was expected, both prediction rate curves are below the corresponding success rate, which were produced using the same landslide data set for modeling and validation. Anyway, the prediction rate curves show fairly acceptable results for both models performed with 4 and 7 variables. For instance, 47% to 50% of landslides occurred after 1980 are within the 10% of the total area classified as more susceptible (Fig. 8). These features grow up to 72% if we consider the 20% of the total area classified as more susceptible by both prediction models (Fig. 8). The prediction rate curves corresponding to landslide susceptibility models produced using 4 and 7

The relationship between the number of predisposing factors within a statistically-based landslide prediction model and the quality of predicted results is not linear. The results obtained here prove that the introduction of additional variables into a landslide prediction model does not generate necessarily better success rates. Moreover, as it was shown in the present analysis, it is possible to obtain an accurate landslide susceptibility map using a limited number of instability predisposing factors in the prediction model (e.g. slope angle, geomorphological units, slope aspect and superficial deposits). However, these ‘‘key variables’’ cannot be extrapolated for other types of landslides within the test site or for other study areas. Therefore, a prudent approach to landslide susceptibility assessment implies, on a first step, the use of a set of coherent and logical landslide predisposing factors as large as possible. On a second step, the landslide predictive models can be simplified, with minor losses, by removing those variables that prove to be irrelevant to the prediction performance of the susceptibility model. ACKNOWLEDGEMENTS The research of R.A.C. Garcia and S.C. Oliveira was supported by the Portuguese Foundation for Science and Technology of the Portuguese Ministry of Science, Technology and Higher Education. REFERENCES Carrara, A., Guzzetti, F., Cardinali, M. & Reichenback, P. 1999. Use of GIS technology in the prediction and monitoring of landslide hazard. Natural Hazards 20: 117–135. Chung, C.F. & Fabbri, A. 1993. The representation of geoscience information for data integration. Nonrenewable Resources 2 (2): 122–138. Chung, C.F. & Fabbri, A. 1999. Probabilistic prediction models for landslide hazard mapping. Photogrammetric Engineering and Remote Sensing 65 (12): 1389–1399. Chung, C.F. & Fabbri, A. 2005. Systematic procedures of landslide-hazard mapping for risk assessment using spatial prediction models. In Glade, T., Anderson, M.G., Crozier, M.J. (eds.), Landslide Hazard and Risk: 139–174. Chichester: Wiley.

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Fabbri, A., Chung, C.F., Napolitano, P., Remondo, J. & Zêzere, J.L. 2002. Prediction rate functions of landslide susceptibility applied in the Iberian Peninsula. In Brebbia C.A. (ed.), Risk Analysis III, Series. Management Information Systems vol. 5: 703–718. Southampton, Boston: WIT Press. Guzzetti, F., Carrara, A., Cardinali, M. & Reichenbach, P. 1999. Landslide hazard evaluation: a review of current techniques and their application in a multi-scale study, Central Italy. Geomorphology 31: 181–216. Guzzetti, F., Reichenbach, P., Cardinali, M., Galli, M. & Ardizzone, F. 2005. Probabilistic landslide hazard assessment at the basin scale. Geomorphology 72: 272–299. Reis, E., Zêzere, J.L., Vieira, G.T. & Rodrigues, M.L. 2003. Integração de dados espaciais em SIG para avaliação da susceptibilidade à ocorrência de deslizamentos. Finisterra XXXVIII (76): 3–34. Zêzere, J.L., Garcia, R.A.C., Oliveira, S.C. & Reis, E. 2007. Probabilistic landslide risk analysis considering

direct costs in the area north of Lisbon (Portugal). Geomorphology. DOI: 10.1016/j.geomorph.2006.10.040 (in press). Zêzere, J.L., Reis, E., Garcia, R., Oliveira, S., Rodrigues, M.L., Vieira, G. & Ferreira, A.B. 2004. Integration of spatial and temporal data for the definition of different landslide hazard scenarios in the area north of Lisbon (Portugal). Natural Hazards and Earth System Sciences 4: 133–146. Zêzere, J.L. & Rodrigues, M.L. 2002. Rainfall Thresholds for Landsliding in Lisbon Area (Portugal). In: Rybar, Stemberk, Wagner (eds.), Landslides: 333–338. Lisse: Swets & Zeitlinger. Zêzere, J.L., Trigo, R. & Trigo, I. 2005. Shallow and deep landslides induced by rainfall in the Lisbon region (Portugal): assessment of relationships with the North Atlantic Oscillation. Natural Hazards and Earth System Sciences 5: 331–344.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Analyses of mechanism of landslides in Tongchuan-Huangling highway L. Zhang & H. He Shaanxi Provincial Highway Survey and Design Institute, China

ABSTRACT: On the basis of researching geology background of Tong-Huang highway, this paper analyzed the characteristics of distribution and occurrence of landslides around the highway, and discussed the disaster mechanism and the characteristic of deformation and failure. The conclusions drawn from this paper may be a lesson for highway building in future.

1

INTRODUCTION

Tongchuan-Huangling Class I Highway (hereinafter as ‘‘THH’’) is located in the Weibei Mountain Area of the northern edge of Guanzhong Basin. It is also the important links between Guanzhong Plain and Loess Plateau in the northern Shaanxi. This highway attracts more attention because of its frequent occurrence of landslide due to the special geography and geological conditions. Large quantities of works on landslide treatment have been done. Currently, THH has already put into operation for 5 years. It shows that treatment of landslide has acquired the better effects. To analyze the distribution law and formation mechanism of slide will play a very important role in the future treatment of landslide.

2 2.1

GEOLOGICAL CONDITION ALONG ALIGNMENT

the top of hills, and the bedrock exposed outside in the middle and lower parts of valley. The width of hill generally is several hundreds meters. The hills incline towards the valley. The undulation between hills ranges 15–30 meters. The valley stepped zone mainly distributes in Wangjiahe River, Qishuihe River and Xihe River, and shapes a widen U-Valley with unsymmetrical joints. The eastern bank is wider, and the stepped land distributes continuously. The western bank is much steeper; the deposit of stepped land mostly concentrates in some place. The bedrock mainly concentrates nearby Haozhuangliang which forms the body of watershed. The exposing rock includes Mesozoic sand rock and mud rock. The height of hill averagely ranges from 1550 to 1600 meters. The rushing ditch is deep with a slope of 30∼50◦ . The height difference ranges from 150 to 200 meters. 2.2 Features of stratum

Features of terrain and physiognomy

The involved area is lower in the both sides, and higher in the middle. The northern and southern parts are located in the gully of Loess Plateau with sea level of 800–850 meters; the middle part passes through the rock mountain at Haozhuangliang with sea level of 1612.6 meters. The difference between gully and plateau averagely ranges 100 to 150 meters; the maximum difference is around 250 meters. There are many gullies in the watershed section together with crashed terrain. The project involved area is classified as loess plateau-hills, river valley and bedrock mountain area. The loess plateau & hills mainly distributes in both banks of Wangjiahe River and Xihe River located in the southern edge of Loess Plateau in the Northern Shaanxi. It has a thin loess overburden lay only in

The upper overburden lay in the project area is classified as the Quaternary loose deposits, including loess, river alluvial deposit and relic slop deposit with a thickness of several meters to hundred meters. The loess mostly distributes in the area of loess hills. In the valley and slope zone, the thickness of loess is 30–40 meters, over 100 meters in the top of hills. The erosion and flood alluvial deposit mainly distributes in the both sides of river valley and huge donga. The upper part is powder clay, and the lower part is the rushed stone and cobble stone layer. The overburden loess layer covers the higher stepped land surface with a thickness of several meters and more. The relic slope deposit mainly locates in the valley and slope zone with exposing rock. Its characteristics are powder clay and more block stone. There are some large block stone in some section with a thickness of 10–20 meters. The thickness is changing to 3–5 metes in the tope of slope. The underlying stratum is Paleozoic and Mesozoic bedrock

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including Permian and Triassic sandstone, mudstone and their inner-bed consisting coal formation, the oil shale and carbonaceous shale, etc. There are a little quantity of Ordovician system and lime system limestone and sandy mud rock.

3

Reservoir landslide belong to this category with enormous scope and difficult treatment. Xihe Reservoir has a thickness of 23–45 m; the slip surface is the thin mud rock among sandy mud rock category. It shows that II and III level slip surface are formed along the different level under the effect of squeezing and rubbing.

FEATURES OF LANDSLIDES 4

3.1 Category of landslides

TRIGGER OF LANDSLIDES

There totally are 27 ancient landslides in various types along THH (shows in Table 1, Zhao Zhisheng, 2002). In accordance with landslide composition and type of landslide interface, there totally are five categories of landslide along the alignment (Zhao Zhisheng, 1998).

There are too many landslides along THH. On the whole, there are the following reasons:

(1) Loess-bedrock Interface landslides The slip surface generates along the interface between loess and bedrock. It is the very common kind of landslides in the loess hill area. They concentrate in the upper range of Wangjiahe River valley. Such as, Fengjiahe Landslide, distributing at the front edge of loess, belongs to the multi level drawing ancient landslides with a loess thickness of 30–42 meters. It slowly slides along the interface of bedrock at the bottom with a width of 0.7 m. The Yansai Lake is formed as the ancient landslides blocks up the Saimiaogou (See Figure 1).

The stratum mainly distributes along Tonghuang Highway along the layers. Rock features are the lamella-medium thick mud rock, sandy rock and mud and calcium cementation. The structure of rock is much loose, and has weak efflorescence resistance. Landslides slip along the efflorescence layer on the top or along the soft interlayer (Figure 2). As a result, it forms the regional easily slip stratum. The soil of slip zone is the efflorescent mass with the colors of dust, celadon and rust yellow. Somewhere there is crushed stone inside soil, or the soil becomes the mud soil in saturation.

(2) Loess and bedrock Mixed Landslides Under effects of efflorescence and structure crash to the surface of bedrock, the slip faces appear inside the bedrock. Taking Wangjiahe Landslide as example, the back slip face is along the interface between loess and bedrock; the middle part cuts cross the bedrock, the loess and bedrock forms the landslide body at the front edge. (3) Alluvial Deposit Landslides The multi level erosion and embedded terrace have formed due to river under the effects of lifting from structure. The alluvial deposit distributes along Banjiegou, Shiyougou and two banks of Xihe River. Various scopes of landslides have been formed along the interface between bedrock and terrace. (4) Relic Slope Deposit Landslides At the back edge or the bank of narrow valley of terrace, the mixed deposit of slope deposit and relic deposit are formed mainly comprising of crushed stone and silt clay. The slide much easily generates along the bedrock interface as rich under ground water. The most landslides along Banjiegou and Yijun Branch line are this kind of landslides. (5) Bedrock Landslides In the exposing bedrock area, the internal soft layer forms the slip surface; the landslides and slip bed are bedrock. Both Jinsuoguan landslide and Xihe

4.1

4.2

Soft rock and soil

Air face

Firstly, the rock erosion face, interface of rock layers and interface covered by the various Cenozoic deposits incline by a certain angle because the alignment is located in the lifting geological structure. Secondly, banks and slopes form the cliffy air face due to the water erosion. These provide many advantages of air face for slipping along these interfaces. 4.3

Underground water

The involved area has a humid climate and strong rainfalls because of its special geography and terrain. The data on landslide surveyed along the alignment shows that substance of most slip belt is wet and is in saturated condition. Based on the indoor experiment, reversed calculation result and large size cut, the value of C generally ranges from 5 KPa to 15 KPa, and the value of  ranges from 5◦ to 13◦ , and the strength is only 40% ( Zhao Zhisheng, 1996) of natural condition. 4.4

Structural crash belt

Though the internal structure activity is very weak, and the trace of structure is not very distinct, the terrace deformation caused by the structure activity

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Table 1. Statistics on features of ancient landslide along the alignment. Distribution of landslide Landslide group at Wangjia river valley

No. 1 2 3 4 5 6

Landslide group at Qishui river valley

7

Landslide group at Banjiegou

9

8

10 11 12

Valley at Shiyou gou

13 14

Landslide group at Xihe river valley

15 16 17 18 19 20 21 22

Landslide groups at the Yijun branch line

23 24 25 26 27

Name

Category

Scope

Thickness

Remarks

Chuankou landslide 2# Mail Box landslide Wangjiahe landslide Bianjiahe landslide Langjiahe landslide Fengjiahe landslide

Loess-bedrock interface landslide Loess-bedrock interface landslide Loess-bedrock interface landslide Loess-bedrock interface landslide Loess-bedrock interface landslide Loess-bedrock interface landslide

Enormous

Thick

Enormous

Thick

Enormous

Thick

Large

Medium

Enormous

Thick

Distributing at the edge of plateau at the banks of Wangjia river valley mainly is the mixed landslide interface and Thickness of 20∼40 meters, with a loess-bed rock and the scope is very large.

Enormous

Thick

Maliantan landslide Jinsuoguan landslide

Loess-bedrock interface landslide Bed rock landslide

Enormous

Thick

Enormous

Enormous thick

Hanjiamao landslide Laoyaozi landslide (I) Laoyaozi landslide (II) Laoyaozi landslide (III)

Residual slope deposit landslide Loess-bedrock interface landslide Relic slope deposit landslide Relic slope deposit landslide

Large

Thick

Enormous

Shallow

Medium

Shallow

Enormous

Thick

Xitai landslide Taishang landslide

Relic slope deposit landslide Alumina deposit landslide

Medium

Medium

Enormous

Thick

Qianjiahe landslide Xihe landslide Shuikuba landslide Liujiaping landslide Houqingping landslide Nangou landslide Caiyuanzi landslide Muguomao landslide

Alumina deposit landslide Bed rock landslide Alumina deposit landslide

Large

Medium

Enormous Enormous

Enormous thick Thick

Enormous

Thick

Large

Medium

Large

Medium

Alumina deposit landslide Alumina deposit landslide

Large

Medium

Enormous

Thick

Alumina deposit landslide Alumina deposit landslide Alumina deposit landslide

Large

Medium

Enormous

Medium

Large

Medium

Alumina deposit landslide

Medium

Medium

Loess-bedrock interface landslide

Large

Medium

Huitouwan landslide Yanqian landslide Landslide at K3+700∼ +840 Landslide at K3+990∼ K4+060 Landslide at K4+330∼ +430

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Distributing in the slope of the east bank of Qishui river, and the scope is very large. Concentrating at the watershed section, the main composition is the residual slope deposit together with loess and flood alluvium. Distributing at the head and opening of trench, and the scope has much difference among landslides.

It extensively distributes along the Xihe river, and the composition mainly is flood alluvium, some part consists of loess and residual slope deposit with medium thickness and thickness. The scope is large.

Ending at top of Yijunliag hill after passing the east bank of Xihe river; the lower part is the flood alluvial deposit and residual slope deposit; and the loess covers on the top of hill. There are many types and the scope is very large.

5.2

Features of deformation during construction

During construction of THH, the measurements of supporting works or realignment were taken at the 4 ancient landslides in order to keep the stabilization of landslide. Other 23 landslides have the resurrected deformation occurred. The deformation has the following laws: 1. The excavation is the major reason caused the resurrected deformation. Among those 23 resurrected deformation, there is one caused by the waste deposit and filling, and the other 22 deformation caused by the excavation of cut. 2. Interfaces are cliffy at the back edge, slow in the middle part and flat at the front edge. There is a little reversed tilt at the front edge of some landslide. 3. Features of ancient landslides resurrect includes: Entirely creeping, intension and crack at the bank edge, and pulling development. 4. Deformation of landslides is categorized to three kinds: integrated slip, multi level slip and loose slip. It is necessary to take different treatment measurements in accordance with different characteristic of each kind of deformation.

Figure 1. Cross section of Fengjiahe landslide.

Figure 2. The soft interlayer.

control formation and distribution of some landslides as well as the scope. Taking Maliantan Landslide as an example, the fault belongs to the normal one, which has a higher angle and inclines toward northwest because of the force from northeast fault at the south edge. The landslide is located on the top part of fault. Under function of fault and fold structure, the rock body is crushed, and the underground water corridor forms, and control the formation and development of landslides.

6 6.1

Changing the geometric form of slope

It refers to reload the deposit in the area of landslides and increase the matter deterring occurrences of landslides. Those measures are simple and easy to implement, and have better effect. It is helpful to entirely keep the small and medium landslides stable. 6.2

5

LANDSLIDES TREATMENT MEASUREMENTS

Supporting works

It covers the following supporting structures: retaining wall, anchored wall, slide-resistant pile, anchorage works, etc. The treatments of landslides focus on choosing the proper supporting work based on the feature and natural condition of landslides.

DEFORMATION FEATURES OF LANDSLIDES

5.1 Deformation feature of ancient landslides The multi level and drawing slip is the dominant type. Generally, the interfaces are very slow, and the major interface has an angle of 5∼10◦ . Along the vertical, the main slip section is very long, and slip resistance section and drawing section is relative short. The soil erosion is serious at the front edge due to the river erosion. Before THH built, the creeping deformation had been occurring in 20 years at ancient landslides of Chuankou and 2# Mailbox due to the human’s construction activities and rainstorm, but the other ancient landslides are keeping stable.

6.3

Strength of soil of slip zone

Strengthening the soil of slip zone refers to the physical strengthening method through baking, freezing and grouting in order to improving the strength of soil of slip zone so as to stabilize the landslides. Those measures have played a very obvious role in treating small and shallow landslides. However, there are many technical problems to popularize these treatment technologies due to difficult implementation, higher cost and the reversed change of strengthening result, etc.

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Table 2. Category of landslides treatment measurements. Category

Effect mechanism

Changing the geometric form of slope

Changing the reinforcement distribution to keep balance

Support and retaining work

To prevent the deformation by using active and passive variable force structure

Strengthening slip belt soil

To improve strength of slip belt soil so as to meet the needs

Drainage work

To reduce the affect to stabilization of slope from under ground water

6.4

Secondary class

(1) Cleaning and reducing load (2) deposit filling and compaction Light support and retaining work Large support and retaining work

(1) Retaining wall (2) dry mortar rubber masonry (3) Small drilling pile (1) Resistant slip pile (2) Resistant slip chain (1) Chemical grouting (2) Block stone pile (3) ash & soil pile

Cut-off flow of surface water Drainage under ground

Drainage works

Underground water has significant roles in the process of generation and development of landslides. The drainage works include underground water drainage and surface water drainage. Comparing with various landslides treatment methods, drainage work is extensively undertaken due to its advantages of easier implementation and lower cost. Referring to large landslides, the water drainage at the deeper part is the necessary measurement.

7

Measurements

THINKING OF TREATMENT

The principle of giving priority to the prevention and combining prevention together with treatment should be insisted in the landslides treatment during highway construction. More attention should be focused on geological survey at the stage of alignment selection. No realization on landslides during the design stage leaded to the increase of investment, delay of the construction and deformation of slope during the construction period. It is necessary to initially analyze and

(1) Cut-off ditch (2) Open channel for drainage (1) supporting sewage ditch (2) Weep hole (3) Seepage blind ditch (4) Upward inclining

identify the harm caused by the landslides. To acquire overall information on rock characteristics, geometrical feature and layout of underground water, analyzing the cause of formation and factor on influence are much important. 7.1 Sensitive analysis Each factor will bring different degrees of impacts to different landslides. During the process of landslides growth, some factors will be much crucial. The measurement undertaken aiming at these crucial factors will ensure to keep landslides more stable, and make the work more economic. Those factors were considered as the sensitive factors of landslides stabilization. During the construction of treatment works, three sensitive factors causing the landslide were found: soft structure surface, higher and cliffy air face (Figure 3) and rich underground water (Figure 4). The aimed treatment measurements have been undertaken based on analysis outcome on sensitive factors of various landslides. Such as, the supporting and strengthening methods aim at the landslide caused due to the

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direction, and classify it as multi levels along the transverse direction. Since the terrain, geological condition and alignment layout always vary along the vertical direction, the handling works should be implemented section by section. Since various protection measurement on the transverse direction have a limitation of height, the protection works could be constructed level by levels to the higher landslides. Based on the location of each part, the treatment works can vividly be described as three types: ‘‘sharpening the top of the slope’’, ‘‘stabilizing the toe of slope’’ and ‘‘corseting the middle part of slope’’. Figure 3. The higher cliffy air face.

7.3

Bright future of anchorage works

The anchorage technology becomes the more popular treatment of landslides due to its advantages of little interruption to the slope body, easier implementation, lower cost and construction by stages. There are 11 landslides to be treated by using anchor and bar. Of which, the pre-stressing anchor has a total length of 42300 meters, and pre-stressing anchor bar is 26100 meters long. Concerning the earth landslide, the force effect of anchor should consider the following factors: 7.3.1 Strengthening effect of anchor to the slope Based on the calculation on current stabilization, the direction of downward force is the same with the slip face of last block. Since the angle between anchor’s force direction and slip downward force ranges from 20◦ to 60◦ , even to 90◦ , the relationship between prestressing force to the anchor and residual downward force from landslides shows as the following: Pt =

E sin θ · tan φ + cos θ

θ – angle between interface and anchor rope; φ – Inner friction angle of slip belt. Bringing the design parameters in common use into the above formula, gaining:

Figure 4. The rich underground water.

soft structure surface; and the supporting and retaining methods applied to the landslides caused by the higher air face together with slowing down the slope degree; drainage works were constructed in order to treat the landslides caused by the rich underground water. The sensitive analysis establishes accurate evidences for working out reliable and economic treatment program. 7.2 Treatment works by stages Concerning treatment for large landslides, it must be abide by to divide it into section along the vertical

Pt = (0.85 ∼ 1.15)E It is necessary to introduce the reduction coefficient during design stage because the above calculation exist much deviation to the loose soil landslide, viz: Pt =

E k1 sin θ · tan φ + k2 cos θ

The value of k 1 and k2 should be varied based on feature of landslides. When the landslides is clay or crushed stone and earth, the experienced value of k1 and k2 range from 0.3 to 0.7.

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leaded by the closed effect in the face of slope makes against the stabilization of slope. The framework gird has a better result because 8–16 anchors inside the framework gird prevents the disadvantage factors existing between ribbed panel and plate wall. 8 Figure 5. The outside anchored works.

7.3.2 Compression deformation of earth All landslides along THH are loess, clay and crushed stone soil. Modulus of compression of Loess of Q2 and Q1 is large, and modulus of compression of Q3 Loess, clay and crushed stone soil generally change from 5 Mpa to 15 Mpa. Under effect of pre-stressing force, the compression modulus from earth is much large, and the order of magnitude is as same as magnitude of elongation of strand. That is means the compression deformation from earth can lead to the loss of most pre-stressing force. In order to compensate the loss of pre-stressing force, it is necessary to construct the second stretching workmanship to the anchor. 7.3.3 Types of external anchorage work The outside anchored work (Figure 5) includes cushion pier, ribbed panel, concrete plate wall, framework girds, and other types. The effects show that the cushion pier is suitable to construct the anchored bar with small pre-stressing force in the area with difficult geography due to easier layout. When any one of anchor doesn’t work, the entire ribbed panel will become unstable because there are only two anchors inside each piece of ribbed panel, and the coordinative deformation is much poorer. Referring to concrete plate wall, the rise of underground water level

CONCLUSION

To reinforce the loess landslides is costly and hardly as special geographical features and characteristic of stratum. So the geological survey must be conducted carefully to the landslides before project construction, which potentially causes heavy impact to the projects. Three mainly sensitive factors causing the landslide were found along the alignment: soft structure surface, higher and cliffy air face and rich underground water. The measurements of reinforced landslide shall against the sensitive factors caused landslides so as the project economically. The measurement of unloading the large quantity of earthwork should apply carefully in treat the loess landslides. REFERENCES Study and comprehensive Analysis on Landslide along TongHuang Highway (R) by Zhao Zhisheng in March 2002. Study on Landslide Mechanism of Landslides along TongHuang Class I Highway (C) by Zhao Zhisheng from the Thesis Collection of Forum on Lanzhou Landslide Debris Flow Held in Lanzhou, Gansu, published by Lanzhou University Press in 1998. Study on the Easy Slip Stratum in Tongchuan City in Geographic Collection of China Highway (C) by Zhao Zhisheng in 1996, published by Shaanxi Science Technology Press.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Treatments of Loess-Bedrock landslides at Chuankou in Tongchuan-Huangling expressway J.B. Zhao Shaanxi Provincial Expressway Construction Group Company, P.R. China

ABSTRACT: The landslide in Chuankou is classified as typical Loess-bedrock landslides. This article analyzes its formation mechanism and distortion feature, and proposes all round treatment measures, such as, slip resistance piles after identifying the sensitive factors affecting landslides. Finally, the finite element method is adopted in the article to analyze and calculate as well as recheck the deformation data. The landslides treatment has a better result. 1

INTRODUCTION

Tongchuan-Huangling Expressway (called hereinafter ‘‘THE’’) is located in the transition belt between Guanzhong Plain and Weibei Mountainous Area in Shaanxi PRC. The construction area is the where landslides occurring more often as special geomorphology, topography and hydrological condition. During highway construction and operation period, large-scale landslides treatment works have conducted which never happen before. Along whole alignment, 33 landslides have been treated by kinds of measurements, such as, reducing the load and removing earthwork, building seepage ditches, slid-resisting piles, prestress anchor piles, anchor framework girds, pre-stress flat, etc. The measures have been assessed through carrying out real-time monitoring, level and vertical displacement method as well as slid belt displacement method. 2 2.1

GEOTECHNICAL ENVIRONMENT Geographic and geomorphic condition

There have been started at Chuankou in Tongchuan City from the south, and ended at Kangyadi of Huangling County to the north. In terms of the geography, it is lower on the both sides and higher in the middle; the loess plateau stands in the north, west and south with a sea level of 800–850 meters. The middle section of road passes through around 1600 m high Stone Mountain at Songzhuangliang, and the river valley fall ranges 100–150 meters with cracked terrain. The area is classified into valley terrain, degraded hill and bedrock mountain area. The loess residual

plateau and hills mainly distributed on both sides of Wangjiahe River Valley and Xihe River Valley which are located at the edge of Loess Plateau in the northern Shaanxi PRC. Its typical features are with thin loess cover layer, which distributes on the top of hills, and with exposed bedrock in middle and lower part of valley. The slope on both sides of loess hills is divided into three types of stable slope, more stable slope and unstable slope. The valley terrain is mainly located in Wangjiahe River Valley, Qishuihe River Valley and Xihe River Valley. The river valleys are ‘‘U-shape’’ valley with unbalanced development.

2.2 Characteristic of stratum The coverage layer in the area passed by the expressway alignment is Quaternary Loose soil. The loess is mainly distributed in loess hills and inclined slope belt, and the thickness generally is 30–40 m, somewhere is higher than 100 m. The flood alluvium mainly concentrates on both sides of river valley and deeper ravines. It is with fine clay in the upper part, block crush stone and pebble gravel bed in the lower part, and with loess covered on the surface with a thickness ranging from several meters to more than ten meters. Degraded slope deposition distribute in the area of valley slop with exposed bedrock. The Lithological Character is classified as fine silty clay with much block crushed stone. Huge block stone exists in some section. The thickness is 10–20 meters. The lower stratum is Paleozoic and Mesozoic bedrock, which mainly is much sandy stone and mud shale larding coal stratum, oil shale etc.

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3

DISTRIBUTION FEATURE OF LANDSLIDES

In accordance with relationships between geotechnical rock, slope structure and landslides, the landslides has the following characteristics: – Landslides of loess and loess coverage. It formed due to thicker deposited loess layer with steeper cliff as well as serious erosion and wash by the rivers. There are 6 landslides along both sides of Wangjiahe River Valley. – Landslides of bedrock. It formed along the crushed belt because of creep between fault and fold, such as, landslides at Jinsuoguan and Xihe. – Landslides of crushed zone. It has a very small scale and is very thin. It forms due to deposited crushed stone belt covering fine arenaceous rock and mud alluvium with loess structure.

4

SENSITIVE FACTORS AFFECTING LANDSLIDES STABILIZATION AND THE FEATURE OF LANDSLIDES

4.1 Formation of ancient landslides Firstly, due to effects of geological structure, the crack of inverted jointing in rock is extremely rich and the terrain easily falls into pieces. Secondly, latest constitution movement forms a steep and higher space. The mountain features are that mountains are very high, the ravine is very deeper, and river flow seriously erode and cut rocks as well as loess perpendicularity and soft and weak soil layer. Taking Guojiahe Landslide as an example, the depth of loess is 80–90 meters, the bedrock is 30–40 meters higher than riverbed, and the relative fall to edge of plateau is 120 meters. Finally, the storm and continuous rainfall promote the formation. The rainfall directly seeks into the landslide mass, flows into the slip zone so as to reduce the loess slope strength and increase the river water dynamical movement. As a result, the slip surface is formed. 4.2 Factors affecting the stabilities of landslide – Groundwater. It is a most sensitive factor, which causes the landslide deformation. Rise of groundwater level in landslide mass increase the pure hydraulic pressure. The landslide mass is immersed into the water, and then both values of C and go down, accordingly. As a result, the landside stabilities are broken. – Weak Structure Plane. The alignment crosses through the front middle part of ancient landslides. The excavation makes the old slip surface exposed. The new cutting outlet exposing at the edge of slope causes the relive of old slip surface.

– Condition of Higher Air Face. The excavation in front of slop shakes the old landslides. The higher air face makes the ancient landslides relived. 4.3 Feature of landslides deformation From the point of macroscopic view, landslide deformation behaves as creeping in the middle part, small size landslide in the front edge and tympanites and upheaval in the toe of slop. However, large quantity of deformation cause by the multi-rip and arccrack, etc. – Creeping deformation. When there is water concentrated in the slip zone, the slip will happen in the middle deeper incline part due to reduction of shear strength from pore pressure and soil mass of slip zone as well as the flow plastic and softening of soil. Its deformation causes rip and sink in the back edge and tympanites in the front edge as well as the crack occurred in the buildings. – Squeezing deformation. It forms due to excavation of slope. The new slope face confronting with space caused by the excavation concentrates the stress in the toe of slope. As a result, the tension in the top of slope was so concentrated that the toes of slope occur deformation due to squeeze. – Tensioning deformation. It mostly happens in the middle part of landslides. It represents as arc tension crack including single arc and double arc crack, small crumble happened in the edge of slope, etc. The steeper the slope is, the more violent the deformation is. 5

TREATMENT OF CHUANKOU LANDSLIDE

5.1 Background THE is located in the slope land, which is distributed in loess plateau and Wangjiahe River Valley in the south of Chuankou Village of Tongchuan City (K94 + 550∼K94 + 980). It has 700 meters from north to west, and 600 meters from east to west. Its average thickness is 21 meters with a volume of 3.6 million cubic meters. The slip surface belongs to contact surface between loess and mud shale. It is classified as the multi pull-type lower and postpones slow loess ancient landslides. The middle part of this landslide eve partially relived in 1970’s, and the peak of activity came in the middle of 1980’s. The relive caused serious personnel casualties and loss of property. The treatment work was implemented during the period from 1987 to 1992, and then the landslide has been keeping stable. The alignment passes askew through the middle part of landslide with an average excavation depth of 8–12 meters. During construction of Ramp A at Chuankou Interchange, the cutting would cross

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+1 0 0

0 + 95

+9

00

+6 5

0

+7 00

+7 50

+8

00

Figure 1. Plan of Chuankou landslide treatment.

Figure 2. Section of Chuankou landslide treatment.

through the southern edge of ancient landslide with a maximum depth of 22 meters. On the basis of analysis on survey conclusion, the alignment will affect the landslides, and it is necessary to conduct further treatment so as to satisfying technical requirement of highway project.

arrangement were designed to support landslide. They were laid along alignment by three sections:

5.2

5.2.4 Refilling crack Excavating crack with a width of 0.5–2.0 meters and depth of 2 meters, and then refilling the plain soil and compacting.

Work measurements

5.2.1 Drainage works a. Surface water drainage. Rebuilding 700 m existing cut-off trench, constructing additional 596 m tree shape intercepting ditch with a cross section of 40×60 cm2 and 60×60 cm2 . b. Drainage for underground water. Recovering 17 existing upward cline drainage holes with a total length of 850 m, rebuilding 200 sink blind ditch. 5.2.2 Supporting works Three different groups of slip-resistance piles (Table 1) with different length and different reinforcement

5.2.3 Landslide resistance retaining wall To build mortar rubble retaining wall with a total length of 200 meters.

5.3 Treatments of weak structure plane—most sensitive factor Ramp A located in Chuankou Landslide stands in the southern edge of ancient landslide. The cutting passes through the middle of ancient landslides with an excavation depth of 22 meters. When the depth was 10–12 meters, the upper part occur obvious displacement, which leaded to rip in house and multi cracks on the

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Table 1.

Slid-resistance pile attribution.

Type of piles Section size Length Number of piles Burying type Interval between two piles Layout

Section I

Section II

Section III

I 2 × 3 m2 16.9 m 19 Full-buried 6.0 m

II 2 × 3 m2 15.7 m 10 Full-buried 6.0 m

III 2 × 3 m2 19.1 m 26 Full-buried 6.0 m

On the same On the same On the same axe axe axe

Figure 3. Elevation view of pile body.

ground surface. Then, construction was stopped, and the residents were asked to leave the dangerous house. Further survey and research verified that the displacement was taken place because Ramp A had higher linear index, the length of alignment inside the landslide was too long, and excavation is too deep. The depth of air face is only 12–14 meters, and designed cutting fully reveals the slip surface, the weak structure plane become the most sensitive factor. Therefore,

Figure 4.

2–2′ Section Diagram.

Figure 5.

Reinforcement drawing for interlock.

Ramp A was relocated so as to reduce the depth of excavation assisting with support work and draining underground water. The detail shows in Figure 6. The major measurements are: (a) Reducing horizontal and vertical design index of Ramp A, such as, Ramp A moving towards to the east, and shortening 40 meters; (b) Maximum excavation depth changes to 13 meters with average depth of 6–10 meters; (c) 18 slid-resistant piles were buried on the toe of cutting with a section of 2 × 3 m2 and 2.5 × 3.5 m2 ; (d) building 13 upwards inclination drainage holes with depth of 40 meters, an interval of 6 meters and inclination angle of 5◦ C; (e) Implementing 180 m slid-resistant retaining wall on the toe of slope with window protection work on the slope face; (f) Compacting and refilling the excavated cutting, and building 90 m intercepting ditch on the top of slope. All those measurement implemented have been keeping the landslide as the status of stabilization. The effects show in Figure 7 and Figure 8.

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Figure 6. Section diagram on landslide treatment work for Ramp A of Chuankou interchange.

Elastic Mechanics Modulus was respectively adopted to verify the calculation result on support works to slid-resistant piles. Of which, Two-dimensional nonlinear Mechanics Modulus was used to verify the slipping resistance function to single pile, and Threedimensional Mechanics Modulus was adopted to analyze integrated slid-resistance function to grouped piles (Figure 9).

7 Figure 7. Image of slid resistance pile treatment slope for Ramp A of chuankou interchange.

Figure 8. Upward incline drainage on the toe of at Ramp A of Chuankou interchange.

6

FINITE ELEMENT ANALYSIS ON TREATMENT WORKS

MONITORING ON LANDSLIDE DEFORMATION

Deformation monitoring consists of two parts: monitoring on deformation on the ground surface and monitoring on deformation underground. The monitoring on surface deformation covers monitoring on crack displacement and surface 3-D coordinates so as to ascertain the stabilization of landslides. Monitoring on underground deformation refers to monitoring on spots’ displacement inside landside in order to justify the accurate location of slip surface and stabilization of landslides. To ensure safe construction, to ascertain accurately displacement and various impact factors, it is necessary to assess the treatment effect of landslides. The real-time monitoring on large landslides at Chuankou has been conducted for 3 years. It carried once a month in the first year, and one each quarter in the second year, and once each half year in the third year. Deep holes monitoring has been implemented for 12 times by using movable deep holes incline tester. Chuankou Landslide started to be treated in March 1998, ended in December 2000. During construction stage, the deformation condition has been monitored constantly. 7.1 Monitoring on surface

In the course of dealing with Chuankou Landslide, Two-dimensional non-linear and Three-dimensional

The horizontal and vertical displacement obtained from two monitoring spots of C27 and C30 (Figure 10

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01.11

01.8

01.3

00.12

00.10

00.9

00.6

00.3

99.11

99.10

99.8 99.9

99.7 99.8

99.6

99.5

99.4

99.3

99.1

98.12

98.10 98.11

Figure 9. Plane finite element model for 15# pile.

10 8 6 4 2 0 -2

Figure 10.

Displacements on C27 observation point at Chuankou landslide.

and 11) show: the landslides were keeping stable condition in the initial three stages because the deformation was slightly. The weak action state during January to June 1999 reflects that the excavation of sub-grade had been completed, the maximum displacement was 6 mm, and the tensive crack appeared surrounding two monitoring spots which were refilled and compacted in time during construction period so as to preventing rainfall from sinking. From June 1999 to January 2000, the horizontal displacement was less than 3 mm, and the vertical displacement was shorter than 2 mm. It represents that the treatment measurement was efficient which shows that the landslides were maintaining stable. 7.2 Monitoring on underground displacement Three monitoring holes were separately placed in the upper, middle and lower part of Chuankou Landslide. Since No.1 monitoring hole was destroyed in November 2000, monitoring results collected from No.2 and 3 Monitoring holes were analyzed, and worked out the displacement-depth graph (Figures 14–15). No.2 Holes. Figure 12 shows, the slip surface at No.2 Hole is 15 meter underground, the maximum slip displacement was 3 mm; the change range between last two monitoring conducted in May and July 2001

was 2 mm which represents that the landslide was maintaining stable. No.3 Hole. Figure 13 shows: the slip surface at No.3 Hole is 10.0 meter underground, the maximum slip displacement was 0.9 mm; the maximum displacement last two monitoring in November 2001 and January 2002 was less than 0.3 mm. The landslide was maintaining stable. 7.3

Application of monitoring to chuankou landslide treatment work

1. The monitoring to deformation of landslide has a leading role in revising and perfecting design plan on landslide treatment work. Chuankou Landslide belongs to the loess ancient landslide. Excavation of sub-grade at Ramp A caused crack deformation on the top of landslide. Afterwards, the slidresistance piles were installed here. In order to inspect the effect of treatment measurement, four deformation observation spots of CZ1–CZ4 were established on the top of slid-resistant piles, and two times of observation were conducted. With analysis on observation data, it was found that there was no distinct change on two observation spots of CZ1 and CZ2; Horizontal displacement of 4.4–4.6 mm occurred at Spots of CZ3 and CZ4.

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01.8

01.11

01.3

00.12

00.10

00.9

00.6

00.3

99.11

99.10

99.8 99.9

99.6

99.7 99.8

99.5

99.4

99.3

99.1

98.12

98.10 98.11

6

7

12 10 8 6 4 2 0 -2

Figure 11.

Displacements on C30 observation point at Chuankou landslide.

(mm)

(mm) 0

5

10

15

20

25

30

0

35

0

1

2

3

4

5

8

0

2 5

4

(m)

(m)

10

6

15

8 20

10 25

12

30 2001

2001

2001

2001

2001

2001

2001

2001

2002

Figure 12. Depth curve on displacement 2# hole at Chuankou landslide.

Figure 13. Depth curve on displacement of 3# hole at Chuankou landslide.

Therefore, design agency ordered to remove the load, to revise the single side slop to multi level low slop and to build upward incline drainage hole on the tope of slope. After implementation of treatment measurement, landslide became stable. 2. Monitoring to deformation of landslide plays a guiding role in construction. During landslides

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treatment, expressway construction leaded to crack appeared in local resident houses. Observation data from monitoring spots of C28, C30, C31, C32 and C33 show that the landslides were maintaining stable, and activities of landslide did not cause cracks of local houses. It provides a valuable evidence for resolve disputes on work construction.

– The dynamic design principle should be followed in the course of dealing with landslides. On the other words, the design must be adjusted in time when the geologic condition changes.

CONCLUSION 1. It is the essential to reinforce alignment geological selection in order to reduce geologic disaster. Alignment selection is very crucial to highway survey and design. First of all is to conduct the geological prospecting with high quality. It is basis of identifying the route trend. The large scale of geologic disaster must be avoided and bypass. If no way, the route must pass through area with the smallest disaster after comparing all disasters. In term of landslides, it is proposed that the route must cross through the upper or lower part of landslides, and the slip surface is forbidden to be exposed. In addition, the embankment must pass through the front part of landslides. 2. Intensifying design management is helpful to minimize the loss caused by the fault design.

3. Taking Rational Measurements

– It is necessary to conduct detail geologic prospecting so as to identify the formation mechanism, location and scope of landslides surface, and analyze sensitive factors of landslides for the purpose of optimizing the location of alignment surrounding landslides. The technical feasibility should be guaranteed as well as the rationality in the term of economy.

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– Removal of load is suitable for medium and small landslides due to its simple construction and lower cost. It should be prudent to deal with the landslide by unloading. It is possible to cause crack due to loss of stability so as to increase difficulty of treatment. In addition, it also causes some problems related to environment protection and soil and water conservation. – Building slid-resistant piles not only is safe but also has obvious effect. – Reducing and dismissing underground water has advantages of lower cost, simplified construction. And the effect is very obvious, such as, upward incline drainage hole. – Reinforcement of deformation monitoring is very important in guiding implementation of landslides, revising and perfecting design as well as verifying treatment effect.

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Types, characteristics and application conditions of anti-slide retaining structures Jing Zheng & Gongxian Wang Northwest Research Institute of China Railway Engineering Corporation, Lanzhou, China

ABSTRACT: Through the full investigation and research on traditional and new existing anti-slide retaining structures, the anti-slide retaining structures are classified into three types according to their characteristics. Their application conditions are researched and summarized from several aspects: landslides, geological structures, sliding surface morphology and sliding-scales according to force, deformation, construction and effect. The results can prove some references for engineering and technical personnel who are engaged in the design and construction of landslide treatment. 1

INTRODUCTION

Anti-slide retaining structure is a structure to maintain the stabilization of landslide by resisting landslide thrust and preventing landslides from moving. During 1950s and 1960s, there were a lot of landslide diseases during the railway construction in mountain areas of China. Through studying, practicing and summing up experience, the railway engineering and technical personnel designed many anti-slide retaining structures at the earliest time. For example, anti-slide retaining walls, bracing blind drains and anti-sliding piles played important roles in railways at that time, and were used to landslides’ treatments in other industries. After the 1980s, with the development of mountainous district construction and the West China Development, and the experience of engineering and technical personnel, some anti-slide retaining structures were proposed such as anchor anti-sliding piles, anchoring cables, and multi-anchor anti-sliding piles and so on. But different anti-slide retaining structures have different application conditions. Only applied in appropriate condition, they can fully play their roles and cut the cost of the project.

2

TYPES OF ANTI-SLIDE RETAINING STRUCTURES

There are many different types of anti-slide retaining structures. For a detailed account please see Reference No 1. Resisting sliding works are classified into four types including anti-slide retaining walls, anti-slide piles, anti-slide cut-and-cover tunnels, pre-stressed

anti-slide anchors according to structure’s formation. (Wang et al. 2004. Zheng et al. 2007) Reference No. 2 classifies anti-slide piles by seven methods: these are pile material, section pattern, pile process, force state, pile body stiffness, pile body’s assemblage form and pile head’s constraint condition. But the artificial rectangular reinfoced concrete bore piles are widely used in landslide treatments in China. Wooden piles, steel plies are hardly used. Circle plies and square piles are less used because their stress is not as functional as rectangular piles. Bore piles are also less used in landslide area due to their limitation of large equipments. Based on the full investigation and research on traditional and new existing anti-slide retaining structures and the practical application, they are classified into three types according to their characters. For a detailed account please see Table 1. 3

CHARACTERISTICS AND APPLICATION CONDITIONS OF ANTI-SLIDE RETAINING STRUCTURES

3.1 Bracing blind drain Bracing blind drain is mainly used to stop landslide from sliding, unplug the groundwater and dry landslide body. Moreover, this method of construction is not only convenient but also with low cost. As the blind drain is buried in the stable stratum under the sliding surface (sliding zone), bracing blind drain changes the material of slide body and partial slide bed into great density pervious material, usually dry laying stones. On one hand, shear strength between slide bed and the bottom of blind drain is greater than the original slide surface. Larger friction generated by blind drain

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Structure type

Characteristics

Application conditions

Single-use type

With double role of retaining and lowering groundwater level, smaller resistance in the landslide thrust, more convenient construction and low cost Having the double role of resisting sliding and lowering the groundwater level, resisting landslide thrust by bracing blind drain and anti-slide retaining wall, convenient construction and low cost

Shallow small landslide having groundwater exposed in the front or having groundwater development, soft-plastic and plastic flow Landslide formed along ditch Ibid, particularly applicable to the project site needing protection in landslide treatment

Gravity anti-slide retaining wall Rib plate anchor anti-slide retaining wall

Clear stress, simple design, adopting grouted rubble stone and rubble stone concrete in region rich in stone, obtaining raw materials locally, low project cost, simple construction Anchor mainly supporting the force, significantly reducing masonry volume and excavation volume compare to gravity anti-slide retaining wall small disturbance to landslides, more complex construction

Applicable to various small and medium-sized landslides that their thrusts are no more than 200 KN/m.

Masonry anchor anti-slide retaining wall Pre-stressed vertical anchor anti-slide retaining wall

Resisting landslide thrust both by anchor and wall, reducing masonry volume and excavation volume, small disturbance to landslides, simple construction, saving investment Resisting landslide thrust both by anchor and wall, reducing masonry volume and excavation volume, small disturbance to landslides, complex construction, saving investment

The type used in conjunction with anti-slide retaining wall

Anti-slide anchor retaining wall Rigid frame anti-slide pile

Anti-slide Pile

Common anti-slide pile

Anti-slide Retaining Wall

General anti-slide retaining wall

Bracing Blind Drain

Table 1. Characteristics and application conditions of anti-slide retaining structures.

Full buried anti-slide pile

Flexible layout, simple construction, short time limit, quick effect, little disturbance to landslide, safe, reliable, large thrust to be resisted, extensive application Cantilevered Relatively flexible layout, the remaining anti-slide pile ibid Buried anti-slide pile Gantry anti-slide pile H-type (chair type) anti-slide pile Bent anti-slide pile Cantilever single anchor anti-slide pile

Ibid, reducing bend distance and economizing material, but relatively large excavation for oversize and thick landslides Inside piles tensed, outside piles pressed, great rigidity, great ability of resisting curve and shear, resisting large thrust, economizing investment, but complex construction Ibid, but inside piles are longer than inside ones, having the action of receiving slope

Applicable to landslides that sliding surface is relatively steep behind the wall, and sliding bed is stable bedrock or compacting soil, and applicable to situation that lacks of local stone or excavates unconditionally in project site Applicable to situation that sliding surface is rather gentle, and sliding bed is stable bedrock or compacting soil Applicable to situation that sliding surface is rather gentle, and there has stable bedrock at the bottom of the wall

Applicable to various middle and large landslide treatment except soft-plastic and plastic flow landslide Applicable to the treatment of landslide which slips off from cut slope toe, scarp and embankment Applicable to oversize and thick landslides’ treatment

Applicable to treat large and extra-large landslides which thrust is large, sliding surface is gentle or sliding bed is steep but stratums are too soft to use anchor anti-slide pile Applicable to treat large and extra-large cut and embankment landslides which thrust is large, need to receive slope and do not fit anchor anti-slide pile Ibid, composed of two vertical piles and Applicable to treat large and extra-large two beams, yet beams constructed according landslides which thrust is large and don’t fit to pilot tunneling method, large construction anchor anti-slide pile difficulty Anchors locating on pile heads, active force Applicable to treat middle and large because both anchors and piles resist thrust, landslide which slip off from cut lope toe and small inner force and deflection of piles, bench, sliding surface is steep after piles, economizing investment, but complex bedrock is integer at sliding bed, thrust is much construction. large (Continued)

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Pre-stressed anchor anti-slide pile

Table 1. (Continued) Full buried single anchor anti-slide pile Cantilever multi-anch or anti-slide pile Full buried multi-anch or anti-slide pile

Pile-slab Wall

Common pile-slab wall Anchor dragged pile-slab wall

Anchor pile-slab wall

Ibid

Applicable to treat middle and large landslide whose sliding surface is steep after piles, bedrock is integer at sliding bed, thrust is much large

Multi-anchors laid on pile heads and pile bodies above ground, active force because both anchors and piles resist thrust, inner force and deflection of piles controlled effectively, economizing investment, but complex construction. Multi-anchors laid on pile heads and pile bodies above ground, active force because both anchors and piles resist thrust, inner force and deflection of piles controlled effectively, economizing investment, but complex construction.

Applicable to treat middle and large, cut and embankment landslide whose sliding surface is steep after piles, bedrock is integer at sliding bed, thrust is much large

Anti-slide piles resisting landslide thrust, slabs stabilizing landslide body between piles, difficult construction Piles and tension rods resisting landslide thrust together, slabs stabilizing landslide body in the soil arch between piles, in the principle of protecting roads, not treating the whole landslide, economizing investment, complex construction Piles and anchors resisting landslide thrust together, slabs stabilizing landslide body in the soil arch between piles, resisting rather large thrust, complex construction

Pre-stressed Anchor

Pre-stressed anchor pad pier

Anchor resisting thrust, pad pier supplying counterforce, simple design, flexible disposition, low smooth degree of earth surface in engineering place, fast construction, but less integer pre-stressed anchor Anchor resisting thrust, ground beam ground beam (rib pile) supplying counterforce, simple design, flexible disposition, symmetrical stress of sloping surface, fast construction pre-stressed anchor Anchor resisting thrust, frame supplying frame counterforce, symmetrical stress of sloping surface, increasing the integer of sloping surface in engineering place, but smooth sloping surface in engineering place, rather difficult construction

resists landslide thrust. In other words, landslide thrust acts on blind drain and then blind drain transfers slide force to stable slide bed. On the other hand, the groundwater table can fall down through the function of the drainage so that landslide body and slide zone can be dried. By doing this, shear strength of slide zone soil can be increased and slide force reduced. Bracing blind drain can be used alone when slide force is relatively weak. But when there is a large slide force, it should be used with an anti-slide retaining wall. For a detailed account please see Table 1.

Applicable to treat large and extra-large, huge thick and multi-layer landslide whose sliding surface is steep after piles, bedrock is integer at sliding bed, thrust is much large

Applicable to treat deep cut and high embankment landslides Applicable to treat embankment landslides that posterior margin is in the road surface width, and that it is stable density stratum outside the cracks in posterior margin Applicable to treat deep cut and high embankment landslides that sliding surface is steep, sliding bed is integer bedrock, and that thrust is much large Applicable to treat high and steep rock landslides and slopes that sliding surface is steep, and that bedrock is integer Applicable to treat landslides that sliding surface is steep, that bedrock is integer, and that ground bearing capacity is low where beam locates Applicable to treat landslides that sliding surface is steep, that bedrock is integer, and that ground bearing capacity of engineering place is low or landslide is loosening

3.2 Anti-slide retaining wall 3.2.1 Gravity anti-slide retaining wall Gravity anti-slide retaining wall resists landslide thrust through the friction generated by its own weight. So it usually takes the form of the large-sized structure, short and thick shape and gentler thoracic slope. Gravity anti-slide retaining wall has the characteristics as follows: its clear stress, simple design, low cost of construction and raw materials locally-produced, particularly in stone-rich region to construct with grouted

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rubble stone and rubble stone. Thus, it has been widely used. The wall is generally set in the retaining part of the front landslides. When there is a stable bedrock fore shaft, it should be set in the fore shaft and be buried in slide bed not less than 0.5 m (rock) or 2 m (soil). And the wall height should guarantee that sliding body would not slide out from the top. Gravity anti-slide retaining wall can be used alone in various small and medium-sized landslides on the condition that their thrusts are no more than 200 kN/m. When ground water is rich in landslides, it should be used with support sewer. For large landslides, it should be used with other engineering measures. 3.2.2 Anchor anti-slide retaining wall Anchor anti-slide retaining wall is a new anti-slide retaining structure which resists landslide thrust by reinforced concrete rib of slab or masonry walls and anchors together. Its mechanical feature is that landslide thrust behind the wall acts on retaining wall rib of slabs or masonry walls, and then reaches anchors by anti-rib of slabs or masonry walls, finally passes to stable layers below sliding surface. Rib anchor anti-slide retaining wall resists landslide thrust mainly by anchoring force, while masonry anchor anti-slide retaining wall stabilizes landslides through anchor anchoring force and frictional resistance generated by the gravity of masonry wall. With the help of pre-stressed anchor (Li Guanghai et al, 2004) and the substitution of the part weight of masonry wall, pre-stressed vertical anchor masonry wall stabilizes landslides by the above force and resistance. Anchor anti-slide retaining wall should adopt better overall wall structure. It means to use ribs and the small section of masonry wall. The latter can be classified into anchor masonry anti-slide retaining wall and vertical anchor pre-stressed masonry anti-slide retaining wall (Figure 1∼Figure 3) according to anchor arrangements. Under appropriate circumstances, the use of anchor anti-slide retaining wall to control landslides can significantly reduce the volume of masonry and excavation compared to the gravity anti-slide retaining wall. By doing this, the disturbance to landslides can be relatively reduced. For a detailed account please see Table 1.

rib bolt head

front panel

landslide

Figure 1. Rib anchor anti-slide retaining wall.

anchor

y

od e b

lid nds

masonry wall

la

anchor

Figure 2.

Masonry anchor anti-slide retaining wall.

anchor head y

masonry wall

od e b

slid

land

anchor

Figure 3. wall.

Vertical pre-stressed anchor anti-slide retaining

3.3 Anti-slide pile Anti-slide pile is a new anti-slide retaining structure developed by the railway sector in 1960s, which belongs to lateral load-bearing pile of large sections. Since the common anti-slide pile was adopted successfully for the first time in Chenkun railway in 1967, it has been widely used in landslide control of railway, highway, hydroelectricity, shipping, mines and factories. Pile section changes from 1.2 m × 2.0 m to 3.5 m × 7.0 m. Pile length ranges from the original 9∼17 m to the present 67 m. The structure of anti-sliding pile is shown in Figure 4. 3.3.1 Common anti-slide pile Common anti-slide pile is a lateral load-bearing structure to resist landslide thrust only by pile body itself (Figure 4a∼c). Pile drills through landslides and

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rammer and filled gravel

landsl

road

anti-slide pile

anti-slide pile

(a)Full buried anti-slide pile

ide

landsl

ide

ide

landsl

(b)Cantilevered anti-slide pile

anti-slide pile

anti-slide pile

(c) Buried anti-slide pile

de

dsli

lan

ide

landsl

de

anchor

(f)Bent anti-slide pile

(d)Cantry anti-slide pile

dsli

lan

anti-slide pile

(g)Cantilevered single anchor anti-slide pile

anchor anti-slide pile

road

anti-slide pile

(e)H-type (Chair-type) anti-slide pile

anchor

(i)Cantilevered muti-anchor anti-slide pile

de

dsli

lan

anti-slide pile

(h)Full buried anchor anti-slide pile

ide

landsl

de

dsli

lan

road anti-slide pile

ide

landsl

anchor anti-slide pile

(j)Full buried muti-anchor anti-slide pile

Figure 4. The structure types of anti-slide piles.

anchors into sliding bed of a certain depth in order to stop landslide sliding. The part buried below sliding surface is called anchorage segment, while the part above is called stress segment. When the space between piles (6∼8 m) is suitable, rock and soil between two adjacent piles can form an soil arching effect in virtue of stress segment and bilateral frictional resistance of piles. Thus, it can stop landslide body from sliding off the piles. Sequentially the landslide thrust behind piles passes to stress segments of piles, and then continues to pass to anchorage segments through stress segments. When anchorage segment deflects, rock and soil on the side of anchorage segment of the pile generates counterforce (lateral resistance of the foundation). On one hand, it embeds plies; on the other, it limits the deflection of piles. If the strength of piles is large enough that it cannot be broken, the sliding of landslides can be prevented. Surely the lateral compression stress generated by rock and soil on side of anchorage segment cannot exceed the lateral permitted stress stratum. Common anti-slide pile has the characters of flexible pile location and pile layout. It not only can be laid intensively or separately but it can be single row planted or multi-emplaced as well. In addition, it can be used independently or integrally with other engineering measures. Especially it can change the design to meet in pile pit according to the practical geological conditions. For a detailed account please see Table 1. 3.3.2 Rigid frame anti-slide pile Rigid frame anti-slide pile is a type of integral antislide retaining structure with two ordinary anti-slide piles and two beams joining together. In fact, it is a combined application of ordinary anti-slide piles (Figure 4d∼f). Its section and length inner and outer can be equal or unequal. Landslide thrust is passed to stable stratum on sliding bed by inner and outer piles. Inner piles are tensed, while outer piles are pressed. It can resist large thrust due to the great rigidity and ability of resisting moment and shear. Therefore, rigid

frame anti-slide pile is applicable to large and extralarge landslide control. In 1980, Xu Fenghe et al adopted gantry rigid frame anti-slide pile to control Luoyixi extra-large landslide on Zhiliu railway successfully. Compared to common anti-slide pile, it saved 49% masonry quantity (Xu Fenghe et al, 1988). At the same time H-type (chair type) anti-slide pile has successfully applied to embankment landslide at K180 on Chuanqian railway and Shirongxi landslide on Zhiliu railway, and bent anti-slide pile has successfully been used to Yutian landslide on Chenkun railway (1992). However, due to the beam construction existing in rigid frame anti-slide pile, especially those constructed based on pilot tunneling method, the construction is more difficult than common anti-slide pile. And it is not widely used like common anti-slide pile and anchor anti-slide pile. The characteristics and application conditions of each pile are detailed in Table 1. 3.3.3 Pre-stressed anchor anti-slide pile Pre-stressed anchor anti-slide pile developed in 1980s is a new type of complex sliding resisting construction, which bears stress through both anti-slide piles and anchors. Pile bodies go through landslide body and then anchor into sliding bed for a certain depth. Its stress is similar to the common anti-slide pile. Generally, anchors act on pile heads, but for cantilever anti-slide pile they can be set on pile bodies (Figure 4g–i). Going through landslide body, it anchors into sliding bed for a certain depth. In recent years, while aiming at treatment of huge thick and multi-layer landslides, a new structure of a full-buried multi-anchor anchor anti-slide pile has been developed and studied (Fig 4j). This anti-slide pile has the advantages as general anti-slide piles. Because of pre-stressed anchors on pile heads or bodies, the deformation of piles is restricted. So it greatly improves stress condition of original cantilever piles. Synchronously the flexural torque of pile bodies and the deflection of piles are greatly reduced due to the passive force of piles. Thus, compared to common anti-slide pile, when

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thrusts are equal, the section and length of pile can be largely reduced and the investment can be economized. Besides, when the section and length are equal, it can endure a greater thrust. But the construction of anchor anti-slide pile is more complex than that of common ones due to its professional nature of pre-stressed anchors. E geci.com specially some anchors are constructed in pile pit, so the construction of full buried multi-anchor anti-slide pile is more difficult than any other anti-slide pile. The characteristics and application conditions of each pile are detailed in Table 1. 3.4 Pile-slab wall Pile-slab wall develops from the anti-slide pile, which is made up of piles and slabs between piles. When cantilever anti-slide pile is used to stabilize cut and embankment landslides, walls or slabs between piles are then applied. It is called Pile-slab wall when slabs are used. In order to solve the problem of stress caused by the high wall, pre-stressed anchor dragged pile-slab wall and anchor Pile-slab wall (Figure 5) are adopted in engineering practice. Beside slabs, the characteristics of force of other parts are similar to common anti-slide pile and anchor anti-slide pile. Here no need to further discuss. For a detailed account please see Table 1. According to the analysis on accidents of Pile-slab wall in recent years, it is the real reason that ground surface water seeps down or ground water cannot drain well because of hanging plates. Consequently ground water table rises and hydrostatic pressure increases, which lead to the collapse or the serious incline of the Pile-slab wall. So during the setting of Pile-slab wall, focus on the drainage of both ground surface water and ground water. Besides, it has to remain enough flush holes for draining groundwater. 3.5 Pre-stressed anchor

The anchor is made up of three parts, including anchorage section, unrestricted section, and anchorage head. Anchorage head is fixed on outside anchor structure. The types of outside anchor structures are pad pier, ground beam, frame and so on. Landslide thrust acts on outside anchor structure while sliding. Then the thrust passes to anchorage section through anchorage head and unrestricted section by outside anchor structure. At last it passes to stable stratum at sliding bed through binding power formed by anchor rods and injected bodies, injected bodies and rock and soil along wall of holes. Thus the sliding is prevented. So the anchor is the major bearing structure, whereas outside anchor structure supplies counterforce only. From Figure 6, the anchor pulling force (Pt) can be discomposed to two forces of parallel sliding surface and vertical one. The force of parallel one is opposite to the direction of sliding, and resists sliding directly. But the force of vertical one equivalently increases normal pressure to sliding surface, and resists sliding through increasing friction of sliding surface. However, the normal pressure generated by anchors at sliding surface is concerned with landslide substance, loosening degree, landslide thickness and so on. Therefore, friction resistance of sliding surface needs to be reduced from 0 to 1 according to concrete situation. Pre-stressed anchor pad pier uses reinforced concrete as the outside anchor structure. It is a single pier independent structure, which fixes anchor heads on pad pier to stabilize landslides mainly depending on anchorage force. Pre-stressed anchor ground beam uses reinforced concrete strip beam as outside anchor structure. It is a structure that fixes strip beams on anchor heads to stabilize landslides. Anchor wire with two or more holes can be fixed on ground beams. Pre-stressed anchor frame uses reinforced concrete frame as outside anchor structure. It is a structure that fixes frame nodes on anchor heads to stabilize landslides. Frame regards a piece as an integral and there

Pre-stressed anchor is a flexible rod component mainly bearing tensile force. It fixes steel wires into stable stratum (sliding bed) in deep by bores and injected bodies. Strengthened body—sliding body surface exerts pres-stressed force to steel wires with stretchdraw. In consequence, landslide body and stable stratum are in deep lock tightly. Thus, it limits the deformation of landslides and stabilizes landslides.

slab

lide

frame slab anchor pier

lide

e

lid lands

road

anchor

lands

anti-slide pile

F

drag bar

slab

lands

(a)Common pile-slab wall

e

anti-slide pile

anti-slide pile (b)Anchor dragged pile-slab wall

W

anc

hor

Pt

id dsl lan y od N b

Pt Pt

Pt

(c)Anchor pile-slab wall

Figure 5. The structure types of pile-slab walls.

Figure 6. landslide.

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Sketch map of pre-stressed anchor treating

are several horizontal and vertical beams in the same piece. For a detailed account please see Table 1. In conclusion, in view of construction feasibility and economy, pre-stressed anchor sliding retain structure is applicable to control landslides that sliding surface is steep, and that sliding bed is integral bedrock. When thrust is not strong and the deformation control is not limited strictly, it can be used alone. Otherwise it needs to be used with rigid support and block contracture together such as anti-slide pile, resisting sliding wall and so on. 4

CONCLUSION

Anti-slide retaining structure is a structure that is used to maintain the stabilization of landslide by resisting landslide’s sliding and preventing deformation. It is mainly used to support thrust. Whether landslides are stable or not after treatment mainly depends on the proper application of anti-slide retaining structure and its exact design when geologic information is accurate. So it is of vital importance to choose proper anti-slide retaining structure according to their characteristics and application conditions. In addition, because of the complexity of landslides and the limitation of

exploration, the construction of anti-slide retaining structure is different from other constructions. As a result, during the construction process, it must carry out the concept of dynamic design and information constructure in order to ensure landslides successfully treated by changing in time according to the actual geological conditions and sliding surface location. REFERENCES First Survey and Design Institute of Rrailways. 1992. Railway engineering design technical manuals (roadbed) [M]. Beijing, Chinese railway press, pp: 160–162. Guanghai Li et al. 2004. Design and engineering examples of new retaining structures [M], Beijing: People’s traffic press, pp: 124–128. Wang Gongxian, Xu Junling, Liu Guangdai & Li Chuanzhu. 2004. Landslides and prevention technology of landslides [M]. Beijing: Chinese railway press, pp: 375–458. Xu Fenghe & Wang Jinsheng. 1988. Designs and constructions of rigid frame anti-slide piles in Luoyixi [C]. Landslide corpuses (No 6), Beijing, Chinese railway press, pp: 1–10. Zheng Yinren, Chen Zuyu, Wang Gongxian & Ling Tianqing. 2007. Engineering treatments of slopes and landslides [M], Beijing: People’s traffic press, pp: 394–419.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

The stabilization of the huge alluvial deposit on the left bank and the high rock slope on the right bank of the XiaoWan Hydropower Project Lichun Zou, Xianliang Tang, Hanbin Feng, Guojin Wang & Hui Xu Kunming Hydroelectric Investigation Design & Research Institute, State Power Corporation of China, China

ABSTRACT: A huge alluvial deposit is located on the left abutment of the Xiaowan arch dam, which has the maximum height difference of about 470 m and the total volume of 5.4 × 106 m3 . The natural slope of deposit is stable, but the properties and mechanisms of the deposit, as well as the boundary conditions are extremely complicated. Some parts of the deposit near the foot must be excavated to arrange the foundation of working machines, the concrete mixing system and the concreting operation platform. The creeping deformations occur during the excavating, and then the deformation gradually increases, which shows gradually extending form down to upward with a drawing model. Based on the dynamic tracing of the developed deformation, a systematic suitable engineering reinforcing treatments is performed, which guarantees the whole stability of the huge deposit. 1

2

GEOLOGY OF THE DEPOSIT

It looks like a tongue from the plane view, with an average slope angle of 32◦ 35◦ from a view of topography. The structure of the deposit is basically compacted with several exceptions. The properties of the deposit are spatially in homogeneous. The deposit are mainly composed of two different types of gradients.One is rock blocks and big lonestones with debris soils, debris and gravel soils, of which the diameters are generally of 0.2 m∼0.8 m, 2 m∼5 m and 5 cm∼15 cm, with the biggest of about 1.0 m, 9 m and 20 cm respectively; The other is deposited materials by collapsing and rockfalling. The debris soils, debris and gravel soils are not formed into layers spatially, but filled into the gaps between the big lonestones and rock locks. The natural overview of the deposit and sketch after the excavation are shown in Figure 1. There exists a layer of materials between the bottom of the deposit and the underlying rocks, This layer mainly consists of relatively fine particles with quite different thicknesses of 0.15 m∼8.05 m, called the contacting soil, such as gravels, silt and sand gravels.The underlying rock of the deposit looks like a nearly U-shaped ancient trough developed along one fault named F7. Figures 2 and 3 show the longitudinal and transverse sections of the deposit.

DESCRIPTION OF DEFORMATION FOR THE DEPOSIT

The deposit slope was stable during the early stage of the excavation (from EL. 1635 m∼1276 m). Several cracks occurred in the some parts of the contact zones (from EL. 1400 m∼1480 m) between the deposit and the underlying rock on the downstream, and the observed data were shown in Figure 4. The deformations were mostly developed within the regions between EL.1245 m∼1460 m. Figure 5 shows the cracking zones during the excavation.Cracks on the construction platform are shown in Figure 6. The monitoring data from the observing equipments at the elevation of 1250 m have indicated the location of sliding and the progressive development. 3

ENGINEERING MEASURES

The deformation mechanism of the deposit slope was deeply carried out for the sake of safety and economy. The engineering measures were performed step by step, which can be primarily divided into three major stages. The first stage is to construct some uncomplicated and easy reinforcements only for safely crossing the flood period as soon as possible, the second is to construct certain proper reinforcements

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Figure 1. (a) The natural overview of alluvial deposit (b) After the excavation of deposit.

(m)

Elevation

1300

original topogarphy

boundary between accumulation and underlying rock 1250

gully

accumulation

F7

underlying rock

1200

f 16

Figure 2. Longitudinal section of the deposit. Figure 3.

Transverse section of the deposit.

Figure 4. The curves of displacement with time from monitoring points at different elevations.

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Figure 5.

Cracks developed during the excavation of the left abutment of Xiaowan arch dam.

Figure 6. Cracks on the construction platform.

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Figure 7. The data obtained from the monitoring equipment at the elevation of 1250 m.

Figure 8. Drilling in alluvium for prestressed cables with a steel pipe that protects the hole.

for guaranteeing the construction safety, and third is to perform the multiple treatments for ensuring permanent safety. Prestressed Cables have been extensively used in China’s hydropower projects. New technology involves corrosion-proved, multi-head anchors and high efficient drilling machines. Figure 8 shows a drilling technique that used in the Xiaoman deposit slope, which enables the protection of the bore holes in alluviums by a steel pipe. This technique allows more 1000 cables to be installed for rescuing the slope. Figure 9 describes the retaining piles that inhibit the increasing large deformations of the deposit slope. A composite structure which combines the cables, piles and buttress has been used in this project. The monitoring results as shown in Fig. 11 indicated that the deposit slope is stable (the Max. value of average speed rate of 0.32∼0.04 mm/d along horizontal direction and 0.34∼0.02 mm/d along vertical one respectively). Figure 12 gives an overview of deposit slope after the excavation.

Figure 9.

4

The retaining piles.

SKETCH OF THE HIGH ROCK SLOPE ON THE RIGHT ABUTMENT

Figure 13 gives the schematic view of rock slope on the right bank of Xiao Wan arch dam. Figure 14 shows an overview of this rock slope. Figure 15 is the cross section of intake vertical slope.

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5

STABILITY ANALYSES AND ENGINEERING MEASURES

For the slopes of the right abutment of the arch dam, different kinds of models for the stability analysis are established in the different portions and areas, according to the boundary conditions, the features of the rock texture and the slope shapes of excavating. For engineering treatments of the rock slopes in the Xiaowan Project, first, taking into account the plan of excavation to be obedient to the natural feature of the slope, second, some portions which have a relatively unfavorable stability are selected with the methods of qualitative and partly qualitative analysis such as the stereographic projection, and then, the detailed study and analysis are carefully carried out based completely on the failure models. A great number of monitoring instruments were buried in the slope area in order to evaluate the feasibility of the engineering treatments. The curves of Load with Time of Prestressed Anchorage are

shown in Figure 16 and Figure 17 gives the curves of displacement with time by Multipoint Displacement Meters. 6

CONCLUSIONS

Based on the aforementioned descriptions about the huge alluvial deposit and the high rock slopes on the left and right abutment respectively of Xiao Wan hydropower project, it comes to the following points:

Figure 12. excavation.

An overview of the deposit slope after

Figure 10.

The composite structure of cable-pile-buttress.

Figure 11.

The curves of displacement with time by monitoring points at different elevation during three stages.

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1. During the excavation of deposit slope of almost 400 height, when the excavation went down from elevation 1645 m to 1245 m, a great number of cracks occurs indicating the slope is in dangerous status.

2. The drilling prestressed cables inhibit the accelerating deformations, which allows the consideration of next plan 3. Fourteen large scale retaining piles on the platform of 1244 m are constructed, which guarantees the stable status of the slope 4. The slope on the right abutment of the Xiaowan arch dam is excavated as high as 600 m and also

Figure 13.

Sketch of rock slope on the right bank.

Figure 15.

Figure 14.

An overview of high rock slope on the right bank.

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Cross section of intake vertical slope.

Figure 16.

The curves of load with time of prestressed anchorage.

Figure 17.

The curves of displacement with time by multipoint displacement meters.

the complicated geological properties adds to the difficulties in excavation 5. A method of stability analysis suitable for the rock slope of with plenty of fissures is presented during the four years of practices

6. A great number of monitoring results have shown that the deformation of the slope increases with the excavation and its increasing rate is gradually decreased after the anchorages are in work and the deformation tends to an invariant as time goes by.

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Risk assessment and management

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Malaysian National Slope Master Plan – Challenges to producing an effective plan C.H. Abdullah & A. Mohamed Slope Engineering Branch, Public Works Department, Malaysia, Kuala Lumpur, Malaysia

ABSTRACT: After numerous high profile slope failures that caused fatalities or extensive property destruction, the Malaysian Government decided that a comprehensive plan had to be established to better manage slopes and developments in hilly areas throughout Malaysia. The Public Works Department of Malaysia (JKR) was instructed to carry out a study to produce a National Slope Master Plan (NSMP) with the goal of producing a comprehensive and effective national policy, strategy and action plan for reducing losses from landslides on slopes nationwide including activities at the national, state and local levels, in both the public and the private sectors. A consortium of consultants was appointed by JKR to carry out the study. The period for the NSMP study is two years. This paper describes the methodology adopted for the NSMP study, the challenges posed to ensure the relevancy and the effectiveness of the plan.

1 1.1

INTRODUCTION Background

A series of landslide (inclusive of debris flow) events since 1993 has resulted in a number of deaths and property extensive destruction. In Malaysia, most of the high consequence slope failures involved man-made slopes. However, there were a few devastating debris flows that may not have been caused by human activities. The first major landslide event, which has since become the landmark landslide was the landslide that brought down an apartment block in an area known as Highland Towers, near the capital, Kuala Lumpur in 1993. In this incident, 48 people perished when the foundation of the apartment block was undermined by the landslide. Fig. 1 shows the apartment building toppling over. After 1993, there was a series of devastating landslide events, such as at Genting Sempah in 1995, that killed 20 people; Pos Dipang in 1996 where 44 people died; at Bukit Lanjan in late 2003 where a major toll expressway to Kuala Lumpur had to be closed for 6 months. The Bukit Lanjan slope failure was the last straw that triggered the Malaysian Government to take an initiative to establish a body that would deal explicitly with slopes. In this incident, there were no casualties. However, the effect on the economy was significant, not to mention the difficulty faced by motorists and the losses incurred by the expressway concessionaire. The repair bill came up to several million US dollars. Fig. 2 shows the slope failure at Bukit Lanjan that caused the Malaysian Government

to take action on slope matters. Finally, in 2004 the Government instructed the Public Works Department of Malaysia to set up a Slope. Engineering Branch (CKC) to look into matters pertaining to slopes. One of the major tasks that was entrusted upon the new CKC was to develop a National Slope Master Plan (NSMP) for the whole of Malaysia. Most of the landslides that affect human activities are

Figure 1. The apartment building in the process of toppling over.

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A consortium of 6 consultants was appointed in March 2006 to carry out the NSMP study. The NSMP consisted of 10 main objectives which were translated into 10 components of the study. The objectives of the NSMP are as follows: 1. Develop an effective policy and institutional framework to minimize risk from landslides on slopes nationwide. 2. Develop a framework of establishing an inventory of susceptible areas and different types of landslide hazards and risk at a scale useful for planning and decision-making. 3. Establish a system for monitoring landslides that pose substantial risk. 4. Compile and evaluate information on the economic impacts and all other relevant information on landslide hazards. 5. Establish an effective system for landslide hazards information transfer. 6. Develop programs for guidelines, training and education for engineers, scientists, decisionmakers, etc. 7. Develop awareness programs of landslide hazards for the general public, developers, engineers, scientists, decision makers, etc. 8. Develop a plan for appropriate mitigation measures. 9. Improve the nation’s ability to respond and recover from landslide disasters. 10. Develop a predictive understanding of landslide processes, thresholds and triggering mechanisms.

Figure 2. The Bukit Lanjan slope failure near Kuala Lumpur. 80 70 60

No. of Landslides

50

Fatality

40 30 20 10 2006

2004

2002

2000

1998

1996

1994

1992

1990

1988

1986

1984

1982

1980

1978

1976

1974

0

Figure 3. Landslides and fatalities due to the landslides between 1974 and 2006.

related to the economic growth of the country or the area. Malaysia experienced a rapid development since the early 1990s and the development continues until the present. Fig. 3 shows the rise in landslide events and fatalities from 1974 to 2006. The figure indicates that landslides began to increase since the early 1990s and taper off slightly after the economic crisis in Malaysia and Asia in 1997. This phenomenon is in line with the observation made by Remendo et al. (2003) who observed that landslide events are related to the development of the area.

From the objectives, it can be seen that the study provides a very comprehensive coverage of the matters pertaining to slope management and planning. Out of these objectives, 10 components under the following headings were derived:

1.2 National slope master plan The NSMP study was officially inaugurated in March 2006. However, the groundwork for the NSMP study began a year earlier. The study is expected to be completed by March 2008 with a total estimated cost of approximately USD 1.8 million. The NSMP’s Terms of Reference (Public Works Department, 2006) were based on the USGS Circular 1244 (Spiker and Gori, 2003) and the work by the Committee on the Review of the National Landslide Hazards Mitigation Strategy Board of the United States National Research Council (2004).

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i. Policies and institutional framework—improve institutional framework ii. Hazard mapping and assessments—develop a plan for mapping and assessing landslide hazards and also standards and guidelines for landslide hazard mapping iii. Early warning and real-time monitoring system to develop a national landslide hazard monitoring, prediction and early warning system iv. Loss assessment—assess the current data on landslide losses and develop a national plan for compilation, maintenance and evaluation of data from landslides v. Information collection, interpretation, dissemination, and archiving—evaluate the state-of-theart technologies and methodologies for dissemination of technical information

Standards and Guidelines

Early War ning System

Engineering Prope rties

Drainage

Public Awareness

Suction

Geology

Ground Water Table

Slope Stability

Research & Development Rainfall

Slope Geometry

Veget ation

Slope Management

Laws and Regulations

Figure 4. Relationship between physical factors, management and control of slopes.

vi. Training—develop training programs for personnel involved in landslides vii. Public awareness and education—evaluate and develop education programs related to predictive understanding of landslides viii. Loss reduction measures—evaluate and develop effective planning, design, construction and maintenance with a view for landslide hazard reduction ix. Emergency preparedness, response and recovery—develop a national plan for a coordinated landslide rapid response capability. x. Research and development—information collection, interpretation, dissemination and archiving—evaluate the state-of-the-art technologies and methodologies for dissemination of technical information The relationship between major physical factors, management and control of slopes can be represented by Fig. 4. The stability of a slope is generally governed by the parameters presented in the inner circle, while at the outer circle are the controls and actions that need to be in place to mitigate slopes failure. The NSMP study was conducted based on the principles presented in the figure. 2

METHODOLOGY OF THE STUDY

To come up with such a comprehensive plan pertaining sto slopes is extremely difficult and complex. One

of the major principles of the study is to identify what are the weaknesses in the system and try to make improvements. Another important matter is to identify the roles and responsibilities of the various parties involved. Finally, lessons learned from other countries is part of the issues considered so that the state-ofthe-art can be considered. Based on the premises mentioned above, the methodology employed is as follows: 1. Identification and compilation of reference materials. 2. Listing down of relevant agencies and organizations, and preparation of questionnaires. 3. Discussion with government agencies at all levels, and other related organizations. 4. Literature reviews and technical visits overseas. 5. Review existing local conditions and practices overseas. 6. Presentation of proposed national plan to stakeholders and obtaining their feedback. 7. Preparation of inception, conceptual, interim and final reports and obtaining comments. 8. Refinement and submission of reports. The NSMP is to be implemented in 3 phases: the first phase is called the short term which would cover the first 5 years; the second phase is known as the medium term i.e. the period of implementation between 5 and 10 years; and final phase which is known as the long term which is the period of implementation of 10 years and beyond.

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A lot of technical and non-technical papers pertaining to slopes were identified, compiled and reviewed. Similarly, the relevant authorities that deal directly or indirectly with slopes were also identified. The authorities involved include local authorities, the Planning Department, the Police Department, Fire and Rescue Department, Environmental Department and others. In Malaysia, Planning, Police, Fire and Rescue and some other Departments are under Federal control, unlike the United States. A local authority is responsible for among other things: local development approvals, providing sanitary services, and collecting local taxes. The relevant departments, the highway concessionaires and some local authorities were included in the technical committee that meets after the conceptual report was produced. Questionnaires were sent to the various federal, state and local stakeholders to examine their responsibilities, capabilities and system that they possess in matters pertaining to slopes. Laws and regulations that are related to slopes were also examined and any deficiencies and overlaps between various parties involved in the implementation of any slope policies were identified. Several dialogue sessions with the stakeholders in the states were carried out in order to obtain better feedback from the personnel who were dealing directly with implementing works pertaining to slopes. From time to time the consultants were required to submit reports which would then be analyzed by CKC and reviewed by local and overseas technical reviewers. The reports would then have to be revised based on the comments from the various parties.

3

PROBLEMS FACED IN THE STUDY

The scope of work for the NSMP was comprehensive and given the relatively short period to cover all aspects of slope management and developments the task of ensuring that the NSMP is relevant and applicable to the nation’s slope management is quite daunting. Many problems were faced by the NSMP study team. Some of the problems are internally generated others and caused by external factors that are more difficult to resolve. The consultants usually work on the individual topics in isolation to the others. Since some of the issues presented under individual topics may also be in the other topics in the NSMP, one of the major problems faced by various components of the study is integration between various topics so that the plan would be consistent. In this case, communication and cooperation between the teams would have to be enhanced such that integration problems can be ironed out without difficulty.

All the plans presented cannot be carried out at the same time due to limited resources. Therefore, one of the key issues of the plan is prioritization of the implementation of the NSMP in terms of the short, medium and long term goals. The short term goals would be to ensure that the impact of the plan not only would address the pressing problems faced by the nation presently, but also ensuring maximum impact is felt once the implementation is in place. Prioritization would in essence consider the essential, important and desirable aspects of the implementation which would then be linked up with other phases of the NSMP. One of the major problems faced by the study team is to acquire reliable data to formulate plans to address some of the problems faced by the country on the matters pertaining to slopes. Some of the available data are inadequate others may be scattered in various departments or with private consultants and highway concessionaires. An example of the problem faced is on the matter pertaining to loss assessment, whereby a general idea of the money spent on repairs and direct damages to properties are difficult to assess due to data that may be available but improperly archived. In the case of the private sector, apart from some of the problems mentioned above, the information is withheld either because they are part of the trade secrets; others due to pending litigation that clients were unwilling to divulge. On the cost of maintenance of slopes, usually, the total cost is combined with that for other maintenance works such as road works, drainage works, and structural works together. Separating the maintenance cost into various elements would be time consuming and require input from the people who have the intimate knowledge on the works done. For the questionnaires that were sent to various agencies, 58% of the reply was received from the federal departments and more than 25% from the state departments. In some cases, the answers obtained were inadequate and further enqueries were made with the relevant parties to obtain the answers. Since some of the questionnaires were not answered by the relevant parties, the study teams have to contact the parties directly or at other times workshops with the parties concerned were carried out at the state level. One of the questions posed to the emergency department was with regard to the essential equipment required during a landslide disaster so that the right equipment can be identified. The equipment identified by the department for the search and rescue operations in the event of landslides are costly. In Malaysia, from 1993 to the present, the average number of human casualties due to landslides is about 13 per year. The benefit of acquiring the specialized equipments may not have the blessing from the Government since the fatalities or damage to property may not be large compared with other disasters such as

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flooding. Consequently the consultants were requested to look at other emergency applications where similar equipment can be utilized. On the matters pertaining to public awareness, Malaysia being a multi-racial country with diverse ethnicity, language and culture the public awareness campaigns and education will have to take these issues into consideration. Therefore, in some cases the public awareness campaign will have to tailor for the targeted community rather than for the general public at large. The public awareness campaigns and education programs to be implemented would have to be multi-faceted actions with due consideration given to the receptiveness of the different communities. A key problem faced by the study team is gauging whether the progress made in coming up with the plan was in the right direction and can be successfully applied when implemented. 4

SUMMARY AND CONCLUSIONS

The preparation of the NSMP for Malaysia is a complex process with many uncertainties and problems. One of the problems faced by the study team is to obtain accurate and reliable data in order to formulate a plan that would be relevant for slope management in Malaysia. This problem can be lessened if all parties are less secretive about their work and data are made easily available. The diverse ethnicity, language and culture bring to the fore the problems related to public awareness campaigns and education program.

However, they can be resolved but with greater resources and different implementation strategies for the different groups of people. The limited resources necessitate prioritization of the implementation of the plan based on short, medium and long term goals. Apart from the importance of the relevancy of the NSMP, the success of the implementation of the plan depends very much on the political will of the Malaysian Government. This may be shaped by public opinion and by landslide events which may take place in the future. It is also important for relevant parties in the nation to understand that the NSMP must be dynamic and evolving to cater for the future changes in the climate, politics and the community. REFERENCES Remendo, J., Soto, J., Gonzalez-Diaz, A., de Teran, J. R. D., & Cendrero, A. 2005. Human impact on geomorphic processes and hazards in mountain areas in northern Spain. Geomorphology, Vol. 66, 69–84. Public Works Department, 2006. Terms of reference for National Slope Master Plan Study. JKR Memorandum of Agreement No:JKR/IP/CKC/23/2006, Public Works Department, Malaysia. Spiker, E. C. & Gori, P. L. 2003. National landslide hazards mitigation strategy—a framework for loss reduction, USGS Circular 1244. National Research Council, 2004. Partnerships for reducing landslide risk—assessment of the national landslide hazards mitigation strategy, The National Academies Press, USA.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Spatial landslide risk assessment in Guantánamo province, Cuba E. Castellanos Abella Instituto de Geología y Paleontología (IGP), Ciudad de La Habana, Cuba

C.J. van Westen International Institute for Geo-Information Sciences and Earth Observation (ITC), Enschede, The Netherlands

ABSTRACT: Within the Cuban national system for multi-hazard risk assessment, landslide hazard and risk have not been properly addressed thus far. This paper focuses on a method for landslide susceptibility assessment, its conversion into hazard, and the combination with elements at risk data for vulnerability and risk assessment. The method is tested in Guantánamo province, one of the areas with the highest incidence of landslides in Cuba. The GIS-based assessment was carried out with input maps at 1:100,000 scale or larger resulting in digital maps with 50 m pixel resolution. For the susceptibility analysis 12 factors maps were considered: geomorphology, geology, soil, landuse, slope, aspect, internal relief, drainage density, road distance, fault distance, maximum daily rainfall and peak ground acceleration. The relationship between these factor maps and the landslide inventory was analyzed using a combination of heuristic and statistical methods (Artificial Neural Network analysis and Weights of Evidence method). Five different landslide types were analyzed separately (small slides, debrisflows, rockfalls, large rockslides and topples), resulting in five susceptibility maps. Success rate curves were generated and analyzed to evaluate the predictability and to classify the maps. The susceptibility maps were converted into hazard maps, using the event probability, spatial probability and temporal probability. Return periods for different landslide types were estimated based on the main triggering events and geomorphological reasoning. The vulnerability analysis started with the generation of a provincial database with five elements at risk maps: number of inhabitants per house, essential facilities and non-residential buildings, roads, agricultural landuse and natural protected areas. The spatial landslide risk assessment was conducted by analyzing the 5 hazard maps and the 5 vulnerability maps. A qualitative risk assessment was carried out using Spatial MultiCriteria Evaluation. Semi-quantitative risk assessment was done by applying the risk equation in which the hazard probability is multiplied with the number of exposed elements at risk and their vulnerabilities. In this paper only the results of the semi-quantitative assessment of population risk are presented and briefly discussed. The study was able to identify high risk areas and the main causes derived either from high landslide hazard or from high spatial concentration of element at risk in Guantánamo province. In order to derive a quantitative estimation of risk more information should be available on temporal probability and vulnerability. Both of these require an extensive landslide database which should be implemented and maintained at the national level.

1

INTRODUCTION

Landslides are one of the hazard types occurring in Cuba. Local authorities have recognized the need for assessing the hazard and risk due to landslides, and incorporate these into a multi-hazard risk assessment. In Cuba, most of the reported landslides are associated with hurricanes, tropical storms or prolonged periods of rainfall (Pérez, 1983, Iturralde-Vinent, 1991, Pacheco & Concepción, 1998, Castellanos, 2000). Landslide damage is normally not recorded separately from the main disaster, so there is no historical landslide database available that could serve as the basis for a landslide risk study.

This research is intended to increase the awareness about landslide problems in Cuba by developing a method for GIS-based landslide risk assessment at different scales, taking into account the specific situation with respect to data availability and landslide types in Cuba. Various methods for landslide hazard and risk assessment have been applied in other countries (e.g. Cruden & Fell, 1997, Dai et al., 2002, Glade et al., 2005), but they need to be translated to the Cuban situation. Results on landslide risk assessment at the national and municipal scale were reported by Castellanos and Van Westen (2007a, 2007b). Here some results of the provincial scale are presented.

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2

CASE STUDY AREA

the Caribbean-North American plate boundary in the south.

The method for landslide risk assessment at a provincial scale was developed and tested in the province of Guantánamo (see Figure 1), the most eastern province of Cuba with a surface area of 6186 km2 , comprising 5.5 percent of the national territory. The population is 511,224 (ONE, 2007), which is 4.6 percent of the national population. About 75 percent of the area is mountainous with the highest point at 1,181 m. The southwest is covered by the large valley, which also forms a separate hydrographic basin. Guantánamo contains both the most humid (in the North) and driest (in the South) zones of Cuba. The province has 10 municipalities (indicated with large dots in Figure 1) and 386 settlements from where 18 are considered urban. Agriculture is the most important economic income for the province which is based on sugar cane, coffee, cacao, wood and coconut. The last four are cultivated in mountainous regions. The industries include an iron foundry, and factories for coffee, agricultural tools, furniture, food, sugar cane and salt. Guantánamo has the national record of 49 devastating hurricanes measured over the period 1789–2003, which are more frequent in September and October. Since 1997 to 2002 there were 93 forest fires reported, affecting an area of 3043 hectares. The landslides resulting from these other disasters are rarely recorded in the official statistics. The province also has a substantial earthquake hazard, due to the presence of

3

LANDSLIDE INVENTORY

The provincial landslide hazard, vulnerability and risk assessment started with a detailed landslide inventory. Landslides were photo-interpreted from 300 aerial photos (format 23 × 23 cm) from the year 2000 at 1:25,000 scale, covering the entire Guantánamo province. The photo-interpretation was transferred from the photos to base maps which were later scanned and digitized. The landslides boundaries were crosschecked using band 8 (15 m) of a Landsat ETM+ satellite image. A spatial database was created with attributes for the size and type of landslide. The database also included only 12 historically reported landslides. Unfortunately no multi-temporal image interpretation could be carried out, which made it difficult to establish the age of the landslides. In total 281 landslides were identified covering an area of about 19.92 km2 . From this inventory, four main types of landslides were determined: rockfalls, debrisflows, topples and slides. Landslides identified as topples were considered in those areas where a number of subsequent detachment blocks could be identified without major downslope movement. A toppling movement may culminate in an abrupt falling or sliding but the form of the movement is tilting without collapse.

Figure 1. Hillshading map the study area, with municipalities, main urban centers and landslides (indicated as black dots).

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Further analysis showed that slide-type movements were basically of two genetically and morphometrically different types: a group of 29 larger landslides located in a high tectonically affected area in the San Antonio del Sur area, and a group of 186 smaller landslides dispersed all over the province. The results are given in Figure 1 and Table 1. 4

METHODOLOGY

A schematic overview of the methodology is given in Figure 2. The method started with a comprehensive landslide inventory, and the collection of input data on landslide causal factors and elements at risk, represented in the upper part of Figure 2. The next step was to generate a number of landslide susceptibility maps for the five different landslide types described above, using a combination of a heuristic approach, and of several statistical methods, such as the Weights of Evidence Modeling and Artificial Neural Network analysis. The susceptibility maps were converted into Table 1. Number of landslide events and areas for different landslide types mapped from photo-interpretation and fieldwork in Guantanamo province. Type

Number

Area (m2 )

Small landslides Large landslides Topples Rockfalls Debrisflows Total

186 29 18 22 26 281

7.72E+06 8.39E+06 1.28E+06 1.29E+06 1.25E+06 1.99E+07

hazard maps, based on the landslide densities of the susceptibility classes and the temporal probability of landslide occurrence. This resulted in five hazard maps (H_slide to H_rockslide), indicated in the middle part of Figure 2. Elements at risk (EaR) data were collected for population, roads, essential facilities and non residential buildings, agricultural land use, and protected areas. In order to estimate the risk to these elements by the five different landslide types, each of the five hazard maps was overlain with the elements at risk maps to calculate the number of elements per hazard class. In the lower part, the method for the risk assessment is presented. Two approaches have been used: qualitative and semi-quantitative methods. A qualitative risk assessment was carried out using Spatial MultiCriteria Evaluation. Each EaR map was standardized to values between 0 and 1, and weights were assigned to each of the maps, depending on their importance, and their estimated vulnerability to the particular landslide types. These maps were integrated into a single vulnerability map, which was combined using a twodimensional matrix with the 5 hazard maps. This resulted in five qualitative specific risk maps (R_slide to R_rockslide). The semi-quantitative risk assessment method is based on the calculation of specific risk as the multiplication of hazard, vulnerability and amount of elements at risk that are exposed. This is indicated by the two maps at the bottom right part of Figure 2. Semi-quantitative assessment was only carried out for population and roads, as there was no cost information available for the other 3 types of elements at risk. The study was based on raster analyses using IL WIS and ArcGIS© GIS software at 50 m resolution taking into account the cartographic rule of a maximum detail of

Figure 2. Flowchart of the method used for landslide risk assessment at the provincial scale. See text for explanation.

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0.5 mm at the scale of the final map (1:100,000 scale in this case), resulting in maps with 2475543 pixels. 5

Debrisflows

HAZARD ASSESSMENT

The casual factor maps were selected based on literature (Carrara et al., 1991, Soeters & van Westen, 1996, Guzzetti et al., 1999) and on the data available in Cuba. They were separated into 3 groups: morphometric factors, ground conditions, distance related factors, and triggering factors. A DEM was created using the ArcGIS© ‘‘topo to raster’’ tool, and four morphometric parameter maps were extracted: slope steepness, slope orientation (aspect), internal relief (vertical dissection) and drainage density. Existing geological, geomorphological and soil maps were used and reclassified by reducing the number of legend units to only those that were considered relevant for landslide susceptibility assessment. The landuse map, which was also obtained from existing maps, was used both as potential causal factor, and as element at risk for estimating the impact of landslides on agricultural production. Two buffer maps were used: distance to main roads in sloping areas, and distance to active faults. Also two triggering factors were used in the landslide hazard assessment. The first of these was a raster map of maximum expected rainfall in 24 hours for a 100 year return period (Planos et al., 2004), and the was a map of the peak ground acceleration (PGA) with a 10 percent exceedance probability in 50 years (García et al., 2003). As part of the hazard analysis two methods were applied for estimating spatial probabilities: Weights of Evidence (WofE) modeling (Bonham-Carter, 1996) and Artificial Neural Network (ANN) analysis (e.g. Lee et al., 2004). The selection of the relevant causal factor maps for each of the five landslide types was made based on initial results of WoE modeling, and expert judgment. The susceptibility maps for 4 out of 5 landslide types were generated using WoE modeling (See Figure 3), as for each of these the main casual factors could be clearly separated, and also because the number of events for each of these was relatively small (See Table 1). For the generation of the susceptibility map of slide-type movements it was decided to use the ANN method, because there were several different causal mechanisms for this landslide type, that were difficult to separate, and also because the number of events was substantially larger than for the other types. The landslide inventory database was randomly subdivided in three subsets: a training set (75% of the landslides) used to optimize the weights, a validation set (12.5%) used to stop the network algorithm before the network starts learning from noise in the data, and a test set (12.5%) to evaluate the prediction capability of the network. An equal number of samples was

Slides

Rockfalls

Large Rockslides

Topples

Figure 3. types.

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Susceptibility maps for five different landslide

Table 2. Results for the annual hazard probability calculated as product of event probability P(E), spatial probability P(S), and temporal probability P(T).

Figure 4. Success rate curves for the five landslide susceptibility maps.

also randomly taken in non-landslide areas. Due to the small scale of the study and the relatively large pixel size it was decided not to include a runout analysis as part of the landslide susceptibility assessment. The five susceptibility maps (See Figure 3) were validated using the landslide inventory and success rate curves were generated (See Figure 4). The results showed generally a very good fit, especially for the topples, large rockslides and rockfalls that occur under very specific conditions. The success rate curves were also used to classify the susceptibility maps with approximately equal percentages of the total number of landslides (e.g. ∼70% of all landslides in the highest class). The next step in the analysis was the conversion of the susceptibility maps into hazard maps. For this three probabilities were calculated for pixels belonging to each hazard class within the five maps: • Event probability, P(E), defined as the probability that if a landslide occurs of a given type, it happens in the particular susceptibility class. • Spatial probability, P(S), defined as the probability that if a landslide occurs within a given susceptibility class, a pixel in this class might be hit. • Temporal probability, P(T), defined as the annual probability of occurrence of a particular landslide type. The event probability and spatial probability were calculated based on the area of landslides within each susceptibility class, in relation to either the total area of landslides (for P(E)) or the total area of the class (for

Rockfall

P(E)

P(S)

P(T)

Hazard

None Low Moderate High

0.00 0.04 0.17 0.79

0.00 6.00E−04 5.20E−03 4.76E−02

0.02 0.02 0.02 0.02

0 5.00E−08 1.71E−06 7.52E−05

Rockslides

P(E)

P(S)

P(T)

Hazard

None Low Moderate High

0.00 0.08 0.21 0.70

0.00 3.40E−03 2.53E−02 2.37E−01

0.01 0.01 0.01 0.01

0 2.70E−07 5.36E−06 1.67E−04

Topples

P(E)

P(S)

P(T)

Hazard

None Low Moderate High

0.01 0.09 0.24 0.66

0.00 6.00E−04 2.90E−03 3.37E−02

0.02 0.02 0.02 0.02

0 1.20E−07 1.36E−06 4.48E−05

Debrisflow

P(E)

P(S)

P(T)

Hazard

None Low Moderate High

0.00 0.05 0.25 0.70

0.00 5.09E−05 5.92E−04 2.71E−03

0.05 0.05 0.05 0.05

0 1.27E−07 7.40E−06 9.48E−05

Slides

P(E)

P(S)

P(T)

Hazard

None Low Moderate High

0.00 0.05 0.25 0.70

0.00 2.49E−04 2.08E−03 8.73E−03

0.05 0.05 0.05 0.05

0 6.24E−07 2.60E−05 3.06E−04

P(S)). Temporal probability was the most difficult to estimate, also in the absence of a historical landslide database. Therefore, based on geomorphological analysis and comparison with return periods for the main triggering events, a return period (RP) of 100 years was selected for large rockslides, a 50 years RP for rockfall and topples, and a 20 year RP for debrisflows and slides. The resulting hazard probability values calculated for each class of the 5 maps are given in Table 2.

6

RISK ASSESSMENT

The risk assessment was carried out for the 5 different landslide types and 5 types of elements at risk, using both a qualitative and semi-quantitative method. Here only results are presented for semi quantitative analysis of the expected number of people that might be killed by landslides annually in the province. This was

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Table 3. Results for the specific risk for population calculated as the product of hazard, vulnerability, and number of persons within a particular hazard class for the 5 different landslide types. Rockfall

Low hazard

Moderate High hazard hazard

Hazard Vulnerability Population Specific risk

5.30E−07 0.6 1616 0.0005

1.71E−05 0.6 522 0.0054

7.52E−04 0.6 200 2338 0.0902 0.0961

2.69E−06 1.0 1265 0.0034

5.36E−05 1.0 453 0.0243

1.67E−03 1.0 184 1902 0.3072 0.3349

1.17E−06 0.2 2139 0.0005

1.36E−05 0.2 745 0.0020

4.48E−04 0.2 12 2896 0.0011 0.0036

Total

Rockslides Hazard Vulnerability Population Specific risk Topples Hazard Vulnerability Population Specific risk Debrisflow Hazard Vulnerability Population Specific risk

1.27E−07 0.4 8047 0.0004

7.40E−06 0.4 553 0.0016

9.48E−05 0.4 0 8600 0.0000 0.0020

6.24E−07 0.7 30490 0.0133

2.60E−05 0.7 2255 0.0410

3.06E−04 0.7 1465 34210 0.3133 0.3676

Slides Hazard Vulnerability Population Specific risk

done by overlaying the hazard maps with a population distribution map, indicating the maximum number of persons in buildings per pixel of 50 by 50 m. Outdoor population and temporal variations of population density were not considered. This results in the number of persons per hazard class as indicated in Table 3. The next step was to estimate the vulnerability of people being killed by a landslide while being indoors, based on the type of landslide and the expected magnitude of the event. These values were based on literature (e.g. Glade et al., 2005) and expert judgment, in the absence of historical landslide damage information.

7

CONCLUSIONS

From Table 3 it can be concluded that the annual population risk for landslides in Guantanamo province is low (0.8 persons/year). As there are no official records

available on landslide casualties it is difficult to validate these results. The method also allows to quantify the risk in monetary values for direct damage to roads, agricultural areas, facilities, and protected areas. The results allow a comparison of annual landslide risk with those from other hazard types, and can form the basis for planning risk reduction measures. In the estimation of the semi-quantitative risk it is important to keep in mind that there are a number of estimated factors that need to be quantified more in detail in future. These relate specifically to the estimation of temporal probability, and vulnerability. Both require the generation and maintenance of a landslide inventory for the province, which also includes actual damage information. Also a more detailed evaluation of the effect of different landslide magnitudes should be taken into account, as well the use of different return periods for the same landslide type and the inclusion of landslide runout assessment.

REFERENCES Bonham-Carter, G.F. 1996. Geographic Information Systems for Geoscientists: Modeling with GIS. Pergamon, Elsevier Science Ltd., 398 pp. Carrara, A. et al. 1991. GIS techniques and statistical models in evaluating landslide hazard. Earth surface processes and landforms, 16(5): 427–445. Castellanos, E. 2000. Design of a GIS-Based System for Landslide Hazard Management, San Antonio del Sur, Cuba, case study. M.Sc. Thesis, International Institute for Aeroespace Survey and Earth Sciences (ITC), Enschede, 108 pp. Castellanos, E. & Van Westen, C.J. 2007a. Generation of a landslide risk index map for Cuba using spatial multicriteria evaluation. Landslides. DOI 10.1007/s10346007-0087-y. Castellanos, E. & Van Westen, C.J. 2007b. Qualitative landslide susceptibility assessment by multicriteria analysis; a case study from San Antonio del Sur, Guant’anamo, Cuba. Geomorphology. DOI: 10.1016/j.geomorph.2006.10.038. Cruden, D. & Fell, R. 1997. Landslide risk assessment. A.A.Balkema, Rotterdam, 371 pp. Dai, F.C., Lee, C.F. & Ngai, Y.Y. 2002. Landslide risk assessment and management: an overview.Engineering Geology, 64(1): 65–87. García, J., Slejko, D., Alvarez, L., Peruzza, L. & Rebez, A. 2003. Seismic Hazard Maps for Cuba and Surrounding Areas. Bulletin of the Seismological Society of America, 93(6): 2563–2590. Glade, T., Anderson, M. & Crozier, M.J. 2005. Landslide Hazard and Risk. John Wiley & Sons, Ltd., Chichester, England, 802 pp. Guzzetti, F., Carrara, A., Cardinali, M. & Reichenbach, P. 1999. Landslide hazard evaluation: a review of current techniques and their application in a multi-scale study, Central Italy. Geomorphology, 31(1–4): 181–216. Iturralde-Vinent, M. 1991. Deslizamientos y descensos del terreno en el flanco meridional de la Sierra Maestra, Cuba

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sudoriental. In: Colectivo de Autores (Editor), Morfotectónica de Cuba Oriental. Editorial Academia, La Habana, pp. 24–27. Lee, S., Ryu, J.-H., Won, J.-S. & Park, H.-J. 2004. Determination and application of the weights for landslide susceptibility mapping using an artificial neural network. Engineering Geology, 71(3–4): 289–302. ONE, 2007. Anuario Estadístico de Cuba 2006. Anuario Estadístico de Cuba. Oficina Nacional de Estadísticas (ONE), La Habana. Pacheco, E. & Concepción, L. 1998. Factores que Originan los Deslizamientos de Tierra. Afectaciones en la Municipio Mariel. In: C.N.d.I.G. (CNIG) (Editor), III Congreso de Geología y Minería. Sociedad Cubana de Geología (SCG), La Habana, Cuba, pp. 528–530.

Pérez, N. 1983. Aspectos ingeniero geológicos del levantamiento geológico de Cuba Oriental. In: Editorial Científico Técnica (Editor), Contribución a la geología de Cuba Oriental. Instituto de Geología y Paleontología. ACC., pp. 173–185. Planos, E., Limia, M. & Vega, R. 2004. Intensidad de las precipitaciones en Cuba, Instituto de Meteorología, Ciudad de La Habana. Soeters, R. & van Westen, C.J. 1996. Slope Instability. Recognition, analysis and zonation. In: A.K. Turner and R.L. Schuster (Editors), Landslide: Investigations and Mitigation. Special Report 247. Transportation Research Board. National Research Council. National Academy Press., Washington, D.C, pp. 129–177.

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Landslide risk management: Experiences in the metropolitan area of Recife – Pernambuco, Brazil Ana Patricia Nunes Bandeira & Roberto Quental Coutinho Civil Engineering Department, Federal University of Pernambuco, Brazil

ABSTRACT: The management system of landslide risk area in Brazil is being improved since the Cities Cabinet setup in 2003. This paper main aim is to be known the model of landslide risk area management of hillside sites applied specifically in Brazilian cities, to introduce the experience in Metropolitan Area of Recife.

1

INTRODUCTION

Urbanization is the urban population increase which happens in a faster cadence than the rural one. Although it is a recent phenomenon (up to the middle of the 19th century, less than 2% of the world-wide population lived in the cities), it is believed that urbanization is an irrevocable process for mankind. According to studies run by the United Nations Population Fund (UNFPA), as of 2008 more than half of the current 6,7 billion inhabitants of the world will live in the cities. Figure 1 below shows the increase of the urban population in the world’s biggest cities and its estimation up to 2020. The urbanization speed, mainly in outskirt countries, came followed by a disorganized urban growth. Therefore, many social and environmental problems came out, such as multiplication of neighborhoods without basic conditions of life, residences built up in risk areas and degradation of natural systems. These

population in thousand

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Figure 1. Urban population increase in the world’s biggest cities from 1950 to 2020 (http://www.unfpa.org.br).

less worthy neighborhoods are, then, occupied by the poorest part of population. In hillside areas, the disorganized occupation is done in an own model, where houses are built on chopped off platforms. The material removed by the gash is thrown over the rim of the hillside without any compacting. So, land slipping is very frequent in the slopes, causing fatal victims. According to Schuster & Highland (2007) the principals triggering mechanism for urban landslides are: • Excessive rainfall (the most common) • Earthquakes • Human alteration of hillsides:

– Cutting slopes at grades that are too steep; – Locating earth fills on top of slopes; – Re-directing rainfall runoff so that flows are concentrated; – Adding water to the slope by landscape irrigation or from septic systems; – Removing trees, shrubs and other woody vegetation.

The accidents for landslides are frequent in almost all the planet, such as Japan, Italy, United States, China and Brazil, representing significant socio-economic impacts (Figure 2). The economic loss in these countries arrives to reach the order of billions of dollar per year (Schuster, 1996). In Brazil, the number of victims for landslides between the years of 1988 the 2006 reached more than 1600 deaths (Figure 3). In the Metropolitan Area of Recife (RMR), from 1984 to 2007, 198 deaths befell (Figure 4). According to Schuster (1996) (see also Cascini et al. 2005), worldwide landslide activity is increasing, this trend is expected to continue in the 21st century, due to

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Figure 4. Fatal accidents in Recife city from 1984 to 2007. (From: Alheiros, 2006 and newspapers).

Russia Filipinas Congo

the increased urbanization and development in landslide prone areas; continued deforestation of landslide prone areas and to the increased regional precipitation caused by changing climate patterns. Schuster & Highland (2007) concluded ‘‘In spite of significant progress in the application of mitigative measures, worldwide population pressures have resulted in increasing landslide hazards on urban hillside slopes. In developing nations, this pattern is being repeated, but with even more serious consequences’’. Having this conscience, it is clear the importance and necessity to manage risk areas.

Vietnã Indonésia Burkina Feso Bangladesh Itália Afeganistão Honduras Nigéria França Turquia Venezuela

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Figure 2. 25 countries with the greatest number of deaths. from: OFDA/CRED (http://www.em-dat.net). OBS: The figure does not include the deceases of the seaquake in Asia, in December 2004.

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Figure 3. Fatal accidents in Brazil from 1988 to 2006 (From: IPT-SP, 2006).

RISK MANAGEMENT STRATEGIES PRESENTED IN THE INTERNATIONAL LITERATURE

Varnes (1984) defines risk as the number of loss of life, damages to the people and properties, and economic activity interrupted due to the phenomenon. For Einstein (1997), risk is a hazard that can lead to entirely different consequences depending on the use of the affected terrain (risk = hazard x potential worth of loss). According to the international literature the risk management process is schematized in three main stages (Figure 5): 1st Risk Analysis; 2nd Risk Assessment; and 3rd Risk Management (Fell and Hartford, 1997 and Fell et al., 2005). Risk Analysis includes hazard analysis and consequence analyses. Consequence analysis includes identifying and quantifying the elements at risk (property, persons), their temporal spatial probability, their vulnerability either as conditional probability of damage to conditional probability of damage to property, or conditional probability of loss of life or injury. Risk Assessment takes the output from Risk Analysis and assesses these against values judgments and risk acceptance criteria.

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Aimed to improve the risk management in Brazil, the government set up the Cities Cabinet in 2003. The Cabinet established the Action of Support to Risks Prevention and Eradication in Precarious Settlement. Due to this Action a new culture of landslide risk management standard procedures had been created, so that they could compare the level of problems among all Brazilian cities in a national scope. The Action is composed by three big activities. (Carvalho e Galvão, 2006):

Figure 5. Risk Management Process (Fell and Hartford, 1997 e Fell et al., 2005).

Risk Management takes the output from the risk assessment, and considers risk mitigation, including accepting the risk, reducing the likelihood, reducing consequences e.g. by developing monitoring, warning and evacuation plans or transferring risk, develops a risk mitigation plan and possibly implements regulatory controls. Landslide risk management involves a number of stakeholders including owners, occupiers, the affected public and regulatory authorities, as well as geotechnical professionals, and risk analysts. According to Schuster & Highland (2007) the mitigation strategies for control of urban landslides are: – technical information related to the hazards and risks; – a technical community of geologists, engineers and urban planners—able to utilize this data base; – a concerned and able municipal government; and a urban population that realizes the value of and supports the hazard reduction program. 3

METHODOLOGY FOR LANDSLIDE RISK MANAGEMENT IN BRAZIL

3.1 Generals aspects Before the world-wide disasters landscape the 90’s were considered, by the United Nations, the International Decade for Natural Disaster Reduction. Among its so many positive results is the shape of four essential strategies directed to the risk management process suggested by UNDRO (1991) and recommended by Brazilian government. Here they are: 1. Identification and Risk Analysis; 2. Physical mitigation measures; 3. Non-physical mitigation measures—preventive plan, monitoring and emergency plans; 4. Educating local and training for prevention.

i. Qualification of the mapping and risk management technicians from municipal civil defense; ii. Development of a Municipal Plan on Risk Reduction comprising proposals of structural measures and non-structural measures towards risk reduction; iii. Development of stabilization projects on hillsides slopes in risk areas. Through the Cities Cabinet Action two National Seminars on Risk Control in Precarious Settlement on Urban Hillside Slopes took place, with special participation of public administrators, municipal, state and federal technicians, specialists and researchers from universities and from research institutions, representatives of professional institutions councils, and Non-Governmental Organizations. During the qualification stage, a total of 396 municipal technicians and administrators, from four different states (Pernambuco, São Paulo, Santa Catarina and Minas Gerais) through extension courses. 1,500 technicians were trained through virtual learning up to 2007. In the elaboration of Municipal Plan on Risk Reduction (PMRR), up to December 2007, fifty-one Brazilian cities started their plans. In February 2006, federal government sent financial support to stabilization projects of hillside slopes in mapped areas. By means of this Cities Cabinet action the risk management process in Brazil is being improved on the platform of the four mentioned strategies of UNDRO, detailed as follow. 3.2

Strategies for risk management process

3.2.1 Identification and risk analysis The first stage, identification and risk analysis, comprise the first stage of International model: Risk Analysis. The identification and risks analysis include identifying and quantifying the elements at risk. A risk diagnosis must inform where evidences of geologicgeotechnical processes exist that, potentially, can cause damages to the population, the constructions or the infrastructure and describe them; to establish some gradation of the situations; and esteem of damage to property.

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Different risk degrees are attributed in the risk analysis, objectifying to present the intervention priorities: R1 (low risk), R2 (medium risk), R3 (high risk) and R4 (very high risk). In Brazil Risk Assessment stage is not disseminated yet. This risk criterion considers that the medium and low risks can be acceptable. After risk identification and analysis it is possible to map the information. The mapping methods’ main aim is to identify and to characterize risk areas in order to implement public politics of management. According to the level of detail the risk cartography can be zoning (1:5.000) or cadastral (1:2.000). 3.2.2 Physical mitigation measures This strategy focus on the execution of a structural interventions plan through engineering measures, in a planned way, aimed to reduce risks. Many municipal experiences have been showing good results since it incorporates the intervention projects to reduce risks among the global urbanization projects of risk areas. On the other hand, it is necessary to enhance the situations as reference to give priority to interventions and also to perform permanent monitoring whilst structural engineering measures are done. 3.2.3 Non-physical mitigation measures— preventive plan, monitoring and emergency plans Non-physical Mitigation Measures stage points out that it is necessary to plan and to implement permanent monitoring pattern and accident prevention measures in risk areas whilst emergency or conclusive interventions are done. The risk control and inspection actions with the best results in Brazilian cities have got the following features (MCidades, 2006): 1. Systematic and periodical inspections; 2. Ongoing record of field-collected information— up-to-date risk mapping; 3. The teams responsible for monitoring each area must be composed by the same agents; 4. Availableness of public attendance duty and other communication channels; 5. Implementation of public equipments in risk areas with high concentration of people; 6. Setup of Civil Defense Community Centers consisted by risk area residents, volunteers and community leaderships, who must be informed and qualified to involve the people in inspection, monitoring and prevention actions of risk areas (Shared Management). Comparing to the international methodology, theses stages non-physical and physical mitigation measures and educating local and training for prevention meet Risk Management stage.

3.2.4 Educating local and training for prevention The Educating Local and Training for Prevention stage may be developed through the following activities: • Lectures for public administrators and the community; • Training courses available for Civil Defense teams, firefighters, and City Hall officers, not only on prevention, but also for emergencies; • Organization of technical handbooks to all the officers involved and primers to guide the population.

4 4.1

THE EXPERIENCE OF METROPOLITAN AREA OF RECIFE Generals aspects

The prevention of accidents connected to landslides must be priority at City Hall. Some Town Halls in Metropolitan Area of Recife took its responsibility and established risk prevention programs, through the setup of Town Hall Civil Defense Commissions, which comprises the community and takes effect action programs. In spite of a few experiences, many towns are not conscious of the importance of preventive public policies of risk management. So, the challenge is to get efforts together from all society due to prop up the governments that have already developed well succeeded policies and to promote knowledge transfer to the most vulnerable towns. Metropolitan Area of Recife has got some metropolitan programs which are applied to get the risk management better. One of them is called ‘‘Live Hill Program’’. It was created in 1997 and has been responsible for many structural and non-structural intervention actions in slope areas. This program leads to important activities comprising not only state and municipal organizations, but also research institutes. Through the metropolitan program it was accomplished in September 2006, in Federal University of Pernambuco, a course on slope restraint for municipal technicians. It was also accomplished in Metropolitan Area of Recife courses on how to setup a Civil Defense Community Center, focused on 14 to 18-yearold teenagers from six different public schools in the nearby risk areas. The program was responsible for the production technical publication (manuals, guides, urban communities education publications), giving information to technicians and to the community in general. The financial distribution from the federal government is also managed by the program, in order to be applied in structural actions in Brazilian cities. Figure 6 describes the number of loss of life caused by landslides on the slopes of Recife. It is noticed a downsize on the incidents due to a better management by the rulers (after

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50

43

45

the rainfall critical that triggering the landslides of hillsides.

loss of life

40

33

35 30 25 20 15 10 5

12 1

5

11

9

8

1

1

2

1

1

1

1984 1989 1990 1991 1994 1995 1996 1997 2000 2002 2003 2004 2005 2006 2007

Figure 6. Reduction of the number of loss of life in Recife after year 2000.

2000) in the three levels of administration: federal, state and municipal. The Metropolitan Area of Recife took the benefit from a federal action on qualifying its technicians and on the management of risk areas. A whole of 155 technicians had been trained in 4 qualifying courses on Risk Mapping coordinated by the Civil Engineering Department of Federal University of Pernambuco. The qualifying of the municipal administration officers is of outstanding importance to burst with the usage of emergency actions. Currently, there is an agreement on the need of higher investments in risk management and risk reduction. Therefore, it is necessary to qualify the biggest number of technicians who act directly in risk areas. In order to organize Risk Reduction Plans, most of the towns comprised by the Metropolitan Area of Recife took federal investments. Through these plans it is possible to make a diagnostic of the risk areas, not only in the town level, but also in metropolitan one. 4.2

CONCLUSION

2

0

Risk management improvement hints for recife metropolitan region

Despite the increase on risk management actions in Metropolitan Area of Recife and the decrease of the number of victims, there is much to be done to improve the efficiency of the process. As far as the dynamic of population enlargement is very fast, and the volume of projected measures is not able to reduce risk in a short spell, it is necessary to take non-structural measures through the organization of Civil Defense Preventive Plans. To improve the risk management through preventive plans, it is necessary, at first, the knowledge of technical patterns of the areas. It must be known the mechanisms of instability in the different soil types and geological arrangements (Barreiras Formation and residual soil of granitic rocks) presenting the region, to understand the faced problems. The rainfall monitoring in the various areas is also important, since the water is one of the main agents responsible to start the mass movements. So, the dynamic of the rainfall water on the different soil types and geological arrangements must be monitored. If possible, it must be known

Brazil is amending its risk management process in the three levels of government: federal, state and municipal. In the federal level it was organized the Action of Support to Risks Prevention in Precarious Settlement, through the setup of the Cities Cabinet in 2003. This federal program focused in three main activities: i) Qualification of the mapping and risk management technicians from municipal civil defense; ii) Development of the risk reduction plan; iii) Development of stabilization projects on hillsides slopes in risk areas; In the state level, Metropolitan Area of Recife counts on the action of the ‘‘Live Hill Program’’ which manages the metropolis actions. In municipal scope, many cities are already conscious of taking permanent actions in risk areas, through the setup of Town Hall Civil Defense. The need is to step forward towards researches on technical pattern which could assist preventive plans and alert systems in risk areas. These technical patterns study will be performed by the engineer Ana Patrícia Nunes Bandeira, through the development of her doctoral thesis, in the Civil Engineering post-graduation program from Federal University of Pernambuco. Comparing the process of risk management in Brazil to the one presented through international literature, it can be concluded that they follow the same working frame, with similar aspects to be developed in order to risk reduction. REFERENCES Alheiros, M.M. 2006. O Plano Municipal de Redução de Riscos. In Guia para Elaboração de Políticas Municipais—Prevenção de Riscos de Deslizamento de Encostas. Brasília, Ministério das Cidades. Cap. 5, pp. 56–75. Carvalho, C.S. & e Galvão, T. 2006. ‘‘Ação de Apoio à Prevenção e Erradicação de Riscos em Assentamentos Precários’’. In: Prevenção de Riscos de deslizamentos em Encostas: Guia para Elaboração de Políticas Municipais. Organizadores: Celso Santos Carvalho e Thiago Galvão—Brasília: Ministério das Cidades; Cities Allience. Cascini et al. 2005. Int. Conf. on Landslide Risk Management—Edited by Hungr, Fell, Couture and Eberhardt. Einstein, H.H. 1997. Landslide risk—systematic approaches to assessment and management. In: International Workshop on Landslide Risk Assessment, Honolulu, Rotterdam: Balkema, pp. 25–49. Epoca Magazine. 2007. Urbanização, sim; favelização, não. No 486, september, Brazil. Fell & Hartford. 1997 Int. Workshop on Landslide Risk Assessment—Edited by Cruden & Fell.

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Fell et al. 2005. Int. Conf. on Landslide Risk Management—Edited by Hungr, Fell, Couture and Eberhardt. http://www.unfpa.org.br. http://www.em-dat.net. IPT. 2006. Instituto de Pesquisas Tecnológicas do Estado de São Paulo. Schuster & Highland. 2007. The Third Hans Cloos Lecture. Urban Landslide: socioeconomic impacts and overview of mitigative strategies. In Bulletin of Egineering Geology and the Environment. Official Journal of the IAEG, Vol 66, No 1, March 2007. pp. 1–27.

Schuster. 1996. Socioeconomic significance of landslide. In Landslides Investigation and Mitigation, National Academy of Sciences, Washington, Transportation Research Board Special Report 247, pp. 12–35. UNDRO. 1991. United Nations Disaster Relief Office. UNDRO’S approach to disaster mitigation. UNDRO News, jan-febr 1991. Geneva: Office of the United Nations Disaster Relief Coordinator. 20p. Varnes, D.J. 1984. Landslide Hazard Zonation: A Review of Principles and Practice. pp. 63. UNESCO, Paris.

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Societal risk due to landslides in the Campania region (Southern Italy) L. Cascini, S. Ferlisi & E. Vitolo Department of Civil Engineering, University of Salerno, Italy

ABSTRACT: This paper deals with the societal risk due to landslides in the Campania Region (Southern Italy). To this end, F∼N curves are generated by using historical data of the fatal landslides which have occurred over the whole Region as well as in three homogeneous geo-environmental contexts, where the most destructive phenomena can occur. The results obtained show that the societal risk due to landslides in the Region is very high. Moreover, the analysis of the incident data clearly highlights the most prone areas to catastrophic events, so turning out to be useful to public Authorities in charge of the landslide risk mitigation strategies by emergency plans and warning systems.

1

INTRODUCTION

In the landslide risk management process (Fell et al. 2005), a fundamental stage of landslide risk assessment is represented by the definition of acceptable and tolerable risk that, in turn, commonly concern individual, societal and total risks. Societal risk is usually measured in terms of the annual cumulative probability F that N or more lives will be lost (i.e. an F∼N plot). In the case of landslides, F∼N plots represent the cumulative probability per year that landslides will cause N or more fatalities versus the number of fatalities resulting from landslides (Fell & Hartford 1997). The F∼N plots are usually obtained from historical data of fatal landslides. According to Düzgün & Lacasse (2005), these plots can be usefully referred to different geographical units (country, region, province, . . . ). In this regard, the quality and reliability of the available incident dataset as well as the completeness of the catalogue need to be ascertained. Referring to the Campania Region (Southern Italy), whose territory is systematically affected by hydrogeological disasters (Cascini et al. 2002, Cascini & Ferlisi 2003, Esposito et al. 2003, Migale & Milone 1998), in this paper an extensive dataset of landslide events with human consequences, spanning from the 5th century up to now, is used to firstly reveal the temporal distribution of both harmful events and related fatalities. Then, the F∼N plots referring to fatal landslides occurred in the whole territory of the Campania Region, as well as in three homogeneous geo-environmental contexts (detected within the Campania Region) are presented. In particular, the F∼N

plot referring to the whole territory of the Campania Region is compared with that of the Italian territory (Guzzetti 2000) in order to assess its societal risk level due to landslides. Finally, the usefulness of the obtained results in the field of landslide risk mitigation at regional scale is discussed.

2

STUDY AREA

The study area (Figure 1) corresponds to the Campania Region (Southern Italy), covering a territory of 13590 km2 and including 551 municipalities.

NAPLES

PIZZO D’ALVANO MASSIF VESUVIUS VOLCANO

ISCHIA

PROCIDA POZZANO

CAVA DE’ TIRRENI

SALERNO AMALFI COAST

Figure 1. Italy).

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The study area of the Campania Region (Southern

N W

E S

Sarno Mounts Vesuvius Volcano

Picentini Mounts

LEGEND Carbonate bedrock (A)

Figure 3. Damage to properties caused by flowslides in Salerno (October 24, 1954).

Tuff and lava deposits (B) Flysch and terrigenous bedrock (C) 0

10

20

30

40

50 km

Figure 2. Map of the areas in Campania Region where pyroclastic soils cover different bedrocks (modified from Cascini et al. 2005).

Due to the complexity of its geological-structural setting, the Region is prone to different types of landslides. However, those associated to the highest consequences, in terms of loss of human life, concentrate in the portion of the study area where pyroclastic soil deposits are present. Within the Region, the abovementioned deposits cover three main peculiar geo-environmental contexts (Figure 2). The first context (A) coincides with large sectors of the Campanian Apennine chain, constituted by Mesozoic carbonate rocks where the presence of thin layers (0.1 to 5.0 m) of pyroclastic covers can be recognized. The second context (B) corresponds to the Phlegraean district, including the city of Naples as well as the islands of Ischia and Procida. This context is characterised by a bedrock of Late Pleistocene volcanic tuffs and lavas, overlaid by pyroclastic deposits that can reach a thickness of several metres (Calcaterra et al. 2004). Finally, the third context (C), mainly located to the north-west of Sarno and Picentini Mounts, includes thin pyroclastic deposits (20–30 >30–40 >40–50 >50 Flat N NE E SE S SW W NW −4–−2 >−2–−1 >−1–1 >1–2 >2–4 >4 −3–−2 >−2–−0.7 >−0.7–0.5 >0.5–2 >2 −2–−1 >−1–−0.4 >−0.4–0.3 >0.3–1 >1–2 >2

0.14 0.06 0.01 0.04 0.08 0.14

Ratio Bedding Slope Slope angle

Aspect

Curvature

Profile curve

Plan curve

3

method based on the percentage of the area covered by failures in each class of the input factors:   Afi Pi = (1) Ai where Afi is the area covered by failures in a given class and Ai is the area of this class. This percentage represents the weight or degree of influence of each class in terrain-failures (Dai et al. 2002, Guinau et al. 2005) listed in Table 1. Combining the weights determined for each class we obtained the landslide susceptibility values (LS):

0.00 0.03

LS = P ([LITHOLOGY ] ∪ [RBS] ∪ [SLOPE] . . . ∪ An ) (2)

Two prediction images are generated and compared: one takes into account DEM derived factors and lithology and Ratio Bedding Slope, the other takes into account only DEM derived factors. In both of cases LS ranges from 0.04 to 0.7. For the calculation vector datasets (i.e. lithology and RBS) were converted to a 10 × 10 m grid, and raster datasets (i.e. slope, aspect, curvature, profile and plan curve) were derived from a 10 × 10 m DEM. In order to compare the two prediction images the success-rate curves of the susceptibility assessment were generated (Figure 5). These curves are computed sorting into descending order the susceptibility values determined for each pixel and comparing the area predicted as susceptible with the landslide data set. For example, in the case of successrate curve 1 of Figure 5, the 19% high susceptible area, with a value of LS > 0.3, includes the 87% of the total landslide area. Although these results are good, the success-rate curves measure a goodness of fit assuming that the models are ‘‘correct’’ (Chung & Fabbri 2003) and can’t be used as a validation of the models, as it is based on the comparison between the prediction images and the landslides used in the modelling.

LANDSLIDE SUSCEPTIBILITY ANALYSIS

3.1 Subaerial Landslide susceptibility analysis for the subaerial slopes has been performed by using a direct estimation

Figure 5. Subaerial slopes comparison between the successrate curve of the prediction image generated taking into account DEM derived factors and lithology and RBS (1), and the success-rate curve of the prediction image generated taking into account only DEM derived factors (2).

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In this case the success-rate curves were used to compare two different models and the results show that the prediction image generated taking into account only DEM derived factors resulted less accurate, but the results were quite similar to the first prediction image. 3.2

Subaqueous

On the basis of the good results obtained for the subaerial slopes taking into account only the DEM derived factors, in a second phase of the research the authors attempted to extend the results obtained from the subaerial spatial analysis to the submerged slopes in order to evaluate the corresponding landslide susceptibility. Thus the calculated weights of each class of DEM

derived factors (Tab. 1) were used to generate a prediction image for the submerged slopes (Figure 6). The latter was compared with the distribution of submerged landslides identified on morphological basis and the results showed a quite good correlation: the prediction-rate curve of Figure 7 provides the validation of the prediction. The 14% of the most susceptible area, with LS > 0.2, includes the 65% of the landslide area. The 28% of the total area classified as more susceptible to slope instability has a value of LS > 0.1, and includes the 88% of the landslide dataset. These results confirm the validity of the prediction. Furthermore, the computation of the prediction-rate curve allows to interpret the landslide prediction (Chung & Fabbri 2003, Zêzere et al. 2004). For the maximum

Figure 6. Landslide susceptibility map of the submerged area.

Figure 7. Prediction-rate curve of the susceptibility assessment of the submerged area.

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precision in the prediction we must consider values of LS > 0.1, and we can also rank the submerged area as high (LS > 0.2), medium (0.2 > LS > 0.1) and low (LS < 0.1) susceptibility. 4

CONCLUSION

The obtained results show that the prediction image of the submerged area generated by using the weights of the DEM derived factors calculated for the subaerial slopes, fits quite good the distribution of the recognized landslides. This evidence imply a similar landslide susceptibility in both subaerial and subaqueous areas. Furthermore, the close relationship between the prediction and the presence of subaqueos landslides shows the efficacy of the morphological survey based on an high resolution bathymetry. The detected drawbacks of the prediction image are probably due to: i) the presence of lacustrine deposits, whose geothecnical behaviour significantly differs from the one of the mainly rocky subaerial slopes and not included in the lithology classes of Table 1; ii) the possible presence of submerged landslides not easily detectable only by means of DEM-derived geomorphological interpretations. This factor can influence the final validation of the analysis, which is based on the comparison between the results of the analysis and the recognized landslides. Nonetheless, this attempt can be considered as a starting point for further and more detailed analyses, by considering that the preliminary results can be judged encouraging. REFERENCES Anzidei, M., Esposito, A. & De Giosa, F. 2006. The dark side of the Albano crater lake. Annals of Geophysics 49: 1275–1287. Bacon, R.C., Gardner, J.V., Mayer, L.A., Buktenica, M.W., Dartnell, P, Ramsey, D.W. & Robinson, J.E. 2002. Morphology, volcanism, and mass wasting in Crater Lake, Oregon. GSA Bulletin 114: 675–692. Baiocchi, V., Anzidei M., Esposito, A., Fabiani, U., Pietrantonio G. & Riguzzi F. 2007. Integrer bathymetrie et lidar. Geomatique Expert 55: 32–35. Bohannon, R.G. & Gardner, J.V. 2004. Submarine landslides of San Pedro Escarpment, southwest of Long Beach, California. Marine Geology 203: 261–268. Bozzano, F., Chiocci, F.L., Mazzanti, P., Bosman, C., Casalbore, D., Giuliani, R., Martino, S., Prestininzi, A. & Scarascia Mugnozza, G. 2006. Subaerial and submarine characterisation of the landslide responsible for the

1783 Scilla tsunami. EGU 2006, Geophysical Research Abstracts 8, 10422. Chacòn, J., Irigaray, C., Fernàndez, T. & El Hamdouni, R. 2006. Engineering geology maps: landslides and geographical information systems. Bull. Eng. Geol. Environ. 65: 341–411. Chung, C.F. & Fabbri, A.G. 2003. Validation of spatial prediction model for landslide hazard mapping. Natural Hazard 30: 451–472. Dai, F.C., Lee, C.F. & Ngai, Y.Y. 2002. Landslide risk assessment and management: an overview. Engineering Geology 64: 65–87. De Rita, D., Funiciello, R. & Pantosti, D., 1986. Dynamics and evolution of the Albano crater, south of Rome. Proc. IAVCEI Int. Conf., Kagoshima: 502–505. Freda, C., Gaeta, M., Karner, D.B., Marra, F., Renne, P.R., Taddeucci, J., Scarlato, P., Christensen, J.N. & Dallai, L. 2006. Eruptive history and petrologic evolution of the Albano multiple maar (Alban Hills, Central Italy). Bull. Volcanol. 68: 567–591. Funiciello, R, Giordano, G. & De Rita, D. 2003. The Albano maar lake (Colli Albani Volcano, Italy): recent volcanic activity and evidence of pre-Roman Age catastrophic lahar events. Journal of Volcanology and Geothermal Research 123: 43–61. Giordano, G., De Rita, D., Cas, R. & Rodani, S. 2002. Valley pond and ignimbrite veneer deposits in the small-volume phreatomagmatic ‘Peperino Albano’ basic ignimbrite, Lago Albano maar, Colli Albani volcano, Italy: infuence of topography. Journal of Volcanology and Geothermal Research 118: 1 31–144. Guinau, M., Pallàs, R. & Vilaplana, J.M. 2005. A feasible methodology for landslide susceptibility assessment in developing countries: A case-study of NW Nicaragua after Hurricane Mitch. Engineering Geology 80: 316–327. ISRM 1978. Suggested methods for the quantitative description of discontinuities in rock masses. International Journal of Rock Mechanics Sciences & Geomechanics 15: 319–368. Longva, O., Janbu, N., Blikra, L.H. & Bøe, R. 2003. The 1996 Finneidfjord slide; seafloor failure and slide dynamics. In J. Locat & J. Mienert (eds), Submarine Mass Movements and their Consequences: 531–538. Dordrecht: Kluwer. Schnellmann, M., Anselmetti, F.S., Giardini, D., McKenzie, J.A. & Ward, S.N. 2002. Prehistoric earthquake history revealed by lacustrine slump deposits. Geology 30: 1131–1134. Zêzere, J.L., Reis, E., Garcia, R., Oliveira, S., Rodrigues, M.L., Vieira, G. & Ferreira A.B. 2004. Integration of spatial and temporal data for the definition of different landslide hazard scenarios in the area north of Lisbon (Portugal). Natural Hazards and Earth System Sciences 4: 133–146.

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Landslide susceptibility zonation of the Qazvin-Rasht-Anzali railway track, North Iran H. Hassani Faculty of Mining, Metallurgical and Petroleum Engineering, Amirkabir University of Technology, Tehran, Iran

M. Ghazanfari Mining Exploration, Geological Survey of Iran (GSI), Tehran, Iran

ABSTRACT: Landslide Hazard Zonation (LHZ) is a method proposed to evaluate risk where there is the potential for landslides. Many factors affecting the occurrence of landslides are analysed. The factors contributing to the hazard in an area can usually be identified, results of the investigations frequently being presented as a landslide hazard zonation map which indicates zones of similar risk of landslide occurrence. Qazvin-RashtAnzali railway track is one of the areas susceptible to landslides in Iran, and in the area several landslides were recorded. Landslides caused damage or disturbance to villages, farmlands and railway, exacerbation of the superficial erosion and consequently the rate of transportation of sediments. The method of landslide zonation used in this study is based on simple grid unit. The factors contributing to the occurrence of landsliding were analysed, including lithology, slope, tectonic and seismotectonic, human activity and hydrology. For each grid unit, the incidence of landsliding and an assessment of the contributory factors were recorded in terms of a surface percentage index. A computer program was written to calculate the LHZ for each unit. This was used to prepare the LHZ map. The study area has been classified into five classes of relative landslide hazard, namely, very low, low, moderate, high and very high. The result shows that 7 % of the study area has a high or very high landslide hazard. 1

INTRODUCTION

The Qazvin-Rasht-Anzali railway area is part of the Alborz structural zone in Iran characterized complex geology (Alavi, 1996), active seismicity and seasonal rainfall. These factors have a nearest influence on the development of natural hazards, including landslides and other types of ground mass movements which may result in loss of life and damage to the railway track. The landslide hazard zonation map is the first step in assess the degree of hazard and to evaluate its potential (Mora & Vahrson, 1993). Qazvin-Rasht-Anzali railway track, an area of about 720 km2 , is situated between Qazvin and Gilan province, in the north of Iran (Fig. 1). It is located between longitude (49◦ and 50◦ 30′ E) and latitude (36◦ and 37◦ 30′ N). The temperature range between −11◦ C in winter and 33◦ C in summer. The mean annual rainfall is 350 mm. The maximum precipitation falling between March and May, coincident with the main landslide activity. The study area drains into the Sefid-Roud river which flows in the Caspian Sea. The aims of the study were to identify and zone the areas susceptible to landslides, to evaluate the hazard,

to determine the main factors involved in the initiation of the instabilities and to consider how their effects may be reduced. 2

GEOLOGICAL SETTING

The general geological setting of the area is shown Fig.1. The study area is located in the Takestan, Roudbar and Rasht quadrangle maps of the northwest of the Alborz Structural Zone. According to Darvishzadeh, 2002, the oldest rock units in the area are limestones and sandstones of Triassic age. The Triassic rocks have a surface distribution of only 3% and over approximately 2% of the area is covered by Cretaceous limestones. The majority of the area (75%) is covered by Tertiary volcanic and plutonic rocks, Neogene red-beds and Jurassic rocks, mainly andesite, tuff, shale, marlstone and sandstone. Quaternary materials, young-medium and old terraces have surface distribution about 20%. 2.1 Tectonic characteristics In the study area three structural zones are identified: 1-Mountain uplift: this zone, in the north-east and

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Leg

Pyroclasi S d

Figure 1. Geological map of Qazvin-Rasht-Anzali railway area.

north-west of the area, is part of the Alborz mountain range. It is characterized by faults and folds, having been elevated by pre-Alpine orogenic activity. Lithologies from Triassic to Cretaceous age are present. 2- Elevated plateau: this zone consists of Neogene red-beds and Jurassic rocks, mainly marl, siltstone and conglomerate. 3- Depressions: this zone is covered by young Quaternary alluvial plain and alluvium terraces sediments, which are very susceptible to weathering and erosion. Several major and minor faults and folds (anticline and synclines) have been identified in the area. The longest fault, the Harzevil fault, extends for about 140 km. The fault is a thrust and has trend of N110◦ . There is evidence for some Quaternary fault activity. 3

LANDSLIDE RECORDS

The study area has experienced many types of ground movement over a long period and more than 14 landslides (Fig. 2) have been recognized. Initially, the

Figure 2. Landslide distribution in the Qazvin-RashtAnzali area.

location of the landslides was plotted onto 1:50000 topographic maps. Using 1:20000 aerial photographs. Site visits were then undertaken to confirm and complete this information and to identify the exact location of the landslides on the ground surface. 4

ANALYSIS

Many factors influence the occurrence of landslides in the area like lithology, tectonics, slope angle, land use, rainfall and seismicity. To evaluate the rate of landslide distribution with regard to each factor, a Landslide Hazard Zonation (LHZ) expression was used. This can be defined as: LHz = (CL × FL ) + (CI × FI ) + (CF × FF )

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+ (CD × FD ) + (CR × FR ) + (CS × FS )

(1)

where:

Table 2. (FI).

FL = Lithology factor FI = Slope factor FF = Fault length factor FD = Road and River length factor FR = Precipitation factor Fs = Seismic factor CL = Lithology coefficient CI = Slope coefficient CF = Fault length coefficient

Slope angle (degree)

45

Effect in instability of amplitudes

No effect

Very low

Low

Moderate

High

Very high

Factor (FI )

0

1

2

3

4

5

Table 1. The amount of relative coefficient factors produced by engineering judgment. CS

CR

CD

CF

CI

CL

2

1

0.75

0.50

1

0.75

The effect of slope angle in instability of slopes

CD = Road and River length coefficient CR = Precipitation coefficient Cs = Seismic coefficient The coefficient, of C, is related to the effect of each factors on instability and produced by engineering judgment (Table 1).

5

GROUND SLOPE

Slope angle is a major factor influencing the development of landslides. As the slope angle increases, the weight and consequently the volume of material affected per unit area will also increases. The type of slope in any area is closely related to the strata and geological history of the region (Varnes, 1984 and ). Topographic maps (1:50000) were used to clarify the slopes in the study area. The aspect and angle of slopes in the area were also analysed (Fig. 3). The area was classified in six classes considering the slope angle using topographic maps and a statistical survey in relation to the slope degree in each area. The results of this classification are recorded in Table 2.

6

Figure 3. Ground slop zonation map of Qazvin-RashtAnzali area.

LITHOLOGY

The geology of the area is controlled by two main factors: the lithology and the bedding. Different lithologies have a different susceptibility to weathering processes and erosion agents and result in different types of weathering products, hence the possibility of different types of slope failure (Parise, 2002). The age and stratification of the rock unit are also important influencing factors. The erodability or the response of rocks to the processes of weathering and erosion should be the main criterion in awarding the ratings for lithology. Rock types such as unweathered quartzites, limestones and granodiorites are generally hard and massive and more resistance to weathering, whereas ferruginous

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Table 3. Landslide Hazard Evaluation Factor (FL) of different rock types in the study area.

II

III

IV

V

VI

VII

VIII IX X

1

2

3

4

7

5

STRUCTURAL SETTING

Structural factors, including layering levels, joints, faults and folds are among the important and influential factors in causing instability in slopes. Considering the role of structural setting in the instabilities in the study area, faults and joints are considered as major structural factors in zoning the landslide risk. Thus, the effect of structural factors in increasing the risk of instability of slope was classified in five categories. The results are shown in Table 4. Figure 4 shows the histogram of the length of faults based on the grid map of the project area.

6

7

8 9

Table 4.

10

The Effect of Fault Length in Instability (FF).

sedimentary rocks are more vulnerable to weathering and erosion. In case of soil-like materials, the genesis and age are the main considerations in

251– 1401

washes away and carries materials away destabilizing the slopes and increasing the slope angle. Considering the instabilities occurring in the area of the project, the factors related to the trenches of road and railways and the length of water canals and rivers are of significant and determining role. On this same basis, the density of length of roads and rivers have been graded and presented in Table 5 and Fig. 5. 9

Figure 4. Histogram of fault density. Table 5. The Effect of Road and River Length in Instability (FD). 501– 1500

1501– 2500

2501– 3500

Very low

Moderately low

Moderate

High

Very high

1

2

3

4

5

Length (m)

< 500

Effect in instability Factor (FD)

>3501

PRECIPITATION

The effect of precipitation on increasing the risk of instability and causing slope movements was also studied. For this purpose, the 50-year (1950 to 2000) precipitation data were gathered from the stations within the study area and neighboring areas and analyzed. The effect of precipitation both in regard to increased soil humidity by increased precipitation and consequently increased potential for landslide and as a starter factor for landslide in the case of heavy and longterm rainfall were studied. The effects of the quantity of precipitation on the instability of slope are shown in Table 6. Figure 6, shows the variogram of precipitation in the north direction. 10

SEISMIC ACTIVITY

Seismic activity is another factor causing slope instability in the study area. Study of the historical Table 6.

The Effect of Precipitation in Instability (FR).

Precipitation 151– (mm) 551

Figure 5. Histogram of road and river length.

8

LENGTH OF ROADS AND RIVERS

In the analysis and assessment of slope factor, excavation and construction of trenches on the route is inevitable due to the restrictions of slopes and necessities of road and railway design. These operations will cause artificial slopes, with a minimum slope of 45◦ along the road and railway route. The rivers are also another factor contributing to the instability of amplitudes. Due to the current water of rivers, along the rivers the valleys undergo erosion and abrasion which

Figure 6.

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Variogram of precipitation in a line to the north.

earthquakes during the 20th century (Ambersis & Melville, 1982 and Nowroozi, 1985) indicates that instabilities are directly and indirectly related earthquakes. Earthquake causes primary tensile fractures in the slope (Slemmons, 1982). The temporary nature of this force causes incomplete instability and imperfect landslides. In the case of repeated earthquakes, the depth and widening of fractures increase and after the earthquake, other factors complete the effect of the earthquake in causing instability. The slopes are directly influenced by the acceleration caused by earthquakes. According to the seismicity studies and field observations of the active seismogenic faults and the zonation map of seismic hazard risk map, the maximum horizontal acceleration resulting from seismic activity is one of the most important factors in causing instability in the landslide hazard. The categorization of maximum horizontal acceleration affective in the study area is presented in Table 7. 11

LANDSLIDE HAZARD ZONATION

After assigning Landslide Hazard Zonation values, a LHZ index was computed for all categories of each factor within each cell. Using the amount of LHZ for each cell, the Landslide Hazard Zonation map of the study area was prepared (Fig. 7). Based on the cumulative value of LHZ obtained, the area was classified into 5 zones of relative instability including, critical (Very High), highly unstable (High), moderately unstable (Medium), moderately stable (Low) and stable (Very Low) classes depending upon the range of cumulative LHZ values (Table 8).

Figure 7. Landslide Hazard Zonation map of QazvinRasht-Anzali railway area. Table 8.

11.1 Critical zone This is a very unstable zone where landslides are likely to occur. The area is degraded to such a state that it is practically impossible to evolve economically and socially acceptable remedial measures which can positively prevent recurrence of the hazard. The area has to be entirely avoided for railway track or other development purposes and preferably left to allow

Landslide Hazard Values of different zones.

Zones

LHZ

Percent area affected

Critical (Very high) Highly unstable (High) Moderately unstable (Medium) Moderately stable (Low) Stable (Very low)

>20.5 20.5–18.5 18.5–16.5

7.5 12.7 50.5

16.5–14.5 S1 or sporadic microseismic strain > S2 or sustained microseismic activity

4 hours 1 hour 15 minutes 3 hours 5 minutes

Vigilance X Alarm

3 hours

X X X X

X X

4.3 Acquisition and monitoring procedure

public safety to set up a permanent early-warning device to ensure anticipation of the collapse. Regarding mine stability, microseismic and deformations are reliable variables to be monitored, as shown respectively by Srinivasan et al. (1999) and Szwedzicki (2001). In the present case, the analysis of the scenario and the established monitoring strategy focused on the potential subsidence phenomena and integrate the consideration of several factors, listed in Table 2. They mainly relate to monitoring: – the flooding of the mining works, the major effect of which is an expected supplementary alteration of the thin inter-layers separating the different mined layers, this by piezometric sensor means; – potential creep of intermediate banks between the mined layers, measured with several convergence; – expansion stations installed before mine closure; – ruptures of the mine pillars, initiating the collapse process which would propagate towards the surface, creating a basin damaging buildings and endangering people’s security. A microseismic array has been set up to get early-warning data while entering in this phase.

4.2

Table 3. site.

Description of the remote measurement system

Thus the early-warning system includes a local microseismic array deployed in the geological overburden and at mine level, as well as deformation and piezometric sensors installed in the old mining works. Three microseismic and geotechnical stations, one dedicated to deep underground measurements and two related to instrumented boreholes are cabled and connected to an acquisition unit installed on site; itself being connected to a monitoring central station located at CENARIS, via secured high speed xDSL.

The main acquisition parameters are described in Table 3. Transmission of measurement data and alarms to CENARIS is immediate in all cases. Upon any change in mode, an information message is transmitted to an on-call person in charge. 5

EXAMPLE OF REMOTE MONITORING OF STAKES

The supervision of a particular stake may be justified when it presents a high level of vulnerability. The necessity of having real-time information for monitoring the stake in the case of crisis management, is then determining. Extension of the early-warning system to monitor stakes may be illustrated by the instrumentation of the namely ‘‘Panoramic tower’’ of Maxéville (France) located above an area with a risk of differed mine collapse from old workings underlying a limestone tableland bordered partially by a steep slope. This 100 meters high residential tower was instrumented at its top level with several high accuracy clinometers in order to directly monitor the impact of a collapse on this building characterized by a low deformation tolerance. Radio transmission was installed to overcome potential difficulty to cable this specific monitoring station to the acquisition unit located far outside the building, (Fig. 4). 6

EXAMPLE OF A DEVICE FOR OBSERVING AN UNSTABLE ROCK SLOPE

In the case of a steep rock slope in the Alps, mechanisms leading to massive rock-fall remain difficult to apprehend. Cyclic deformations of thermal origin seem to play a predominant role in the quasicontinuous surface erosion of this type of geostructures. New fragments and small blocks of rock, very regularly found at the bottom of the slope, confirmed

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Of course beyond scientific objectives potential validation and transfer of new measurement techniques to operational early-warning systems is part of the project.

7

Figure 4. Schematic of the global monitoring system: the case of Maxéville, France.

this degradation. In a global issue concerning the impact of climate changes on natural hazards, INERIS sets up a monitoring system, intended for scientific observation, having several cells for measuring 3D deformations and temperatures profiles (for a total of 120 gauges), a weather station as well as acoustic sensors, according to the factors identified in Table 4. Such an observation network has the purpose of investigating and quantifying this superficial ≪driving force≫ and its possible impact on pre-existing fractures at a depth ranging from a few centimetres to a few meters (Dünner et al., 2007) Here ≪smart≫ acquisition of fully synchronised data on a unique versatile acquisition unit between recorded microseismic ruptures and immediate triggering of deformation measurements–scheduled by default to a dozen cycles of measurement per day– provides a significant benefit for the scientist in charge of establishing a possible cause-to-effect link.

Table 4. Worsening factors and precursory signs usually considered for a landslide. Phenomenon Earth quakes Cyclic change in temperature Rainfall Water flow Micro-seismic internal rupture

Worsening factor

Precursory sign

X X X X X

X

CONCLUSIONS

Optimal risk management of ground failures through early-warning systems imposes a global strategy based on a design specific to each situation. Extensive risk assessment expertise – hazard and vulnerability – is required for an optimized design of the system and a correct operational procedure. Facing the field situation reveals quite often as strong requirements and specifications, from the expectations of the expert responsible for following up the measurement data, and from the decision makers in terms of inherent economical constraints. Nevertheless a strategy thought globally from the very early design and based on stateof-the-art technologies is clearly valuable, in a general societal environment where geohazards have less and less acceptance. In fact, reciprocally, integration of new technological capabilities and performances has a direct structuring influence on the design and quality of the alarm device, upstream, as well as on the warning procedures managed downstream; the latter proving to be simpler for specialists and managers and therefore optimising the decision-making process when handling a crisis.

REFERENCES Bigarré P. Lizeur A., Bennani M. & Felt T. 1995. SYTMIS: software for real-time microseismic monitoring systems, FMGM 95–4th International Symposium, Bergamo, 407–414 Couffin, S., Bigarré P., Bennani M. & Josien J.P. 2003. Permanent real time microseismic monitoring of abandoned mines for public safety. In Myrvoll (ed.), Swets & Zeitlinger, Field Measurements in Geomechanics, Lisse. Dünner C., Bigarré P., Clément C., Merrien-Soukatchoff V. & Gunzburger Y. 2007. Natural and thermomechanical stress field measurements at the Rochers de Valabres Pilot Site Laboratory in France Proc. 11th congress Intern. Society for Rock Mechanics, Lisbon, 9–13 July 2007: 69–72. Leiden: Taylor & Francis. Khazai B. & Sitar N. 2003. Evaluation of factors controlling earthquake-induced landslides caused by Chi-Chi earthquake and comparison with the Northridge and Loma Prieta events. Engineering Geology (71): 79–95 Kilburna C.R.J. & Petley D.N. 2003. Forecasting giant, catastrophic slope collapse: lessons from Vajont, Northern Italy. Geomorphology (54): 21–32. Kolesnikov, Yu. I., Nemirovich-Danchenko, M.M., Goldin, S.V. & Seleznev V.S., 2003. Slope stability monitoring

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from microseismic field using polarization methodology, Natural Hazards and Earth System Sciences (3): 515–521. Leroueil, S., 2004. Geotechnics of slopes before failure. Proc. 9th Intern. Symposium on Landslides, Rio de Janeiro, June 28–July 2, 2004: 863–884. London: Taylor & Francis. Lollino G., Arattano M., Allasia P. & Giordan D., 2006. Time response of a landslide to meteorological events. Natural Hazards And Earth System Sciences 6 (2): 179–184. Olalla, C., 2004. Recent developments in landslide monitoring. Proc. 9th Intern. Symposium on Landslides, Rio de Janeiro, June 28–July 2, 2004: 549–555. London: Taylor & Francis. Srinivasan C., Arora S.K. & Benady S., 1999. Precursory monitoring of impending rockbursts in Kolar gold mines

from microseismic emissions at deeper levels. Intern. J. Rock Mech. and Mining Sciences 36 (7): 941–948. Szwedzicki T. 2001. Geotechnical precursors to large-scale ground collapse in mines. Intern. J. of Rock Mech. and Mining Sciences (38) : 957–965. Tastet J., Contrucci I., Klein E., Bigarre P. & Driad-Lebeau L., 2007. Large-scale field experiment to calibrate microseismic source parameters applied to real-time monitoring of post-mining instabilities.). Proc. 11th congress Intern. Society for Rock Mechanics, Lisbon, 9–13 July 2007: 1147–1150. Leiden: Taylor & Francis. Senfaute G. & Gourry J.C., 2004. Auscultation microsismique appliquée à la détection des éboulements des falaises crayeuses. Journées Nationales de Géotechnique et Géologie de l’Ingénieur (JNGG), Lille, 28–30 juin 2004, pp. 275–284.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Regional slope stability zonation based on the factor overlapping method Jinfeng Liu, Guoqiang Ou & Yong You Key Laboratory of Mountain Surface Process and Hazards, CAS, Chengdu, Sichuan Province, China Institute of Mountain Hazards and Environment, CAS, Chengdu, Sichuan Province, China

Jinfeng Liu Graduate University, CAS, Beijing, China

ABSTRACT: Regional slope stability is to study the possibility of slope deformation and instability from the view of regional scale. The paper took the Maerqu watershed from Makehe to Moba County in the upper reaches of Dadu River as the study area. Gradient, fault, slope type, elevation difference, earthquake intensity, lithology, rainfall and land use were chosen as the main factors to affect slope stability. These 8 factors were classified, quantified and scored firstly. Then the original maps of the 8 factors were overlapped by using the factor overlapping method. Consequently, the regional slope stability of the study area was divided into three types of regions: the stable region, the middle-stable region and the unstable region. The area of the unstable region is 810 km2 which occupies 51.9% of the total study area. It is indicated that the possibility of slope deformation and instability of the study area is high. 1

INTRODUCTION

Slope failures are significant natural hazards in many areas throughout the world. Generally, a slope failure can be defined as a downward movement of a large amount of material. Slope failure hazard mainly includes collapse, landslide and so on (Wu 2004). Regional slope stability is to study the possibility of slope deformation and instability from the view of regional scale. With the increase in human activities, especially the implementation of Chinese Western Development Program, the engineering construction in mountain area is intensifying. Consequently, slope failure hazard happens frequently. It usually causes casualties, burial of railway and highway, threat to the transportation safety, damage to hydraulic engineering and impediment to the economic development in mountain area. So, the regional slope stability study is very important and essential to engineering construction, regional planning, disaster mitigation and economic development in mountainous area. Nowadays, the researches on slope stability mainly focus on three scales: single slope, small watershed and regional slope (De Jong et al. 1999). The slope stability analysis of these researches aims at a specific collapse and landslide, and the scale is focused on single slope and small watershed (Wieczork 1984; Ellen 1988; Mello & Pratson 1999; Kentli & Topal 2004; Ercanoglu & Gokceogl 2004; Qiao 2001). In terms of regional slope, since the topography, geology,

lithology and some other factors differ in thousands of ways in one region; the stability analysis is different from the analysis of single slope and small watershed. In order to replenish the current few researches on regional slope stability, this paper discusses a method for delineating areas liable to slope failure hazards based on GIS. 2

STUDY AREA

This paper takes the Maerqu watershed from Makehe to Moba County in the upper reaches of Dadu River as study area. The study area lies in the southeast Qinghai Province and the northwest Sichuan Province of China (Figure 1). Its area is 1560 km2 . The area lies in the southeast part of Qinghai-Tibet Plateau. The geologic conditions in this area are very complicated. The strata belong to Triassic System which is characterized by steep bed, strong fold, developed active fault, strong neotectonic movements. The geomorphology is characterized by high mountains and deep valleys. This area also lies in the Kekexili-Jinshajiang seismic belt which is one of the strongest seismic belts in the Qinghai-Tibet Plateau. The permafrost and seasonal frozen ground are very developed in this area. Due to the complicated natural conditions and the influence of human activities, the slope failure hazards are rather serious in this mountain-hazard-prone area.

1933

fault, rainfall and land use) are selected as the zonation factors to analyze the regional slope stability. The 8 factors are classified and quantified based on the current researches and the practical conditions in study area. 4.1

Slope gradient is the main factor influencing regional slope stability. JSECE analyzes the relation between slope gradient and the frequency of 1000 slope failure hazards (JSECE 1992). The result indicates that the sequence of frequency from high to low with slope gradient is: 20◦ ∼45◦ , 10◦ ∼20◦ , >45◦ , slope type > slope aspect > elevation difference > lithology > earthquake > fault > land use. After determining the sequence, we use the AHP method to calculate the weight of each factor (Table I). The calculation formula is F=

n  i=1

ri × bi .

gradient is below 10 degree, its quantification score is zero. The RSSI value is equal to zero after overlapping. That is to say, the regional slope stability is stable in the area which slope gradient is below 10 degree. Therefore, this overlapping method can make the analyzing result more realistic. After calculation, the regional slope stability zonation map is obtained based on the RSSI values (Figure 2) and the study area is divided into three areas: stable region (0–40), middle-stable region (40–70) and unstable region (70–100). The researching result shows the stable region is 100 km2 , which occupies 6.4% of the total study area; the medium-stable region is 650 km2 , which occupies 41.7% of the total study area; the unstable region is 810 km2 , which occupies 51.9% of the total study area. There are no historical data in the study area. So, we use the data from field investigation to test the zonation results. We use slope failure hazards as main index and debris flow basin as the secondary index when testing the zonation results. The statistic data is shown in Table 2.

(2)

where F is the zonation index; ri is the zonation factor and bi is the weight of the zonation factor. 6

Figure 2.

FINDINGS

In the research, after classified and quantified, the 8 factors form 8 basic maps. Then the 8 basic maps are converted to grid maps which take 100 m as cellsize. At last, the regional slope stability index (RSSI) of each grid is calculated as RSSI = r1 b1 ∗ r2 b2 ∗ r3 b3 ∗ r4 b4 ∗ r5 b5 ∗ r6 b6 ∗ r7 b7 ∗ r8 b8 .

(3)

In order to analyze the influence of each factor on regional slope stability exactly, we adopt multiplication overlapping method. For example, when slope

Table 2.

The zonation map of the study area. The test of the zonation results.

Zonation results Stable area Middlestable area Unstable area Summation

1936

The amount of slope failure hazards

Ratio of total amount (%)

The area of debris flow basin (km2 )

Ratio of total area (%)

1

12.5

3.7

1.9

2

25.0

68.5

34.9

5 8

62.5 100

124.0 196.2

63.2 100

The statistics shows that 87.5% of total slope failure hazards and 98.1% of the total debris flow area distribute in the middle-stable area and unstable area, and the ratio of slope failure hazards and debris flow area which distribute in the unstable area is the biggest. It indicates the zonation result of the study area is reasonable. 7

CONCLUSIONS

In mountain area, slope failure hazards usually cause great damages. Therefore, it is very important to recognize mountain-hazard-prone area for the economic development in mountain area. This paper discusses a method for delimiting the areas liable to slope failure hazards based on GIS. The zonation results indicate that unstable region account for large part of the study area (51.9%). It is because the study area is located in the northeast edge of Qinghai-Tibet Plateau. In this area, the geologic structure, neotectonic movement and earthquake are very active. The slope failure hazards are liable to develop in this mountain-hazard-prone area due to the complex natural conditions. In result testing, there is still 1 slope failure hazard distributing in the stable region (account for about 12.5% of the total amount). It is mainly because of lacking of data; consequently, some factors are neglected, such as the trend of rock stratum, rock fabric, and ground water, etc. ACKNOWLEDGEMENTS The work was supported by the Chinese National Key Technology R& D Program (2006BAB04A08-03) and the Foundation of Knowledge Innovation Program, CAS (KZCX2-YW-302).

David, K.K. 2002. Investigating Landslides Caused by Earthquakes-A Historical Review. Surveys in Geophysics 23 (6):473–510. De Jong, S.M. & Bertolo, F. 1999. Regional Assessment of soil Erosion Using the Distributed Model SEMMED and Remotely Sensed Data. Catena 37:291–298. Ellen, S.D. 1988. Landslide, Floods and Natine Effects of the Storm of January 1982. In the San Francisco Bay Region, California V.S.S: 3–5. Ercanoglu, M. & Gokceogl, C. 2004. Use of Fuzzy Relations to Produce Landslide Susceptibility Map of a Landslide Prone Area. Environmental Geology 75:229–250. GAO, H.X. & YIN, K.L. 2007. Discuss on the correlations between landslides and rainfall and threshold for landslide early-—warning and prediction. Rock and Soil Mechanics 28 (5):1057–1060. JSECE. 1992. The movement phenomenon of slope. Erosion control engineering lecture 3:154. Kentli, B. & Topal, T. 2004. Assessment of Rock Slope Stability for a Segment of the Ankara Pozanti Motorway , Turkey. Engineering Geology 74:73–90. Mello, U. & Pratson, L. 1999. Regional Slope Stability and Slope Failure Mechanics from the Two Dimensional State of Stress in an Infinite Slope. Marine Geology 154:339–356. Qiao, J.P. & Zhao, Y. 2001. Review on Rick Degree Regionalization of Landslide. Journal of Mountain Science 19 (2):157–160. Shan, X.J. & Li, Zh.F. 2002. Prediction method of dangerous areas of regional landslide based on GIS and its application. Chinese Journal of Rock Mechanics and Engineering 21 (10):1507–1514. Wieczork, G.F. 1984. Evaluating Danger Landslide Catalogue Map. Bulletin of the Association of Engineering Geologists 11 (1): 337–342. Wu, L. & Zhang, W.Ch. 2004. Remote Sensing& GISBased Distributed Hillslope Stability: Quantitative Evaluation Model. SCIENTIA GEOGRAPHICA SINICA 24 (4):458–464. Zhong, Ch. & He, Z.Y. 2005. Evalution of EcoEnvironmental Stability Based on GIS in Tibet , China. Wuhan University Journal of Natural Science 10 (4):653–658.

REFERENCES Brown, T.C. & George, L.P. 2005. The judged seriousness of an environmental loss is a matter of what caused it. Journal of Environmental Psychology 25 (1):13–21.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Landslide hazard and risk assessment in the areas of dams and reservoirs of Serbia P. Lokin & B. Abolmasov Faculty of Mining and Geology, University of Belgrade, Belgrade, Serbia

ABSTRACT: The influence of landslides, rockfalls and other instability phenomena on dams and reservoirs presents a very complex problem, caused by the specific interaction between these constructions and the terrain. This problem has an undoubtedly economic and environmental significance, considering that the constructions in question not only have great value, but also carry risks. This paper will present qualitative landslide hazard and risk assessment methodology applied during the investigation of dams and reservoirs of Serbia.

1

2

INTRODUCTION

The inventory, understanding and characterization of all the instability phenomena on the interaction area of the dams and reservoirs are extremely important. We can find these problems at all levels of the designing process and especially during the usage of these constructions. The process of activation/reactivation of a landslide and other terrain instability phenomena in the interaction area of the dams and reservoirs is very complex. Basically, the causative factors of slope instability, generally, are geological settings, morphological, hydrogeological, hydrological, engineering geological and similar characteristics of the slope. On the other hand, the direct influence of the activity of these events is closely connected to oscillations of the water levels in the reservoirs during their regular usage. During the years 2002–2005, the public company ‘‘Elektroprivreda Srbije’’-EPS (Serbian Electricity Board) financed a study that was aimed at assessing the hazards and risk of instability processes on the interaction area of the dams and reservoirs in Serbia. The study encompassed all existing hydroenergy objects and constructions. It was conducted by the Faculty of Mining and Geology, Department for Geotechnics, ‘‘Energoprojekt-Hidroinženjering’’ and ˇ ‘‘Jaroslav Cerni’’. This paper will present applied qualitative methodology of landslide hazard (debris slade only) and risks assessment, which is the most important factor influencing instability in the investigated area of interaction of dams and reservoirs in Serbia.

THE GOALS OF THE STUDY

The study of influences of landslides (debris slade) and other types of instabilities in the interaction area of dams, near-dam constructions and reservoirs in Serbia was conducted in order; – To register all landslides, potential rock falls, erosion areas and other instability occurrences, – To determine the circumstances of instability and erosion occurrences and causes of process activations, especially with the view of their interactions with the regular regimes of hydro-energy objects (primarily the reservoirs), – To produce a cadastre of landslides, rock falls, torrential floods and other instability occurrences, – To form a uniformed database of all dams, reservoirs and related risk assessments, – To evaluate their technical, ecological and economical aspects of the surrounding area during usage, – To estimate the hazard and risks according to the uniformed criteria for all hydro-energy objects in Serbia. 3

METHODOLOGY OF HAZARD AND RISK ASSESSMENT

In order to complete the Study, a uniformed investigation methodology and criteria for data processing and analysis was used (Lokin, 2003). The First Phase of the Study conduct involved the following activities:

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– Analysis of all available existing documentation; – Analysis of aerial photo data; – Engineering geological mapping of the terrain scaled 1:25 000 with creating cadastre of instability phenomenon; – Creating a database in MS Access which integrates all contents in Arc View 8.1; – Data analysis, classification and then hazard and risk assessment and determining landslide categories; – Plan of the hazard and risk treatment—additional detailed investigations, monitoring, designing and realisation of the preventative and remedial measures depending on the assessed risk category. Further investigations proposed as a result of the Study could be conducted during the second (II) and third (III) phases of Study. Methodology of the hazard and risk estimation had to fulfil a string of objective specific factors that influenced the choice of the methods and analysis criteria as well as the hazard and risk assessments, some of which were: – Hydro-energy constructions included in the Study, dams, near-dam constructions and reservoirs were constructively different, thus the technological process of it’s usage was different. – Hydro-energy constructions were of a different age, which particularly influenced reservoir basins. – The objects themselves (dams and near-dam constructions), as well as the reservoirs, were constructed and formed in different regions of Serbia. The geological settings, geo-morphological, engineering geological, hydro-geological, hydrological and climate conditions were different. – Social infrastructure as well as the land usage were also different. – Available previous documentation of the objects and instability occurrences data were of differing levels of reliability. – Preliminary check of all documentation pointed to different types, magnitude and stage of activity of instability processes. Hazard and risk estimation has been performed as an expert, qualitative method. As a criterion for analysis and landslide hazard estimation, a qualitative product of supposed magnitude and landslide activity is taken. Magnitude occurrence is expressed through a potential area (m2 ) and depth to sliding surface (m), e.g. volume of the mass that can be moved (m3 ) and which is estimated directly in the field based on the known data (in case of a landslide reactivation), or by direct estimation (Lokin, 2005). Based on the accepted magnitude values, landslide classes have been formed. During the process of classification, a difference in

magnitude (volume of the debris slade) that can potentially jeopardise dams and near-dam constructions or the reservoirs themselves has been observed (Table 1). According to the stage of activity, landslides are divided into: Aa – Active landslides with present sliding process; Ap – Active landslides with dormant sliding process (between two phases of activity); Sn – Landslides with remedial measures where it is judged that the measures are not adequate or sufficient; Su – Landslides with remedial measures where it is judged that the measures are sufficient and provide permanent terrain stability; U – Dormant landslides, stabilisation is the result of the natural processes and reactivation is not expected; F – Fossil landslides, covered with younger sediments so the reactivation process is not expected; USp – conditionally stable slope, no sliding process, but the slope inclination and rock-mass properties are such that the sliding process could be activated in the extreme environmental or techogenic circumstances. Based on the classifications mentioned above and the analysis of gathered data, the landslide hazard has been assessed and expressed in hazard degrees. Hazard degrees represent qualitative expressions where, for instance, IV degree of hazard denotes active deep debris slade over 5 metres in depth and area over 1000 m2 (see Table 2). Based on the hazard degrees and possible damages incurred following the activation of a landslide near dams’ constructions, reservoirs and other objects in their vicinities endangered by those landslides, landslide risk assessment has been performed (see Table 3). Using the grades of risk levels determined in this way, a categorisation of landslides has been performed, Table 1. Landslide class I II III

Classes of landslides (only debris slade). Area (m2 )

Depth (m)

Volume (m3 )

25.000

8

200.000

Note: Figures in top row relate to landslides jeopardising reservoirs alone, figures in bottom row relate to landslides jeopardising dam and near-dam constructions.

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where a landslide category essentially presents risk level related specifically to a dam and reservoir, gives recommendations and program of further research in ensuing phases. Table 4 represents landslide categories in relation to their risk grades, as well as suggestions for additional investigations, preventive, remedial and

Table 2. Level of hazard. Degree of landslide hazard IV

III

II

I

Depth

Area

II III III II II III I II I I I

III II III II III II I II III I I, II

State of activity Aa Ap Ap Ap, Sn Aa, Sn U, F, Su Ap U, F, Su, USp

Note: Classes of depth and area were taken from Table 1, and activity was given in the textual part.

monitoring measures for conduct of the Phase II and Phase III of the Study. 4

DISCUSSION

In a study including all dams, near-dam constructions and reservoirs of Serbia, 300 different types of instability occurrences have been registered during the Phase I of the Study. Certainly the most influential were landslides (debris slade), with 192 of them being inventoried. Out of that number, 100 are active landslides (17 with current active landslide process and 83 with processes that are temporarily dormant). Based on the magnitude and estimated level of activity, landslide hazards have been assessed, and taking into account the type and importance of the endangered objects-constructions, risk assessment has been performed listing the possible consequences for those objects-constructions. All landslides have been categorised. A total of 33 landslides belonging to the III and IV category have been submitted for detailed investigation and instrumental monitoring (Phase II), as well as designing and planning of the remedial measures (Phase III). Landslides from I and II categories, so called monitoring measures with instrumental recordings or periodic visual observations have been recommended.

Table 3. Risk level. Risk grade Hazard Element of risk IV a b c

IV IV IV

d III a b c d

IV III III III IV

e a

IV I

b

II

c

III

II

I

Possible damages

Dam and dam-related objects/constructions Reservoir, and indirectly the dam Capital objects (infrastructure, public function constructions, main water supply), landslides caused by the work of the reservoir Ditto, without the reservoir influence Dam and dam-related constructions Capital objects, landslides caused by the work of the reservoir Ditto, without the reservoir Reservoir Local infrastructure, individual water supply objects Dam and dam-related constructions, capital objects related to the reservoir Reservoir

Possible long-period stoppages of work, massive functional damage, victims are possible Possible short-period stoppages of work, medium to massive damage Considerable lowering of the reservoir capacity Possible collapse No damage

d

IV

Lowering of the reservoir capacity, minor to insignificant Local infrastructure, individual water supply objects, smaller public Possible minor and medium objects damage Without significant objects Minor or no damages

a b

I II

Reservoir and capital objects Local infrastructure and less significant objects

c

III

No objects-constructions

Note: Level of landslide hazard was taken from table 2.

1941

Practically no damages Practically no damage or minor damage No damage

Table 4. Landslide categories. Category

Risk grades

Investigations

IV

IV-a IV-b IV-c III-a III-b III-d III-e II-c

detailed investigations investigations conduct geotechnical report detailed investigations investigations conduct geotechnical report monitoring recommendations suggestions of preventive measures observation recommendations suggestions of preventive measures

III

II

II-a II-b

I

I-a I-b

Uncategorised

IV-d III-c II-d I-c

Precautions

Remedial measures

Monitoring

undertaking preventive measures for III-d

designing remedial measures realisation of remedial measures designing remedial measures for III-e and II-c realisation of remedial measures

control monitoring according to the plan (instrumental surface monitoring and inclinometers) periodic instrumental monitoring for III-d control monitoring for III-e and II-c

undertaking preventive measures

undertaking possible remedial measures

periodic visual monitoring for II-a and instrumental monitoring

undertaking preventive measures

periodic visual monitoring

Suggestions of a need to inform the authorities and their understanding of the facts about the landslide activities, possible consequences and measures that need to be taken are communicated to the Investors.

Note: Risk grades have been taken from Table 3.

5

REFERENCES

CONCLUSION

The Study of landslide hazard and risk assessment in the interaction area of dams and reservoirs of Serbia has been conducted for the first time by the methodology presented in this paper. The contemporary way of processing the collected data, its analysis and hazard and risk assessments as well as the possibility of using a data base for all EPS system users, allows the Investor (EPS) a very flexible package of all necessary information needed for modern functioning and usage of the hydro-energy system.

Lokin, P. et al 2003: Methodology of the research of landslide influence and other terrain instability on dams and reservoirs. Proceedings of the II Congress of the Yugoslavian High-dam Society, Kladovo, 2003. 677–684, (on Serbian). Lokin, P. & Abolmasov, B. 2005. Methodology of estimation of hazards and risks from landslide occurrences in the interaction area of dams and reservoirs of Serbia. Proceedings of Advisory Meeting on Geotechnical Aspects of Civil Engineering, Kopaonik, 2005, 257–262, (on Serbian).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

The evaluation of failure probability for rock slope based on fuzzy set theory and Monte Carlo simulation Hyuck-Jin Park Department of Geoinformation Engineering, Sejong University, Republic of Korea

Jeong-gi Um Department of Environmental Exploration Engineering, Pukyung National University, Republic of Korea

Ik Woo Department of Ocean System Engineering, Kunsan National University, Republic of Korea

ABSTRACT: Uncertainty is pervasive in rock slope stability analysis due to various reasons and sometimes it causes serious rock slope failures. Therefore, since 1980’s the importance of uncertainty has been recognized and subsequently the probability theory has been used to quantify the uncertainty. However, not all uncertainties are objectively quantifiable. Some uncertainties, due to incomplete information, cannot be handled satisfactorily in the probability theory and the fuzzy set theory is more appropriate. In this study the random variable in rock slope stability analysis is considered as fuzzy number and the fuzzy set theory is employed. In addition, the Monte Carlo simulation technique is utilized to evaluate the probability of failure for rock slope. This overcomes the shortcomings of the previous studies, which are employed vertex method, first order second moment method and point estimate method. Since the previous studies used only the representative values from membership function to evaluate the stability of rock slope, the approximated analysis results were obtained in the previous studies. With Monte Carlo simulation technique, more complete analysis results can be secured in the proposed method. The proposed method was applied to the practical example. According to the analysis results, the probabilities of failure obtained from the fuzzy Monte Carlo simulation coincide with the probabilities of failure from the probabilistic analysis.

1

INTRODUCTION

One of the difficulties in slope stability analysis is uncertainty inevitably involved in the variability of the material properties and the geotechnical model. The natural materials comprising most slopes have an innate variability difficult to establish and to predict and therefore, the variability of geologic material is one of the major sources of uncertainties. In addition, insufficient amount of information for site conditions and incomplete understanding of failure mechanism are also another sources of uncertainties. Therefore, the presence and the significance of uncertainties in slope stability analysis has been appreciated for long time. Consequently several approaches such as observation method (Peck, 1969) have been suggested to deal properly with uncertainty. The probabilistic approach has been proposed as an objective tool for representing uncertainty in failure model and material characteristics. Many probabilistic analyses have

been published in literature (Einstein and Baecher, 1982; Mostyn and Small, 1987; Mostyn and Li, 1993; Nilsen, 2000; Park and West, 2001; El-Ramly et al., 2002; Pathak and Nilsen, 2004; Park et al., 2005). However, the probabilistic analysis requires the statistical parameters and distribution type for random variables in order to quantify the uncertainty. That is, the mean, standard deviation and probability density function for uncertain parameters are prerequisite in order to carry out the appropriate probabilistic analysis. However, a large amount of information and data are required to obtain statistical parameters and distribution type for random variable but in many practical conditions, the amount of data is frequently limited. Consequently it is difficult to secure statistical parameters and distribution type of the uncertain variable, and this situation makes the application of probabilistic analysis difficult. The uncertainties caused by limited or incomplete information cannot be handled satisfactorily

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2

FUZZY MONTE CARLO SIMULATION METHOD

2.1 Fuzzy set theory In classical set theory, an element either belongs or does not belong to the set. That is, the membership of classical set theory is defined in strict sense. When a certain element x belongs to set A, x is a member or element of a set A and can be written x|A

(1)

x|A

(2)

Whenever x is not an element of a set A, we write

1.0

Membership function

in the probability theory and the fuzzy set is more appropriate (Dodagoudar and Venkatachalan, 2000). Therefore, the present study proposed the utilization of fuzzy set theory in order to overcome the limitations of the probabilistic approach. Fuzzy set theory has been proposed by Zadeh (1965) and it has been known as appropriate approach for dealing with uncertainty mainly caused by incomplete information. Consequently, fuzzy set theory has been employed in many slope stability analyses (Juang & Lee, 1992; Lee and Juang, 1992; Davis and Keller, 1997; Juang et al., 1998; Dodagoudar and Venkatachalan, 2000; Giasi et al., 2003; Li and Mei, 2004). However, the previous studies combined the fuzzy set theory with the approximate method such as point estimate method or first order second moment method. Since the approximate methods use only few representative values from uncertain parameters, the analysis cannot provide accurate analysis results. Therefore, this study proposed the new approach incorporating the Monte Carlo simulation which provides complete analysis results with fuzzy set theory.

0.5

CORE

0.0 SUPPORT

x

Figure 1.

For each x | A, when µA (x) = 1, x is declared to be a member of A. When µA (x) = 0, x is declared to be a nonmember of A. However, in fuzzy sets, which is introduced by Zadeh (1965), more flexible sense of membership is possible. That is, the membership function can be generalized such that the values assigned to the elements fall within a specified range. In fuzzy set, the degree of membership to a set is indicated by a number of between 0 and 1. In fuzzy set theory, each fuzzy set is uniquely defined by a membership function. Since an element’s

Concept of membership function.

membership function in a fuzzy set may admit some uncertainty, its membership is a matter of degree. The membership function can be manifested by many different types of function and different shapes of their graphs. Triangular and trapezoidal shapes are most common types in the membership function. Fig. 1 shows the concept for support, core and height in a trapezoidal shaped fuzzy set. The support is the set of all elements of set x that have nonzero membership in A. In addition, core is the set of all elements of x for which the degree of membership in A is 1. The height of a fuzzy set A may be defined as the largest membership grade obtained by an element in that set. If the height of a fuzzy set A is 1, set A is called normal and otherwise, it is called subnormal. There are two commonly used ways of denoting fuzzy sets. A = x, µA

or

A set can be defined by membership function that declares which elements of x are members of the set and which are not.  1, x | A µA (x) = (3) 0, x | A

HEIGHT

x 1

A=

µA (x) x

2.2

Fuzzy Monte Carlo simulation

(4)

The probabilistic analysis has been known as an effective tool to quantify and model uncertainty. However, limited information for uncertain parameters makes the application of the probabilistic analysis difficult. This is because the probabilistic analysis is carried out on the premise that the precise mean and standard deviation and the appropriate probability density function for uncertain parameter can be obtained. However, in order to obtain the adequate statistical parameters and distribution function for uncertain parameter, a large amount of data is required but it is often not practically possible. Frequently only the maximum and minimum values for uncertain parameter can be obtained and therefore, uncertain parameter can be expressed only with interval between minimum

1944

and maximum. Under this condition, uncertain parameter may be expressed as a fuzzy set, if there is some reason to believe that not all values in the interval have the same degree of support (Juang et al., 1998). Since uncertainties due to incomplete information are pervasive in the procedure of slope stability analysis, several researches utilized fuzzy set theory in slope stability analysis (Juang et al., 1998; Dodagoudar and Venkatachalan, 2000; Giasi et al., 2003). However, the previous researches utilized the vertex method (Dong and Wong, 1987) to evaluate fuzzy input parameters in slope stability analysis. The vertex method is based on the α-cut concept of fuzzy numbers and involves an interval analysis. The basic idea of the vertex method is to discretize a fuzzy number into a group of α-cut intervals. By replacing fuzzy numbers in the slope model with intervals, the fuzzy computation obtains factor of safety in the deterministic slope model. However, when the factor of safety is evaluated from the deterministic slope model using two interval values, the first order second moment method (Giasi et al., 2003) or point estimate (Dodagoudar and Venkatachalam, 2000) has been applied. According to Harr (1987), the first order second moment method and point estimate method are considered as approximate method since the methods do not utilize complete information for random variables to evaluate performance function. The approximate method has been proposed to evaluate the probability using simple calculation with only few representative values of random variable without distribution information. However, since the previous researches used incomplete information in the analysis, there is a possibility that approximate results would be obtained instead of the precise analysis results. Therefore, this study proposed the new approach evaluating the reliability of rock slope with fuzzy number and Monte Carlo simulation. The Monte Carlo simulation is the most complete method of the probabilistic analysis since all the random variables are represented by their statistical parameters and probability density function. In addition, the complete information is employed to evaluate performance function in Monte Carlo simulation. In order to combine Monte Carlo simulation with fuzzy set theory, uncertain parameter is considered as fuzzy number and its membership function is decided by means of available information and engineering judgment. Then Monte Carlo simulation is utilized to evaluate the probability of slope failure from fuzzy numbers of uncertain parameters. In most rock slope stability analyses, the friction angle of discontinuity is considered as uncertain parameters. This is because the number of the direct shear tests which are carried out to acquire shear strength of discontinuity is always limited and therefore, the true value of friction angle cannot be evaluated. Consequently, in the present study the friction

angle is considered as fuzzy number and its membership function is decided on the basis of analysis for laboratory test results. However, in the Monte Carlo simulation, the cumulative density function for uncertain parameter is required. In the present study, the membership function is adapted to cumulative density function in the calculation of performance function. Then in Monte Carlo simulation, the process takes a single value selected randomly from its cumulative distribution. The randomly selected parameter is used to generate a single random value for factor of safety. By repeating this process many times to generate a large number of different factors of safety, a cumulative density function for factor of safety can be obtained and then probability of failure is evaluated.

3

CASE STUDY

The proposed method in the present study has been applied to practical example in order to check the feasibility and validity of the proposed approach and compare with the probabilistic analysis results. A slope has been selected and the detailed field investigation has been carried out. The dip direction and dip angle of the slope are 325 degree and 65 degree, respectively and its height is 40.8 m. The slope is composed of Precambrian metasedimentary rock. Approximately 350 discontinuity data has been obtained on scanline survey and 6 discontinuity sets were identified by means of clustering process (Table 1). Among 6 discontinuity sets, 2 sets (set 2 and set 4) are analyzed as kinematically unstable for planar failure. In this study, the analysis for only planar failure is performed since the analysis results of planar failure are easy to compare to other analysis results. In addition, the direct shear test is carried out in order to acquire the shear strength parameter for discontinuity. Based on the 19 direct shear test results, the friction angle ranges from 20.9 to 46.3 and their mean and standard deviation are 34.6 and 8.2, respectively (Fig. 2). However, even if 19 tests were performed, the probability density function cannot be determined due to severe scattering as can be seen in Fig. 2. Even the previous Table 1. gation.

Discontinuity sets observed from field investi-

Discontinuity sets

Representative orientation

Set 1 Set 2 Set 3 Set 4 Set 5 Set 6

217/77 320/30 061/66 311/40 196/56 183/05

1945

0.12

5

0.10

4

0.08

Frequency

Frequency

6

3

2

0.04

0.02

1

0

0.06

0.00

20 - 25

25 - 30

30 - 35

35 - 40

40 - 45

0

45 - 50

1

2

3

Factor of safety

Friction Angle

Figure 3.

Figure 2. Results of direct shear tests.

Results of probabilistic analysis for joint set 2.

0.16

3.1 Results of probabilistic analysis In order to compare to other analysis results, the deterministic analysis based on the limit equilibrium approach has been carried out for joint set 2 and 4, which are analyzed as kinematically unstable on the stereonet analysis. This analysis has been performed with same input values for all the deterministic parameters used in the probabilistic analysis and mean value of the distribution for random parameter. The factor of safety for set 2 is evaluated as 1.20 and the factor of safety for set 4 is 0.82. That is, joint set 2 has been analyzed as stable in the deterministic analysis but unstable for joint set 4. The probabilistic analysis is also carried out for set 2 and set 4 using the procedure proposed by Park et al. (2005). In the present study, the orientation of discontinuity is taken into account for the deterministic parameter and therefore, the single fixed dip direction and dip angle for discontinuity orientation is employed in the probabilistic analysis. On the other hand, the friction angle for discontinuity is considered as the random variable. The mean value and standard deviation have been used and subsequently normal distribution

0.12

Frequency

researches proposed normal distribution for probability density function of friction angle (Mostyn and Li, 1993; Nilsen, 2000; Pathak and Nilsen, 2004; Park et al., 2005), it is not easy to decide normal distribution as probability density function for friction angle in this study due to uncertainty. Therefore, the friction angle is considered as fuzzy number whose support is between 20.9 and 46.3 in this study. The triangular shape is chosen for membership function of the friction angle and the core value of membership function is decided to 34.6 which is mean value of the test results. In addition, on the basis of Hoek’s suggestion (1997) in rock slope stability analysis, cohesion is not considered in slope stability analysis.

0.08

0.04

0.00 0

1

2

3

Factor of safety

Figure 4.

Results of probabilistic analysis for joint set 4.

has been chosen for probability density function of friction angle. In order to evaluate the probability of failure, the Monte Carlo simulation approach is employed in the probabilistic analysis. Total 16,000 repeated calculations are carried out. Figs. 3 and 4 show the results of analysis, which show the distributions of the factor of safety. For joint set 2, the probability of failure is evaluated as 29.3% and the probability of failure for set 4 is 73.5%. In case of set 2, the result of the deterministic analysis indicate stable but the result of the probabilistic analysis shows quite high probability of failure. This is because the deterministic analysis does not reflect the variability and uncertainty in input parameter. However, the coefficient of variation for friction angle used in this study is calculated as 23.3% and this is quite high value compared to the other previous researches, which show 10% (Park and West, 2001). It means that the dispersion of direct shear test results used in the present study is too large. That is, the randomly generated friction angle from Monte Carlo simulation ranges from 10 to 59.2 in the confidence interval of 99.8%. Consequently the uncertainty of friction angle is too large and subsequently in

1946

1.0

0.21

0.8

Membership function

Frequency

0.28

0.14

0.07

0.6

0.4

0.2 0.00 0

1

2

3

0.0

Factor of safety

10

Figure 5. Results of probabilistic analysis for joint set 2 when COV = 10%.

30

40

50

Internal friction angle

Figure 7.

0.36

Triangular membership function.

0.16

0.27

0.12

Frequency

Frequency

20

0.18

0.09

0.08

0.04

0.00 0

1

2

0.00

3

0

Factor of safety

1

2

3

Factor of safety

Figure 6. Results of probabilistic analysis for joint set 4 on when COV = 10%.

Monte Carlo simulation, too small value of friction angle could be generated and used in factor of safety calculation. This could cause serious error in the evaluation of the probability of failure. Therefore, in order to check out the influence of uncertainty in input parameters, the dispersion of friction angle is reduced to 10% of C.O.V (coefficient of variation) and the probability of failure is recalculated (Fig. 5 and 6). Figures 5 and 6 show that the dispersion of factor of safety is reduced comparing to Figures 3 and 4. The probability of failure for joint set 2 is reduced to 8.8% but the probability of failure for joint set 4 is increased to 94.0%. This is because the lower half of the dispersion for factor of safety in joint set 2 is reduced but the upper half of the dispersion in joint set 4 is reduced. This shows that uncertainty and dispersion of input parameter affect the analysis result. Consequently if the number of data is limited and subsequently the random properties cannot be recognized precisely, the results of the probability analysis can be affected by the dispersion of the input parameters.

Figure 8.

Results of FMC analysis for joint set 2.

3.2 Fuzzy Monte Carlo simulation As can be seen previously, the friction angle obtained from direct shear test includes a large amount of uncertainty. This uncertainty is usually caused by a lack of test results. A lack of test results prevents the precise understanding of random properties for uncertain parameters and also makes the application of the probabilistic analysis difficult. Therefore, in the present study the friction angle is considered as fuzzy number. The friction angle is considered as triangular fuzzy number and the minimum and maximum values of the membership function are decided as 20.9 and 46.3, respectively on the basis of test results. In addition, the mean value, 34.6 is decided as core in the membership function (Fig. 7). This means that the randomly generated value from Monte Carlo simulation ranged from 20.9 to 46.3. This shows the dispersion of fuzzy number is much smaller than the dispersion used in the probabilistic analysis. The C.O.V of the fuzzy number is calculated as 13.3% on the confidence level of 99.8% and this value is smaller than the C.O.V used in the probabilistic analysis. The distribution of

1947

0.12

Frequency

0.10

0.08

0.06

0.04

0.02

0.00 0

1

2

3

Factor of safety

Figure 9. Results of FMC analysis for joint set 4.

factor of safety evaluated from fuzzy Monte Carlo simulation is given in Figures 8 and 9. Comparing the factors of safety distribution obtained from fuzzy Monte Carlo simulation with the factors of safety distribution obtained from the probabilistic analysis, the dispersion of factor of safety distribution in fuzzy Monte Carlo simulation is reduced. In accordance with the analysis results, the probability of failure for joint set 2 is 33.5% and the probability of joint set 4 is 72.9%. In case of joint set 2, the probabilities of failure evaluated from the probabilistic analysis and fuzzy Monte Carlo simulation are 29.3% and 33.5% respectively and the analysis results are somewhat different. But in case of joint set 4, the probabilities of failure evaluated from the probabilistic analysis and fuzzy Monte Carlo simulation are 73.5% and 72.9% respectively and the analysis results are quite similar. Consequently, even if the application of fuzzy set theory reduced the dispersion in input value, the probabilities of failure obtained from two different approaches are similar. As a result, the application of fuzzy set theory manages the uncertainty of input parameter effectively.

4

CONCLUSIONS

Uncertainty is pervasive in rock slope stability analysis due to various reasons and sometimes it causes serious rock slope failures. Therefore, the probability theory has been used to quantify the uncertainty. However, not all uncertain-ties are objectively quantifiable. Some uncertainties, due to incomplete information, cannot be handled satisfactorily in the probability theory and the fuzzy set theory is more appropriate. In this study the random variable in rock slope stability analysis is considered as fuzzy number and the fuzzy set theory and Monte Carlo simulation are employed. In order to verify the feasibility and validity of the proposed approach, the proposed method was applied

to the practical example. In the deterministic analysis results, joint set 2 is analyzed as stable but joint set 4 is analyzed as unstable. On the contrary in the probabilistic analysis results, the probability of failure for joint set 2 is 29.3% and the probability for joint set 4 is 73.5%. The data used in the probabilistic analysis are widely scattered since the COV of friction angle is evaluated as 23.3%. The widely scattered data may cause serious miscalculation in the evaluation of the failure probability since impractical data could be used in the calculation. Therefore, the probability of failure is recalculated with the modified data whose COV is reduced to 10.0% and the probabilities has been changed. The analysis results of the proposed method using fuzzy Monte Carlo simulation are 33.5% for joint set 2 and 73.5% for join set 4 and the COV of the data that considered as fuzzy number is 13.3%. The probabilities from the probabilistic analysis and the proposed method are somewhat similar but the COV of data in the proposed method is smaller than the COV in the probabilistic analysis. Therefore, the fuzzy set theory managed uncertainty in data more effectively than the probabilistic analysis.

REFERENCES Davis, T.J. & Keller, C.P. 1997. Modelling uncertainty in natural resource analysis using fuzzy sets and Monte Carlo simulation: slope stability prediction. Int. J. Geographical Information Science. 11(5): 409–434. Dodagoudar, G.R. & Venkatachalam, G. 2000. Reliability analysis of slope using fuzzy sets theory. Computers and Geotechnics. 27: 101–115. Dong, W.M. & Wong, F.S. 1987. Fuzzy weighted averages and implementation of the extension principle. Fuzzy Set and System. 21: 183–199. Einstein, H.H. & Baecher, G.B. 1982. Probabilistic and statistical methods in engineering geology. Rock Mechanics, Supplement. 12: 47–61. El-Ramly, H., Morgenstern, N.R. & Cruden, D.M. 2002. Probabilistic slope stability analysis for practice. Can. Geotech. J. 39: 665–683. Giasi, C.I., Masi, P. & Cherubini, C. 2003. Probabilistic and fuzzy reliability analysis of a sample slope near Aliano. Engineering Geology. 67: 391–402. Harr, M.E. 1987. Reliability based on Design in Civil Engineering. New York. Hoek, E. 1997. Rock Engineering; Course note by Evert Hoek[Online], http://www.rockeng.utoronto.ca/ Hoekcorner.htm. Juang, C.H. & Lee, D.H. 1992. Mapping slope failure potential using fuzzy sets. Journal of Geotechnical Engineering. 118(3): 475–494. Juang, C.H., Jhi, Y.Y. & Lee, D.H. 1998. Stability analysis of existing slopes considering uncertainty. Engineering Geology. 49: 111–122. Lee, D.H. & Juang, C.H. 1992. Evaluation of failure potential in mudstone slopes using fuzzy sets. Proc. of Stability and Performance of Slopes and Embankments. 1137–1151.

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Li, W.X. & Mei, S.H. 2004. Fuzzy system method for the design of a jointed rock slope. Proc. of SINOROCK 2004. 2B-18. Mostyn, G.R. & Li, K.S. 1993. Probabilistic slope analysis— state of play. Proceeding of Conference on Probabilistic Method in Geotechnical Engineering: 89–109. Mostyn, G.R. & Small, J.C. 1987. Methods of stability analysis. Soil Slope Instability and Stabilization. Balkema, 71–120. Nilsen, B. 2000. New trend in rock slope stability analysis. Bull. Eng. Geol. Environ., 58: 173–178. Park, H.J. & West, T.R. 2001. Development of a probabilistic approach for rock wedge failure. Engineering Geology. 59: 233–251.

Park, H.J., West, T.R. & Woo, I. 2005. Probabilistic analysis of rock slope stability and random properties of discontinuity parameters, Interstate Highway 40. Engineering Geology. 79: 230–250. Pathak, D. & Nilsen, B. 2004. Probabilistic rock slope stability analysis for Himalayan condition. Bull. Eng. Geol. Environ., 63: 25–32. Peck, R.B. 1969. Advantages of limitations of the observational method in applied soil mechanics. Geotechnique, 19: 171–187. Zadeh, L.A. 1965. Fuzzy sets. Information and Control, 8: 338–353.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Macro-zoning of areas susceptible to flowslide in pyroclastic soils in the Campania region L. Picarelli Research Center in Environmental Engineering, C.I.R.I.A.M., Seconda Università di Napoli, Aversa, Italy

A. Santo & G. Di Crescenzo Department of Geotechnical Engineering, Università di Napoli Federico II, Napoli, Italy

L. Olivares Research Center in Environmental Engineering, C.I.R.I.A.M., Seconda Università di Napoli, Aversa, Italy

ABSTRACT: A great part of Campania Region, whose capital is Naples, is mantled by non lithified unsaturated granular soils of pyroclastic nature which cover even very steep slopes. These are subjected to different types of rainfall-induced landslides, including catastrophic flowslides. Bearing on the results of recent research on the mechanics of rainfall-induced landslides and on a detailed survey of the geomorphological and lithological features of pyroclastic deposits, is proposed a preliminary macro-zoning of areas which are susceptible to flowslide. 1

2

FOREWORD

A great part of Campania, one of the most crowded regions in Europe, is covered by non lithified pyroclastic soils. Generally, these are unsaturated, mantling even very steep slopes, and can experience rainfallinduced failure which occurs when suction drops to a critical value depending on slope and friction angle. In some cases, slope failure gives rise to a fast catastrophic flowslide or debris flow, whose velocity can reach a few tens of metres per second. In other cases, induced landslide is a slide or a debris avalanche (Hungr et al., 2001). According to experience, these present much lower velocities (slides) and smaller run-out (either slides or debris avalanches), thus prediction of the type of landslide which can occur is very important for risk assessment and land management (Picarelli et al., 2007). Recent research on the mechanics of rainfallinduced landslides provides useful information for prediction of the type of landslide which can take place as a consequence of slope failure. Bearing on such results, on detailed geomorphological and lithological data as well as on sound experience, in this paper is proposed a macro-zoning of the parts of Campania Region which are exposed, more than others, to the risk of flowslide.

GEOLOGICAL FRAMEWORK

During Quaternary the area occupied by Campania (Fig. 1) experienced distensive tectonic phenomena which caused uplift of the Apennines chain and formation of a wide depressed graben area (Campanian Plain). Those events were accompanied by an intense explosive volcanic activity. Neglecting the marine environment, different volcanoes were active in the continental area, covering with pyroclastic products a significant part of the region. Figure 1a reports a simplified geological map of Campania with location of such deposits. The thickness of these depends on the distance from the vent and on the slope of the depositional surface, thus it is higher in the vicinity of the volcanic centres and on plains, and is lower in the distal areas and on steep slopes. Moving from North to South, the first volcanic centre is Roccamonfina, whose activity can be divided into two phases: the first one started about one million of years ago, ending 400 ky ago; the second one developed between 250 ky and 170 ky ago. The volcano is now definitely inactive. The volcanic centres located to South are younger and still active. The Phlegrean Fields district consists of several volcanic edifices located in the middle of the Campanian Plain graben. The activity of Phlegrean

1951

Figure 1. a) Geological map of the Campania Region; b) pyroclastic macro-areas. 1. Pyroclastic air-fall deposits (Quaternary). 2. Alluvial deposits (Quaternary). 3. Lavas, pyroclastic flows and tuffs (Quaternary). 4. Arenaceous conglomerates (Pliocene). 5. Marly arenaceous terrigenous deposits with clay interbeds (Tertiary). 6. Carbonate rocks (Mesozoic). 7. Volcanic centres. 8. Rivers. 9. Pyroclastic air-fall deposits of Phlegrean Fields and Somma-Vesuvius.

Fields started in the upper Pleistocene; about 45 ky ago (Di Girolamo et al., 1984). The main product is the so called Ignimbrite Campana (about 39 ky) amounting to about 150 km3 , which was essentially deposited by pyroclastic flows; the eruption which produced the Ignimbrite Campana led to formation of the phlegrean caldera, a collapse structure having a diameter of 14 km: Another important product of Phlegrean Fields is the well known Yellow Neapolitan Tuff. Starting from about 12 ky ago until to 1538 AD, the Phlegrean district produced non lithified pyroclastic deposits through several cones and rings which originated within the caldera: these deposits were essentially spread towards East (Fig. 1a). The Mt. Somma-Vesuvius system consists of an older volcano, the Mt. Somma, whose activity was characterised by explosive plinian phases (Rolandi et al., 2000), and of the Vesuvius volcano which formed during the Middle Age within the Mt. Somma caldera The main historic eruptions of Vesuvius occurred from 79 AD to 1944 (Lirer et al, 1973; Rosi et al., 1993; Rolandi et al, 2000). The Mt. Somma-Vesuvius system has produced alternating deposits of pyroclastic soils and lavas. The air-fall products consisting of

pumices, scorias and ash have been mainly spread towards East, where they mostly mantled calcareous mountains (Fig. 1a).

3

MAIN NON LITHIFIED PYROCLASTIC DEPOSITS IN CAMPANIA

Non lithified pyroclastic deposits can be differentiated according to age, deposition mechanism, grain size and nature of the bedrock. In Campania can be recognized the six following macro-areas (Fig. 1b). Roccamonfina volcano area (A). In this macro-area the bedrock is constituted by lava. The pyroclastic products are very old (more than 150 ky) and consist of fine-grained or humified ash. Flowslides are practically absent. Phlegrean Fields and phlegrean inlands (B). The pyroclastic soils were deposited on tuff through flow, surge or fall. The average slope angles are quite high (about 35◦ ) and tuffaceous cliffs are frequent. This area is subjected to small to medium-size flowslides.

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Northern side of the Mt. Somma-Vesuvius system (C). Coarse pyroclastic materials (pumice and scoria) are spread (proximal deposits). The slope angles are quite gentle: values higher than 35◦ can be found only along main drainage channels around the volcano. Only small flowslides can be recognized. Matese Mt., Maggiore Mt., Massico and Tifatini Mts. (D). The calcareous slopes are mantled by air-fall products erupted by Roccamonfina and Phlegrean Fields. On the southern slopes, where the smallest amount of vegetation favours run-off, the largest part of the oldest deposits (about 100 kys) has been almost completely eroded. The northern slopes present some tens of centimetres of weathered ash. No flowslides can be recognized. Marzano Mt., Cilento and Vallo di Diano (E). Because of the large distance from vents, only thin deposits of pyroclastic soil (a few decimetres) are present, mostly on the northern slopes. Pumices are almost absent, while ash is generally weathered. No flowslides can be recognized. Lattari Mts., Sorrentina Peninsula, Picentini Mts., Pizzo d’Alvano Mt., Avella Mts. (F). These mountains are the closest ones to Somma-Vesuvius system. In the last 15 ky they have been mantled by the products of several eruptions, thus the maximum theovretical thickness of primary deposits should range between 4 and 7 m (De Vita & Celico, 2006). Steep slopes have been repeatedly subjected to flowslides. 4

SHORT REMARKS ON THE MECHANICS OF LANDSLIDES IN PYROCLASTIC SOILS

is worth mentioning that this idea is incorporated into special classifications of flow-like landslides (Hungr et al., 2001; Hutchinson, 2004). However, liquefaction can happen only if the soil is saturated and has some special features concerning index and state properties as well as state of stress. Hence, since cohesionless pyroclastic soils on slopes are generally unsaturated, the occurrence of liquefaction is not obvious. If this does not occur, a different type of landslide is triggered: for instance, small falls involve rather tall cliffs in highly unsaturated material, debris avalanches mostly occur on very steep slopes covered by unsaturated soil, slides occur on quite gentle slopes mantled by dense saturated soil. These remarks come from experience but can be supported by simple theoretical considerations. In fact, cliffs and very steep slopes are stable because of suction. As a consequence of the decrease of the saturation degree caused by infiltration, the soil can fail; however, since at rupture it is still unsaturated, liquefaction cannot occur and a flowslide cannot develop: cliffs are subjected to formation of tension cracks giving rise to falls, while very steep slopes fail by sliding giving rise to debris avalanches. In case of gentler slopes, the soil can fail only after complete saturation: if it is liquefiable, rupture triggers positive excess pore pressures and a flowslide (or a liquefied debris flow) can develop; if the soil is not liquefiable, a slide takes place.

5

The main cause of slope failure in Campania is rainfall. However landslides in non lithified (cohesionless) pyroclastic soils display quite variable features: in fact, small falls, slides, debris avalanches and flowslides can be equally triggered as a function of even small geomorphological and geotechnical details. The magnitude of these landslides is highly variable, because of different size, velocity and run-out. The most catastrophic movements are flowslides. In the last tens of years these have provoked almost two hundreds of victims. In particular, on May, 5, 1998, 159 people living in four different towns located at the foot of the Pizzo d’Alvano Mt. (macro-area F), were killed by a number of flowslides triggered by the same meteorological event (Cascini et al., 2000): the largest number of victims (137) occurred in the town of Sarno. According to some Authors which have investigated the mechanisms of landslides in granular soils (Sladen et al., 1985; Eckersley, 1990), Olivares & Picarelli (2001, 2006) stress that liquefaction caused by building up of positive excess pore pressures is the main cause of flowslide in cohesionless pyroclastic soils. It

MAIN FACTORS GOVERNING FLOWSLIDE GENERATION AND ZONING CRITERIA

According to previous considerations, Picarelli et al. (2007) propose a simplified approach for zoning the potential sources of flowslides accounting for geomorphological and geotechnical features of the outcrops. Such a criterion can help in land management, providing information about the areas which are exposed to catastrophic slope movements. The key points of the proposed approach concern: 1) the conditions for full soil saturation before slope failure; 2) the susceptibility of soil to liquefaction. As discussed above, the first condition does not apply to very steep slopes in cohesionless soil (c′ = 0) which are stable because of apparent cohesion due to suction, and fail as a consequence of rainfall, when the mobilised cohesion is still higher than zero, i.e. for a degree of saturation less than one (Picarelli et al., 2007). In the simplified hypothesis of infinite slope subjected to vertical infiltration, the critical slope angle, i.e the maximum angle compatible with the assumption c′ = 0 (full saturation), is equal to the friction angle of soil. For a lower slope, rupture can happen only for u > 0, i.e. for groundwater level

1953

located above the base of the layer: for initially unsaturated soil this can occur only for impervious bedrock which allows formation of water ponding at the bottom of the layer. Under the above conditions (full saturation), if the soil is susceptible to liquefaction a flowslide can take place. In the geomorphological contexts of Campania Region, the assumption of infinite slope is generally correct because of the high ratio between length of the slopes and thickness of the pyroclastic cover. In addition, according to experience, the shear strength parameters of stratified deposits, as alternating ash and pumice layers, are quite similar: therefore, the cover can be roughly considered uniform, at least in terms of resistance. Assuming a friction angle of 33◦ –40◦ , the typical range of values for considered materials (Picarelli et al., 2006), also the critical slope angle, β, should range between 33◦ and 40◦ : For vertical seepage, which can take place only for highly pervious bedrock, the critical slope angle is also the minimum value for which slope failure can occur: if soil is susceptible to liquefaction; a flowslide can be triggered. For impervious bedrock the minimum value of the slope angle may obtained assuming the groundwater level located at the ground surface and seepage parallel to slope: for γsat = 15 kN/m3 , the slope is definitely stable only for angles less than 13◦ –15◦ . Therefore, if soil is susceptible to liquefaction, a flowslide is theoretically possible for β comprised between 13◦ –15◦ and 33◦ –40◦ . These considerations match experience. Figure 2 reports the slope angle of the sources of flowslides and liquefied debris flows occurred in Campania in the last years. A major factor is the nature of bedrock, which can be constituted by either fractured limestone (pervious bedrock) or flysch (impervious bedrock). For pervious bedrock, about 90% of flowslides occurred for angles in the range 30◦ –50◦ , while for impervious bedrock, about 95% of flowslides occurred for angles in the range 10◦ –35◦ . Summing up, previous theoretical considerations and field data provide a first important geomorphological element for zoning long slopes covered by thin pyroclastic soils, accounting for the nature of the bedrock. In particular, for vertical seepage, field data

Figure 2. Slope angle in source areas of flowslides in pyroclastic soils.

suggest a range of values of the critical slope angle of 30◦ –45◦ . The maximum values of 45◦ can be justified accounting by the shape of the water retention curves of soil (Picarelli et al., 2007): in fact, for slope angles only slightly higher than the friction angle, the mobilised cohesion (and suction) is very small and the degree of saturation only slightly less than 100%. The second element to consider for zoning is the susceptibility of soil to liquefaction. For accurate zoning at a detailed scale, it can be checked only through accurate soil testing. However for large areas for which even a first macro-zoning of flowslide prone slopes can be useful, all available data along the lines discussed above, including geological and geotechnical data and information on historic landslides, can be of great help. According to present knowledge the susceptibility of soil to liquefaction depends on index and state properties as well as on the old and recent stress history. However, the prominent factors are grain size, plasticity and initial density. The highest liquefaction potential is possessed by loose uniform silty sand with non plastic fines (Yamamuro & Lade, 1997). Hunter & Fell (2003) report the grain size curves of soils which proved to be liquefiable, which are quite in a good agreement with data concerning air-fall deposits present in all macro-areas individuated above, apart macro-area C, which are quite uniform regardless of the site of deposition (Picarelli et al., 2006). These materials are non plastic, with the exception of old weathered deposits present in the macro-areas A, D and E. Concerning density, it is well known that liquefaction takes place in those soils which have a void ratio at rupture well above the Steady-State Line of soil (Castro, 1969). For thin covers of air-fall ash, Picarelli et al. (2007) notice that a void ratio comprised between 1.5 and 1.8 should separate liquefiable soil (e > 1.5–1.8) from non liquefiable soil (e < 1.5–1.8). Collected data throughout the Region show that airfall ashes are very loose, presenting values of the void ratio up to 3–4. In contrast, a quite lower void ratio and lower susceptibility to liquefaction features materials deposited by flow and surge, generally close to the volcanic centres, weathered and altered deposits, as well as reworked material accumulated at the foot of slopes (secondary deposits). Finally, also slightly lithified materials as altered ash, should present a lower susceptibility to liquefaction because of a more stable structure. All these considerations inform the proposed macro-zoning of areas susceptible to flowslide, which includes all covers having as main component volcanic ash with the following features: 1. a grain size falling in the range of silty sands (mostly primary air-fall deposits); 2. absence of plasticity (unweathered ash);

1954

3. absence of true cohesion (non altered ash); 4. low density (primary deposits of air-fall ash).

6

MACRO-ZONING OF AREAS SUSCEPTIBLE TO FLOWSLIDE

Previous data can explain why the macro-areas A, C, D and E (Fig. 1b) are only marginally affected by flowslides. In contrast, these are widespread in the macro-areas B and F, even though with quite a variable size. Major flowslides are listed in Table 1. In the sectors B and F, macro-zoning can be further developed accounting for the slope angle and excluding those covers which have a thickness less than 50 cm. By means of a GIS application, have been selected all those slopes with impervious bedrock which have an angle comprised between 13◦ and 45◦ , and those with pervious bedrock which have an angle comprised between 30◦ and 45◦ . Accounting for the nature of the pyroclastic cover, further sub-areas have been then individuated (Fig. 3). Phlegrean area (Ba). Pyroclastic covers generally do not exceed 2–3 m and mostly consist of ash. In this area mostly debris flows, but also small flowslides can occur: the largest ones (Tab. I) attain tens thousands of cubic metres. Caserta Mts. and southern slope of the Avella Mts. (Fa). These calcareous mountains are located at the boundary of the dispersion area of the Somma–Vesuvius deposits (Fig. 1a, 3). Pyroclastic covers, mainly made up by weathered ash and pumices, reach thicknesses in the range of 1 m. Only small flowslides have been identified. Table 1. Features of the main historic flowslides. Sector Site Ba Fb Fb Fb Fc Fc Fc Fc Fd Fe Ff Fg Fh Fi

Ischia Cervinara Avella Cancello Sarno Bracigliano Siano Quindici Gragnano

Date

2006 1999 1998 1998 1998 1998 1998 1998 1764– 1997 Maiori 1954 Massalubrense 1973 Avellino 2005 Montoro Inf. 1997 Salza 1970

Victims

Length (m)

Volume (m3 )

4 5 – 1 137 – 6 11 2– 153 >300 10 1 – –

450 2∗ 103 15∗ 102 8∗ 102 2–4∗ 103 1–2∗ 103 14∗ 102 1–4∗ 103 10∗ 102

3∗ 104 4∗ 104 2∗ 104 3∗ 104 5∗ 105 15∗ 104 4∗ 104 5∗ 105 1–6∗ 104

103 3∗ 102 4∗ 102 2∗ 103 4∗ 102

5∗ 104 7∗ 103 2∗ 104 3∗ 104 20∗ 103

Avella, Roccarainola and Cervinara Mts. (Fb). This sector is located in the distal axial area of the air-fall deposits of the Somma-Vesuvius and Phlegrean Fields systems. The pyroclastic cover overlie fractured limestone and consists of the products of several eruptions reaching a maximum total thickness of 4 m. This zone experienced several large flowslides as the Cervinara one (Tab. I). Pizzo d’Alvano, Monteforte and Mugnano Mts. (Fc). This zone is located in the distal axial zone of the dispersion products of the Somma-Vesuvius system. The thickness of the pyroclastic cover which rests on fractured carbonate rock, range from 4 to 7 m. In 1998 this area was subjected to a lot of large flowslides (Tab. I). Northern sector of Lattari Mts. (Fd). The sector is located in the axial and proximal areas of the air-fall products of the 79 AD Vesuvius eruption, thus the pyroclastic cover (maximum thickness 2 m), is the result of only one eruption. In the past this sector, whose bedrock is constituted by fractured limestone, experienced large flowslides. Southern sector of Lattari Mts. (Fe). This zone is located in the distal axial area of the air-fall products of the 79 AD eruption. The thickness of the pyroclastic cover, which overlie fractured limestone, can reach only 1 m, thus only small flowslides (tens thousands of cubic metres) can take place. Sorrentina peninsula and Capri island (Ff). Also this sector is located in the marginal area of the air-fall deposits of the 79 AD eruption. The pyroclastic cover is discontinuous with a thickness less than 1 m. Only small local flowslides can develop. Irpinia hills (Fg). The bedrock consists of flysch or clay. The slopes, which are quite gentle (15–25◦ ), are located in the axial zone of the air-fall products of Somma-Vesuvius and Phlegrean Fields systems. The cover is generally thicker than 2 m. In this context can be generated flowslides having a moderate size (less than 10, 000 m3 ) but a long run-out. Salerno Mts. (Fh). This sector is located in the marginal dispersion area of the air-fall deposits of Somma-Vesuvius system. Discontinuous pyroclastic covers up to 1 m thick rest on fractured carbonate rocks. Isolate small flowslides can take place. Picentini Mts. (Sector Fi). The zone is located in the very distal dispersion area of the air-fall deposits of Somma-Vesuvius and Phlegrean Fields. The pyroclastic covers have an high clay content and include ancient paleosoils reaching a total thickness of about 2 m. Flowslides are not usual and can reach a volume of a few tens thousands of cubic metres. North-eastern sector of the Irpinia hills (Fl). It is located in the distal area of the dispersion zone of the

1955

Figure 3. Zoning of the macro-areas B and F. 1. Potential sources of flowslides in pyroclastic materials. 2. Sector; 3. Flowslide or group of flowslides. 4. Boundary of municipality. 5. Pyroclastic air-fall deposits of Somma-Vesuvius and of Phlegrean Fields.

air-fall deposits of Somma-Vesuvius system. A thin pyroclastic layer (max 0.5 m) crops out. Flowslides are small (a few hundreds of cubic metres).

5

CONCLUSIONS

Rainfall-induced landslides in non lithified pyroclastic deposits can display different movement patterns and present very different magnitudes, thus a distinction among the types of landslides which can be provoked by slope failure is of prominent importance. This can obtained by an intelligent mix of geomorphological and geotechnical data. For large areas, even an approximate macro-zoning of flowslide prone slopes using all available information, including geological and geotechnical data and historic data on old landslides, can be of great help. This is the key idea which informs such a paper whose principal scope is to draw a general overview on the potential impact of flowslide in Campania Region. A more detailed zoning at a larger scale can be obtained through geotechnical testing on each pyroclastic level, as shown by Picarelli et al. (2007)

who calibrated such a model in an area subjected to previous landslides. This approach can be supported by accurate surveys concerning geomorphological predisposing factors, which allow to calibrate, and possible improve, the geotechnical model (Di Crescenzo & Santo, 1999; 2005). REFERENCES Cascini, L., Guida, D., Romanzi, G., Nocera, N. & Sorbino, G. 2000. A preliminary model for the landslides of May 1998 in Campania region. In A. Evangelista & L. Picarelli (eds.), The Geotechnics of Hard Soils—Soft Rocks; Proceed. int. symp, Napoli, 3: 1623–1649. Balkema, Rotterdam. Castro, G. 1969. Liquefaction of sands. Ph.D. Thesis, Harvard Soil Mechanics Series n.81, Harvard Un., Cambridge Mass. De Vita, P. & Celico, P. 2006. Distribuzione delle coltri piroclastiche sui versanti carbonatici perivesuviani e suscettibilità a franare. Giorn. di Geol. Appl. 3: 145–151. Di Crescenzo, G. & Santo, A. 1999. Analisi geomorfologica delle frane da scorrimento-colata rapida in depositi piroclastici della Penisola Sorrentina (Campania). Geogr. Fis. e Din. Quat. 22: 57–72.

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Di Crescenzo, G. & Santo, A. 2005. Rapid earth flows in the carbonate massifs of the Campania region (Southern Italy): morphological and morphometric data for evaluating triggering susceptibility. Geomorphology 66: 255–276. Di Girolamo, P., Ghiara, M.R., Lirer, L., Munno, R., Rolandi, G. & Stanzione D. 1984. Vulcanologia e Petrologia dei Campi Flegrei. Boll. Soc. Geol. It. 103: 349–413. Eckersely, J. 1990. Instrumented laboratory flowslides. Géotechnique 40(3): 489–502. Hungr, O., Evans, S.G., Bovis, M.J. & Hutchinson, J.N. 2001. A review of the classification of landslides of flow type. Environmental & Engineering Geoscience 7 (3): 1–18. Hunter, G. & Fell., R. 2003. Mechanics of failure of soil slopes leading to ‘‘rapid’’ failure. In L. Picarelli (ed.), Fast Slope Movements. Prediction and Prevention for Risk Mitigation; Proceed. int. conf,. Napoli, 1: 283–290. Patron, Bologna. Hutchinson, J.N. 2004. Review of flow-like mass movements in granular and fine-grained materials. In L. Picarelli (ed.) Occurrence and Mechanism of Flows in Natural Slopes and Earthfills;Proc.Int.Work.; Sorrento, 3–16.Patron, Bologna. Lirer, L., Munno, R., Petrosino, P. & Vinci, A. 1993. Tephrostratigraphy of the A.D. 79 Pyroclastic deposits in perivolcanic areas of Mt. Vesuvio (Italy). Journ. of Volc. and Geoth. Res. 58: 133–149. Olivares, L. & Picarelli, L. 2001. Susceptibility of loose pyroclastic soils to static liquefaction: some preliminary data. In M. Kühne, H.H. Einstein, E. Krauter, H. Klapperich & R. Pöttler (eds.) Landslides—Causes, Impacts and Countermeasure; Proceed. Int. Conf., Davos: 75–85.

Olivares, L. & Picarelli, L. 2006. Modelling of flowslides behaviour for risk mitigation. Phys. Mod. in Geot., Proc. 6th int. conf.; Hong Kong, 1: 99–112. Taylor & Francis, London. Picarelli, L., Evangelista, A., Rolandi, G., Paone, A., Nicotera, M.V., Olivares, L., Scotto di Santolo, A., Lampitiello, S. & Rolandi, M. 2006. Mechanical properties of pyroclastic soils in Campania Region. In T.S. Tan, K.K. Phoon, D.W. Hight & S. Leroueil, Characterisation and Engineering Properties of Natural Soils; Proceed. Int. Symp, Singapore, 4: 2331–2384. Taylor & Francis, London. Picarelli, L., Olivares, L. & Avolio, B. 2007. Zoning of slopes susceptible to liquefaction in pyroclastic soils of the Campania Region. Engineering Geol., submitted for publication. Rolandi, G., Bertolino, F, Cozzolino, G., Esposito, N. & Sannino, D. 2000. Sull’origine delle coltri piroclastiche presenti sul versante occidentale del Pizzo d’Alvano (Sarno—Campania). Quaderni di Geologia Applicata: 7–I. Rosi, M., Principe, C. & Vecci R. 1993. The 1631 eruption of Vesuvius reconstructed from the review of chronicles and study of deposits. J. Volcanol. Geotherm. Res. 58: 151–182. Sladen, J.A., D’Hollander, R.D. & Krahn, J. 1985. The liquefaction of sand, a collapse surface approach. Canadian Geotechnical J., 22: 564–578. Yamamuro, J.A. & Lade, P.V. 1997. Static liquefaction of very loose sands. Canadian Geotechnical J., 34: 905–917.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Zoning methods for landslide hazard degree Jianping Qiao & Lili Shi Inst. of Mountain and Environment, CAS, Chengdu, China

Lili Shi Graduated University of Chinese Academy of Sciences, Beijing, China

ABSTRACT: The methods used practically in zoning landslide hazard degree, are summarized that include the comprehensive factor analyses and index system of bottom factors, the superposition of comprehensive factors, the fuzzy determination of bottom factors, and the superposition of weights of bottom factors. Three primary problems are also discussed: the choice of evaluation index, the establishment of weight, and the classification of hazard degree. 1

INTRODUCTION

Zoning degree of landslide danger involves analysis of the environment conditions, the triggering factors and the situation of the landslide, which together determine the danger risk. Several steps are concerned with the zoning: 1) choosing evaluation indexes; 2) establishing models; 3) classifying hazard degrees; 4) testing the result; 5) quantifying the degrees. Despite the literature on landslide hazard degree is extensive, accepted standard and methods are still not yet any general. In generally, three methods are mostly used. one is single factor method that is to choose one of the factors which is most correlation of landslide (such as terrain or gradient) in research area to evaluate landslide hazard degree (Larsen & Torres-Sanchez 1998, Gao 1993). Another is multi-factor method that is to evaluate landslide hazard degree of area through fuzzy determining internal factors and a few effect factors that are all correlation of landslide (Yin, 2001; Wang et al. 1992, Zhao, 2002, Wu et al. 2002).The other is superposition of internal factors of slope that is superposition valuated factors correlation of landslide, such as distribution density, terrain condition and stratum lithology (Zhu et al., 2004, Dai & Lee, 2002, Finlay et al., 1997, Miller, 1998, Carrara et al., 1991).

2 2.1

Three-class-assessment system was established (Figure 1) (Qiao, 2000). In the diagram, the number from 1 to 12 stand for the elected factors: geomorphology, formation lithology, geological Structure, cut density, rain intensity, human destruction, earthquake strength, erosion strength, distribution density, distribution age, scale Type and hazard degree. The first class indices (LDD—I): These are type indices for categorizing the structure of factors so that categorize types of structure. The second class indices (LDD—II): These are sub-type indices for classifying individual factors under the first class. The third class indices (LDD—III): These indices assign values for the 12 factors in the second class according to the danger degrees. Index system of landslide hazard degree

LDD- -ECi Controlling factor

Triggering factor

Hazard factor

LDDSecond class index

INDEX SYSTEM OF ZONING LDDThird class index

comprehensive factors analyses

2.1.1 Index system Controlling factor, triggering factor and hazard factor are mostly factors in assessment of landslide hazard.

Figure 1. Block diagram of the index system for danger degree division of landslide.

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2.1.2 Classification of hazard degrees According to the five-level classification, the hazard degrees are listed in table 1:

Step 1 Meth od

Ui (U 1

Model

2.2.2 Classification of hazard degrees According to three-level classification, risk evaluations are listed in table 2: Table 1. Classification of the danger degrees of landslide. level

Classification

Hazard evaluation

I

High

II

Middle-high

III

Middle

IV

Middle-low

V

low

Giant Landside (V≥1000 Mm3 ), high frequency Large Landside (100 Mm3 ≤V< 1000 Mm3 ), very high frequency Middle scale landslides (50 Mm3 ≤V1500 m Fuzzy 0.02 0.64 0.02 0.02 0.02

Class

Figure 4 shows the landslide susceptibility map in the study area, expressed as probability of occurrence with the value range from 0 to 1.

3

Legend 0.5 m 1 km 1.5 km 2 km 2.5 km 3 km others Fuzzy 0.9 0.76 0.5 0.18 0.14 0.05 0.02

Table 5. slope.

4

2

Fuzzy membership value for evidential map of

1

2

Legend 0–5 5–20 Fuzzy 0.05 0.1

1988

◦

3 ◦

20–35 0.1

4 ◦

35–50 0.25

5 ◦

50–65 0.05

6 ◦

65–80 0.01

7 ◦

80–90◦ 0.01

Table 6. Fuzzy membership value for evidential map of aspect. Class

1

2

3

4

5

6

7

8

Legend Fuzzy

0–45◦ 0.01

45–90◦ 0.25

90–135◦ 0.20

135–180◦ 0.10

180–225◦ 0.10

225–270◦ 0.08

270–315◦ 0.05

315–360◦ 0.10

Table 7. Fuzzy membership value for evidential map of NDVI. Class

1

2

3

4

Legend Fuzzy

0–105 0.05

105–160 0.10

160–210 0.30

210–255 0.05

Table 8. Fuzzy membership value for evidential map of drainage network density. Class

1

2

3

4

5

6

7

8

Legend Fuzzy

5–262 0.05

262–519 0.05

519–776 0.05

776–1033 0.05

1033–1291 0.17

1291–1548 0.07

1548–1805 0.6

1805–2320 0.18

Landslide susceptibility map

operator

OR operator

OR operator

lithology

fault

rock types & structure

aspect

slope

landform

elevation

OR operator drainage distance network to river

NDVI

triggering factors

Figure 3. Fuzzy inference network designed in the study.

approach is appropriate for landslide susceptibility assessment. 5

Figure 4. Landslide susceptibility map in the study area, expressed as probability of occurrence.

DISCUSSION AND CONCLUSION

In this study, fuzzy logic has been used for assessing landslide susceptibility considering such factors as rock types and structure (lithology, fault), valley slope geometry and landform (aspect, slope and elevation ), and triggering factors (drainage network, NDVI and distance to river) in the study area. The research result is very coincident with the actual field conditions. The final map, which represents landslide susceptibility map in terms of susceptibility classes, shows that the occurrence of high susceptible zones are concentrated in the three areas along Yangtze river. Comparison between spatial distribution of these classes and the occurrences of the known landslides shows the results are very accurate and of good quality. Fuzzy

logic could determine the fuzzy membership value for evidential map of the main factors based on the actual occurrence of the known landslides and the experiences and knowledge of the experts on the regulation of the landslides in the area. So the methodology is very useful for landslide susceptibility assessment.

ACKNOWLEDGEMENTS This work was supported by the special research program of the Ministry of Science and Technology of the

1989

People’s Republic of China (No.2002DIB10044) and the research program of Changjiang River Scientific Research Institute of China (No.YWF0722/AQ02). REFERENCES An, P., Moon, W. M. & Rencz, A. 1991. Application of Fuzzy Set Theory for Integration of Geological, Geophysical and Remote Sensing Data[J]. Can. Jour. Exploration Geophysics., 1991, 27 (1):1–11. Bonham-Carter, G. F., Agterberg, F. P. & Wright, D. F. 1988. Integration of Geological Datasets for Gold Exploration in Nova Scotia[J], American Society for Photogrammetry and Remote Sensing 1988. 54(11):1585–1592. Bonham-Carter., 1994., Geographic Information Systems for Geoscientists: Modelling with GIS. Pergamon, Oxford, 398 pp. Du, R. 1990. Landslide and Debris Flow of the Reservoir Region of the Three Gorges Project on the Yangtze River. Geological Press, Beijing (in Chinese).

Huang, R, Li, Y. 1992. Logical model of slope stability prediction in the Three Gorges Reservoir area, China. Proceedings of the Sixth International Symposium on Landslides—Glissements deterrain. Balkema Publishers, A.A., Christchurch, pp. 977–981. Wu, S., Shi, L., Wang, R., Tan, C., Hu, D., Mei, Y. & Xu, R. 2001. Zonation of the landslide hazard in the forereservoir region of the Three Gorges Project on the Yangtze River. Engineering Geology 59, 51–58. Zimmermann, H. J. 1991. Fuzzy Set Theory andand Its Applications: 2nd edn. [M]. Boston: Kluwer Acad. Pub1. Dordrecht, 1991: 299. WANG Zhiwang, LI Duanyou, CHENG Qiuming. Zonation of the landslide hazards based on weights of evidence modeling. The proceedings of the 11th congress of the internation society for rock mechanics, 2007. 7:1321–1324

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Prediction of the spatiotemporal distribution of landslides: Integrated landslide susceptibility zoning techniques and real-time satellite rainfall Yang Hong NASA Goddard space Flight Center, Laboratory for Atmospheres, Greenbelt, MD, USA School of Civil Engineering and Environmental Sciences, University of Oklahoma, Norman, OK, USA

Robert F. Adler NASA Goddard Space Flight Center, Laboratory for Atmospheres, Greenbelt, MD, USA

George J. Huffman NASA Goddard Space Flight Center, Laboratory for Atmospheres, Greenbelt, MD, USA Science Systems and Applications, Inc., Lanham, Maryland, USA

Dalia Bach Columbia University, USA

ABSTRACT: Predicting global landslide occurrences is very difficult and expensive in terms of time and money. Drawing upon recent advances of satellite remote sensing technology, an experimental landslide prediction model is developed to identify the timing for landslides induced by heavy rainfall, the primary trigger. This system includes three major modules: (1) zoning the global landslide hotspots from a high-resolution geospatial database; (2) a real-time multi-satellite precipitation estimation system at fine spatiotemporal scales; and (3) a simplified decision making procedure which integrates the landslide susceptibility zoning map and the rainfall information to locate the spatiotemporal likelihood of landslide occurrence. A trial version of the system is updated every 3-hat NASA Website. Validation is underway through comparison with various inventory databases and news reports of landslide disasters. Success of this prototype system bears promise as an early warning system for global landslide disaster preparedness and mitigation given the fact that landslides usually occur after a period of heavy rainfall. Additionally, it is possible that the warning lead-time can be extended by using rainfall forecasts (1–7 days) from operational numerical rainfall forecast. Ultimate goal of this work is to provide landslide decision support tools that rapidly disseminate landslide potential alerts for disaster mitigation activities on a global basis for end users. 1

INTRODUCTION

Rain-induced landslides rank among the most devastating natural disasters, causing billions of dollars in property damages and thousands of deaths in most years around the world. Society has dealt with landslide hazards primarily by trying to site development away from hazard zones based on experience; but such an approach has limitations. Growing populations plus related environmental impacts such as deforestation have put a growing number of people at risk from landslides. At the same time, the required data infrastructure and analysis capabilities required to minimize injuries and deaths due to landslides are not yet practical in most developing countries in the tropics that need them the most. The challenge facing our science community is to better understand the surface and

meteorological processes leading to landslides and determine how new technology and techniques might be applied to reduce the risk of landslides to people across the globe. This article discusses the use of information from satellite remote sensing in the study of rain-induced landslides on a global basis, with an eye toward developing a system to detect or forecast such events on a global basis. Landslide warning systems can save lives and reduce damages if properly implemented in populated areas of landslide-prone nations (Sidle & Ochiai, 2006). Such systems usually map landslide hazard zones first and then attempt to predict probabilities of landslides and associated consequences. Today, the possibility exists to take advantage of advances in satellite remote sensing and other global data sets in the development of: 1) global landslide susceptibility

1991

maps based on satellite-based Digital Elevation Maps (DEM), satellite land-cover information and digital maps of soil characteristics; and 2) high time resolution, multi-satellite precipitation analyses with sufficient accuracy and availability to be useful for detecting the heavy rainfall events that provoke landslides. The combination of these products potentially provides information on the ‘‘where’’ (susceptibility) and ‘‘when’’ (rain events) of landslides and the potential to detect or forecast landslide events. It remains a matter of research to implement these concepts into a cost-effective method for capacity building in landslide risk management for developing countries. 2

A GLOBAL LANDSLIDE SUSCEPTIBILITY MAP

Landslide occurrence depends on complex interactions among a large number of factors, which Dai et al. (2002) broadly classify into two categories: (1) preparatory variables that make the land surface susceptible to slide (slope, soil properties, lithology etc); and (2) the triggering variables that induce landmass movement (e.g. rainfall). Field surveys are the most exact method to assess landslide susceptibility. However, performing such surveys in data-sparse or mountainous regions is very difficult and in many countries, remote sensing information may be the only source available for this purpose (Catani et al. 2005, Nadim et al. 2006). Recent advances in remote sensing techniques contribute to determining landslide susceptibility by providing information on land surface features and characteristics. This global view takes advantage of high resolution DEM data from the NASA Shuttle Radar Topographic Mission (SRTM; http:// www2.jpl.nasa.gov/srtm/). The 30-m SRTM data, used to derive topographic factors (slope, aspect etc), provide a major breakthrough in digital elevation mapping of the world. In addition, digital maps of soil characteristics prepared by the Food and Agriculture Organization (http://www.fao.org/AG/ agl/agll/dsmw.htm) and satellite-based land cover information (e.g., from NASA’s Moderate Resolution Imaging Spectroradiometer [MODIS; http:// modis.gsfc.nasa.gov]) are combined with the information from the SRTM to estimate a static landslide susceptibility index for each point on the globe over land. The satellite precipitation information in this study includes the Tropical Rainfall Measuring Mission (TRMM) Multi-satellite Precipitation Analysis (TMPA; Huffman et al., 2007; http://trmm.gsfc.nasa.gov). As needed, the various land datasets are downscaled by linear interpolation to the SRTM full resolution grid to provide the susceptibility information at the finest

Figure 1. (a) Global Landslide Hazard Index and Hotspots and (b) landslide occurrences collected from news reports and other sources during period of January 2004 through September 2006.

resolution. A global landslide susceptibility map is then derived following Hong et al. (2007a) from these geospatial data based on each factor’s relative significance to the sliding processes using a weighted linear combination approach; slope and lithology are the primary factors, while land cover type and soil properties are secondary in importance. Fig. 1a shows the resulting global Landslide Susceptibility Index (LSI) map with a descriptive scale ranging from ‘‘negligible’’ to ‘‘high’’. Excluding permanent snow/ice regions, Fig 1a shows that the low LSI areas cover about half of the land (52%), while the areas of high LSI (4%) are mostly located in tropical or sub-tropical regions: the Pacific Rim, the Himalayan belt, South Asia, the Maritime Continent, Central America, Northwestern USA and Canada, Rocky Mountains, the Appalachian Mountains, the Caucasus region, the Alps, and parts of the Middle East and Africa. Figure 1b shows the spatial distribution of major landslide occurrences collected from news reports and other sources during the period of January 2004 through September 2006. The distribution of landslide occurrences in Fig. 1b generally confirms the regions identified by the derived LSI map, although both susceptibility and a trigger are required for each event. 3

RAINFALL AND LANDSLIDE

The spatial distribution, duration, and intensity of precipitation play important roles in triggering landslides.

1992

Comprehensive modeling of the physical processes involved in landslides helps pinpoint causes of landmass movement (Keefer and Wilson, 1987; Iverson et al., 2000) in relation to rainfall. However, data requirements for implementation of such models can often be prohibitive, leading to simplification of the processes for practical use (Gritzner et al. 2001). In practice, landslide occurrence has been related empirically to rainfall intensity-duration statistics from rain gauge information for specific regions (Larsen and Simon, 1993; Godt et al., 2006) and on a quasi-global basis (Caine, 1980). The recent development of high time resolution, multi-satellite precipitation analyses has provided the potential of detecting heavy rain events associated with landslides in tropical and temperate latitudes without regard for the availability of rain gauges, an issue which frequently limits the application of the previous studies. By using the precipitation information from TMPA, Hong et al. (2006) derived the first satellite-based rainfall intensity– duration threshold curve from landslide cases in various climate and geological locations (Figure 2), in parallel to the previous rain-gauge–based studies. Note that the TMPA-based threshold falls below Caine’s threshold, likely because the TMPA is an area-average value, rather than a point accumulation. Knowledge of landslide susceptibility as displayed in Fig. 1 (the ‘‘where’’ of the problem) and the ability to detect heavy rain events that meet threshold conditions as shown in Figure 2 (the ‘‘when’’ of the problem) provide the basis for exploring the potential and limitations of such approaches for analyzing and studying the occurrences of landslides, and even possibly forecasting them. A preliminary evaluation using the information in Figs. 1 and 2 demonstrates the potential effectiveness of this approach, at least for the 25 large events examined (Hong et al. 2007b). Taking the landslide cases in Fig. 2 and only keeping the ones with greater than moderate susceptibility in Fig. 1, the probability of detection (POD) is 0.76, 19 successful detections out of 25 occurrences. However, the results also indicate that this first-generation system fails to identify landslides triggered by short-duration

Figure 2. Satellite-based rainfall intensity-duration threshold curve for triggering landslides (red; Intensity = 12.45 Duration−0.42 ) for landslides (squares) that occurred around the globe in the period 1998–2005 and rain-gauge–based threshold curve from Caine (1980), in blue (After Hong et al. 2006).

heavy rainfall events ( 1) or unstable (K < 1) two conditions which can not reflect the meticulous extent of unstable slope conditions. As a result, LEP model was unable to satisfy

the present demand on typical landslide quantitative analysis. Quantitatively conducting typical landslide hazard degree assessment obtains more attention with the development of landslide assessment research, Previous studies about typical landslide hazard assessment mainly use factors superposition of the qualitative and semi-quantitative models (Qiao. 1991, Wang & Kong 2001), By using AHP method, Subsequent studies have gradually introduced index weights based on expert experiences, and utilized this model to appraise hazard degree of Baota landslide in Three Gorges area. (Fan, et al. 2004) The entropy originated from thermodynamics in 1948. Information entropy is a quantitative degree of system confusion, and it can forecast the development trend of system, so It was quickly applied in valley System Simulation (Ai 1980). At present, information entropy method has been widely used to determine the weight of index in natural hazard and environment integrated assessment such as debris flow, drought, sandstorm, etc. (Li, et al. 2002, Ren 2000, Yi & Shi 1994, Mon, et al. 1994). Landslide is an open system of exchanging materials and energy with the environment, so it can be measured and described by information entropy method. The entropy of landslide refers to the extent of various factors influence the landslide development. Important factors provide more entropy in the index system. As a result, the value of entropy can be used to calculate objective weights of index system. This paper set up an index system of typical landslide hazard degree assessment. Based on information entropy method, subjective weight of each index factor

1995

was calculated by the basic data of 21 typical landslides in Wushan County, Three Gorges, and then a qualitative typical landslide hazard degree assessment model was established. Afterwards this model was applied in evaluating the hazard degree of the landslides in Wushan County, eventually the results was carried on contrast and confirmation through field investigation. 2

METHOD

2.1 Index selection Hazard analysis formally consists of determining the probability that a damaging event, of intensity equal to or greater than a threshold value (Varnes 1984). From the statement above landslide hazard assessment analysis should requires: (1) the intensity assessment of the landslide event and (2) its probability of occurrence (Romeo, et al. 2006). Therefore the index group should be composed of factors reflecting the possibility and intensity of landslides.

ri,j pi,j = m

i=1 ri,j

Where Ej = entropy value, pi,j = specific density of eigenvalue, ri,j = eigenvalue, K = (ln m)−1 . 2.4

Weights of index

The weights has attributed the role the factors plays in the synthesis assessment, bigger value indicated the factor’s function is more important in index system. Vj W j = m

j=1

Positive effect: ri,j =

⎧ ⎪ 1 ⎪ ⎨ x−x

min

⎪ x − xmin ⎪ ⎩ max 0 ⎧ ⎪ 1 ⎪ ⎨ x

−x Negative effect: ri,j = ⎪ xmax − xmin ⎪ ⎩ 0 max

x = xmax

xmin x xmax

x = xmin

x = xmin

xmin x xmax

x = xmax

(2)

Afterward, the eigenvalue matrix R was formed by m landslide samples and n index system: ⎡ r 1,1 ⎢ r2,1 R=⎢ ⎣ .. .

rm,1

2.3

r1,2 r2,2 ··· rm,1

· · · r1,n ⎤ · · · r2,n ⎥ ⎥ ⎦

· · · rm,n

Entropy calculations

Ej = −K

m 

After weights of index factors has been determined, typical landslide hazard degree assessment model can be define as:

i=1

Hi =

(3)

n  j=1

wj · ri,j

(6)

Where Hi = landslide hazard degree, Wj = weight of index factors, ri,j = eigenvalue.

3.1

APPLICATION The study area

Wushan County lies in the hinterland of Wushan Mountain and on the east edge of Sichuan Basin. It is on the north of Daba Mountain. The length from north to south of the whole area is 80.3 km, and the width from east to west is 61.2 km. Wushan County is located in the eastern section of Three Gorges Reservoir, and it is the key city that the Three Gorges Projects immigrants relocate. Because of the particular geological environment, Wushan County is threatened by the geological calamity of landslide more seriously. 40% of the landslides in Three Gorges Reservoir centralize in Yunyang to Wushan district. After the impoundment of the reservoir, the highest water level will go to 175.4 m alongside the Yangtze River. That will aggravate landslide and the instability of slopes. So the situation of disaster prevention and reduction in Wushan County is very severe. 3.2

pi,j lnpi,j

(5)

2.5 Model

3

(1)

Vj

Where Wj = weight of index factors, Vj = 1 − Ej

2.2 Data standardization Various factors in index system are not the same attributes and dimension, and therefore unable to directly compared, so it’s necessarily to conduct standardization before comparing calculate.

(4)

Data and index selection

This study selects 21 typical landslides data in Wushan County from the Landslide Database of

1996

Institute of Mountain Hazards and Environment, Chinese Academy of Science. After correlation and cluster analysis, 9 independent landslide susceptible factors were selected into the index system (See table 1 for basic data and index factors).

the basic data. Equation (1) and (2) were employed to standardize the basic data into eigenvalue matrix R.

Intensity factors: Volume (V ): total volume of landslide material; Height difference (Gd ): the height from the vertex to the bottom of landslide; Rock content (K): the percentage ratio of rock of landslide material; Probability factors: Gradient (G): the slope angle of the landslide; Orientation difference (Od ): the discrepancy between slope and stratum orientation; Structure (S): the discrepancy between slope and stratum angle; Slope form (Sf ): the different types of slope section shape; Strata (St ): the different epoch of strata; Lithology (L): the different types of lithology of landslide material 3.3 Model established Contributing ration method (Qiao, et al. 2006) was employed to assign the values of factor Lithology, Strata and Slope form (see figure 1), others remain Table 1. Landslides database in Wushan County, Three Gorges. Factor signs Landslide

Gd /m

St

L

O d /◦

Sd /◦

G/◦

Sf

V /m3

K/%

Renzicun Majiawu Houping Houpinglu Jiguanshibei Longmenqiao Longmenqiaoyi Yangoupo Tongzipingnan Fengtousi Guanjiaqi Caiziba Dawanli Zhangjiawn Weiqiangcun Baizhangpo Dabaoxun Caiyuanba Zhongnancun Wangjiawan Shaojituo

310 80 80 15 90 145 120 50 50 70 55 125 100 200 500 200 300 500 500 80 80

T1j T2b T2b T2b T2b T1j T1j T2b T1j T2b T2b T2b T2b T2b T2b P1 D3 T1j T2b P2 m P1

dolomite mudstone marlite marlite marl limestone limestone marl limestone marl mudstone marl marl marlite mudstone marlite marlite limestone sandstone limestone limestone

70 5 1 10 5 0 20 160 22 70 100 85 60 158 30 85 120 155 5 75 75

−4 8 −9 0 0 3 −8 15 1 −14 −22 −8 −25 23 −30 15 21 45 16 7 14

31 29 26 45 40 35 24 31 33 26 18 30 35 40 20 40 35 60 38 25 45

convex convex convex straight stair concave straight convex straight stair straight convex convex stair stair convex convex convex convex convex stair

270,000 320,000 360,000 12,000 70,000 120,000 170,000 12,000 21,600 300,000 24,000 144,000 72,000 500,000 60,000 3000,000 400,000 800,000 250,000 840,000 200,000

45 85 65 55 40 40 60 60 40 40 60 80 25 20 50 80 60 70 30 25 50

1997

The equation (5) shows that Volume (V ) weight biggest, followed by Slope form (Sf ) and Height difference (Gd ). The volume directly reflected the intensity and the harm scale of landslide, and should occupy the dominant position in the hazard recognition. Slope form provided the slope free surface, which is an important index to reflect the possibility of landslides occurrence, and the height difference indicated potential energy of the slope, is also the principal index of landslide. As a result, the objective weight of index conducting by information entropy method is consistent with the general rules of discrimination and development of landslide. The grades of landslide hazard were in common divided into 3 or 4 grades; in this study it was divided into 4 grades by tolerance 0.25 (see Table 2).

60.00% 52.38%

Contributing ration

50.00% 40.00% 30.00% 23.81% 20.00%

19.05%

10.00% 4.76% 0.00% concave

stair

convex

straight

Slope form 30.00%

Contributing ration

25.00%

28.57% 23.81%

23.81%

Table 2. Landslide characteristic.

20.00% 15.00%

14.29%

10.00% 5.00%

4.76%

4.76%

0.00% marl

dolomite

limestone

marlite

sandstone

mudstone

Lithology 60.00%

57.14%

degree

and

activity

Score range

Hazard grade

Hazard evaluation

0–0.25

low

0.25–0.50

middle

0.50–0.75

high

0.75–1.0

extreme

minor scale landslides (V < 1 Mm3 ), occasional occurrence middle scale landslides (V < 100 Mm3 ), the moderate frequency. Large Landside (100 Mm3 ≤ V 2.0, M is increasing with the decrease of m2 , that is the hydraulic condition of crosssection is becoming better with the increase of m2 . For example, when m1 = 0.5, M decreased with m2 and the hydraulic condition of the cross-section is improving. When m2 = 2.0, M has the minimum value of 6.67 and the hydraulic condition of cross-section becomes the optimal. After this, M increases with m2 and the hydraulic condition becomes worse. 3.2 The change of hydraulic condition relate to the side slope coefficient m1 when the groove transverse slope coefficient m2 is definite Figure 3 shows the relationship curve of M and m1 when m2 is 0.5, 0.6, 0.8, 1, 4, 6 and 10 separately. From Figure 3, when m2 is unchanged, the crosssection configuration parameter M is decreasing with the increase of m1 and the hydraulic condition is better. When m1 > 0.2∼0.5, M is increasing with the

2003

increase of m1 , that is the hydraulic condition is worse. For example, when m2 = 4, M decreased with m1 and the hydraulic condition becomes better; and M has the minimum value 6.69 and the hydraulic condition comes the optimal when m1 = 0.4. As the further increase of m1 , M becomes greater and the hydraulic condition becomes worse. 4

CONCLUSION

1. Assuming h1 and h2 are characteristic lengths of cross-sections, and let β √ = h1 /h√ 2 . And  it can be deduced as β =

m2

m1

√

1+m21 −

1+m22 −m2

1+m22

√

1+m21

under

the optimal hydraulic condition of the ‘‘trapezoidV’’ shaped drainage canal. And the cross-section configuration parameter of ‘‘Trapezoid-V’’ shaped debris flow drainage canal (M ) is deduced as 2  √ √ 2β 1+m21 +2 1+m22 M = under the optimal m1 β 2 +2m2 β+m2 hydraulic condition. 2. Based on the analysis of hydraulic condition, when the side slope coefficient m1 of the ‘‘trapezoid-V’’ shaped drainage canal is definite, the cross-section configuration parameter M is gradual decreasing with the increase of m2 and the cross-section hydraulic condition is becoming better; when m2 > 2.0, M increased with m2 and the hydraulic condition is worse. 3. When the groove transverse slope coefficient m2 of the ‘‘trapezoid-V’’ shaped drainage canal is unchanged, the cross-section configuration parameter M is gradual decreasing with the increase of m1 and the cross-section hydraulic condition is becoming better; when m1 > 0.2∼0.5, M increased with m1 and the hydraulic condition is worse.

REFERENCES Chen Xiaoqing, Wang Shige, Li Deji, et al. 2001. A comparison of two main types of debris flow drainage grooves. Journal of Catastrophology, 16 (3): 12–16. Fei Xiangjun, Shu Anping. 2004. Movement mechanism and disaster control for debris flow. Beijing: Tsinghua University Press, 262–267. Li Deji, Fang Guoqing, Chen Yuesheng. 1995.Debris flow drainage canal in Dongchuan City, Yunnan Province. In: Institute of Mountain Hazards and Environment, Chinese Academy of Sciences and Ministry of Water Conservancy ed. Debris Flow (4). Beijing: Science Press, 62–65. Shen Shouchang. 1994. Designing method of drainage channel for debris flow. In: The Seminar of Landslide and Debris Flow de. Proc. of the 4th National Symposium on Debris Flow. Lanzhou: Gansu Culture Press, 434–455. Wang Jikang, Huang Rongjian, Ding Xiuyan. 1996. The prevention engineering technology of Debris Flow. Beijing: China Railway Press, 93–102. Yang Wenke, Yin Chongqing. 2000. Techniques of guide channel in Dongchuan. In: Institute of Debris Flow Prevention in Dongchuan City ed. Proc. of the Study on Debris Flow Prevention. Kunming: Yunnan Science and Technology Press, 45–52. Youyong. 1999.Optimal hydraulic condition of debris flow drainage canal. Journal of Mountain Science, 17(3): 255–258. Youyong, Liu Jinfeng, Ou Guoqiang. 2006. Comparison of hydraulic conditions among usual debris flow drainage canal. Chinese Journal of Rock Mechanics and Engineering, 25 (supp.1): 2820–2825. Zhang Kaiping. 1997. Study on diversion flume of debris flow and its optimization design. The Chinese Journal of Geological Hazard and Control, 8 (4): 73–77. Zhang Kaiping, Lv Taineng. 2002. Study on scouring and deposition laws of debris flow drainage trough. Research of Soil and Water Conservation, 9 (4): 61–63. Zhou Bifan, Li Deji, Luo Defu, et al. 1991. Guide to prevention of Debris Flow. Beijng: Science Press, 125–129.

ACKNOWLEDGEMENTS The work was supported by The Chinese National Plans for Science and Technology Support (2006B AC10B04) and by The Important Heading Item of Knowledge Innovation Project of Chinese Academy of Sciences (KZCX2-YW-302).

2004

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Practice of establishing China’s Geo-Hazard Survey Information System Kaijun Zhang China University of Mining and Technology (Beijing), Beijing, China China Geological Survey, Beijing, China

Yueping Yin China Geological Survey, Beijing, China

Hui Chen China Institute of Geo-Environment Monitoring, Beijing, China

ABSTRACT: On the basis of the geo-hazards survey on seven hundred counties in China, with each county as a unit, a Geo-Hazard Survey Information System was developed. This system gathers national geo-hazard survey information and provides the functions of statistical and analysis assessment. The system contains a foundational geo-hazard database which is the most complete in the country. The database includes geo-hazard information such as landslides, rockfalls, debris flows, ground fissure and unstable slope. The amount of foundational survey information of geo-hazard locations and hidden danger spots in the database is more than 10,000. Therefore, this system can reduce the geo-hazard loss as much as possible by finding out the current situation and the development trend, by assessing the situation of geo-hazard and danger, by identifying the susceptible regions and the dangerous region, and hence by suggesting the prevention plans and arrangements.

1

INTRODUCTION

With the development and application of information technology, geologic information can be used not merely for intuitionistic browsing and simple visual judgment but for comprehensive analysis and application. The traditional information communication mode is unable to adapt to the requirement for information gathering. It requires an entire solution, including data processing, data management, information analysis and information publication. By the means of the information technique, the solution can manage and analyze the information combining data and geologic information, which can make decision intuitively by adding geographical analysis to various information systems and discovering implicit relation, rule and changing trend. The Geo-Hazard Survey Information System is an efficient method for achieving this. Since the geo-hazard warning project was launched, we have carried out geo-hazards surveys and developed regional plans in seven hundred counties. The main purpose is to reduce furthest the geo-hazard loss by finding out distribution of sites with potential geologic hazards, and the conditions of threats to population and property in these sites, forecasting the development trend, assessing the geo-hazard danger,

zoning map the susceptible region and the dangerous region, suggesting the prevention plan and arrangements, advancing the monitoring network combining experts and mass. The Geo-Hazard Survey Information System, which is based on the survey and GIS platform, gathers national geo-hazard survey information and provides the functions of synthetical statistic and analysis assessment. The system results in a geo-hazard database that includes geo-hazard information such as landslides, rockfalls, debris flows, ground fissures and unstable slopes and so on. The amount of foundational survey information of geo-hazard spots and hidden danger spots in the database is more than 100,000. The establishment and improvement of the database provides rapid and valid information services for geo-hazard prevention and control and national geo-environment management. Meanwhile, it accelerates geo-hazard survey, monitoring, prevention and control. 2

MAIN FRAMEWORK OF SYSTEM

Main framework of the system is made up of information gathering, information transmission, information

2005

Web Publish System

Achievement Demonstration System

Synthesis Achievement System

Data Statistics

Data Query

Data Distribution System

Data Service Module

Data Management

National Synthesis Database

Data Summary Data Control System Data Management Module

Figure 2. The geo-hazard distribution map of developmental degree.

Foundation Survey Database

Data Entry System

30%

Field Collection System

25%

Data Collection Module

20% 15%

Figure 1. The main framework of system.

10% 5% 0%

processing and information services. The system provides the corresponding functions according to the application demand of the users. The main functions include data input, field data collection, data quality control, data summary, data management, data query, data statistics, data publishing and data display and so on. The system has three modules: data collection module, data management module and data service module. The main framework of the system is shown in Figure 1. Every function is designed for specific users and usage stages. The data collection module is mainly used to collect data, edit data and enter data according to system requirements and a unified data collection standard, in order to ensure data consistency and integrality. The data management module can be mainly used at the stage of summary and application of data. It can summarize survey data of different regions and provide the function of query, statistics and auxiliary processing for data managers and users. It manages data effectively and provides powerful technology support for data synthetic application and development. The data service module provides services for different data users. It can publish synthesis data, show thematic data and distribute custom data. The query and statistics functions provide powerful technology support for comprehensive analysis of geohazards. The auxiliary processing can generate many types of statistics distribution maps. An example is the geo-hazard distribution map of developmental degree, shown in Figure 2.

Jan

Figure 3. time.

Feb Mar Apr May Jun Jul

Aug Sep Oct Nov Dec

Statistical diagram of the geo-hazard occurrence

80% 70% 60% 50% 40% 30% 20% 10% 0% huge

Figure 4.

big

midem

small

Statistical diagram of the geo-hazard scale.

The system can take statistics conveniently and reproduce the result data as a set of diagram. Two examples are shown as follows. 1. Statistical diagram of the geo-hazard occurrence time is shown in Figure 3. 2. Statistical diagram of the geo-hazard scale is shown in Figure 4.

3

SYSTEM DATA MODEL

To collect, manage and apply geo-hazard survey production data is the main functions of the system. The

2006

Table 1.

landslide Rock fall Geo-hazard foundation Sink background information

Ground Fissure Debris Flow Image Data

Synthesis data ID Coordinate Occur Time Hazard Type Scale … …

Spatial Data Distribution Geographic Element Layer Spatial Data Hazard Spot Distribution Layer Distribution of county Degree of Occurred Easily Layer

Preventing hazard scheme

Foundation Geological Layer Survey Report

Figure 5. The geo-hazard data model.

data includes landslide data, rock fall data, debris flow data, sinkhole data, ground fissure data, hidden hazard data (unstable slope) and so on. It also includes production report and correlative production diagrams. The core of the system is data. Data modeling is the key for how to organize and manage data so that the system can express the geo-hazard phenomena completely, then, users can query, process and analyze data conveniently. The design of geo-hazard data model adopts a method combining classic modeling and object-oriented modeling. The data model allows different thematic elements to be abstracted from some thematic layers in the form of point, line and polygons. By internal attribute correlation, the geo-hazard object-entity attribute is correlated. Then both organization and management of space data and consistency of multi-data are assured, so that GIS and database system shows the advantages for spatial information management. By researching the application characteristic of GIS in professional field, the system modeled national geo-hazard data and provided solid foundation for professional fields. The geo-hazard data model is shown in Figure 5. For designing the data model, we considered the following characteristic of geo-hazard survey information sufficiently. (1) Performance Characteristic of Analyzing and Processing The purpose of the geo-hazard survey information is the comprehensive analysis and utilization of data, especially of spatial information. For data related spatial analysis, the size is very huge and the source is multiple. Spatial analysis takes up lots of time, but not simple data query. Therefore considering availability of data sufficiently is essential. (2) Integration Characteristics of Multi-data The geo-hazard survey information contains the vector, which describes spatial distribution, the geohazard spot attribution, unformatted documents and images. For valid analysis and decision it is the initial

Landslide survey main data structure table.

Field name

Data type Description

ID Name Field ID Location longitude latitude Slide time Occurring time Landslide type Landslide property Stratum property Ground water type Annual rainfall Max day-rainfall Max hour-rainfall Flood level Slope type Length Width Thickness Area Volume Slope gradient Slope aspect Sloil property Structural area property Deformation property Geological factor Topography factor Physical factor Man-made factor Dominator factor Present state Developing trend Death Damage Hazard type Influential range Threaten population Threaten wealth

char char char char single single char char char char char char single single single single char single single single single single integer integer char char char char char char char char char char integer single char char integer single

Unit

degree degree single select single select single select multi select multi select mm mm mm m single select m m m m2 m3 degree degree single select multi select multi select multi select multi select single select single select single select person RMB person RMB

premise that the multi-data is integrated and managed. The more completely the related data collected and the closer the connection is, the more credible the result is. (3) Dynamic Characteristics of Data The managed data is finished by survey only once, but geo-hazard spot may change with time. Thus the dynamic characteristic of the data should be considered when designing data model. (4) Comprehensive Characteristics of Data The current survey achievement is mainly foundational information, including a great deal of data which reflects geo-hazard feature details. But these detail data is not analyzed. It is necessary to synthesize detail data or get useful information before analysis.

2007

Table 2. Factor system of data quality. Level 1 quality elements

Description

data integrality

entity attribute, entity relationship

Level 2 quality elements

Description

redundant

redundant degree of data in a data set absence degree of data in a data set

absence logic consistency

spatial location accuracy

consistency degree of logic rule about data structure, attribute and relationship

Accuracy of spatial entity location

concept consistency domain consistency format consistency topology consistency math foundation accuracy emendation accuracy collection accuracy

thematic data accuracy

map decoration suitability

Accuracy of ration attribute, accuracy of qualitative attribute

Configure of color, pattern, symbol and line type, Specification of map name, map number, legend, figure and inlay

classify accuracy

qualitative attribute accuracy ration accuracy symbol suitability line suitability color suitability pattern suitability graph structure rationality annotation suitability

Therefore, to synthesize data and extract data conveniently are necessary, meanwhile data mining and data aggregation should be supported. An example is the landslide main data structure table as show in Table 1.

4

consistency degree to structure design consistency degree of value to domain match degree about data storage to physical structure of data set accuracy of topology feature Accuracy of map contour, coordinate, height datum and parameter rationality of number and distribution of reference spot accord degree of spatial entity location to acceptable value accord accuracy of entity and its attribute classify to real value or a reference data set such as input accuracy of hazard spot attribute such as accuracy of value correct symbol, accuracy location, reasonable denotation correct line type, lubricity line appropriateness of professional color palette standard and rule pattern type, color, height, width is correct or not reasonable map structure Correct and readable annotation, rationality of parameter and boundary relation

DATA QUALITY GUARANTEE

With the geo-hazard survey, we have developed the information system gradually since 1999. There are

2008

great many data. So the quality of the system is particularly important. The outcome of having incorrect or inaccurate information is incorrect or inaccurate result and corresponding decision. As a result, loss could be huge. Therefore the data quality guarantee is very crucial. The system implementation conforms to ≪the work guide of geo-hazard survey information system construction and The standard of geo-hazard data quality control≫, by projects IS0 9000 19113 and 19114, and by referring DZ/T 0179–1997, DZ/T0160–95 and ≪the work guide of geo-hazard survey information system construction and The standard of geo-hazard data quality control≫ The work guide of geological map spatial database construction (and edition). Based on geo-hazard data characteristics, the factor system of data quality was established, including Level 1 quality elements, such as data integrality, logic consistency, spatial location accuracy, thematic data accuracy and map decoration appropriateness, and Level 2 quality elements. The factor system of data quality is shown in Table 2. The data check and quality evaluation methods result from the factor system of data quality. The data quality control software is developed for improving the accuracy and efficiency of data check.

5

CONCLUSIONS

Construction of the Geo-hazard Survey Information System is a complicated system engineering task, which is based on the geo-hazard survey data and applying information technology. The information technology relates to digital production technology, data quality control technology, spatial database technology, mass multi-map sheet data organization and management technology, data share and publishing technology. The construction of the system is not only a simple digital and software development, but also a huge innovational project utilizing information integrated technology. The achievement reflects systemic, professional, authority, reliability and superiority. REFERENCES Yang, Q.M. 2006. Study of geological hazard information system based GIS. West-china Exploration Engineering 2006(6):283–285. Wu, S.R. 2003. Study of geological hazard information system study: an example from the Fengdu county in the Chongqi city. Quaternary Sciences 23(6):683–691. Wei, J.L. 2002. Data preparation and database design for information system on geological disaster. Coal Geology & Exploration 30(1):30–32.

2009

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

A XML-supported database for landslides and engineered slopes related to China’s water resources development Yufei Zhao & Zuyu Chen Department of Geotechnical Engineer, China Institute of Water Resources and Hydropower Research, Beijing, China

ABSTRACT: China is bestowed with abundant water energy resources, ranking at the first in the world in terms of hydropower. Stability of high slopes is an extremely prominent problem when developing these power stations. In this paper, the authors introduced briefly the database of slopes related to the water conservation and hydropower construction, including general information, the classification of the slope and the content of the database and so on. This slope database can provide important references for the design and construction of slopes in similar places in future. At present, the XML (extensible markup language) technology developed fast and is applied widely. XML supports the data exchange between databases on web; it can provide the network share of the database of slope login using XML. This paper introduces the work of establishing the web database for landslides and engineered slopes using the XML technology.

1

INTRODUCTION

China is bestowed with abundant water energy resources, most of which has not yet developed. Todate, the installed capacity of hydropower accounts for about 24% of the state total power capacity (Lu Youmei, 2005). In the next 20 years, China will develop more large-scale hydropower projects. With the increasing of the number of the huge dams and large-scale reservoir; stability of high slopes will be one of the prominent concerns (Huang Runqiu, 2007). Collecting the geology and the geotechnical mechanic characteristics of engineered slopes related to hydropower development, has been believed to be of utmost important to understanding the landslide mechanism, establishing the reasonable analysis and the control methods. The number of documents that recorded landslide is extremely huge, but their styles, contents and levels are different. It is very difficult to utilize these record documentations if they are not undertaken in accordance with a unified methodology and format. Therefore, as early as in 1987, the work of the landslide to register had started. The landslide inventory requests the complete technical information of the landslides or slopes in according with the specifically designed tables in order to build a database. In the past several years, China’s water resources and hydropower workers had recorded 117 slopes with complete technical information, and published these documentations in the web. These documentations will provide the reference for the design and the control of the related slope project. Figure 1

shows the distribution of the 117 slopes and landslides related China’s hydropower development (Chen Zuyu et al. 2005). With the advent of internet it is possible to create an internationally shared database for engineered slopes. Under such background, further work towards a webbased landslide and slope inventory has started. In past several years, XML (extensible markup language) technology has rapidly developed and found wide applications. It advantages of having unique standardization, might expand, have overcome HTML’s shortcomings. It is more important that XML supports

Figure 1. The distribution of the 117 slopes and landslides in China (the number in circle show the amount of slope in the area).

2011

the data exchange between databases on web; we can enjoy the slope database on web by using the XML technology (Gu Bing, 2007). Now, in China, under the auspice of ISRM Commission on Case Histories in Rock Engineering, some pioneered works has begun. Introduce the database briefly.

g. The recording of slope destruction and destabilization: general information, the loss situation, the landslide geometry shape and the size, destruction situation, adopt engineer measure, the landslide plane figure (sectional drawing ) and so on; h. References: inner literature, published literature and some picture and so on.

2

These contents, reflect various aspects of slope engineer more comprehensively, according to these contents, we can conveniently to obtain the material of slope which needs.

INTRODUCE THE DATABASE BRIEFLY

2.1 Introduction The work of landslide inventory started in 1987, organized by a world working whose report provided the information, software and terminology of documenting landslides (Chen Zuyu. 1995, Cruden, D. and Brown, W. 1992). From 1995 to 2000, under the background of ‘‘85’’ national key scientific and technological project, the work of recording engineered slopes was carried on among China’s water resources and hydropower community. This work record the basic information of engineered slope case histories using standard methods, that include geology, geotechnical properties and the slope material, design parameters, construction details, and instrumentation and monitoring information, etc (China Institute of Water Resources and Hydropower Research, 1995). 2.2 The database The content of the registered slopes covers the comprehensive information that includes: a. Basic information: the type of slope, the geography information, the main character, the stability situation, the factor of destruction, the general description and so on; b. Engineering geology: rock mass structure type, lithologic character, joint, ground water, decency, earthquake, plane figure, sectional drawing, joint statistics, and rock mass quality and so on; c. Geotechnical mechanics character: interior soil (rock mass) physical and mechanical properties test, field test, crustal stress test; d. Design of the slope: Stability analysis, design parameters, government situation and so on; e. Excavation the demolition of slope: the spots of slope excavation, the demolition technology, the situation of projectile filling, the parameters of powders hole, the security measure, the security situation of slope, the actual ground speed, the acceleration and so on; f. Instrumentation and monitoring: Monitor project, instrument and arrangement, observation data and so on;

2.3

Classifications of the database

2.3.1 Classifications based on the types of the slopes According to the slope property, the water conservation and hydropower engineered slope can be divided into three kinds: excavation slopes, reservoir slopes and the river bank slopes. The number of each kind slope registered in table 1. 2.3.2 Classifications based on the height of the slopes Table 2 shows the classification and statistics of the slope registered in the documentation of the slope database based on the height of the slope. From table 2, one may find that in water conservation and hydropower engineering, most slopes have heights more than 100 meters, which is a prominent factor that affects the stability problems. 2.3.3 Classifications based on the rock mass structures The rock mass structure of a slope can be divided into 6 types that are: massive, layered (including synthetic layered, reversal layered and oblique layered), the disintegration and loosening. The statistical situation is shown in Table 4. The statistics result of each kind of rock mass structure of unstable slope is shown in Table 3. According to the statistical result, the layered structure slope in the slope engineer occupies the majority. The layered rock mass structure is further divided into three kinds of structures, namely: Synthetic layered, reversal layered, and obliquely layered Table 1.

The statistical table according to the slope property.

Slope classification

The excavation The reservoir slope slope

The river bank slope

Amount Total

42 117

29

2012

46

Table 2. The statistical results based on the height and stability status of slopes. h < 10 m

Slope classification

10 m < h < 50 m

50 m < h < 100 m 100 m < h < 200 m h > 200 m

Stability Instability Stability Instability Stability Instability Stability Instability Stability Instability

Excavation slopes 0 Reservoir slopes 1 River bank slopes 0

0 0 0

0 0 0

2 0 0

2 1 1

9 0 1

8 6 1

15 2 4

2 18 6

4 18 16

Table 3. The statistical results of unstable slopes based on rock mass structures.

Destruction type

Massive

Synthetic layered

Avalanche Slide Burst Topple Crack Flow Compound Statistics

1 8 0 0 1 0 0 10 14.3

1 10 2 0 0 0 1 14 20.0

Total Percent

Statistics

Reversal layered

Oblique layered

Disintegration

Loosening

Amount

Percent

2 2 0 4 0 0 2 10 14.3

1 6 1 0 1 0 1 10 14.3

0 2 0 0 0 0 0 26 37.1

1 23 0 1 0 0 0

6 51 3 4 2 0 4 70

8.6 72.9 4.3 5.7 2.8 0 5.7 100 100

Table 4. The statistical results based on different rock mass structures. Rockmass construction

Massive

Synthetic layered

Reversal layered

Oblique layered

Disintegration

Loosening

Amount Total

14 117

26

15

17

3

42

Table 5. The statistical results based on the failure modes of the synthetic layered slopes. Destruction type Slide Stability situation Instability Stability after reacting Creep deformation Total

Scale Cambered volume Amount Percent Avalanche Plane surface Wedge Topple Crack Burst Compound Flow (104 m3 ) 7

50

2

14.3

5 4

35.7 100

1

2

4

165!∼4200 1

1

2 4

1 5

1

2013

1 2 2

22.64∼30 7.0∼1800

1

Table 6. The statistical results of massive slopes based on the failure modes. Destruction type Slide Stability situation Instability Stability Afterreacting Creep deformation Total

Scale Cambered Volume Amount Percent Avalanche Plane surface Wedge Topple Crack Burst Compound flow (104 m3 ) 3

30

6

60

1 3

10 30

1 1

1

3 1

1

5∼9

2

0.01∼140

3 1

1

1

15.6 5∼9

Table 7. The statistical results based on the triggering factors of failures. Failure modes Factor

Distortion and Amount Stability Percent destruction Percent Comment

Water 62 Storm rainfull 32 Reservoir filling 18 Groundwater 3 Rain induced groundwater 6 Erosion 3 Human activity 44 Excavation 41 Mining 3 Others 11 Gravity 7 Earthquake 4 Total 117

30 15 10 1 3 1 12 12 0 4 3 1 46

48.4 46.9 55.6 33.3 50 33.3 27.3 29.3 36.4 42.9 25 39.3

32 17 8 2 3 2 32 29 3 7 4 3 71

slope, for details, see table 5. the destruction statistic of massive structure slope to see table 6. The factors that trigger the failure can be divided into natural and man-made. The statistical results of 117 slopes in the water conservation and hydropower projects are shown in Table 7.

3

THE APPLICATION OF XML IN THE DATABASE OF SLOPE RECORD

3.1 Introduction of XML XML is a markup language and similar with HTML in programming, a subset of SGML (the Standard Generalized Markup Language to generate the standardize document of ISO8879 published in 1986). It inherits the self-defined markup character, and

51.6 53.1 44.4 66.7 50 66.7 72.7 70.7 100 63.6 57.1 75 60.7

Primarily large and middle scale or giant landslides

Primarily middle and small scale wedge slide, crack and the Large-scale avalanche Topple, the avalanche and burst, slide.

changes the deficiency of HTML in function, to have more extendable character (Gu Bing, 2007). XML has some features: 1. Extendable. XML is a language to create markup, to create new markup to use. Thus its use level can be extended finitely; 2. Simple to understand. XML code is text-based, unlike other ASCII language. So it can be edited by usual edit software. And it expresses directly and easily understands; 3. Information exchanged with different platform. As XML is simple to understand, its format can be used to mark different data type. Only if there has a XML decode system between the exchanging plats, the right information can be obtained by interpreting the marked data;

2014

Figure 2. The relationship of XML file, XML Schema and XSL document.

Figure 4.

The display of ‘manwan.xml’ file in IE6.0.

XSL document can be displayed clearly in the user’s browser. 3.3 XML slope data file example Figure 3. The structure of ‘slopedata.xsd’ file.

4. International. At the beginning of XML proposed, the international was considered. So it is founded on Unicode. XML’s files can be viewed by IE with the aide of CSS (cascading style sheets) and the extensible stylesheet language. Because XML-file is only used to store data, not including other information such as format, et. al., it is generally used to process the data. First an ‘.xsd’ XML Schema file is created, to be judged between the style and element character. So that to determine the requirement of the XML document, that is to say the ‘.xsd’ files descript the character of them. It is necessary to use same style material file in database storage. Another ‘.xsl’ file is defined how the XML document is viewed in browser. The relation of XML file, ‘.xsd’ and ‘.xsl’ file is expressed as Figure 2: 3.2

XML format of slope data file

Based on the standard document of slope login, the structure file ‘slopedata.xsd’ is created, including every main factor. The structure of slopedata.xsd file is shown in Figure 3. In the tree-based structure, the related information of individual structure is included of each slope case. Thus using the XML managing technology, the information can be described in details. The combined

Figure 4 shows the XML data file ‘manwan.xml’ for the left bank slope in Manwan waterpower project using XML technology. By loading slopex.xsl, manwan.xml is displayed in IE6.0. HTML file is as Figure 4.

4

USE OF XML DATABASE NETWORK SHARE

XML has many advantages in database application. First, Cross platform. XML file is text-based file, not only restricted to OS and software plat. Second, it is simple and straight-forward. XML has the ability of Schema’s self-description which can be understudied and auto-processed by computers. Third, XML describes not only the structural data, but also the sub-structural, even non-structural data. Now we are constructing a slope database based on Internet share, combining the SQL Server 2000 and XML, incorporating the network programming technology. Figure 5 is the develop structure of SQL Sever 2000 and XML combination system. According to the different system of structure and service, different visit component or protocol is used. Considering the capability and programmable reason, it is often used to couple the logic and data visit. Thus we can using the standard components to realize it, such as OLEDB, ADO and NET API and so on (Xing Chunxiao et al. 2006). Through above analysis, based on the database of slope login, we can add, search or browse the slope data file by internet. From the internet, we can share

2015

engineering database under the help of international committees. We believe, in the near future, the information of geotechnical engineering over the world can be searched and browsed on the internet with the slope login files. REFERENCES

Figure 5. SQL Server’s XML visit system.

the cherish resource of our world, and every engineers of each country can obtain their slope data file concerning. This job is complicated, much quality, and many people to cooperate. It has been carried on. Near the future, we believe it can share on the internet. 5

CONCLUSIONS

China is developing its water power quickly, especially in Southwest areas. Stability of high slopes has been a main concern. Collecting slope data files can provide valuable references for solving high slope problems. The advanced network and computer technology have made a database of collected slope information possible. Now the generalized XML technology can realize the data transfer cross platforms. It has simple readable and extensible characters. By the relate knowledge of database, it can provide the network share of the database of slope login using XML. The Chinese committee of rock mechanics and engineering is working on the establishment of slope

Youmei Lu. 2005. Hydropower and sustainable development in China [J]. Water Resources and Hydropower Engineering. 36 (2):1–4 (in Chinese). Runqiu Huang. 2007, Large-scale landslides and their sliding mechanisms in China since the 20th century. Chinese journal of rock mechanics and engineering. 8909∼8913. Zuyu Chen et al. 2005. Rock slope stability analysis-theory, methods and programs. Beijing. China Water Power Press. Bing Gu. 2007. XML practical technology course. Beijing. Tsinghua university press. Zuyu Chen. 1995. The working group party on China’s water resources related landslide inventory. Landslides, Bell(ed.), Balkma, Rotterdam, ISBN 905410032X. 2011∼2013. Cruden, D. and Brown, W. 1992 Pregress towards the world landslide inventory. Prec. Symp. 6th Int. Conf. Landslides. Christchurch, Balkema. 59∼64. China Institute of Water Resources and Hydropower Research. 1995. 8.5 Chinese National Programs for Science and Technology Development-the collection of research report of the stable analysis method of rock high slope and the software system[R]. Beijing: China Institute of Water Resources and Hydropower Research (in Chinese). Chunxiao Xing et al. 2006. XML data management. Beijing. Tsinghua university press.

2016

Landslide and engineered slopes in China

Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Failure and treatment technique of a canal in expansive soil in South to North Water Diversion project Y.J. Cai, X.R. Xie, L. Luo & S.F. Chen Changjiang Institute of Survey Planning Design and Research, CRWC, Wuhan, China

M. Zhao Changjiang Geotechnique Engineering Corporation, Wuhan, China

ABSTRACT: Expansive soil problem is one of the most complex engineering geologic problems associated with the middle line of the South to North Water Diversion project. Landslides and lining failure of canals often occur in surrounding expansive soil or rock. Deformation behavior of the expansive soil and failure mechanisms of the canal during construction are studied. The canal failure is revealed to be induced by four main resources including the deformation due to expansion-contraction of the soil, the surface erosion, the shallow and deep landslides. Based on the study, practical techniques are proposed for treatment of different positions of the canal.

1

INTRODUCTION

The expansive soil problem is one of the major engineering geologic problems for canals in the middle line of the South to North Water Diversion project of China (SNWD project). There is 386.8 km of expansive soil (rock) along the canal line (Cai et al. 2007). It shows that the strong and medium expansive soil slope is still unstable if the canal bank is not specially treated although the slope is gentler than 1:3∼1:3.5. The stability of the weak expansive soil slope is relative better than others, but the expansive deformation will still cause the canal lining structure failure. In addition, it’s also possible that the canal excavation meets some soft intercalations in the soil and rock from Quaternary to Neogene. The creep deformation will take place while the soft intercalation appears at the slope or at the bottom. The deep sliding will be possible during the grave rainfall season. The expansive soil is normally of super-solidification. But it will become very loose by several expansions-contractions at ground surface environment. And the erosion gully will easy take place, which cause the loss of the soil and water, the canal deposition, and the generation of the shallow sliding. To prevent these problems, the treatment is often expensive and complex. Also, the different methods are necessary for the different parts of canal. Selecting an economic, safe, and friendly environmentally plan is an important task for the investigation and design of the canals in the middle line of the South to North Water Diversion project.

Since the 50’s of last century, a lot of studies have been realized. The three basic features, the soil substitution and modification treatment, and more than ten reinforcement methods have been proposed. But a systematic expansive soil failure theory and treatment design method are still absent. According to the slope failure investigation, the field observation and the mechanism study, the design pansy of the expansive soil in the South-North water transfer middle canal is proposed. 2

EXPANSIVE SOIL IN SNWD PROJECT

Along the canal, the expansive soil and rock arrive in 386.8 km. The most excavation depth of the canal is less than 10 m, but local depth will surpass 30 m. The length of the expansive rock arrive in 169.7 km, and the expansive soil 279.7 km. Among these, the strong expansive rock reaches to 34.2 km, the medium expansive rock 58.73 km, the weak expansive rock 76.79 km, strong expansive soil 5.69 km, the medium expansive soil 103.5 km and the weak expansive soil 170.5 km. The age and the origin of the expansive soil are not the same in different regions. 2.1 Taocha–Shahenan canal section There are 180.15 km of expansive soil and rock, which occupies 75% of the total length of the canal. The expansive rock includes Neogene grey mudstone,

2019

silt mudstone and argillo-limestone. The expansive soil includes the lower Pleistocene (Q1 ) red silt clay, the medium Pleistocene (Q2 ) grey silt clay and clay, the superior Pleistocene (Q3 ) yellow silt clay and clay. The Q1 formation is normally of weak to medium expansibility. The Q2 silt clay is normally of weak to medium expansibility and the clay is of medium expansibility. The Q3 formation is normally of weak expansibility. 2.2 Shahenan–Huanghenan canal section There are 63.48 km of expansive soil and rock, which occupies 27% of the total length of the canal. The expansive rock is mainly the Neogene red mudstone, and locally argillo-limestone. The expansive soil includes the Q2 silt clay and the Q3 silt clay. 2.3 Huanghebei–Zhanghenan canal section There are 71.69 km of expansive soil and rock, which occupies 30.2% of the total length of the canal. The expansive rock includes Neogene red mudstone and grey-yellow argillo-limestone. The expansive soil is mainly the Q2 silt clay. 2.4 Zhanghebei–Guyunhenan canal section There are 71.49 km of expansive soil which is mainly the Q1 clay. In the SNWD project, the expansibility is mainly determined on the free expansion ratio. The analyze reveals that there is a good relation between the free expansion ration and the expansion force. In Henan province, the Q2 formation is of typical expansive soil feature. And in Hebei province, the Q1 formation is of typical expansive soil feature. 3

ENGINEERING GEOLOGIC PROBLEM OF EXPANSIVE SOIL CANAL

Because of the sensibility to water, the expansive soil canal slope and basement problem is more complex than the common canal, and the treatment is much more difficult. The expansive is of three features, the expansibility, the super-solidification, and the fissuring. Nevertheless, the soil structure zone division, non-saturation characteristic, and the strength dynamic change have more important influence to the expansive canal design[2] . During the excavation, the protection treatment is necessary to assure the slope stability and to prevent the soil inside the design profile from the atmosphere influence. For decreasing roughness, the canal slope below the first bench is often lined with the thin concrete plate, which lets the soil to be in a relative closed environment. The formation of the atmosphere influence zone will be limited. But

the concrete plate can’t prevent the canal water into the inside soil. The rainfall water could also penetrate into the soil behind the concrete plate through the fissures. The engineering experience shows that the expansion deformation of the soil behind the lining and the concrete plate crack are not avoidable if there is not the special protection. The engineering treatments propose is to prevent the soil from the deformation for the part below the first bench, to assure the slope stability and to prevent the slope surface from the erosion for the part above the first bench. 3.1

Canal bank failure

In 1970, while the excavation of the entrance of the Taocha canal reached to the Q2 formation, the rainfall season began and 14 landslides taken place. These landslides may be divided to two types, one is the relative deep slide taking place along the limit between the Q4 and Q2 , and another is the shallow slide taking place inside of the Q2 formation. The depth of the deep sliding is normally 2 to 5 m, maximally 8 m. The shallow sliding depth is only 1 to 3 m. The construction of the Purna canal in India began in 1955 and finished in 1968. Part of the canal is located in the expansive soil region, which causes a lot of sliding every year. After the first repair with the masonry, the sliding continued (Kulkarni & Sawaleshwarkar 1987). The Yindan canal in China is finished in 1974. It was in good situation of function until 2005. In October 2005, a large deep sliding taken place after one month of rainfall, with a maximal depth of 19 m and a volume of 35×104 m3 , which is the largest expansive soil landslide in China (Cai et al. 2007) (see photo 1). The expansive sliding could be divided to two types, the shallow one and the deep one (Cai et al. 2006). The shallow sliding, normally small but frequent, takes place not only during the excavation stage, but also in the function period. The deep sliding, normally

Photo 1.

2020

Taocha 2005 landslide.

large but not frequent, takes place particularly in the function period, but also during the construction stage.

4

3.2

4.1 Formation mechanism of the erosion gully

Canal deformation

The Frint-Kern canal, 245 km of long, 6–6.7 m of high, meets the expansive soil nearby the city of Porterville. After 3 years of function, the lining bending and crack were found, the bank sliding is also grave[5,6] . The expansive soil is sensible to water. If the water content in expansive soil changes, the concrete lining, often 8 to 10 cm thick, will be destroyed by the expansion and contraction. The lining cracks will help the water penetrating into the soil and make the soil water content change gravelly. Since 50’s to 80’s of last century, almost all of canal in the expansive soil region has met the basement deformation and the lining crack problem. 3.3 Canal slope surface erosion The expansive soil, a super-solidified soil, is of high strength when it has not been influenced by the outside environment. 3354 groups of shear test in the natural water content displays that, in Nanyang basin, the cohesion of the weak-medium expansive soil arrives in 63 to 78 kPa, the correspondent friction angle arrives in 16 to 18 degree. With such a high strength, it seems difficult to be eroded for the super-solidified soil. However, this type of slope displays a fragile feature on surface erosion. After 30 years of function, there appears a lot of gully at the bank toward to south in the Yindan canal, whose depth arrives in 1 to 2 m and the interval is often 1 m. This gully is particularly frequent in the Q2 formation (see photo 2). Relatively, the gully is less frequent at the bank toward to north, whose depth is mainly less than 0.5 m. It shows that the large change of temperature and moisture at southern bank is favorable to the formation of the atmosphere influence zone. The atmosphere influence extent depends also on the soil expansibility.

Photo 2. Slope surface erosion at Yindan Canal.

FAILURE MECHANISM OF EXPANSIVE SOIL CANAL

The field investigation reveals that the gully is formed by the surface flow and the groundwater. The gully situation is different at different bank and at different soil formation. (1) Weather factor The gully appears mainly at the southern bank, which is probably related to the temperature variation, the rainfall and the evaporation. The fundamental influence of the weather condition is reflected by the change extent and frequency of soil water content. The more the water content changes, the more intense the expansion and contraction will be, and the more grave the soil structure damage will be. The exploration well shows that there is a good relation between the soil structure and the atmosphere influence zone. Vertically, the soil structure could be divided to 3 zones, one is the intense influence zone where the soil structure is completely destroyed and the soil is loose, another is the transition zone where the origin structure remains almost and the fissure is developed, the third zone is the normal soil where the soil is relatively closed to the outside environment. The thickness of the first zone is often 1 to 2 m. In the transition zone whose thickness is normally 2 to 3 m, there exists often the perched groundwater whose level fluctuation is less than 3 m. In dry season, the perched water could be dried-up. The normal soil, super-solidified and no saturated, is of high strength with less fracture. Evidently, the first zone is easy to be eroded. The gully depth is also correspondent to the thickness of the intense influence zone. The erosion gully is discontinued following the slope direction. The gully trailing edge is steep, and the groundwater spring is often found there. It shows that the gully is caused not only by the surface flow erosion, but also by the groundwater permeation action. The gully develops with the tendency of backward. (2) Lithologic factor The Yindan canal basement is composed of Q3 , Q2 , Q1 formation. Among these, the Q2 soil is typical of expansive soil with high solidification and high expansive force. Although with large free expansion ratio, but the Q1 soil is of the red clay feature with great pore ratio and weak solidification. The Q3 soil, with weak solidification and low expansion force, is normally of low expansibility. The development of erosion gully has a good relation with the formations and the expansibility. The gully in Q2 formation is dense (with mean interval 1m) and deep (normally 1 to 2 m). The gully in Q1 formation is sparse (with mean interval 3 to 5 m) and relatively shallow (normally less than 1 m). And in

2021

the Q3 soil, the gully is the most shallow and scattered (with the depth normally less than 0.5 m). It shows that the atmosphere influence depth and extent is not the same at different expansive formations. For the SNWD project, the Q2 in Henan province and the Q1 formation in Hebei province are the study emphasis.

4.2 Shallow sliding mechanism The shallow sliding appears not only during the canal excavation, also in the period of function. (1) Shallow sliding during excavation stage There exists naturally certain fissure in the expansive soil. And the montmorillonite richness phenomenon is often fund in the fissure (see photo 3). At the atmosphere influence range, the fissure is normally characteristic of extrusion smooth surface because of the expansion and contraction (see photo 4). These smooth fissures, with low strength, exist as natural weak discontinuity.

The saturation degree of the normal soil is mainly near 85% to 90%. Once the soil exposes at ground surface, the groundwater will lose rapidly and the crack takes place soon afterwards. In exploration well, the contraction crack appears 1 to 2 hours after the excavation. The combination of the new cracks with the existing fissures could constitute the small potentially unstable bloc. When the rainfall season comes, these cracks provide a good penetration passageway for the rainwater and form the perched aquifer at the transition zone, which will cause the expansion deformation and soil deterioration. This is probably the direct factor inducing the shallow sliding. Although the shallow sliding volume, normally 101 to 102 m3 , is limited, but it’s of high frequency and difficulty to foresee, having a great influence to the construction. (2) Shallow sliding in the function period The shallow sliding of function period takes place normally at the slope above the first bench, where there is not lining. The sliding depth is mainly less than 5 m. This type of sliding is closely related to the formation of the atmosphere influence zone. The canal slope above the first bench is often nude, or there are only some light protection treatments. As time goes on and the outside environment acts, the soil structure changes gradually. Once the influence zone forms, the shallow sliding could be induced by a grave rainfall. The sliding surface corresponds well to the perched aquifer. The soil softening by the perched water, the expansive force induced by the water content rising, and the water pressure in fissures could promote together the shallow sliding. The volume of the function period sliding could arrive in 103 m3 . 4.3

Photo 3. Montmorillonite richness in fissure.

Photo 4. Smooth fissure and friction trace.

Deep sliding mechanism

The deep sliding is related to the intercalation, whose volume varies from 102 m3 to 105 m3 . When there exist two layers of formation with evidently different strength, the intercalation appears easily along the limit. And in the expansive soil, the montmorillonite enriches often in the intercalation, which makes the limit to be a surface controlling the slope stability. As the intercalation is exposed while excavation, the sliding could take place with the help of rainfall. Several landslides appeared during the excavation of the Taocha entrance canal, disturbing gravely the construction. The Taocha landslide appeared in 2005, 30 years after the construction. The landslide, maximally 19 m deep and 35 × 104 m3 , is the largest one in the expansive soil region. The sliding surface, with 5 to 8 cm of thickness, pursues the intercalation between the Q1 and the Neogene (Cai et al. 2007). Before the sliding, the rainfall continued more than one month. The sliding mechanism is similar to the horizontal sliding

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often fund in the three-gorge region. The static water pressure play a important role of promoting the sliding. The sliding 30 years after the construction shows that the slope has undergone a long process when the cracks develop toward to bottom and the potential sliding zone strength decays gradually. After the excavation, the un-loading and the expansion will lead to an accumulative displacement at the existing soft surface, which will cause the directional arrangement of mineral and decay the shear strength up to the residual one. The shear strength obtained from the back analysis of the Taocha landslide is almost the same as the residual one obtained from laboratory (Cai et al. 2007). 4.4

Failure mechanism of the lining

The lining failure is related to the expansion-contraction deformation of the soil behind the lining. In most case, this failure has no relation with the slops stability. The soil expansion-contraction means that the water content has changed. There are three origins of the soil water. One is the canal water seepage through the lining plate. Another is rainfall penetration through the soil fissures. The third is the groundwater accumulation under the concrete lining by the capillary action. The soil water under the lining drains mainly by the evaporation and the capillary action toward to the ground surface. After the concrete lining coverage, the water content in the soil under the lining has a rising tendency, which could be the direct factor causing the lining plate crack. K.R.Datye fund that, at the same time of preventing the canal water seepage, the geomembrane can limit the groundwater movement and evaporation, causing a water accumulation at the boundary of the geomembrane and the soil, and inducing easily the soil weakness and the slope sliding (Liu 1988). In Taulierville canal of Tunis, the soil groundwater accumulation under the lining and the concrete crack phenomena have also been fund [7] . This mechanism has been examined by a field monitoring in Israel. The water content rising under the bitumen can arrive in 15% (Liu 1997). These studies show that the factor causing the soil water content change is not only external, but also internal. The treatment of seepage prevention can eliminate the external cause, but not the internal one. 5 5.1

EXPANSIVE SOIL CANAL TREATMENT Protection during the excavation

The excavation period protection is for preventing the shallow sliding, the deep sliding and the soil damage inside the design profile. (1) Selecting the right excavation time There is a small risk in the dry season for the construction. The soil water content lost and crack will

take place while it is exposed in the air. And the slope sliding is normally impossible if it doesn’t rain. In Nanyang basin, when the soil state evolves from nosaturation to saturation, the cohesion changes from about 80 kPa to 20 kPa. (2) Remaining protection layer In the excavation period, the sliding is normally shallow and the air influence depth is also limit. Remaining the protection layer, usually 50 to 70 cm, could prevent the soil inside the design profile from the damage. (3) Holding soil moisture Holding the excavation surface moisture can prevent the soil water lost and the crack. (4) Temporary slope protection There are several temporary protection methods, such as jetting mortar, laying geomembrane or sand layer et al. (5) Resistance for anti-sliding If there exist the potential deep sliding, the antisliding pile could be used. 5.2

Protection of canal lining

The key of preventing the lining deformation and crack is preventing the soil expansion and contraction. Through the treatment of the soil under the lining, it’s possible to eliminate or reduce the soil expansibility. Even if there exist still the expansion, it could be well-distributed by the cushion of the treatment layer. (1) Substitution treatment The soil substitution can cut off the water exchange between the air and the expansive soil. This method is widely used in China, India, South Africa and America. The substitution material is often the no-expansive clay. The substitution thickness is normally 1 to 1.5 m. This thickness is suitable for the SNWD project. When the loading is 25 kPa, the expansion deformation of the Nanyang expansive soil remains only 5% to 15% of the one without loading (Liu 1988), When the loading arrives in 50 kPa, the deformation remains only 0 to 3% (Table 1). The study by Katti (1987) shows when the substitution thickness exceeds 60 cm, the lateral force decrease evidently. When the thickness is greater than 100 cm, the situation is similar to the normal soil (Katti 1987). For the practical need, the treatment thickness of 0.6 to 2.0 m could be selected depending on the soil expansibility. (2) Feature-changing treatment Mixing the feature-changing material into the expansive soil can eliminate the expansibility. There are three types of feature-changing material, cement or chalk, sand and chemical product.

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Table 1. Soil expansive ratio at different pressure. δep / %

geomembrane

Formation

δef /% P/kPa δe /% 25/kPa 50/kPa 100/kPa

Q3 Q2 Q2 Q1 Q1

60 66 72 72 73

silt clay silt clay clay silt clay clay

58.8 6.4 1.04 0.21 −0.24 64.3 10.02 1.32 0.24 −0.13 162.5 18.2 1.83 0.97 0.39 38.8 4.84 1.2 −0.48 −1.23 48.9 7.98 −0.07 −0.98 −1.52

10cm concrete plate geomembrane 10cm sand

δef —free expansive ration P—expansive force δe —expansive ratio without loading. δep —expansive ration with loading.

When the cement or chalk is mixed with the expansive soil, the basic icon exchange will take place and the molecule texture of clay mineral will be changed. Meanwhile, the soil strength will be raised. This method is very common in the canal project. The cement is mixed from 3% to 5%, and the chalk is mixed from 7% to 9%. When the expansive soil is mixed with sand, the free expansion ration can decrease evidently. When the mixing percentage arrives in 30% to 50%, the expansive soil can normally be changed to no-expansive soil. (3) Geosynthetics treatment There are several geosynthetics and several treatment ways. The methods often used are geogrid, geofabriform, and geocell. The geogrid can restrain the soil deformation and raise the soil strength. This way is effective in the landslide treatment, but there is not the example of controlling the canal lining crack. The geofabriform can limit the soil expansion and provide a weight for the soil under the geofabriform. Related to the geogrid, the geocell can directly restrain the lateral expansion deformation, being of better effect. But it’s relatively expensive and difficult to compaction. Depending on the economy, the safety, the construction condition and the lining failure mechanism, the substitution, the feature-changing and geosynthetics treatment are selected in the expansive soil canal design of Nanyang basin. (4) Canal geohydrology The groundwater influences the canal deformation, the slope stability and the lining safety. The understanding to the canal soil geohydrology determines directly the canal safety evaluation, the engineering treatment design and the project cost. The typical design profile in Nanyang basin canal is shown in the Fig. 1. The expansive soil permeability is very slight, although the air influence zone is a little permeable.

1m changed soil Figure 1.

Typical structure of the middle SNWT canal.

The movement of the perched water is normally slow. The soil under the air influence zone is no-saturated. It means that there is no groundwater flow, no water pressure to the building. With help of the geomembrane at the slope top, the rainfall infiltration will be limited. To control the deformation, it’s asked to prevent the direct contact of the geomembrane with the expansive soil. The geomembrane arrests also the capillary water movement and evaporation; meanwhile it cuts off the canal water penetration. That the water content of the soil under the geomembrane goes up causes the soil expansion and softening. 5.3 Canal slope stability reinforcement For the slope above the first bench, if there is not the protection, the slope stability will decrease gradually following the formation of the air influence zone, causing the shallow sliding. If there exist the deep intercalation at slope or near canal bottom, the integral sliding is possible. If there exist the potential deep sliding, it’s necessary to take the reinforcement treatment in advance. In the SNWD project, the limit of Q4 /Q2 , Q4 /N, Q1 /N need pay close attention. In addition, the reinforcement of the southern bank of canal must be strengthened. The treatment of the bench and the slope shoulder is the same important as the slope. The treatment purpose is to prevent the water penetration. According to the analysis, the treatment width of the slope shoulder should be greater than 5 m. 6

SLOPE STABILITY ANALYSIS

There are three difficulties for the expansive slope stability analysis, one is the soil parameter determining, another is the groundwater consideration, and the third is the analysis method.

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For the soil parameters determining, three factors must be considered, one is the soil structure. The dimension effect should be studied for the expansive soil due to the existence of the discontinuity. Another is the water content or the saturation degree, which influences the test method selection and the soil cohesion. The third is the slope protection effect. The soil strength changes depending on the slope protection method and effect. The perched water exists in the transition zone. The soil under this zone is normally not saturated. No matter the shallow sliding or the deep sliding, the traditional circle arc stability calculation is not suitable because the two types of sliding are not in circle arc form. And the present calculation method can’t consider the perched water. The expansive soil parameters should be determined considering the structure difference of the three zones, the dynamic change in the atmosphere environment, and the engineering protection method. The groundwater of expansive soil exist normally only in the transition zone with the form of perched water. The soil can’t be considered as saturated one for the stability calculation. At last, the no-circle arc method should be selected for the slope stability analysis. 7

CONCLUSION

The expansive soil distributes widely along the SNWD line. The different treatment way should be adopted according to the lithology, the expansibility, the failure mechanism and the natural construction materials situation. There are mainly four types of the expansive soil canal failure, the shallow sliding, the deep sliding, the

lining crack and the slope surface erosion. Each type displays a different formation mechanism. So in the practical engineering, the different geologic condition needs a different design mentality. REFERENCES Changjiang Institute of Survey, Planning, Design and Research. The feasibility study of the first stage of South-North Water Transfer Middle Project, Appendix 2: Engineering geology [R]. Wuhan, 2005. (in Chinese). Cai, Y.J., Zhao, M. & Yang, Y.H. 2006. Study on expansive soil canal stability and mechanical parameters [J]. Resource environment and engineering, 20(4): 373–376. (in Chinese). Kulkarni, D.N. & Sawaleshwarkar N.R. 1987. Expansive Soils in Canals: Purna project—A Case Study. Proceedings of the 6th International Conference on Expansive Soils. Cai, Y.J., Zhao, M., Yang, Y.H., et al. 2007. Taocha head expansive soil landslide formation mechanism in the South-North water transfer middle canal. Blasting, 24(104): 83–86. (in Chinese). Byers, J. 1980. Treatment of Expansive Clay Canal lining. Proceedings of the 4th International Conference on Expansive Soils. Holtz, W.G. & Bara, J.P. 1972. Comparison of the Expansive soils in the Middle Californian Basin. Proceedings of the 2nd International Conference on Expansive Soils. Liu, T.H. 1997. The expansive problem in the engineering construction. Beijing: China Architecture and Building Press. (in Chinese). Liu, T.H. 1988. Engineering geology of the Nanyang basin expansive soil. Wuhan: Survey team of CWRC. (in Chinese). Katti, R.K. 1987. Effect of CNS On Active Pressure Development in Expansive Soil. Proceedings of the 6th International Conference on Expansive Soils.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

High slope engineering for Three Gorges ship locks G.J. Cao & H.B. Zhu China Three Gorges Project Corporation, Yichang, Hubei, China

ABSTRACT: The double-line five-step ship locks of the Three Gorges Project (TGP) are built in deep rock excavation. The main part of the ship locks stretches 1,621 meters, with totally 37 million m3 of earth and rock excavated; the highest slope stands at 170 meters, and the vertical walls of the ship lock chambers range from 45 meters to 67 meters in height; A 57-meter wide rock pillar is left between the two lock lines serving as the middle-isolated pier. The ship locks are massive in scale, facing highly complex engineering conditions and unprecedented technical difficulties with no prior experience in either China or the rest of the world to draw upon. In order to resolve a series of technical difficulties in the rock engineering for the ship locks and to complete this extremely challenging project, the China Three Gorges Project Corporation, with the assistance of universities and research institutes, organized close collaboration among design institutes, research institutes, construction units and construction supervisory institutions. Together they conducted comprehensive, systematic research, adopted state-of-the-art technologies and made constant innovations. Digging started in 1995 and was completed in 1999. It was followed by the installation and commissioning of metal structures for the ship locks. In June 2003, the ship locks were put into operation, thus signaling the high-quality completion of this permanent ship lock project. During this entire process, a complete system of technologies for rock excavation and for forecasting and controlling the stability of steep artificial slopes have been developed, with significant innovations and breakthroughs having been made.

1

INTRODUCTION

The double-lane five-step continuous ship lock (DFCSL) constitutes one of the most essential structures in the Three Gorges Project. They are located in the mountain body on the north of Tanzi Ridge, which locates on the left shore of the dam. The locks were constructed by digging deep into the mountain body of granite. They comprise upstream and downstream navigation channels, the ship lock chambers, the water conveyance system and the drainage system located in mountain around. The total length of the ship lock from the entrance of the upstream approach channel to the exit of the downstream approach channels is 6442 m. The main structure of the ship locks (lock chambers) is 1,621 meters in length. It is the area with the highest concentration of high slopes, with the highest slope standing at 170 meters. The slopes exceed 120 meters in height continuously extend about 460 meters in length. Altogether, 41.96 million m3 of open cut earth and rock and 980,000 m3 tunnel earth and rock were excavated. A 54–57 meter wide rock pillar is left between the two lock lines serving as the middle-isolated pier. Therefore, there are two steep-channel slopes on the main structure (Chaoran

Zhang, 2001). In an effort to take full advantage of the strength of the rock itself and to reduce engineering costs, vertical slopes were excavated in the steep channels; the side walls of the ship lock chambers are thin concrete lining anchored on the vertical slopes, and the rock mass also forms part of the ship lock structure. Compared with regular high slopes, the high slopes of permanent ship locks have the following characteristics: they are steep, high slopes deeply cut from the mountain, featuring significant height, complicated form, and wide scope and completely release of geostress, and exhibiting remarkable unloading characters and heterogeneous features. There are strictly requirements regarding slope stability, especially deformation attributes. The Yangtze River is China’s golden waterway. As for the high slopes of permanent ship locks, the overall and local stability must to be assured and the rheology of the slopes has to be strictly controlled in order to meet the requirements of the normal operation of the miter gates of the ship locks. As such, construction faced extreme technical difficulty, numerous interferences, and a tight schedule. The ship lock engineering involved a high intensity of ground construction, and it was hard to perform digging for the vertical walls of the narrow, deep and

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steep ship lock chambers. Furthermore, a large quantity of underground tunnels and vertical shafts had to be dug simultaneously with the digging of the vertical walls. Thus, there were numerous difficult issues to be resolved, such as how to eliminate the impact of excavation blasting on one another, minimize damage to the rock mass and ensure construction safety (Changjiang, 2004a, 2004b). Figure. 1 shows an overview of the double-lane five-step continuous ship locks. A typical cross-section is shown in Figure 2.

Figure 1. Typical profile of the double-lane five-step continuous ship locks.

2

KEY TECHNOLOGIES FOR THE HIGH SLOPES OF PERMANENT SHIP LOCKS

2.1 Engineering geology of high slopes The geology of the rock mass in the slopes of the Three Gorges ship locks consists of amphibole granite (PreSinian Period), occasionally of schists, with 4 dominating sets of joints dipping at 50◦ ∼75◦ . Pegmatitic dikes are commonly encountered, which may also contribute to wedge failures. The axis of the ship lock line strikes at 111◦ . The dip directions of the left and right walls are 201◦ and 21◦ respectively. Figure 3 shows the statistical charts of the main joints in South slope. The stability of the rock masses was divided into overall stability and local stability for research and assessment. On the strength of macro geological analyses and assessments of the attitude of structural planes and the mechanical property of the rock mass, it was concluded that the slopes were overall stable and the scheme of deep slope excavation was feasible. The conclusion cast out the doubts in the engineering community about the stability of the high slopes and laid down the groundwork for a decision to be made on the scheme. Research found that local stability was a key engineering geological issue for the high slopes of the ship locks; the local stability issue had to be resolved by advance forecasting of block dynamics in conjunction with construction geology. The advance forecasting of block dynamics involved a complete set of ideas, approaches, methods and procedures; it was a scientific and rigorous task running through the slope excavation process. Over the six years of excavation and retaining of the slopes of the ship locks, more than 1,000 blocks of various types were forecasted and about 400 forecast alerts were sent out. Thanks to such forecasts, not a single engineering accident or casualty occurred as a result of overlooking an alert. Thus, the forecasting has ensured slope stability and safety as well as the long-term operation of the ship locks. The forward-looking nature and accuracy of the

Inclination slice

Distribution slice

Rose map of Distribution

Drainage hole Drainage cave

Drainage hole

Rose map of inclination

Drainage cave

Water conveying tunnel

Water conveying tunnel Isopycnal map

Figure 2. Typical profile of the double-lane five-step continuous ship locks.

Figure 3.

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Inclination-angle histogram

Statistical charts of the main joints in south slope.

forecasting and its seamless integration with design and construction have set a new precedent never seen before in the history of hydropower engineering. The geo-stress is a key factor in slope stability and deformation. Geo-stress was measured in nine deep holes and three shallow holes in the ship lock area and the total depth of geo-stress measurement reached 300 m; a three-dimensional regression analysis of ground stress was conducted as well. With these efforts, it was concluded that the maximum principle horizontal geo-stress in the ship lock area intersects with the main slope at a small angle (about 30◦ ), the horizontal geo-stress within the excavation zone in depth was at a relatively small level of 4–11 MPa, and that the direction of the stress was favorable for the slope-stability. These conclusions provided a comprehensive and reliable basis for engineering design. With regard to the research into rock mass permeability, based on the hydro-geological research findings obtained over many years, a generalized model was established for the hydro-geological structure of the crystalline rock mass at the Three Gorges. With integration of three-section water pressurization and cross-hatch water pressurization, the anisotropic hydro-geological parameters of the rock mass were obtained, which provides a scientific basis for analyzing the permeation of slope drainage. During the course of high and steep slope excavation, the stress and strain of the high slope changed sharply, resulting in a massive ‘‘tensile stress zone’’ and ‘‘plastic zone’’ in numerical analysis. The engineering significance of the plastic zone of the rock mass and the methods for reinforcing the zone have always been controversial topics in the engineering community. A typical profile with a middle-isolated pier was selected, and seven holes were drilled for comprehensive research using the combined methods of coring, sectional water pressurization, hole color videotaping, hole sonic testing, electromagnetic wave CT, hole elasticity modulus measuring and detailed geological documentation. The research revealed that the rock mass at the core of the middleisolated pier mainly remained intact in its original state; and only a part of the structure-surfaces in the top and two lateral unloading relaxation belts of the middle-isolated pier splayed; the number and density of structural surfaces and the structure-type of the rock mass mainly remained unchanged, which meant that the rock mass was still usable for engineering purposes; and that, only deformation modulus had dropped to deferent extent and the shearing strength of the structural surfaces had been weakened somewhat. On the strength of a large body of information and data on the inside-observation of the rock mass and macro geological investigations, the retained rock masses of the high slopes were divided into several unloading belts, and their corresponding unloading

rock mechanical parameters were proposed. All these provided a reliable basis for the stress and deformation analysis, the block stability assessment and the optimization of the slope reinforcement design schemes. In the research of the engineering geology of the high slopes of the ship locks, a series of new technologies and methods were developed and adopted, including a high-definition comprehensive drill hole color videotaping and interpreting system, a full hole wall color digital videotaping and interpreting system, electromagnetic wave and elastic wave tomography, geological radar, quick catalog system, fault microstructure research and activity research, isotopic measuring of the age of various types of rocks in the dam area and research on weathering mechanisms and characteristics. All of these provided strong backup for the engineering geological research on the high slopes. Figure 3 gives the statistic chart of fissures in South slope. 2.2 Research to the mechanical properties of the high-slope rock mass In view of the mechanical characteristics of the slope rock-mass of the Three Gorges ship locks, various methods, including rock-mass mechanical testing, were employed for systematic integrated research. A great deal of valuable results were achieved in studying the mechanical features of rock mass in a stretched state and in a three-way stressed state, the mechanical property of rock-mass fracture, the rheological characteristics of rock mass, the criteria of the shearing strength of rock mass, the weakening degree of the mechanical nature for unloading rock mass, the size effect of the macro mechanical parameters of rock mass. These research achievements provided an important basis for furthering the research of the mechanical nature of the rock mass and the reinforcement design of the rock-mass slopes. Following total-process testing of the granite stress and strain at the Three Gorges, the Three Gorges granite elasticity and plasticity models were established. The mechanical characteristics and yield criteria of the rock mass under shearing stress were studied by means of experiment in laboratory and field testing. In light of the restriction of the amount of allowable time-sensitive deformation of the rock masses after the high slope excavation of the Three Gorges ship locks, large scale field rock-mass single-shaft and triple-shaft compression creep tests as well as rigid structural surface and rock-mass shearing creep tests were conducted, with highly valuable research achievements attained, including the fractured rock mass creep model and the creep parameters of rock masses and structural surfaces. With respect to the high slopes and the unloading rock masses of middle-isolated pier of the ship locks,

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various approaches and methods were adopted for the testing and research of the rock mechanical properties, and the degree of weakening of the mechanical properties of the rock masses in the excavation disturbed zone was analyzed and assessed. It was found that after excavation and unloading, the mechanical nature of the rock masses underwent a numeric change within different ranges, and compared with the slightly unloaded zones, the deformation modulus of the rock masses in the strongly unloaded zones and in the weakly unloaded zones dropped by about 60% and 30%, respectively. An integrated research approach was adopted to examine the size effects of the macro mechanical parameters of the hard rock mass in the slopes of the Three Gorges ship locks. Through rock-mass mechanical testing, engineering rock mass grading, computer simulated testing and displacement back analysis, the correlation between the rock-mass deformation modulus E and the rock-mass dimensions D was established, the typical unit dimensions of the rock mass were determined and the macro mechanical parameters of the rock were sampled. Rock-mass quality was assessed by means of engineering rock-mass grading in the unloading zone and testing of rock-mass mechanical properties. On the strength of this assessment, the mechanical characteristics of the engineering rock mass were evaluated in an integrated manner with research achievements from testing, inversion and rock-mass grading taken into consideration. A combined rock-mass mechanical parameter sampling method was adopted, which encompassed statistical analysis of testing achievements, the E∼Vp correlation, computer simulated testing, engineering rock-mass grading, the Hoek-Brown criteria and back analysis of monitoring information and data. With this method, recommended mechanical parameters of rock masses in the unloading zone of the slopes of the Three Gorges ship locks were established (Xia-Ting Feng, 2004). The strength parameters based on the traditional approaches and Hoek-Brown criterion are summarized in Table 1 and Table 2. They are used in the numerical analysis.

Table 1. The strength parameters based on the traditional approaches. Rock

Weathering

Unit weight (×9.8 kN/m3 ) φ (◦ )

c (MPa)

Granite

Completely Heavily Moderately Slightly Slightly

2.50 2.65 2.68 2.70 2.68

0.05 0.20 0.50 1.50 0.15

Schist

35.00 45.00 52.40 60.90 35.00

Table 2. The strength parameters based on Hoek-Brown criterion. Rock

Weathering m RMR σc (MPa) φ (◦ ) c (kPa)

Granite Moderately 25 57 Granite Slightly 25 77 Schist Slightly 17 57

50 100 50

41.4 60.01 57.7 199.5 37.7 57.8

2.3 Research to the underground permeation and drainage measures of high slope 2.3.1 Methodology for analyzing the saturatedunsaturated permeation field of the high slopes In light of the rock-mass characteristics and hydrogeological attributes in the ship lock area, rock-mass saturated-unsaturated parameter testing and research, field monitoring and research of the effect of rainfall on the underground water of the high slope rock masses, research on saturated-unsaturated permeation model, engineering calculation and analysis of saturated -unsaturated finite element programming were conducted. These testing and research procedures were valuable in guiding the research of the fractured rock-mass permeation field on the high slopes of the ship locks and in facilitating the design of the underground drainage scheme for high-slope rock masses. 2.3.2 Rock-mass infiltration and supply process of the high slopes in rainstorms Through experiments, the characteristic parameters of the saturated/non-saturated seepage flow of the fully weathered zone under rainfall infiltration were determined. Following analysis of the parameters, an analytic model was established to interpret unsaturated hydraulic conductivity, providing important basic information for the research of the underground permeation field of the high slopes under rainfall and the effects of its drainage system. On the strength of the achievements from experiments, the rainfall infiltration and supply patterns and the characteristics of the changes of the permeation field were analyzed. The conclusions were highly useful for improving the high-slope rock-mass permeation model and for analyzing drainage. 2.3.3 Design scheme for underground drainage of the high slopes The results of the analysis of the engineering geology and hydro-geological characteristics of the high slopes, the field rock mass drainage tests and 2D/3D underground permeation fields were examined in an integrated manner, and the basic scheme for the

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underground drainage system design of the high slopes was drawn up. Through field inspection and monitoring information and data it was confirmed that underground drainage is remarkably effective and plays an important role in ensuring the stability of the high slope. 2.4

Summary of factors of safety for cross-section

By By design HoekNo. Condition Calculation site Cables rule Brown 1 2 3

Research to the design criteria of high slope and dynamic design theory

Through an integrated analysis, the geological conditions of the high slopes, the mechanical performance of the rocks, the primary scientific research achievements and the operating requirements were studied. On the basis of these studies, the design criteria for the high slopes of the ship locks were established. These criteria included design principles, design standards, design conditions and calculating and analytical method. They provided guidance for the formulation of the basic scheme for the high-slope design. In view of the characteristics of the high slopes of the ship locks and actual construction conditions, a dynamic design approach and methodology was adopted based on research and was successfully applied to actual construction. This approach was able to resolve various technical difficulties and ensured the successful completion in the high slope construction of the Three Gorges ship lock. The approach of this design and the actual experiences can act as a reference for other projects. 2.5

Table 3. 20–20.

4 5 6 7 8 9 10 11 12 13 14 15

No Phreatic Line

Phreatic Line

Phreatic Line Faults Invalid drainage

Vertical walls yes Upper vertical walls Upper wide flat Total Vertical walls yes Upper vertical walls Upper wide flat Total Vertical walls yes Total Vertical walls yes

2.009 2.240 6.429

1.895 2.058 7.166

7.814 3.255 2.011 2.251 6.073

8.809 3.523 1.895 2.058 6.848

7.900 2.936 2.510 2.011 1.685 0.682 0.875

8.738 2.308 2.119 1.949 1.415

Stability analysis of the high slopes

In an effort to gain a clear understanding of the basic characteristics of the stability and deformation of the rock masses in the high slopes, numerical simulation analyses of the rock masses in the high slopes (including the middle-isolated pier) were conducted with models of various materials, such as 2D and 3D elastic-plastic, elastic-brittle-plastic, elasticbrittle-plastic damage, and viscous elastic-plastic materials. With these analyses, the basic patterns of the stability and deformation of the rock masses in the high slopes were determined. These results provided an important basis for the design of the high slopes. A large amount of stability calculations and analyses showed that the construction stage constituted the controlling engineering conditions in overall slope stability; under the impact of a VII-magnitude earthquake and considering the dynamic amplification effect of the slope, the safety coefficient of the stability dropped by about 15% but still remained above 1.5. This is adequate for the design requirements. The conclusions drawn are consistent with the results of the macro geological assessment. Table 3 summarizes the factors of safety for crosssection 20–20 by the conventional limit equilibrium

Figure 4. Typical calculation sketches for slope stability analysis by the limit equilibrium method.

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method. Figure 4 gives some typical calculation sketches. Through calculations and cross-collations with various models, the patterns of the high slope displacement, stress and plastic zone distribution were found to be consistent. The patterns and characteristics of the high slope stress and strain were completely revealed. Rheological analyses of the rock masses in the high slopes revealed that the top of vertical slope had a displacement of only 4.16 mm over a 50-year calculation period, and that, the bulk of the displacement occurred in the slope construction stage. These analyses have led to the basic conclusions that the rheological characteristics of the rock masses in the ship lock slope are weak, which means that the rheology is stable and that there will be no long-term deformation affecting the safety of the ship lock operation. Numerical analyses of simulating the slope reinforcement showed that the anchorage cables had better results for the reinforcement of slope rock masses with fractures, crevices and other discontinuous geological structures. It was shown that they could greatly reduce excavation displacement as well as being effective on improving the stress conditions of the slope rock mass. However, they had limited reducing effects on the plastic zone. This finding was highly valuable in guiding the design of the slope reinforcement. Back analyses of rock mass parameters, tests employing multiple survey methods and field rockmass mechanical tests all showed that, after the ship lock excavation was completed, the slope rock masses could be divided into a blasting-unloading relaxation belt, an unloading relaxation impact belt and a non-relaxation belt. The plastic zone had largely occurred in the blasting-unloading relaxation belt and the unloading relaxation impact belt; the slope displacement, stress, plastic zone distribution and the drainage zone were largely consistent with the results of numerical analyses and the findings of monitoring. These results were important in verifying the basic characteristics of the high slope stress and strain. 2.6 High slope anchorage technology With large-scaled testing and research on anchorage structure and construction technologies, various problems were successfully resolved, such as grout stopping difficulty and poor grouting compactness of long horizontal anchorage cables of a large tonnage. Overloading and destructive tests also verified the safety and reliability of the pre-stressed anchorage structure. Figure 5 shows the structural details of the cables. Field tests revealed that the stress in the inner section of the pre-stressed anchorage rods and prestressed anchorage cables were concentrated on the tip of inner section and then diminished rapidly. It was

Figure 5.

The structural details of the cables.

also shown that the tensile load was primarily supported by the front 2.5 meter part of the inner section. Tests in the drilled holes showed that the precision of hole inclination could be effectively controlled by welding centering devices on the drill rods at certain intervals, and that it was feasible to drill a 70 meter horizontal hole in the gallery, with all indicators meeting requirements. Through repeated tests in the lab and in the field, R3 350# and R7 350# early-strength cement mortars were successfully developed. Furthermore, they were used in high slope engineering, delivering good results. Research on the corrosion resistance and durability of the anchorage cables showed that the underground water in the Three Gorges ship lock area had a pH value close to 8, which is slightly alkaline. Additionally, the underground water contained relatively low levels of Cl− and SO2− 4 and therefore had no corrosive effect on the steel strands in the solidified cement stones. The alkaline environment of the cement slurry provides long-term well protection for the steel. Actual measurement showed that the anchorage cables of the ship lock slope suffered an average locking loss of 2.87%. The loss averaged at 10.58% in two years after the anchorage cables were locked. The loss generally ranged from 5% to 15% after locking, and 70.5% of the anchorage cables had a loss within this range. The overall pre-stress of the reinforcing anchorage cables of the slope had a small loss and design requirements were satisfied. Field tests of the 1000 KN double-shielded (with corrugated pipe) non-bonded anchorage cables on the Three Gorges ship lock slope proved that overall compensatory stretching and single-cable stretching were both feasible, adequate to adapt to rock mass deformation and to adjust the stress level by compensatory stretching in the vertical slope section of the ship locks, thereby meeting the requirements of the nonbonded anchorage cables. On the strength of these tests, double-shielded 3000 KN terminal non-bonded anchorage cables were developed. This kind of anchorage cable with advanced structure is well suited to use in China. Such cables have been successfully used in the Shuibuya project on Qingjiang River and in the Zaoshi, Goupitan and Pengshui projects. They enjoy favorable prospects for application.

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2.7

Optimization construction methods and blasting excavation technology of high slope

The ship lock slope was an enormous project with a complicated structure. Therefore, it was essential to establish sound construction procedures. On the strength of the analyses of numerical simulations of the ship lock excavation and reinforcement processes as well as the impact of blasting excavation of tunnels and shafts and open excavation around, it was determined that: (a) the two ship lock channels should preferably be excavated downward parallel to each other and the excavation could be performed in zones divided along the central line of the ship locks. However, the difference in the height of downward excavation between the two ship lock channels or the upper and lower adjacent zones should not exceed one step section; (b) tunnel-excavation should have 20–30 m ahead of open-excavation at corresponding area. Vertical shaft excavation should have at least one blasting beach ahead of its corresponding part in elevation; (c) anchorage bracing should be performed in time just following the excavation. Drilling and blasting was the main methods of excavation of the ship lock. On the basis of test results, the technical requirements of blasting on the slope was proposed as follows: the pre-splitting blast was used for excavation of the slope surfaces; horizontal pre-splitting blast was used for excavation of the packway; the wall-top platforms of the ship lock chambers was excavated with the procedure of pre-placed protective layers; a combination of pre-splitting blasting and smooth surface blasting was used for excavation of vertical slope sections. By flexibly adopting the multi-minisecond delay bench blasting technology and smooth surface blasting, pre-splitting blasting, buffer blasting and other controlled blasting technologies, the shock impact of the blasting operations were minimized and the fragmenting results were improved, thereby ensuring the quality and progress of slope excavation. Research on the dynamic effects of the excavation blasting shows that by reducing the charge amount per delay, the peak of vibration induced by blasting could be reduced and the main frequency of the vibration could be increased, thereby the impact of the blasting on slope stability is minimized. Thus the controlling of the charge amount per delay is a key factor for minimizing the impact of blasting on slope stability.

3

anchorage rods and shot Crete support were adopted. Throughout the process, construction procedures were placed under stringent control and a complete set of blast-controlling technologies were adopted. Moreover, prototype monitoring and feedback analysis were strengthened and the dynamic design were implemented as well. All of the above mentioned constituted the comprehensive control measures. During the process of ship lock excavation, pre-splitting and smooth surface blasting technologies were adopted and enabled the slope to have a good shape. It was found that there were 52 blocks larger than 1000 m3 , 107 blocks larger than 500 m3 , and 360 blocks larger than 100 m3 , all of which were reinforced with prestressed anchorage cables and rods. The 3000 KN pre-stressed anchorage cables had a reinforce-depth of 60 meters. More than 4000 meters of pre-stressed anchorage cables were used, and as many as 100,000 high-strength 600 KN anchorage rods were used. An integrated analysis of the data obtained from prototype monitoring (as many as 3,268 instruments of various types embedded throughout the ship locks were used) shows that the deformation of the rock masses has tended to be stabilized since April 1999, when the excavation was largely completed. Field surveys showed that up until July 20,2007, the maximum cumulative displacement of the south slopes and north slopes towards the central line of the ship lock chambers were 72.07 m and 52.96 m, respectively, averaging at a monthly displacement of 1.64–1.96 mm. The maximum displacement of the vertical walls on the south and north slopes was 36.93 mm and 30.08 mm respectively. The maximum cumulative displacement of the south and north sides of the middle-isolated pier was 23.16 m and 31.50 m, respectively. The top of the lock head had a maximum displacement of 4.73 m towards the ship lock chambers and had maximum

(a)

MAJOR MONITORING RESULTS

In view of the uniqueness and importance of the high slope of the permanent ship locks, mountain water drainage, ground water blocking, interception and drainage, pre-stressed anchorage cables, high-strength

(b)

Figure 6.

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Measured displacements on the slope surface.

displacements of 2.75 m towards upstream and 2.84 m towards downstream. Deformation had largely ceased. The lock head had a displacement of no more than 0.5 mm during the lock chamber filling process after the ship locks had been put into operation, and it was completely adequate for the requirements of the normal operation of the miter gates of the ship locks and within the range of design forecasts Figure 6 shows the displacements of various surface monitoring points in association with the excavation process.

4

MAJOR INNOVATIVE TECHNOLOGIES AND ACHIEVEMENTS

In addition to basic prospecting methods such as drilling, adition and trenching, a number of high-tech technologies were also adopted, including drilling hole elasticity modulus testing technology, drilling hole full hole-wall color digital videotaping technology, electromagnetic CT tomography, earthquake tomography, geological radar and sound wave testing. During the slope excavation process, a complete set of ideas, approaches, methods and procedures for the advanced forecast of geological dynamics were proposed and developed. With the comprehensive forecasting of all kinds of geological conditions and issues, more than 1,000 blocks of various types were successfully forecasted, and not a single engineering accident or casualty occurred as a result of missing a forecast. The forward-looking nature and accuracy of the forecasting and its seamless integration with design and construction have set a new precedent never before seen in the history of hydropower engineering. Total-process stress and strain tests, singleshaft and triple-shaft compression creep tests of rock masses, tensile and shearing creep tests of rock masses, rigid structural surfaces and shearing creep tests of rock masses and various other methods were adopted to conduct systematic research on the mechanical characteristics of the rock masses in the high slopes. A great deal of valuable results were achieved in probing the mechanical nature of rock masses in a stretched state and in a three-way stressed state, the mechanical property of rock-mass fracture, the rheological characteristics of rock masses, the criteria of the shearing strength of rock masses and the sampling of mechanical parameters of rock masses. These research achievements provided a solid foundation for furthering the mechanical research and the stability analysis and assessment of the high slope rock mass. Various surveying approaches and integrated testing and monitoring technologies were adopted to conduct systematic research of the high slopes and the unloading zone in the middle-isolated pier rock mass. For the first time, the principles and methods

for dividing unloading zones into blasting-unloading relaxation belts, unloading relaxation impact belts and non-relaxation belts were proposed. Unloading relaxation belts mainly exhibited the stretching and expansion of some original structural surfaces, with slightly reduced mechanical strength in the rock mass. The limiting equilibrium method and models with various materials, such as elastic-plastic, elasticbrittle-plastic, elastic-brittle-plastic damaged and viscous elastic-plastic material were adopted to perform separate 2D and 3D numerical analyses. Integrated analyses and assessments were performed to examine the stability and deformation characteristics of the high slopes, providing a theoretical basis for establishing the basic scheme for the design of high slope excavation reinforcement. These efforts led to the conclusions that the bulk of the slope deformation occurred during the construction stage and the amount of time-sensitive deformation was minimal, thereby creating no impact on the normal operation of the ship locks. These conclusions were verified by prototype monitoring information and data. For the first time, a 68 meter vertical slope (with a total slope height of 170 meters) and a 48 meter threesided unloading vertical rock-mass of middle-isolated pier were used in the deep cutting slope excavation of hard rocks and system anchorage rods, large-tonnage pre-stressed anchorage cables and a 60 meter threading anchorage cable for the middle-isolated pier were adopted. Key technical measures, including slope surface water interception and drainage and setting-up of seven-level drainage tunnels inside the rock masses on both sides of the slopes, were adopted. These technologies have provided an effective solution to the stability and deformation problem of the high slope.

5

CONCLUSIONS

The Three Gorges permanent ship locks were formed by deep cutting in the mountain, with a middle-isolated pier rock mass reserved in the middle. The slopes are mostly suspended and the redistribution of stress and strain of the rock masses are extensive and change sharply. Slopes are high (170 meters), steep (vertical slopes being 70 meters high) and long (total length exceeding 7,000 meters). As the rock masses in the slopes form part of the ship lock structure, there are stringent requirements in regard to deformation. The success of the Three Gorges ship lock project depends on the long-term stability of the high slopes. The key technologies for the high slope engineering of the Three Gorges ship locks contain various research achievements obtained over a period of more

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than 10 years since the national key technological breakthrough program of the 8th Five-Year Plan was launched. The project is a comprehensive application of multidisciplinary advanced technologies and methods and an integration of slope dynamic design theory and practice combining geologic surveying, rock mechanics testing and research, numerical simulation analysis and rock slope engineering design. Since the ship locks went into operation in June 2003, all monitoring data has been within forecasts and deformation has tended to be stabilized. This research achievement has high theoretical and practical values and has generated remarkable economic and social benefits. It is capable of being applied in similar projects.

REFERENCES Chaoran Zhang, 2001. Practice and Experimental Verification of TGP’s Permanent Ship lock Slope, Engineering sciences, (5) 22–27. Changjiang Institute of Survey Planning Design Research. 2004. CWRC, The Key Technology of TGP Ship lock [R]. Changjiang Institute of Survey Planning Design Research. 2004. CWRC, The Key Technology of TGP Ship lock Slope [R]. Xia-Ting Feng, Zhiqiang Zhang, Qian Sheng, 2000. Identifying Rockmass Mechanical Parameters of Three Gorges Project Permanent Ship lock Using Intelligent Displacement Back Analysis Method, Int. J. Rock Mech. Min. Sci, 37(7):1039–1054.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Large-scale landslides in China: Case studies R.Q. Huang State Key Laboratory of Geohazards Prevention, Chengdu University of Technology, Chengdu, China

ABSTRACT: Landslides occurred frequently in China and they are characterized by their large scale, complex mechanism of formation and serious destruction. Described and investigated in this paper are nine landslides as well as their geological conditions and triggering mechanisms. Occurrence of the large-scale landslide is shown to be dependent on topographical and geomorphological conditions. About 80% of the landslides were found in the eastern margin of the Tibet plateau, which exhibits active tectonic activities. Intensive interactions between the endogenic and epigenetic geological processes cause serious dynamic change of high steep slopes. Large-scale landslides in this region are attributed to strong earthquake, extreme weather conditions and global climatic changes. In southern China, heavy storm with an intensity of 200–300 mm/d is also a crucial factor to triger large-scale landslides. Frozen-thawing of soils in spring is another main cause of many landslides in the loess aera of northwestern China. Recently, global warming-caused temperature rising induces glacier retreat and glacier lake collapse, etc. In general, more than 70% of the large-scale landslides were closely related to the human activities.

1

INTRODUCTION

Since the 20th century, with increasing population in the world and gradual expansion of human activities spaces, geological environment has been continuously impacted by greater and greater engineering activities which are based on the conditions of technology and economy. In addition, climate changes of the whole world, such as El Niño, landslide hazard, especially large-scale landslide hazard, becomes more frequent and causes more and more economic losses and person death (Au, 1998; Yin et al., 2000; Schuster et al., 2001). By far, landslide disasters almost have happened in all mountains in the world where human lives and engineering projects are being constructed. So nowadays, the landslide hazards stand the second geological hazards following earthquake (Li et al., 1999; The U.S. Geological Survey, 2000). In Europe, partial Scandinavia region occupied by Sweden and Norway, partial Alps mountainous region occupied by the East of France, Switzerland, Austria and the North of Italy, Appenines throughout Italy, and most of England are all high frequence landslide hazard areas (Parise & Wasowski, 1999; Collison, et al. 2000; Mauritsch, et al., 2000; Staub, 2001; Raetzo et al. 2002). Rocky Mountains lying at the edge of America Plate, Madre, Andes, which constitute the barriers of the western American mainland, and when they go across Canada, Western United States, Mexico, Salvador and Chile and so on, they make the

most landslide hazards in the world (Radbruch-Hall, et al. 1983; Li et al., 1999; Parise & Wasowski, 1999; Collison, et al. 2000; Mauritsch, et al., 2000; The U.S. Geological Survey, 2000; Schuster et al., 2001; Staub, 2001; Raetzo et al. 2002). Japan and Taiwan, Tibetan Plateau, Nepal and North India on the south of Himalayas are also a highly-happening region of landslide hazard in Asia (Yamagishi, 2000; Lin. et al., 2002; Bhasin et al., 2002). In the world, China is one of the areas where the most severe landslide hazards happen. Since 1980s there has been a trend to aggravate the landslides due to high-speed development of engineering constructions and change of corresponding natural characteristics. At present, except for Shandong Province, serious landslides happened in the other provinces of China, and the most severe areas included Yunnan, Guizhou, Sichuan, Chongqing, Tibet in Western China, the Western Hubei, western Hunan, Shaanxi, Ningxia and Gansu. According to statistics more than 400 cities and 10,000 villages suffered from severe landslide disasters, in other words, more than 0.41 million landslide events happened in those areas above, and the total area is 1.7352 million m2 , which account for 18.10 percent of the whole area of China (until 2000). Since 1995 average person death ratio caused by landslides has exceeded 1000 persons per year. Grievous landslides gave rise to great loss of local residents’ properties, and destroyed a large quantity of factories, mines. Landslides severely affect the security of

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railway, highway and hydropower stations (Li, 1992; Schuster, 1996; Duan, 1999, 2000; Wang, 1999; Jiang, 2000; Yin, 2001). Large-scale landslide has a large incidence due to its instability, great potential, and always breaks away from mother rock, and then forms high-speed, long distance collapse-slide-fluid complex hazard geological bodies (Brückl, 2001; Crosta, 2001), causing destructive breakage and heavy death. Therefore research on the landslides is very important. And it becomes an emphasis of several reports and studies (Schuster, 1996; Voight & Faust, 1992; Baum et al., 2001; Wu & Wang, 1989; Zhong, 1999; Chen & Kong, 1991; Jin, 1998; Xu, 1992; Zhang & Liu, 2001; Sun, 2000; Yin, 2000; Huang et al., 2005; Lin et al., 2002). For example, Surte earth landslide in Sweden in September 1950 held the volume of 400 × 104 m3 , resulting in forty buildings destroyed and traffic interrupted (Jacobson 1952; Caldenius, 1966; Johnson & Fleming, 2001). Vaiont reservoir landslide in Italy in October 1950 caused the death of 2500 to 2600 persons with the total volume of 2.7 × 108 m3 (Voight & Faust, 1992). Helens volcano at Washington in United States broke out in May 1980, as gives rise to large collapse-slide-fluid disaster with the total volume of 600 × 104 m3 . Heavy rainfall by Mitch hurricane in August 1998 induced El Berrinche deep-sited landslides on Honduras (Baum et al., 2001). Earthquake landslides in Salvador on January and February 2001 with the total volume of 75 × 104 m3 , caused more than 500 people deaths (Baum et al., 2001). Since the twentieth century, China has suffered from a series of large and huge landslides early or late. For example, earthquake landslide at Haiyuan, Ningxia in 1920, landslide at Diexi, Min River in 1933,

large landslide at Lugong, Yunnan in 1965 (Li, 1992; Li, 1999; Bhasin et al., 2002). Since nineteen eighties landslides in China have entered into a new period of activities with the rapid development of social economy. Jipazi landslide happened in Yangtse River in July 1983, and Sanleshan landslide in Gansu in March 1983, Xikou landslide in Huayingshan, Sichuan in July 1989, Touzhaigou landslide on Zhaotou, Yunnan in September 1991, Laojinshan Landslide on Yuanyang, Yunnan in June 1996, Badu Landslide on Nankun railway in July 1997, Yigong Landslide in Bomi, Tibet in April 2000, and Xuanhan Landslide in Sichuan in July 2004. The total volume of these large landslides is more than 200 × 104 m3 , which caused heavy loss for national economy construction and social development. 2 2.1

TYPICAL LARGE LANDSLIDE EVENTS IN CHINA SINCE 20TH CENTURY Diexi earthquake landslide in Min River

Diexi is located on the left bank of upstream Min River, Mao county, Sichuan, which is 249 km away from Chengdu. It is an old city near the mountain. Diexi is a key road to Songpan Plain, Qinghai, and Gansu. It experienced Han, Tang, Min and Qing dynasties. During the People’s Repbulic of China it belonged to Wenchuan County. August 25th, 1933, at 15:53, Diexi experienced a Magnitude 7.5 earthquake, epicenter Intensity X. The earthquake spread from Xi’an to Zhaotong (from North to South), and from Wan county to Maerkang (from East to West). The area above Intensity VIII covered 14,000 km2 (See Figure 1). In the epicentral

Min

N

Xiaohaizi

Dahaizi i

Songping Ditch

Jiaochan Landslide hazard point River and its flow direction

Diexi

Town and village destoyed

River

Lake

4 km

Figure 1. Landslide disaster distribution induced by the earthquake from Diexi to Songping Ditch after Diexi Earthquake in 1933 (from Sichuan Earthquake Bureau).

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region such as Shawan, Diexi, Jiaochangba, Houerzhai and Longchi, the sky became dark everywhere, mountains collapsing and rivers being blocked. Instantly, this ancient county was buried by rock-falls from collapse of WN Strike Mountain. Because the earthquake happened in mountainous region, it induced intense collapses and landslides of the river bank slopes and gulch ramps (See Figure 1). The disaster destroyed Diexi ancient county, Shawan, Jiaochangba, Houerzhai, Longchi and nearby 21 Qiang Villages, and made 6,865 persons die and 1,925 persons wounded. The landslide dam formed by the slope collapses produced 11 barrier lakes with different scales in which Dahaizi and Xiaohaizi are the biggest. Guanyin cliff and Yinping cliff, which were respectively located on the two sides of Min River and to the north of Jiaochangba, were confronted with each other. During the earthquake the two cliffs fell into the Min River from the hill top, so that Dahaizi river-blocking dam was formed, 800 meters long, 170 meters wide, 255 meters high. At present, there is still 98 meters deep water upstream the dam, and the storage capacity of 7300 × 104 m3 is kept (Zhong, 1999). Xiaohaizi dam is at the downstream of Dahaizi dam, and they are linked end to end. October 9, 1933, 7 pm, i.e., 45 days after the earthquake, intense after-shock gave rise to breach of 7 places, such as songpinggou and Bainazhai. Owing to much rainfall in the region, the water level of the Min River rose suddenly and sharply, so Dahaizi dam was ruined, and lake water effused, water head above Dadian arriving at more than 60 meters high, which was still 12 meters when reaching Guanxian. Flood flux at Doujiangyan was up to 10200 m3 /s. After fadeaway of the flood human and animal bodies could be discovered along the River sides. Death of more than 8000 persons were caused by the flood (Li, 2002). Dixie is located at a trigonal belt, which is composed of fold zone of Songpan-Ganzi geosyncline, parageosyncline foldbelt of Qinling and Longmenshan fracture zone. In the region there were a series of compact linear and arc overturned folds and corresponding reserve faults. In the region the exposed bedrock is metamorphite from Devonian, Carboniferous, Permian, Triassic Systems, of which main rocks are sandstone, marble limestone, phyllite and slate. In the slope calamity induced by Diexi earthquake, Jiaochang landslide is the biggest in scale. The landslide deposit lied on the left of Xiaohaizi, and its fronter was 1400 m long, along Ming River. Its altitude was from 2110 m to 2425 m. Mean length along the slope trend was 1400 m; Mean width along the slope strike was 900 m; And the mean area covered about 1.5 km2 . Average thickness of the slide was 170 m, and

its volume was 2.1 × 108 m3 . The landslide showed a wide front-edge and a narrow after-edge, and the slide gradient is from 15◦ to 35◦ with 40◦ to 50◦ of the front edge slope. In terms of substance composition, Jiaochang landslide could be divided into two parts, of which the tail presented dualistic structure. The top was lake deposit of Quaternary, and the bottom was triassic metasandstone, crystalline limestone and phyllite with fracture structure. The foreside of the slide mainly showed collapsing deposit consisting of block and sub-sandy soils. The front part overturned above pebble alluvial deposit of Ming River, and Xiaohaizi slide dam was formed (Sun & Chen, 2001).

2.2 Yalong River Tanggudong Slide June 8th, 1967, 9 Am, large-scale shore slope collapseslippages happened at Tanggudong happened on the right side of Yalong River, which was 1 km far away from Xiari Village, Posihe, Yajiang, Ganzi (Cai, 1988, 1989; Wu et al., 1996; Leng et al., 2002). The height difference between the landslide after-edge and the slope foot was 1030 m. The maximum level length was 1900 m, and maximum width was 1300 m. The area coverage of the landslide was 1.7 km2 with the volume of 6800×104 m3 . Due to high-cold mountainous area and uneasiness of vegetation discovery, huge ‘‘wound’’ formed by the slide could be still found in recent planet photos. Tanggudong gliding mass slipped into Yalong River at high speed, and rushed at the opposite bank. Therefore huge coast dam was formed, with the length of 200 m. Its base width along the River was 3050 m, and the left and right heights were 355 m and 175 m respectively. Hereafter, the water level in front of the dam rapidly jumped. The end of backwater area extended upstream continuously, finally reached Luocuo 30 km away from Yajiang county. The length of backwater area was up to 53 km, and water storage was 6.8 × 108 m3 . However, the downstream flow of dam markedly decreased, even dry. The lowest water of all year appeared in the range of 200 km to 300 km (See Figure 2). June 17th, 8 Am, the reservoir water began to turn over the dam. At 2 Pm, the dam was ruined. Extraordinary flood was formed. Water level, 10 km away from the dam, raised 48 m. The water level, at Jinhe, Yayuan, raised 30 m. Water level of hydrological station in Luning increased 20.4 m. And water level of hydrological station at Miyi increased 16.6 m (See Figure 2). The dam failure caused sudden rise of water level in Yalong River, and furtherly, influenced a trunk stream of Jinsha River and upstream Yangtse River. Water level of a hydrological station at Huili County raised 12.4 m. The water level raised 6.87 m at Pingshan Hydrological Station near Xiangjiaba and raised 2.86 m at Yinbin (Figure 2).

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Yalong River

Abeijie

Shanbeihou

Wata Andejie Mazishi Sezhiha Aluozong Laqituoka Rangong Ericuo Marihe Chengzhang Dedaola Damsunduo Bangsha’e Luoxi Milong Gedou

Wog

Xiyigan Zhuluocha

Naiyi

Figure 3.

Yaba

Malang Tang’e

Nuoluopga

Jipazi landslide in Yangtse River.

Guyi

intense physical weathering, bad quality of the slope rock, river side erosion and strong flushing at the slope toe, the slope finally lost its stabilization.

Muzirong Rega Duoruogeng

Tanggangg Pinjiujiang Zhangzhixia Zhe’a Egu Tangjia

Geyixi

2.3 Jipazi Slide in Yangtse River

Cimarong Kouduo Gaduo Bayirong

Ci’a

Hengbucuola

Xiamyixi

Luke

Shanmaxiyi Cajieqi

Riyi Muhui

Yalong River Sajiagong Bosihe Jiangyixi Nangen

Yayihe

Laka Lenggu Luodi Caiyu Jida

Songyu

Mu’en Kusi

Xiari

Mutuo Jiahe Jiju

Backwater Original riverway

9 km

Figure 2. Tanggudong landslide dam and its backwater area.

Owing to huge water volume, coast and dam failure led to sharp rise of water level downstream, and huge flux. Maximum flux was 57000 m3 /s when dam failure. The peak discharges measured at lower Guili, Nuning, Deshi and Pingshan hydrological stations were 23600, 23100, 19900 and 6091 m3 /s respectively. The whole dam failure process approximately lasted 12 hours. The sliding material of Tanggudong was composed of black argillaceous slate and heavy-film sandstone from Xikang Formation, Trias System. Owing to

In July 1982 in Yunyang, Chongqing, rainstorms continuously fell, and monthly rainfall was 633.2 mm. At 8 Pm of July 17th, Jipazi slope on the left bank of Yangste River began to become unstable. At 2 Am of July 18th, the slope rapidly slips, maximum speed 12.5 m/s. The front of the landslide was pushed into the River, and then onto the opposite bank. The maximum landslide length was 200 m. Finally a huge coast was formed, 350 m long. On the east and west sides of the landslide, the length was 1.4 km and 1.6 km respectively. The whole area covered 0.77 km2 with the volume of 1500 × 103 m3 · 230 × 104 m3 of the whole coast entered into sea-route of the River (Figure 3). In spite of no person death in the slide, the slide ruined 1730 houses. Agricultural and industrial losses were up to 6 millions of Chinese RMB. More important, owing to many rock blocks falling into the river, riverbed raised 40 m, and 700-m-long torrent and cascade were formed. Therefore ships in the Yangtse River had been forced to be canceled for 7 days. Recovering the sea-route needed 80 millions of Chinese RMB. Indirect economical loss was 100 millions of Chinese RMB. Jipazi Slide was a part of Baota old landslide. There was a natural drainage ditch on the right side of Baota landslide. The landslide’s thickness ranged from 0 to 93.7 m. Its cover layer was silty clay, and middle layer was composed of clay including block rocks and gravels, and the bottom was bedded cataclastic rock mass. The slide mass was debris of Quaternary, and slid along weak structural plane of lower bedrock, which was a renewed old coast. At the same time, the bedrock slide was induced. The slide belt consisted of amaranthine silty clay including sandstone offal, 0.2 to 1.0 m thick. Under natural conditions, it presented soft or

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Landslide border

2200

J3

Motion vector

2100

2150 250 192

Tension Low-lying pit

Landslide Qde accumulation Penglai group, J3P upper Jurassic

Primary natural riverway

2050

2000

Water level when slip occurred 370 53 2000

Qdel

420 53

760 105 1990 1990

1990

1980 1970

850 13

Current manpower riverway

River

Baxie Landslide border Original place of the house

Yangtse River

200 m

Original position of the trees House place after slide

Feather-like crack Tree place after slide

Level dist. (m) 370 Trees, house Level dist. (m) 53

160 m

690 37

1090 17

1990

2010

Contour (m)

Figure 5. Sanleshan landslide shape plan and slippage vectorgraph (after Varnes & Savage, 1996).

Figure 4. Jipazi landslide plan (after Zhang, 2000).

hard plastics, and clear scrapes could be seen in the surface of the slide with the dip angle of 6◦ ∼8◦ . The slide bed was made up of feldspar-quartz sandstone and mudstone from Penglai Formation with the dip direction of east-west direction (See Figure 4). Strong rainfall was a direct reason for the Jipazi landslide. The rainfall was 240.9 mm during 24 h before and after happening, maximum 38.5 mm per hour. During primary deformation, the soil and rock of 17×104 m3 slid from the after-edge into Shiban Ditch, and jammed it. Thus this led to strong rainfall infiltration went into the slide mass along the after-edge, as resulted in climbing of the water level in the slide mass. Meanwhile, vertical rainfall infiltration was also an important factor to make the water level rise in the slide mass, and make corresponding mechanical properties rapidly get worse. Powerful dynamic and static water pressure forced the west of the slide to fail in plastic debris flow. Hereafter, the middle and east parts slip and failed.

2.4 Sanleshan Slide in Gansu March 7, 1983, large scale high-speed slide happened at Sanleshan, located in the north of Sanle Village, Dongxiang County, Gansu. The ridge at the altitude of 2283 m slid into the valley of Baxie River at the height of 2080 m. The front of the slide got across the River valley with the width of 800 m, and passed across the River and stopped at the opposite shore slope at 10 m high. The total volume formed was 3100×104 m3 . The huge slide only spent one minute. The fatal disaster destroyed three villages near the slope toe and Baxie River valley, and caused death of 237 persons. In the slide region original slope plane was steep. The slope angle at the elevation of 2100 m ranged from 30◦ to 35◦ . There was no plant, and a series of gulleys developed. Baxie River (intersecting Yellow River at Liujiaxia Dam) is a branch of Tao River, the first-order branch of Yellow River, which is a seasonal stream in Dongxiang County with flow direction from east to west (Figure 5). With semi-arid climate annual rainfall

2041

m 2275

Main cliff

2175

S Original ground line Collapse and accumulation from main cliff Slip body of the back

2075

Baxie River

Landslide of the front

1975 1500

1250

1000

750

500

250

0m

Figure 6. The profile of Sanleshan Landslide after slipping (after Zhang, 1996).

of the area is 485 mm, 80 percent of which focuses from July to September. In 1979 it rained heavily in the region, and rainfall of that year was 650 mm, in the same year tension cracks occurred in the ridge of Sanleshan. The slope was composed of loose Malan loess, semi-hard loess and horizontal mudstone of the Tertiary from the top to the bottom. In the mudstone structural fissions developed, strike east-west, dipping to south with dip angles from 73◦ to 86◦ . The middle and bottom parts were covered by the terrace deposit, old slide debris and residual (See Figure 5). Plan of Sanleshan Slide was similar to a palm. The east-west width ranged from 700 to 1100 m, and the south-north length ranged from 1100 to 1630 m. The thickness of the slide ranged from 5 to 75 m, and the area covered 1.3 km2 (See Figure 6). Main broken cliff was 750 m long, and its height was from 220 to 240 m. Its dip angle ranged from 75◦ of the top to 45◦ of the bottom, average 55◦ . Lengthways, the slide mass was divided into two parts. The back part (north part) was composed of a series of spindly east-westtrend massifs. In the region the original land surface was covered with mudstone, offal and loess debris with the height of 8 to 50 m. Sand pebble layer on the old 1st terrace was hardly disturbed, and front-edge of fracture plane was higher than the level of the 1st terrace. This showed that the part of the slide mass moved in flow type rather than in slip type. The slip surface of the Sanleshan Slide presented chair form, which was composed of highly steep unloading fracture and horizontal mudstone layer. The slip plane cut through 120-m-high loess, and then entered horizontal mudstone and mudstone bedding plane developed at the bottom (See Figure 6). Therefore, Sanleshan Slide was developed at the mudstone slope covered with 120-m-high loess. Cutting layer and rotational slide were showed at the back of the coast, and the front moved in highspeed debris flow. Although it happened abruptly, it experienced a long-term incubation process. Gravitational creep along the layer surface at the slope toe and tension fracture at the after-edge contributed two

important phases to the landslide formation. Based on it, ‘‘locked section’’ in the slope was finally cut off, and high-speed landslide was formed. 2.5 Xikou Slide in Huayingshan July 10th, 1989, Xikou landslide induced by heavy rainfall was the greatest slip among Chinese geological hazard events in 1980s, which resulted in death of 221 persons. Its direct economic losses were more than six millions of Chinese RMB. In July 1989, infrequent heavy rain occurred in the region of Xikou, and monthly rainfall was 222.9 mm, and the maximum rainfall intensity was 88.6 mm/hour from July to October. At noon of July 10th, rainfall intensity rose and the block rocks rolled and hit the farmer houses. Henceforth, with boom from the slide source zone, the ground began to uplift and swell, and Maanping village began to incline. After this, the slide mass was turned into debris flow and run toward the Xikou Ditch, and rapidly rushed to the north of Xikou town. The whole slide event only lasted about 60 seconds, which ruined Xikou cement mill, Hongyan coal mine, Xikou Grain Barn and several villages on the way. This caused great death and heavy economic losses. The elevation of the slide original zone between the after-edge and sheared edge ranged from 848 to 655 m (See Figure 7). The zone between 790 and 848 m was broken cliff of secondary landslide, and the other one between 695 to 790 m was main broken cliff. The main broken cliff showed trapezoid form in plane, and the upper width and lower one were 75 and 110 m. It inclined northwestern with dip angle of 47◦ . Sliding plane extended from intensely weathered carbonatite into colluvium-deluvium of Quaternary, and finally got through, as a arc curve. The whole sliding plane was like a spoon, lowest elevation 655 m. Upper bare slip bed was 210 m long, and lower one was 165 m. Main and secondary slide masses volumes were 18 × 104 m3 and 2 × 104 m3 respectively. The longitudinal section of the landslide presented double-deck structure, namely, hard upper layer and

2042

290

(m)

o

Secondary glide mss 1

2

3

4

5

6

I II

Qinglongzui

II

835 810

735

III

7

IV

VII

785 760

710

V 685 Maanping

O3

VIII

660

Q

P

F6

S

F7

635 100 m 610

Figure 8.

Figure 7. Structural sketch map of Xikou landslide source zone 1 Original topography line 2 Topography line after slipping 3 Sliding plane 4 Intensely weathered zone 5 Calcareous breccia 6 Colluvial deposit from Quaternary 7 Subarea of rock structure (I: blocky cataclastic texture; II: Huge layer strucuter; III: layer strucuter; IV: layered cataclastic texture; V: cataclastic texture; VI: brecciated texture; VII: layer limestone of Ordovician; VIII: mudstone and shale of Silurian).

soft lower layer. From the side wall of the landslide, upper slide surface developed along boundary of strong and weak weathering belt. The lower part was shale, mudstone and sandstone of Silurian System. The slide developed along the colluvial deposit above the mud shale. According to rock structure characteristics, the slide source zone was divided into six subregions: layered block-crack structure—thick-layer structure—layered structure—layered shattered structure—shattered structure—brecciated texture from top to bottom. It should be pointed out that the rock mass with brecciated texture was in fact the breccia formed by renewed cementation of the fault crush zone. Owing to huge potential, the slide mass showed several phases of high-speed clastic flow, debris flow and rapid pathway. Finally the volume of accumulating coast-debris flow zone was up to 100 × 104 m3 . The landslide mainly originated from geological structure conditions of hard supper and soft lower. During geological history, the fault blocked the upper slope body from deforming. In the part, there was obvious stress concentration. Enduring creeping of the lower soft base forced the stress concentration to rise. Under particular rainfall, the deformation of the soft base increased, and locked section became unstable. This led to the occurrence of rapid coast-debris flow. 2.6

Zhaotongtouzhai slide

September 23th, 1991, 6 Pm, king-sized landslide happened at the Toutonggou village, situated on Panhe 30 km away from Zhaotong, Yunnan. The slide mass

Touzhai Landslide—earth flow accumulation.

was sheared outwards at the elevation of 2300 m, and slipped into Touzhai Ditch at a high speed. Finally, it was rapidly turned into earth and rock flow to run along the Ditch. The Ditch channel and the houses near it were buried. The slide mass stopped at the mouth of the Touzhai Ditch after several high-speed impingements and changes of direction. Finally 400 × 104 m3 soil and rock deposit was formed with 3000 m long, 130 m wide, and 10 m high. This process only took 3 minutes. The great slide calamity led to death of 216 persons, and buried 252 cattle, and ruined farmer field 2.00 × 103 m2 . The direct losses were 12 million Chinese RMB (Zhong, 1999; Chen & Kong, 1991) (See Figure 8). From local geological structure, Touzhai coast was located in the west wing of Panhe syncline. The lithology exposed near the slide zone belonged to Emeishan basalt (P2 β) of lower segment from the Permian, sand shale of Xuanwei formation of upper segment from the Permian (P2 x), and sand shale of lower Triassic (T1 ). The coast slide happened in P2 β (See Figure 9). The slide source region showed a rectangle with 300 m wide and 400 m long. With similar parallelogram in vertical section, the total volume was 900 × 104 m3 . On the surface of the slide there was intensely weathered basalt slip zone with 1 m thick. The slip belt consisted of soil and shiver particles. The main space of the slip bedding was occupied by residual slide mass and secondary side slide ones. The longitudinal surface of the slip plane could be divided into three sections: the first section (AB) was interlaminar slip surface, and remaining slide belt on its surface was the product of weathered volvanic tuff; The second section (BC): its surface was hard and smooth rock structural plane, 110◦ /∠38◦ ; The third section (CD) was locked one. The surface of basalt was coarse, weakly weathered, with clear stratified structure, and the dip angle was 5◦ . Several factors resulted in the occurrence of the Touzhai Coast Landslide. Under the control of erosion datum plane of the Jinsha River region and local

2043

113o

Height (m) 2600

142o

A

2400

B

2200

117o C

D

Yuantouzhai Village

P2

Pan River

2000 1800

P2x 0

1600

800

T1

P2 x

3200

2400

P2 Distance

Figure 9. Longitudinal profile of Touzhai landslide.

Slide zone (I)

Current landform Original landform

5000

Slide zone (II) II-2

II-1

4000

Accumulation zone (III)

F

1000

2000

3000

4000

5000

6000

7000

8000

3000

9000

2000

Figure 10. The profile of Yigong oversize Landslide—landslide area, slide area, accumulation area, original landform line, present landform line.

erosion surface of Panhe, the source of Touzhai Ditch was strongly eroded. However, owing to relatively high intensity of the basalt, the erosion was hindered. Then at the mouth of the Ditch, 400-m-high slope was formed with three free faces, dip angle 40◦ . This provided condition of the landslide. Meanwhile, strong unloading of the basalt rock mass happened because of good free faces. Annual rainfall of 1100 mm and favorable storage condition made much intensely chemical weathering, and forced the rock qualities to be more deteriorated. Thus before the occurrence of the landslide the rock mass had become loose media. Long-term creeping induced the locked section to be cut off, and high-speed catastrophic landslide was formed. 2.7 Yigong Landslide in Tibet April 20th, 2000, at 20:05, world-shaking Yigong Landslide happened at Zamulong Ditch, Yigong, Pomi, Tibet. The rock mass of 30 million m3 collapsed

from the top of mountain at the altitude of 5000 m, and fall distance was up to 1500 m. Strong impulse force of the landslide stroke detrital substance that was deposited at Zalmulong Ditch for more than a century, and broke them up, so as to form high-speed landslide only in three minutes. Very soon, clastic grain flow including block rocks was formed which ran and swept the rock and soil mass on the both sides of the mountain valley. It moved 8 to 10 km and stopped at the exit of Yigong Lake, as jammed Yigongzangbu. Finally 4.6 km long, 3 km wide, and 60–110 m high natural dam was formed with the total volume 0.3 billion km3 . The vertical distance of the Landslide was 3 km, and horizontal one was 8.5 km, and maximum rate was more than 44 m/s. This seldom happened in China, even in the world (Figure 10) (Yin, 2000). According to substance movement of different parts and deposit characteristics during the disaster, the landslide was divided into three regions (Figure 10), namely, collapse zone, landslide zone and accumulation zone. The collapse zone (I) was situated in the

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source region of Zamulong Ditch valley, and its elevation ranged from 4300 m to 5500 m with 1200 m height difference. Owing to incision of two sets of NEE-direction structural planes inclining vis-a-vis, wedge-like landslide with the volume of 0.4 × 108 m3 was formed. In terms of landslide process, landslide zone (II) was divided into instant high-speed landslide region (II-1) and high-speed block rock clastic grain flow region (II-2). Accumulation zone of Yigong Landslide showed obvious zoned characteristics. Pyramidal deposit was distributed on the surface of the front belt of Yigong Landslide accumulation zone, and some pits similar to liquefaction of sand due to earthquake could be seen at the side edges of accumulation area full of sand and dust. Many trees were uprooted outside the accumulation zone. In addition, thin-layer mud of 0.5 cm to 1 cm was attached to the trees in the margin belt of this landslide. This showed that during the rapid movement the landslide material drove the frontal air to flow at high speed, and to produce high-speed gas wave. The river-blocking event of the extraordinary Yigong Landslide included collapse-landslide clasticgrain flow (debris flow) blocking river-the lake breaking-flood-remodeling terrain of the river valleysecondary collapse and slide of the valley, which was a very complete geological hazard chain. Mechanism of Yigong landslide was as follows: first, relatively high temperature caused superfluous thawing of the ice and snow, and the surface runoff entered the mountain fissures, which was a factor inducing the abrupt slide and collapse of wedge-like rock mass at the elevation of 5000 m. The abrupt destabilized mountain dropped more than 3000 m under the action of the gravity, and owned huge kinetic energy. It impinged the loose deposit of more than 100 years old. This kind of ‘‘super dynamic consolidation’’ and highstrength extrusion effect forced the saturated clastic material in the Ditch to produce instantaneously exceptional moment more than pore water pressure, which resulted in liquefaction, and made instant loss of shear strength, which further induced high-speed landslide. This kind of high-speed landslide mechanism was called ‘‘dynamic consolidation liquefaction-inducing mechanism’’. 2.8

Xuanhan Landslide in Sichuan

Since september 3rd, 2004, heavy rainfall had fallen in many places of Xuanhan, Sichuan. The rainfalls in March, April and May were 15.9 mm, 122.6 mm and 257.0 mm separately. On May 15, Nanxiang Road, located on the riverside of Qian River—a branch of Qu River—in Tiantai Village, began to rupture. Subsequently roadside houses broke down and fell into the River. Afterwards, the front of the slope slowly slipped at all times, and deformation area gradually extended

from the front to the back; At that night, the main landslide mass of the slope foreside started to enter the Qain River; and the back of the landslide mass instantly slipped and the great landslide was formed. This landslide destroyed 1736 houses, and caused no homes of 1255 people. It led to traffic interruption, and damaged communication. In addition, a 23 m high dam was formed due to the front of the landslide mass slipping into the Qian River, and it blocked 1.2 km long channel, which gave rise to flow cutoff of the River for 20 hours. The dammed lake formed had the storage capacity of 6000 × 104 m3 , and the water level rose from 20 m to 23 m, which caused 2 towns to be submerged. More than 10 thousand persons had no houses to live in. The slope angle of the unstable slope ranged from 10◦ to 33◦ . The mountain top elevation was 1100 m. The front near the valley was a scarp of 30 to 40 m high. The altitude of Qian River valley was 356 m, and the valley width varied from 80 m to 100 m. Four natural ditches, Yujiahe, Maliushu, Liangshuijing and Dahe, developed from south to north (See Figure 11). The slope was composed of red siltsandy mudstone, silty siltstone, and gray fine-particle quartz-sandstone from Suining formation of Mid Jurassic, and its total attitude was 110∼120◦ /∠5∼10◦ . The rock formation was similar to the bank slope in the strike, belonging to bedding bank slope. The slide surface of Tiantai Landslide was a contact belt of mudstone and sandstone, and the main slip direction varied from 97◦ to 107◦ . Its slide bed was composed of black gray fine-particle quartz sandstone, hard, weakly weathered. The dip angle of the slide bed was from 8◦ to 10◦ . The main soil of slide belt was brown silty clay, and the thickness was commonly from 20 to 30 cm (See Figure 12). The back slide mass was mainly composed of cataclastic rock mass from the red mudstone, and blocky gravel including silty clay. The front of the slip mass was mainly from silty clay. The landslide showed a circle chair shape in plane, and the altitude of the after-edge ranged from 520 to 570 m, and its cliff height was from 10 to 30 m. The Landslide mass was 350 m to 1100 m long (from east to west), and its width was 1100 m to 1500 m (from south to north), and its thickness varied from 15 to 35 m. Its volume was about 2500 × 104 m3 . Sheared edge height of the front ranged from 380 to 424 m, which was 30 to 35 m higher than the one of the river bed. Serious deformation could be seen on the slope surface, and most houses had already collapsed. Tension cracks and fissures were distributed on the ground with the length of 30 to 70 m, and maximum crack depth was up to 2 m, and its width varied from 0.2 to 1.2 m (See Figure 12). Tiantai Landslide presented several characteristics as follows: 1) gentle dip angle of the slide surface; 2) Not too thick, 10 to 20 m on both sides of the landslide,

2045

Landslide border

Sheared edge

Motion vector

Gully

Gully position before slide

Qian River before slide

Qian River after slide Dahe Ditch

J2sn J2sn

Liangshuijing Ditch

Maliushu Ditch Yujiahe Ditch

J2sn

Qian River

200 m

Figure 11.

Engineering geological plan of Tiantai Landslide.

200 m Landform line before slide

550 500 450

J2sn(Mudstone)

Maliushu Ditch Maliushu before slide Ditch after slide

95O

550 500

Landform line after slide

J2sn (Sandstone)

450 Qian River

400

400

7O

350

350 200 Main slip belt and its slip direction

Figure 12.

400 600 Secondary slip surface and its slip direction

800 Shattered mudstone

1000 Silty clay

Landslide accumulation

Engineering geological section profile of Tiantai Village Landslide.

and 20 to 30 m in the middle; 3) Clear massive characteristic; 4) Long-term deformation process, however quick slide; 5) No obvious gradient changes before and after the slide; 6) Special movement style—each slide block presented difference in slip speed and movement direction; 7) Waveform pushing characteristic. High-strength rainfall was a factor resulting in the Tiantai Landslide. Much rainfall infiltrated the slope, which accelerated the formation of the landslide by uplifting, splitting and pushing. The landslide process included ‘‘groundwater converging-uplifting, pushing-splitting, collapse-slide’’ three phases. In the landslide region tectonic joints and weathered cracks of the original rock developed, which was available for infiltration of rainfall. The upper layer of the

original slope was water-tight mudstone, so groundwater mainly flew along the interface between sandstone and mudstone, or flew in the fissures. High-intensity rainfall induced that the groundwater converged could not be drained immediately, so confined water pressure was formed. Not too thick slope mass was floated by the high water head. ‘‘water cushion effect’’ make friction decrease rapidly. Meanwhile, because the groundwater pushed the cracks, high-pressure water flow was formed in the cracks, which split main structural fissures. When pore-water pressure increased up to some extent, the slope mass fleetly was disaggregated, and the confined water pressure was released. This drove the blocks of the slope to move downwards or disjointed them during the movement.

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S Q4col+dl

Q4del

II Q4del

Smx4 Q4 Q4del

Figure 13.

del

Smx4

I

Danba Landslide.

III

2.9

Danba Landslide in Sichuan

Danba County in Ganzi, Sichuan, was situated in a narrow valley, on the right side of Dajin River, altitude 1864 m, which was in the east of Qinghai Tibet Plateau (See Figure 13). The city zone area covered 2.5 km2 , and the population was 11 thousand. It was political, economical and cultural center of the whole county, and an important gate to Ganzi. In August 2002, after Danba County, a high and deep slope began to deform, with the side height of 200 m, and mean slope gradient of 32 degrees. In October 2004 the deformation became more serious. The deformation had been speeded up four times from January to March. The daily displacement increased from 6 mm to 8 mm on Feb 3rd, and the rate arrived at 17 to 33 mm/d. On March 14th the slope deformation was again accelerated, and local parts slide happened, so some houses were destroyed, which led to losses of 10.66 million Chinese RMB. Here, the accumulative deformation was from 70 to 80 cm, maximum 1 m. The slope was almost run through by the slide cracks. At that time, the Danba Landslide was ultimately formed, with the volume of 220 × 104 m3 . The landform of Danba, which was marked by strong uplift due to neotectonic movement and sharp downcutting of the Dajin River, provided favorable conditions for development of the landslide. In the region, mean annual rainfall was 605.7 mm. In the landslide area the bedrock was composed of garnet two-mica schist including biotite leptite from the Fourth Formation-complex (S4mx ) of the Maoxian Formation, Silurian System. The cap rock was formed by ancient coast colluvial (Qdel 4 ), rockfall fgl col+dl deposit (Q4 ) and glaciofluvial accumulation (Q4 ) (Figure 14). According to borehole data, there was no groundwater exposure in the landslide region due to quite thick overburden. According to corresponding data, it could be inferred that the front and middle parts of the

Smx4 Jianshe Street

Landslide border

Figure 14. landslide.

Tension crack

Zoning

Engineering geological sketch of Dangba

landslide was situated at the interface of colluvial stratum and the bedrock. The slide belt of the back part was located in the colluvial stratum. Dip angle of the landslide surface was 30◦ , almost consistent with the dip angle of the slope. The thickness of the landslide mass ranged from 20 to 35 m. On the surface of the landslide tension cracks were developed in different parts and on different scales, total length more than 1500 m. In terms of deformation characteristics the whole landslide could be divided into three regions (Region I, II and III in Figure 14). As a main part of the landslide, Region I was 200 m wide for the front-edge, 150 m wide for the backside, and 290 m long. The mean gradient of the slope was more than 30◦ . It covered 0.055 km2 , and the volume was 170 × 104 m3 , Main slip direction 353◦ . The front of the landslide was 6 to 28 m high, like steep ladder. Its whole gradient varied from 50◦ to 70◦ . Strong shearing of the slide mass led to damage of the houses and streets located at the slope foot. The back of the landslide showed higher topography in the west and lower topography in the east, but its overall landform was relative plat. Tension cracks and stagger benches developed there. Region II was located on the left side of the back of the landslide, and covered 6000 m2 . Main dip

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direction was 20◦ , and dip angle varied from 20◦ to 30◦ . The volume was 15×104 m3 and thickness is from 15 to 20 m. The deformation mainly included tension fissures in the back part and on the both sides edges, besides rockfall and collapse in front. The region was more unstable than Region I. Region III lay on the right of the landslide with the front 50 m wide, and the back 85 m wide. Its length was 180 m, which covered 0.014 km2 . The mean thickness of the slide mass was 25 m with the gradient of 30◦ to 45◦ , and the slide volume was 35 × 104 m3 . The landslide mainly consisted of big rock blocks. Slide bed was situated on the surface of bedrock. The left side of Region III was linked with Region I. Main deformation presented at the back of the landslide and both side edges, 5 m to 15 m long, 10 cm to 50 cm deep, with 5 cm to 40 cm staggered distance. Counter pressure loaded at the slope foot, prestressed anchor cable protection, were used in earlier protection of the Danba Landslide. There were 269 prestressed anchor cables, and prestressed force of everyone was 1300 kN with the length from 40 m to 52 m, and the anchor segment length was from 8 m to 10 m. The volume of cutting slope was 1800 m3 . These methods presented marked reinforcement effect. The Danba Landslide belonged to reactivated ancient landslide. Its reactivation had several reasons. Firstly, huge ancient landslide deposited on relatively stable highly steep slope composed of metamorphic rock. There were greatly different properties between two kinds of medium. There was a trend to slip for the accumulative formation along the surface of the bedrock. And human activities accelerated the revival process of ancient landslide. Since 1998 many buildings had been made along the slope. Excavation of the slope and cut slope had enlarged the free surface of the frontal part of the slope, as resulted in deformation of the slope. The deformation of the slope was extremely accelerated in 2003. From March to October 2004, a large-scale reconstruction of Jianshe Street further inlarged the free surface of the steep slope. This led to the supporting force of the slope foot to decrease, and caused the slope extension, creep and deformation toward the free surface. Therefore the slope structure was damaged, and the strength was constantly dropped. At last, shearing-creeping at the front of the ancient landslide were formed.

3

CONCLUSIONS

According to findings mentioned above, conclusions can be drawn as follows: 1. China is a nation where landslide hazards frequently happen, of which large and middle scales occupy an important position (See Figure 15). In

the southwest of China, landslides are famous due to large scale, complex mechanism, great harm, and huge difficulty in prevention and treatment, which are also typical and representative in the world. 2. The best essential reason for development of the large-scale landslide is complicated landform and relief characteristics. From east to west there are Qinghai-Tibet Plateau, Yun-Gui Plateau, mountain zone of the middle part, and eastern offing plain, in turn, which compose three levels of great terraces. A huge mainland gradient region is formed among them, and mainland landform in China sharply falls from west to east, especially in gradient zones among Qinghai-Tibet Plateau, Yun-Gui Plateau, and Sichuan basin, where several rivers originated, such as Jinsha River with its branches (Yalong River, Dadu River), and Lantsang. These Rivers deeply cut river valley, so that high mountains and gorges were generated, as established the morphological base of generation of large-scale landslides. Figure 15 indicates that most landslides in China happened around the first gradient zone and the second level of terrace of eastern Qinghai-Tibet Plateau. 3. China is the best active zone of plate tectonics activities in the world. Intensive collision between Indian Ocean Plate and Eurasian Plate, made continual uplifting of Qinghai-Tibet Plateau; Endogenic geology power of the crust transfers from west to east, even extending up to the middle and eastern parts, which is the most active in the west. As a result, high crustal stress, strong activities of tectogenesis and incidental strong shock process generate prominent internal power of this region. 4. Extreme climate conditions and global climate change are major factor inducing large-scale landslide occurrence (See Figure 16). Formation of Qinghai-Tibet Plateau and obstruction of Qinling Mountains create different climate zones from south to north. Southern weather system in China is controlled by warm and wet air flow from Indian Ocean. There are locally heavy rainfalls in summer, and rainfall intensity arrives at 200 to 300 mm/day. This kind of climate induces largesized landslide disasters, especially in Yunnan, Guizhou and Sichuan. However, in the northwest of China it is cold in winter, which is controlled by northwestern monsoon climate. Seasonal frozen crust is formed due to capillary water level rising in the broad loess region. In the coming spring the frozen layer is thawed, and large-scale loess landslides are induced. In high altitude region such as Qinghai-Tibet plateau, and Tian Mountain in Xinjiang, increasing temperature of global climate, makes snow line move upwards, and glacier retreat. These provide

2048

Haerbin Yulumuqi Changchun Shenyang Huhehaote

Second-order stair Yinchuan

Beijing

Taiyuan

Xining

First-order stair

Yellow River

Lanzhou

Xi’an Zhenzhou Huai River Nu River Lhas a River Yaluzhangbu

Jinsha River Chengdu Chongqing Yibin

The boundary of landform element

Shanghai Wuhan Yangtse River Changsha Nanchang

Guiyang

Third-order stair

Kunming

Large-scale landslide point The boundary of Yellow River, Yangtse River, Huai River valley

Taiwan

Guangzhou Lantsang Yuan River Pearl River Naning

200 km Hainan

Figure 15.

The map of large-scale geological hazards of China since 20th century.

Earthquake

Figure 16. China.

Rainfall

Thaw

Creep

Artificial intervention

Triggering factors of large-scale landslides in

direct conditions for occurrence of large-sized landslides. According to statistic data since 1980s, 70 percent of large-scale catastrophic landslides are related with extreme climate conditions or climate change, and 50 percent of them are caused by direct rainfalls (See Figure 16).

5. Extensive humanity activities have been main factors inducing large-sized catastrophic landslides in mainland of China since 1980s. With development of humanity activities, occurrence frequency of the landslides in mainland of China has a trend to ascend. The data from Table 1 show that 50 percent of the large-sized landslides are directly or indirectly related with human activities. For example, since 1990s when Development of the West Regions policy was being carried out, in these areas, large-scale engineering activities have happened most frequently, so these areas have become their centre. In fact, landslide disasters also happened frequently in those areas.

ACKNOWLEDGEMENT Great thanks are due to Prof. Z.M. Xu for his helping arrange some parts of representative landslide cases. The writer is grateful to Prof. Q. Xu for his offering some of landslide cases.

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Table 1. Geological calamity events of larger collapse and landslide in China since 1900s. Date (dd-mm-yyyy)

Volume (104 m3 )

Slope type

Factors

Loss

Haiyuan, Ningxia

16-12-1920

–

Loess slope

Marineoriginated earthquake

Diexi earthquake landslide

Maoxian, Sichuan

25-08-1933

21,000

Trias metamorphic rock

Diexi earthquake

Chana landslide

Gonghe, Qinghai

07-02-1943

25,000

Thaw

Lugong landslide

Gonglu, Yunnan

22-11-1965

39,000

Lacustrine Stratum of Tertiary Basalt of Permian

Tanggudong landslide

Yalong River, Sichuan

08-06-1967

6,800

Weathered slate of Trias

Side-corrosion and total creep of the slope

Yanchihe rock-fall

Yichang, Hubei

03-06-1980

150

Near level layered slope

Underground mining

Jipazi landslide

Yunyang, 18-07-1982 Chongqing

1,500

Ancient landslide

Rainstorm

Sanleshan landslide

Dongxian, Gansu

07-03-1983

3,100

Creep, frost thawing

Xietan landslide

Zigui, 12-06-1985 Chongqing

3,000

Zhongyangcun Wuxi, 10-01-1988 landslide Chongqing Tiexi Xide, 02-09-1988 landslide Sichuan Xikou Huaying, 10-07-1989 landslide Sichuan

765

Loess coverburden, mudstone of Tertiary Ancient landslide and colluvial deposit Limestone

675 Large-scale landslides induced, forty barrier lakes formed, many villages destroyed, and 0.1 million people death Towns and villages destroyed, death of 6800 persons, barrier lake formed, and later dam failure leading to death of 8000 persons Zana Village destroyed, death of 114 persons 5 villages buried, and death of 444 persons Jamming Yalong River for 9 days, causing flood peak of 57000 m3 /s after dam failure Destroying mining area, and death of 284 persons Intermission of Yangtse River sea-route for 7 days, and economic loss of 100 million Yuan Death of 237 persons

Case title

Location

Group earthquake landslide

Touzhai landslide

Zhaotong, Yunnan

23-09-1991

4 100

900

Accumulative formation Intensely Weathered carbonate rock Intensely Weathered basalt

Total creep

Rainfall

Migrating instantly

Rainstorm

Death of 33 persons

Rainstorm

Overthrowing train

Rainstorm

Death of 221 persons

Total creep

Death of 216 persons

(continued)

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Table 1. (continued) Date (dd-mm-yyyy)

Case title

Location

Jiguanlin rock-fall

Wulong, 30-04-1994 Chongqing

Huangci landslide Erdaogou landslide

Volume (104 m3 )

Slope type

Factors

Loss

424

Limestone, Steep obsequent slope

Underground mining, rainfall

Lanzhou, 30-01-1995 Gansu Badong, 10-06-1995 Chongqing

200

Loess slope

A 10 m high water level fall was formed; Wave height varied from 1 to 5 m; Flow intermitted for half an hour; 5 boats sank; direct economic loss was 100 million Yuan Migration of 1000 persons Death of 5 persons

Oldjinshan landslide Yankou landslide

Yunyang, Yunnan Yinjiang, Guizhou

01-06-1996

500

18-07-1996

1,500

Shachonglu landslide Yigong landslide

Guiyang, Guizhou Bomi, Tibett

04-12-1996

2

09-04-2000

28,000

Shuangliu landslide Yingjiang landslide

Shibao, Gulin Ying River, Yunnan Lanping, Yunnan Three Gorges Reservior region Tiantai, Xuanhan, Sichuan

06-06-2000 14-08-2000

Danba, Sichuan

Lanping landslide Qianjiangping landslide Tiantai landslide

Danba landslide

60

Intensely Weathered slope Broken slope

Mining

Oblique bedding slope of limestone

Quarrying at the foot of slope

Layered slope Base rock, broken body

Cutting the slope foot Snow dissolving

2

Loose bodies

Rainstorm

Barrier lake formed, reservoir region submerged Death of 10 persons

0.2

Mixed granite residual soils

Rainstorm

Death of 13 persons 5000 persons removed Death of 14 persons, loss of 57350 thousand Yuan

Flood

03-09-2000

2,000

Dumping slope

Rainstorm

13-07-2003

2,400

Reservior impounded

05-09-2004

2,500

Bedding landslide of sandstone and siltstone Gentle bedding slope of sandstone and siltstone

01-03-2005

220

Accumulative formation landslide

Total reep and disturbances from mankind

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Rainstorm

Death of more than 200 persons Blocking river dam was 65 m high; A 8-km-long barrier lake was formed Death of 35 persons

Migration of 1255 persons; the dam deriving from landslide was 23 high. A 20 km long barrier lake was formed. Buildings destroyed, loss of 10.66 million Yuan was caused, the whole county safety was threatened

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for International Development. U.S. Geological Survey Open-File Report 01-0119, Version 1.0, 22 p., 2001. Bhasin R, Grimstad E, Larsen JO et al. (2002) Landslide hazards and mitigation measures at Gangtok, Sikkim Himalaya. Engineering Geology, 64(4): 351–368. Broadbent CD, Ko KC (1971) Rheology aspects of rock slope failure. Proc. of 13th Symp. on Rock Mechanics, Illionis, pp. 537–572. Brückl EP. (2001) Cause-Effect Models of Large Landslides. Natural Hazards, 23(2–3): 291–314. Cai ZX (1988) On characteristics of soil and water loss and related policies of control in upper reaches of Changjiang River. Discovery of Nature, (4): 97–104 (Chinese). Cai ZX (1989) The study on slide disaster and its countermeasure. Journal of Catastrophology, (1): 74–77 (Chinese). Chen YB, Wang CH & Pan XY (2003) Landside feature and the genesis of Qianjiangping in Hubei Province. Journal of Mountain Research, 21(5): 633–634 (Chinese). Chen ZS, Kong JM (1991) A catastrophic landslide of Sept. 23, 1991 at Touzhaigou of Zhaotong, Yunnan Province. Journal of Mountain Research, 9(4): 265–268 (Chinese). Collison A, Wade S, Griffiths J et al. (2000) Modelling the impact of predicted climate change on landslide frequency and magnitude in SE England. Engineering Geology, 55(3): 205–218. Crosta GB (2001) Failure and flow development of a complex slide: the 1993 Sesa landslide. Engineering Geology, 59 (1–2): 173–199. Duan YH (2000) Present state, trend and countermeasure of geological hazards in Chinese West. Economic research reference, (58): 12–18 (Chinese). Huang RQ, Zhao SJ & Song XB (2005) The formation and mechanism analysis of Tiantai landslide, Xuanhan County, Sichuan Province. Hydrogeology and Engineering Geology, 32(1): 13–15 (Chinese). Huang ZZ, Tang RC & Liu SL (2002) Re-discussion of the Seismogenic Structure of the Diexi Large Earthquake in 1933 and the Arc Tectonics on Jiaochang, Sichuan Province. Earthquake Research in China, 18(2): 183–192 (Chinese). Jiang CS (2000) Present state and prevention of China’s geological disasters. Chinese Geology, (4): 3–5 (Chinese). Jin DS (1998) Laojinshan Landslide in Yuanyang, Yunnan Province. The Chinese Journal of Geological Hazard and Control, 9(4): 98–101,80 (Chinese). Leng L & Leng RH (2002) Flood in Yalong River and its historical lesson. Sichuan Water Conservation, (2): 42–44 (Chinese). Li YC (2002) On-the-spot record of Diexi Earthquake. Literary History World, (6): 39–42 (Chinese). Li N (1992) Landfall-landslide blocking river disasters and its prevention measures in Yunnan Province [C]. Commission of Proceedings of Landslides ed. Proceedings of Landslides (No. 9). Beijing China Railway Publishing House, 50–55 (in Chinese). Li TB, Chen MD & Wang LS (1999) Real-time following prediction of the landslides. Chengdu: Chengdu University of Technology Press. Liao QL, Li X, Li SD et al. (2005) Occurrence, geology and geomorphy characteristics and origin of qianjiangping landslide in three gorges reservoir area and study on ancient landslide criterion, Chinese Journal of Rock Mechanics and Engineering, 24(17): 3146–3153.

Lin DM, Pi LY, Huang HL et al. (2002) Study on engineeringgeological conditions for landslide at Badu k343 section Nankun railway and control on it. Journal of Engineering Geology, 10(2): 220–224 (Chinese). Lin PS, Lin JY & Hung JC et al. (2002) Assessing debris-flow hazard in a watershed in Taiwan. Engineering Geology, 66(3–4): 295–313. Mauritsch HJ, Seiberl W, Arndt R et al. (2000) Geophysical investigations of large landslides in the Carnic Region of southern Austria. Engineering Geology, 56(3–4): 373–388. Parise M & Wasowski J (1999) Landslide Activity Maps for Landslide Hazard Evaluation: Three Case Studies from Southern Italy. Natural Hazards, 20(2–3): 159–183. Radbruch-Hall DH, Colton RB, Davies WE et al. (1983) Landslide Overview Map of the Conterminous United States. United States Geological Survey Professional Paper 1183. Raetzo H, Lateltin O, Bollinger D et al. (2002) Hazard assessment in Switzerland—Codes of Practice for mass movements. Bulletin of Engineering Geology and the Environment, 61: 263–268. Schuster RL (1996) The 25 most catastrophic landslides of the 20th century, in Chacon, Irigaray and Fernandez (eds.), Landslides, Proc. Of the 8th International Conf. & Field Trip on Landslides, Granada, Spain, 27–28 Sept. Rotterdam: Balkema. Schuster RL & Lynn MH (2001) Socioeconomic Impacts of Landslides in the Western Hemisphere. U.S. Geological Survey Open-file Report, 01-9276. Staub I.B (2001) A methodology for the mapping and analysis of ‘‘debris-flow initiation’’ hazard—application to the Bragousse torrent (France). Bulletin of Engineering Geology and the Environment, 59(4): 319–327. Sun DY (2000) Project regulation of Badu Landslide in NanKun Railway. Beijing: Chinese railway press. Sun T & Chen Y (2001) Application of finite element technique to analyzing the slope stability—evaluation the stability of the Jiaochang talus slide in Tianlonghu Hydropower Station. Sichuan Water Power, 20(1): 28–31 (Chinese). The U.S. Geological Survey. (2000) Landslide hazards. USGS Fact Sheet Fs-071-00. Varnes DJ & Savage WZ (1996) The Slumgullion Earth flow: A Large-Scale Natural Laboratory, U.S. Geological Survey Bulletin 2130. Washington: united states government printing office, 1996 (available at: http://pubs.usgs.gov/bul/b2130/). Voight B & Faust C (1992) Frictional heat and strength loss in some rapid landslides: error correction and affirmation of mechanism for the Vaiont landslide. Geotechnique, 42: 641–643. Wang SJ (1999) Tasks and future of engineering geology. Journal of Engineering Geology, 7(3): 195–199 (Chinese). Wang YH & Yang RH (2005) The activity characteristics and movement style of Qianjiangping Landslide in the Three Gorges Reservoir region. The Chinese Journal of Geological Hazard and Control, 16(3): 5–11 (Chinese). Wu C, Ran HX, Zhen YH, et al. (1996) Hydrograph of the dam-break flood of the reservoir formed by mountain collapse in Ya Longjiang, Journal of Hydrodynamics (A), 11(6): 646–652 (Chinese).

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Wu WJ & Wang SY (1989) The mechanism of Saleshan Landslide. National landslide conference—landslides. Chengdu: Sichuan Science and Technology Press. pp. 184–189. Xu BD, Pan HT (1992) The effect of coal mining on Kengkou and Hancheng Power Plants slope[C]// Commission of Proceedings of Landslides ed. Proceedings of Landslides (No. 9). Beijing China Railway Publishing House, 1–9 (in Chinese). Yamagishi H (2000) Recent Landslides in Western Hokkaido, Japan. Pure and Applied Geophysics, 157(6–8): 1115–1134. Yin SL, Han ZS, Li ZZ (2000) Progress of landslide researches in the world. Hydrogeology and Engineering Geology, (5): 1–4 (Chinese). Yin YP (2000) The research on characteristics of rapid huge landslide in Yigong River in the Bomi, Tibet and disaster relief. Hydrogeology and Engineering Geology, (4): 8–11 (Chinese).

Yin YP (2001) A review and vision of geological hazards in China. Management Geological Science and Technology, 18(3): 26–29 (Chinese). Zhang ZY, Wang ST, Wang LS et al. (1994) Analytical theory of engineering geology. Beijing: Geology Press. Zhang ZY (2000) The present status, technical advance and development trends of landslide remedial measures. Journal of Geological Hazards and Environment Preservation, 11(2): 89–97 (Chinese). Zhang ZY, Liu HC (2001) A special case history of the environmental impact by undergound mining—the mechanism and control measures of the ground upheaval deformation of hancheng power plant. Earth Science Frontiers, 8(2): 285–295 (Chinese). Zhong LX (1999) Case study on significant geohazards in China. The Chinese Journal of Geological Hazard and Control, 10(3): 1–10 (Chinese).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Early warning for Geo-Hazards based on the weather condition in China C.Z. Liu, Y.H. Liu, M.S. Wen, C. Tang, T.F. Li & J.F. Lian Institute for Geo-Environment Monitoring, Beijing, China

ABSTRACT: According to national early warning practice for geo-hazards from 2003 to 2005, it is systematically concluded that the basic characteristics of geo-hazards, early warning method and forecast result based on the geological maps of China in a scale 1:6,000,000. With the contrast of different characters between sustained rainfall and typhoon rainfall inducing geo-hazards, the disaster reduction result and some problems are preliminarily analyzed. Some basic recognition is that early warning to geo-hazards is feasible, national scale forecast is only to call attention, but can’t immediately be used to disaster reduction decision-making. And, the future direction is to build a united disaster reduction framework of early warning system including national, provincial and county levels based on weather factors in different scale of area.

1

INTRODUCTION

In order to raise the disaster prevention and reduction consciousness of the public as soon as possible, to reduce effectively geo-hazards inducing by weather factors, to boost disaster reduction work of all levels of government, China Ministry of Land and Resource and China meteorological administration are united to carry on early warning work for geo-hazards of national scale in the mainland of China from 2003 at every flood season (5–9 month). At the same time, two departments direct provinces and counties to carry on this project gradually. The basis in law of this project is national rule for mitigation of geological hazard (Open-file, 2003). 1. Forecasting zone is continent region of People’s Republic of China (beside Hong Kong, Macao special administrative areas and Taiwan province at present). 2. Forecasting season is from May 1 to September 30 every year. It begins from June 1, the first year, 2003. 3. Forecasting time is from 20:00 on that same day to 20:00 the next day. 4. Forecasting classes include 5 grades: First grade, possibility is little; Second grade, it is less; Third grade, it is middle; Fourth grade, it is big; Fifth grade, it is biggest. 5. Forecasting result achieving fourth grade is released in CCTV-1 News at 19:30 every evening, and that achieving third grade is released in China Web-net of geo-environment information (http://www.cigem.gov.cn).

National early warning project for geo-hazards has been carried on three years since 2003, every province (beside Shanghai municipality) also starts similar event gradually. So, it has been of scientific basis to analyze its result, and this analysis can provide support to develop the second generation of early warning system. 2

BASIC CHARACTERISTICS OF PAROXYSMAL GEO-HAZARDS

Geo-hazard forecasting object is mainly regional or drainage paroxysmal geo-hazards such as rockfall, landslide and soil-rock debris flow etc., induced by terrific weather conditions such as rainfall, frost or melting etc. ‘‘Chain type’’ slope geo-hazards, with distinct characters, is the emphasis and difficulty in geo-hazard prevention.

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1. Regional distribution: It often occurs in the range of hundreds to thousands square kilometer. 2. Multitudinous: A group of hazards occurred in a region. 3. ‘‘Chain type’’ action: Rockfalls, landslides, soilrock debris flows occur and transform successively in one place, which has larger strength destructive force. 4. Break out simultaneously: Falling-landslide-soilrock debris flow occurs in several ten minutes to several hours successively or simultaneity. It has the characters of break out simultaneously, the whole slope collapse suddenly or flush. 5. Break out in the condition of super intense sustained rainfall: With regional super intense

6.

7. 8.

9.

10.

3

rainfall, larger flushing strength induces incompact accumulation materials to slide. Steep terrain, and upper soft but lower hard in geologic structure: Collapsing slope often has some characters as following: 25◦ –70◦ slope angle and 100–400 meters drop in level; double layer of structure in geology, which is incompact accumulation material in upper layer, but hard base rock in lower layer; incompact accumulative layer is thin, 0.5–5 m in general. Special vegetation form: Some shallow root tree such as bamboo, fir, etc., will aggravate destroy of slope surface. Occurred in lower development region: In these regions, house, road and cutting slope are lack of maintenance; residents are lack of geo-environment protect consciousness. Occurred lately: Large landslide often take place after rain, even several days after rain, such as China Xintan landslide (June 12, 1985), Philippines oversize landslide (February 17, 2006). Casualty and loss for whole region is larger caused by landslides, but in which a single hazard is less in scale.

Figure 1. Critical criterion drawing curve of early warning based on rainfall in a regional geo-hazards. Zone A—no forecast zone; Zone B—forecasting zone; Zone C—warning zone; Line α—forecast critical line (dividing line 2 to 3 grade); Line β—warning critical line (dividing line 4 to 5 grade).

line α and line β is forecast area (3 or 4 grade). And zone C above line β is geo-hazard warning area (5th grade). 3.2 Geo-hazard early warning practice from 2003 to 2005

NATIONAL EARLY WARNING OF GEO-HAZARD BASED ON WEATHER FACTOR

3.1 Critical rainfall criterion method (Liu Chuanzheng, 2004) According to statistic analysis of historic geo-hazards and 15-days rainfall before hazard, geo-hazard forecasting criterion mode chart based on weather factor to be established, and critical rainfall criterion chart in typical regions geo-hazards is preliminary gotten. Then, some forecasting region adjusting criterion chart is built with existing data. Then, with national rainfall data and image from National Weather Center everyday, we can forecast in half an hour if there will be geo-hazard and its spatial area and its harm. According to the research of relationship between geo-hazards and rainfall, scatter plot of geohazards and different period critical rainfall can be gotten. With the plot, we can find: scatter point is distributed in zone, its upper limit is line β, and its lower limit is line α. Therefore, using processing rainfall such as 1-days, 2-days, 4-days, 7-days, 10-days, and 15-days, etc., inducing geo-hazards, early warning criterion mode charm can be built (Fig.1). With Fig.1, lateral axis is the number of raining day, longitudinal axis is its corresponding processing accumulate rainfall. Line α and line β are critical rainfall (it is curve in practical application), Zone A below line α is no forecast area (1 or 2 grade). Zone B between

With geo-environment conditions and weather factors inducing geo-hazards, the mainland of China was divided into 7 parent zones and 28 forecast zones utilizing geological maps of China in a scale 1:6,000,000 in 2003, and divided into 74 forecast zones in 2004 and 2005. Considering geo-environment characters and weather factors of every zone, its corresponding critical rainfall criterion is built respectively. Zone partition uses the method from high to low, which is acceptable for little scale geo-hazard spatial partition. Test run shows that geo-hazards early warning based on weather factor is feasible in science and technology, which can contribute to initiative disaster prevention. With two years practice, using rainfall criterion mode method, geo-hazard forecast result is better. So, we continue to use rainfall criterion mode method in 2005. 3.2.1 Geo-hazard early warning practice from 2003 to 2004 (CIGM, 2003 & 2004) With undercount, there are 264 paroxysmal geohazards in the whole country in flood period (6–9 month) of 2003, in which there are 101 (878 point at least) geo-hazards in the early warning range. Forecasting success ratio is about 38 percent. In the period of 6–9 month 2003, 56 geo-hazard forecasting information are released in CCTV-1, and 109 geo-hazard forecasting information are released in China geoenvironment information Web-net. Practice prove that the mode and technical method of national geo-hazard

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early warning based on weather factor in 2003 is availability and feasible, and its result is better. In 2004, the whole country was divided into 74 early warning zones, and the forecasting criterion for every zone is built. In the period of flood period (5–9 month), 2004, 83 geo-hazard forecasting information are released in CCTV-1, and 107 geo-hazard forecasting information are released in China geo-environment information web. According to collective feedback information, there are 382 geo-hazards in the whole country, in which there are 163 geo-hazards in the forecasting range (Fig. 2).

Frequency (%)

3.2.2 Geo-hazard early warning practice and comparison month by month in 2005 (CIGM, 2005) With undercount, there are 1399 paroxysmal geohazards induced by rainfall, in which there are 706 geo-hazards in forecasting range. With geo-hazards forecasting and occurring in fact, the success ratio of august is lower, July and September are relative higher, May and June are same as the whole year by and large (Fig.3). Typhoon rainfall is the main factor inducing geohazards. Geo-hazards induced by typhoon ‘‘Haitang’’ (July) are in Fujian, Zhejiang, Hubei, Sichuan, Chongqing province, etc. Typhoon ‘‘Maisha’’ (August) affects Zhejiang, Jiangsu, Shandong, Liaoning

2003

70 60

50

46.9

50 40 30

2004 46.9 48.1 44.8 42.7 38.3

50

37.6

31.6

26.5

20 10 0

(month)

Figure 2. Geo-hazard early warning success comparison between 2003 and 2004 in China. Number of geo-hazard

Number of successful forecast

Number 1400 1200 1000 800 600 400 200 0

5

6

7

8

9

sum

Figure 3. Geo-hazard early warning number success contrast from May to September in 2005 in China.

province, etc. Intense tropical storm ‘‘Shanhu’’ affects Fujian, Jiangxi, Hunan and Hubei province, etc. And, typhoon ‘‘Taili’’, ‘‘Kanu’’, ‘‘Dawei’’ (September) induces many geo-hazards in Fujian, Zhejiang and Hainan province, etc. 3.3 Basic experience and example Analysis found that the geo-hazards induced by sustained rainfall and typhoon rainfall behave as different characters: 1. With intense rainfall (such as rainfall bigger than 70 mm), geo-hazards has the character of synchronous happened. 2. With less intense but sustained rainfall, geohazards behave as some hysteretic effects, i.e., inducing geo-hazard need some rainfall accumulation. 3. After geo-hazard happened on a large scale, even there is sustained rainfall, it still need some rainfall accumulation to induce geo-hazard again. 3.3.1 Geo-hazard inducing by sustained rainfall Sustained rainfall induces cluster geo-hazards in Guangdong and Fujian province, at June 18–23, 2005. Compare geo-hazards distribution with that very day rainfall, it is well consistent in the north of Fujian, but not consistent in the west of Fujian and in the north of Guangdong. Compare geo-hazards distribution with 3-days accumulative rainfall. There are many geo-hazards in the north of Fujian, where 3-days accumulative rainfall is over 200 mm, and it is the same instance in the middle-east region of Guangdong. That is, there are many geo-hazards in rainfall center of 3-days rainfall. Then, we compare geo-hazards distribution with 5-days rainfall (Fig. 4), more than 1000 geo-hazards in the north of Fujian province are all in the area which 5-days rainfall over 500 mm. And it is so in Guangdong and in the east of Guangxi province where geo-hazards distribution is not consistent with rainfall in one day or 3-days. That is, geo-hazards and 5-days rainfall center is consistent. 3.3.2 Geo-hazard inducing by typhoon rainfall July 18, 2005, typhoon ‘‘Haitang’’ landed on Yilan, Taiwan, and landed again on Huangqi, Fujian province, at 17:00, July 19. While landing, maxim wind near typhoon center is 12th grade (wind speed is 33 meter per sec.), center air pressure is 975 hundred Pascal. Typhoon ‘‘Haitang’’ have the characters as following: larger range, larger wind, intense rainfall, complex path, and gravity influence, etc. A great deal of rainfall is induced by typhoon in Fujian and Zhejiang province. Maxim processing rainfall (671 mm at Zherong) and maxim day rainfall (472 mm at Zherong) are all break through historical record since 1949.

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rainfall, geo-hazard is mainly affected by prophase 5-days rainfall. But, with typhoon weather, geo-hazard is mainly affected by rainfall in that day. The relationship of geo-hazards and rainfall accumulative days is controlled by rainfall process, that is, is relative with rainfall and rainfall intensity.

4

DISASTER MITIGATION RESULTS AND PROBLEMS

4.1

Figure 4. Correlation between rainfall for five days and geohazards distribution in southeast of China.

Figure 5. Contrast among rainfall forecast, early warning of geo-hazards and its actual distribution in southeast of China.

Many geo-hazards are induced by rainfall, which is more influence on Zhejiang and Fujian province. The contrast of Geo-hazard forecast, rainfall forecast and geo-hazard occurring during typhoon is showed as Fig. 5. There are 101 geo-hazards in the range of 5th grade early warning zone, and less geohazard in the range of third grade or 4th grade, which indicate geo-hazard forecast is relative right. When geo-hazards happened, day rainfall is 50–100 mm, and it is 100 mm in Fujian, but 50–100 mm in Zhejiang province. It is almost consistent that the distribution of 3-days rainfall or 5-days rainfall and day rainfall. It is clear that typhoon rainfall inducing geo-hazard is different from sustained rainfall. With sustained

Disaster mitigation results macroscopic analysis

According to statistic, there is 11654 person death or missing for paroxysmal geo-hazards, such as rockfall, landslide and soil-rock debris flow, etc., in 11 years from 1995 to 2005. And there is 1059 persons dieing or missing on average per year (Fig. 6). The general trend is obvious, that is, the number of dieing or missing for geo-hazards is falling trend year by year, since ‘‘resident self-understanding and selfmonitoring’’ is carried on in 2001 and regional geohazard forecast is practiced in 2003, although the range and strength of human activity is still developing. At the same time, after general rainfall and flood disaster of south and north in China, 1998, geologic material store is relative decrease, which may be the reason of less geo-hazard in 1999. Now, there is yet no authentic economic loss statistic data only by reason of geo-hazards. With the statistic analysis of 715 geo-hazard success forecasting and keeping away instances, 2004, in China, success keeping away hazard can be divided into three classes, that is, resident decision by themselves, resident self-understanding and selfmonitoring, critical rainfall forecast, which account for 3.5%, 86.7% and 9.8% respectively (Fig. 7) (Liu Chuan-Zheng, 2006). Obviously, resident selfunderstanding and self-monitoring is an effective disaster mitigation method at present and a long period in future of China, which should be extend vigorously and advance its work system in geo-hazards prone or susceptible area.

number 2000 1800 1600 1400 1200 1000 800 600 400 200 0

1573 1288

1270

1160

1080 864

1995

1996

1997

1998

1999

2000

1049

2001

962

2002

868

858

2003

2004

682

2005

year

Figure 6. Contrast of death or missing people made by geohazards from 1995 to 2005 in China.

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frequency (%)

level, emergency response from province level and united behavior in county level with resident self-understanding and self-monitoring in lot of villages.

86.7

90.0 80.0 70.0 60.0 50.0 40.0 30.0

5

9.8

20.0 3.5

BASIC CONCLUSION AND FUTURE

10.0 0.0

resident decision by themselves

resident selfunderstanding and self-monitoring

5.1

critical rainfall forecast

Figure 7. Ratio balance about success keeping away from geo-hazards of China in 2004.

The class of resident self-understanding and selfmonitoring keeping away geo-hazards successfully is as following: self-understanding and self-monitoring system is established at known geo-hazards hidden danger point, where simple monitoring or patrol is carried on. When geo-hazards sign is found, geohazards prevention scheme in advance is started in time and kept away from geo-hazards successfully. Alternatively, when resident found some microscopic signs such as local slide or crack, etc., they report to autochthonous government department in charge, who patrol and decide those possible hazards and organize resident leaving from danger through some means such as broadcast, telephone, cell note or yell, etc. Then geo-hazards loss can be avoided successfully. 4.2 Discussing problems 1. The early warning method of geo-hazards based on the critical rainfall at present can not sufficiently reflect the change of geo-environment conditions and key factors. 2. The study of geo-hazard early warning accuracy evaluation in time, space and strength is yet not carried on. 3. The mathematic analysis and quantizing expression of geo-hazard early warning grade is not enough. 4. Investigation and analysis to typical geo-environment condition and inducing factors, in geohazards prone or susceptible area, is not enough, which is disadvantage to sum up geo-hazards development trend. 5. Forecasting system sustained upgrade idea is not definite or its consciousness is not intensity. With infirm lateral communication, and insufficient longitude guide, forecasting operation and science study are yet lack of united science theory method and work platform at present, which results in the shortage of consult language together. 6. Serve orientation of geo-hazards early warning is not clear, such as warning action from nation

Basic conclusion

1. Geo-hazards can be pre-warned and forecasted, although it is controlled by many factors. This early warning is more geo-hazards risk forecast than assured geo-hazards affair forecast. 2. National even provincial scale geo-hazard early warning is only to call attention, and can not be used to disaster reduction decision-making. 3. Different scale forecasting systems should cooperate with each other, and upgrade in time. And then, geo-hazard forecast could reflect the change of geo-environment conditions and meet disaster prevention of different scales. 4. Disaster mitigation is a systematic engineering that involves sciences, governments, social publics and enterprises, especially emergency response and self-understanding and monitoring from the local residents.

5.2

Thought for tomorrow

In view of regional or drainage geo-hazards such as rockfall, landslide, soil-rock debris flow, induced by the coupling action among some terrific natural conditions such as rainfall, freeze thawing, and earthquake, etc., and irrational exploitation of geo-environment, we should make efforts to study regional geo-hazard classified early warning system, implement classified early warning scheme, and provide technical support to different classified government manage and resident self-understanding and self-monitoring. According to analysis on different classified geohazard early warning technical method, we should integrate sufficiently the survey, monitoring and study results of geo-environment and geo-hazards at present, advance the theoretic depth and technologic maturity of regional geo-hazard early warning. Then, forecasting accuracy such as spatial limit, time range and outbreak intensity can be obviously advanced. And, disaster mitigation emergency reaction will be more quickly and more pertinent. For example, geo-hazard forecasting spatial accuracy of national scale will be advanced to 1:2,500,000–1:1,000,000, and provincial scale accuracy will be 1:500,000–1:100,000, and county scale accuracy will be 1:10,000–1:50,000. Geo-hazard forecasting time accuracy will be consistent with weather forecasting. Geo-hazard forecasting

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intensity accuracy will approach to its inducing factors. At the same time, on the part of information system and share application, geo-hazard information platform including collection, explanation, transmission, filing, processing, modeling, analysis, sharing and multimedia issue, etc., will be found to meet different need. REFERENCES China institute for geo-environment monitoring 2003. National geo-hazards early warning technologic summary report, 2003.11. China institute for geo-environment monitoring 2004. National geo-hazards early warning technologic summary report, 2004.12.

China institute for geo-environment monitoring 2005. National geo-hazards early warning technologic summary report, 2005.12. Liu, C.Z., Wen, M.S. & Tang, C. 2004. Meteorological early warning of geo-hazards in China based on raining forecast. Geological Bulletin of China, 2004, 23 (4): 303–309. Liu, C.Z., Zhang, M.X. & Meng, H. 2006. Study on the geohazards mitigation system by resident self-understanding and self-monitoring, Journal of Disaster Prevention and Mitigation Engineering, 2006, 26 (2): 175–179. Open-file 394 of the State Council of P.R. China 2003. National rule for mitigation of geological hazards, 2003.11.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Slope engineering in railway and highway construction in China Gongxian Wang & Huimin Ma Northwest Research Institute Co., Ltd. of China Railway Engineering Corporation, Lanzhou, China

Maorun Feng Ministry of Communication, Beijing, China

Yuan Wang Highway Research Institute of Ministry of Communication, Beijing, China

ABSTRACT: Slope engineering is one of the most important tasks in railways and highways construction in mountain areas in China in which slope failure are frequent due to complex topographical and geological conditions. The paper briefly reviewed the development history of slope engineering in railway and highway construction in China, and the achievements in research of slope problems, such as slope deformation and failure mechanisms, stability analysis and slope evaluation methods, slope deformation monitoring and forecast, as well as slope control techniques. The slope problems need to be study in the future were also discussed. With the rapid development in highway and railway construction in China, slope problem will be still a big issue. Therefore, technique development and scientific research in slope engineering are still a special and important task for professionals.

1

HISTORY OF TECHNIQUES DEVELOPMENT AND SCIENTIFIC RESEARCH IN I SLOPE ENGINEERING

Slope stabilization has been one of most major problems encountered in railway and highway construction in mountain areas of China since 1950’s, with the rapidly development of the national economy. For design and construction of stable slopes in railway and highway construction, many techniques have been developed and a great deal of scientific research has been carried out. The history of technique development and scientific research in slope engineering in China can be divided into three periods. 1.1

1950’s, the period of suffering from frequent slope failures

At early stage of the People’s Republic of China, occurrence of slope failure was quite frequent in railway construction because of lack of experience and techniques for slope stabilization, resulting in many disasters. Slope problems in some sections along some railway have been one of major safety problems for the railway operation for nearly 30 years. For example, Baoji to Chengdu Railway along which 2136 natural and cut slopes have been under deformation, among which there were 76 large-scale landslides,

337 rockfalls, 34 unstable rockmass [1] . Many railway stations of this railway were located on ancient landslides. These ancient landslides were reactivated due to subgrade excavation. These slope problems resulted in an extension of the period for the railway construction and an increase in investment. Only for the section of 247 km long from Baoji to Guangyuan, the investment in for slope control was about four hundred and seven millions Yuan (RMB), about 1.6 times as many as the cost planned[1] . Baoji to Tianshui section of Longhai Railway, about 150 km long, was built during the period between 1939 and 1945. Owing to limited experience at that time and in order to decrease the investment, the railway was constructed by cutting slopes, building high embankment, short tunnel and small bridge and culvert. Some part of the railway was located on a large fault zone. Rockfall and landslide disasters were very serious along the section. As a result, these disasters often destroyed subgrade, bridges and tunnels, and badly influenced safety operation of the railway. This railway was thus named as ‘‘blind gut’’ in China due to its severe slope problems. Since 1950’s 259 unstable slopes have controlleds, and 100 open tunnels have had to be extended to prevent rock fall and fall stone in exit of tunnel. K1358 loess landslide, occurred in 1963 and 40×104 m3 in volume, stopped traffic for 169 hours; K1357 loess landslide, occurred in 1981 and 60×104 m3 in volume, destroyed

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an open tunnel 75 m long and stopped traffic for 314 hours; Putaoyuan 1# landslide, occurred on Sept. 1985, pushed away the railway’s subgrade and rail for several ten meters and stopped traffic for 66 hours. The above mentioned two sections had to change the line. The investment for controlling various disasters along the railway was about five hundred and seven million Yuan (RMB) and it was 6.8 times as many as the cost planned[2] . The Yingtan to Xiamen railway constructed in the 1050’s, 693 km long, there were various disasters from 3000 sites among which 100 were very serious. In 1962 the railway stopped traffic for 3 months in total because of avalanches and collapses landslides and flood. The investment for controlling various disasters was about four hundred and six million Yuan (RMB) and it was 1.15 times as many as the cost planned[2] . In same period slope disasters was very serious in highway construction. For example, (1) Sichuan to Tibet highway (northern line), constructed in 1950’s, the total length was bout 2155 km. Sichuan to Tibet highway southern line was constructed during the period of 1954 to 1969, and there is 1286 km within the boulders of Tibet. The highway goes through mountains. Owing to complex natural factors, disasters often occurred, there were 389 landslide, avalanche, debris flow, sand slip, snow fall and snow berry, the traffic was often stopped by these disasters. (2) Qinghai to Tibet highway, constructed by the P.L.A and local persons, about 1160 km in total length, it was the first highway that paved black grade surface in high cold region in the world and it was also one of the highest highways above the sea level. Slope thaw slumping and snow fall and berry disaster along the highway are serious. (3) Xinjiang to Tibet highway, started from Yecheng of Xinjiang in the north to Lsha in the south, goes over Jieshandaban pass at 5406 m above sea level and Kuda’enbu pass at 5432 m above sea level and other 11pass, it is often suffered from snow storm, snow fall and snow berry disasters. Pic. 1. The reasons for more disasters in this period are: (1) lack of recognition of complex of geological condition; (2) limited stabilization measures on many high slopes; (3) deterioration of rock slopes’ stability by some large exploration construction. 1.2

1960’s—1970s’: the period of learning from experiences and lessons, and paying attention to scientific research

At the end of 1950’s and in the early 1960’s, it is realized that experiences and lessons in railway and highway construction in mountainous area must be learned, it was also recognized that it is very important to strengthen geological work and training of qualified geologits for survey and design projects.; In 1959 Collapse scientific and Technological research Institute

Pic. 1.

The montane- highway.

of M.O.R was founded in Xi’an (previous Northwest Research Institute Co., Ltd. of China Railway Engineering Corporation), a special research institute in slope disasters mechanism and engineering control. In 1060’s the M.O.R issued a (Regulations for Railway Survey and Design), framing a series of technical policies, such as geological work should take part in line selection, tunnel should be longer rather than shorter, slope height more than 30 m should be compared with tunnel plan and embankment height more than 20 m should be compared with bridge plan. This effectively decreased high slopes deformation and ancient landslide reactivation. For example, Chengdou to Kunming Railway in selecting line avoided landslide more then 100 sites; Baoji to Lanzhou railway in second selecting line once went over Wei River for 55 times to avoid worse geology section and large landslide section. It ensured railway traffic safety. In this period the state has constructed several key highways in the western. (1) China to Nepal highway, constructed by China and Nepal and started from Yangbajin of Dangxiong County in the north to the end at Kathmandu in Nepal, was about 850 km in total length among which 736 km was in China. There were 6 rock landslides, 26 soil landslides, 60 various types of large, medium and small collapse and avalanches along this section. (2) Yunnan to Tibet highway, started from Xiaguan of Dali in the south

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to Mangkang of Tibet in the north and linked to Sichun-Tibet highway, was about 800 km in total length. Landslide, avalanche, debris flows, sand-slip and snow fall disasters often occurred along this highway. (3) Tiansan highway, constructed from 1974 to 1983 and went through Xinjiang, it started from Dushanzi in north to Kuche in the south and was about 562.74 km in total length. The highest section along road was 3800 m above sea level and it often suffered from snow disasters. In this period, many research projects were carried out, involving in high slope stability, slope formation conditions and failure causes, failure and movements mechanisms of landslides as well as new techniques of slope control measures, such as variation law of residual shear strength of slip zones in cohesive soils, design and theory of reinforced concrete anti-slide pile, horizontal and vertical drainage hole group and etc. Results of these research played an important role in controlling slope disasters. 1.3

Since 1980’s: the period of disaster prevention

With development of the national economy and largescale infrastructure construction, unstability of slopes along expressway are still a big problem in expressway construction in mountainous area, owing to implementation of high standard and width road surface of expressway although the length of bridges and tunnels in some section have taken up 30∼40% of the total length. For examples, within 300 km along the section in Guangdong Province along Beijin-Zhuhai Expressway there are 200 slopes with height of more than 30 m. Investment for controlling unstable slope and landslides increased to ten hundred million (RMB). Yuanjiang to Mohei expressway in Yunnan Province is about 147 km in total length, owing to its special landforms and geology condition, there are 337 slopes with height of more than 30 m, among which 66 slope are higher more than 100 m, more than 130 slope occurred deformation or landslides, increased investment for controlling unstable slope and landslides was six hundred million. Tonghuang Expressway in Shanxi Province is about 93.8 km in total length, there are many high slopes along 119 sections in which the highest slope is about 88 m. 32 landslides have to be controlled, 11 landslides were controlled by anti-slide piles, the cost for controlling landslides was about 20% of the total engineering cost; In railway engineering slope an ancient landslide was reactivated in Badu railway station along Nanning to Kunming Railway, controlling cost of the landslide was about ten hundred million (RMB). Unstable slopes and landslides are important factors leading to increase fin investment, extension of construction period and safety operation. Pic. 2. Unstable high slopes and landslides have drawn attention of various leaders and technical staff. How to

Pic. 2.

The landslide in Badu.

effectively prevent unstable high slopes and landslides has become an important research subject. Institutes in survey, design, construction, scientific research and universities have carried out many research projects related to railway and highway construction. As far as the special geological condition is concerned, the state have invested twenty hundred million (RMB) to support the research of key techniques in communication construction in the western area and have obtained obvious achievements. For example remote sensing and GIS system are widely used and principle of ‘‘geological select line’’ was carried out, therefore serious geological worse section was avoided; for potential unstable slopes, comprehensive survey techniques and engineering geological mechanics investigation analysis methods were used and fussy evaluation, mutation theory and neural network and the other methods were used to evaluate and predict slope stability. In order to prevent unstable slope after excavation ‘‘prestabilization’’ measure was used; pre-stressed anchor wire anti-slide pile measure was also used to control deformed slope and landslide. This has saved 30% compared with the common cantilever piles. Pre-stressed anchor framework, earth beam, anchor block, anchor rod, soil nail for surface prevention and whole stabilization technique have been widely used in high slope stabilization.; Mini-pile, churning pile and various grouting technique have also been widely used.; Geo-synthetics have been gradually accepted and widely used in slope surface protection and high backfill reinforced and stabilized engineering.; Surface and subsurface drainage have been a common measure in slope engineering. Consequently, slope engineering has gone up to a new stage. 2

RESEARCH ACHIEVEMENTS RELATED WITH SLOPE PROBLEMS

2.1 Classification of types of slope deformation Types of slopes in terms of deformation forms and slope materials are classified into six types and ten

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Table 1. Classification of slope deformation. Materials Types of slope deformation

Rock

Soil

Avalanche Landslide Slump Fault fall Topple Complex type

Various rock avalanche Bedding landslide, insequent landslide Weathered fracture rock mass collapse Weathered fracture rock and fault fall Topple in medium thick rock layer combination of the above mentioned tow

Loess avalanche, semi-rocky avalanche Various soil landslide Various soil collapse / / Combination of the above mentioned tow

Table 2. Slope deformation according to its deformation depth. Depth of slope deformation H

Types of slope deformation fault fall mass

H6 m

Wash, slope failure, desquamation and fall stone Landslide, slope failure, avalanche within slope range Avalanche, landslide, fault fall, collapse and topple over slope

structural plane

soft layer

Figure 1.

subtypes: avalanche, landslide, fault fall, collapse, topple and compound type, as shown in Table 1 Types of slope deformation can also be divided into four types according to depth of slope deformation, which is related to difficulty of slope control, as shown in Table 2. Slopes in weathered fracture rock and structural fracture zones subjected to squeeze underlying soft layer often result in cracks of the upper slope part, often leading to slope deformation in the form of vertical sinking with little horizontal deformation. This phenomenon is named as ‘‘fault fall’’. It is much different from general landslides that slide along shearing zones and horizontal displacement is main deformation. Its diagram is shown in Figure 1. 2.2 Slope mass structures and slope failure modes Landslide and high slope deformation have been studied by Northwest Research Institute Co., Ltd. of China Railway Engineering Corporation. It was found that slope mass structures have controlling function for slope failure types, position, dimensions and failure modes. Slope mass structure can be divided into V types and 12 quasi-types which are shown in Table 3. Academy of Sciences of Ministry of Communication

Diagram of deformation of fault fall.

caries out special research on rock slope engineering design and its disasters control along Tonghuang highway in Shanxi Province. Based on field test of rock structural surface and analysis and research of rock structural network, prevention and stabilization design of rock slope have been established on the tow technical indexes of fracture degree and general integrity of rock structural surface, thus rationality and reliability of design and control could be ensured. 2.3 Research on landslide mechanism Based on research of slope mass structure, landslide characterized by main slide zone along geological soft and slide surface with broken line can be divided into main slide section, drag section and anti slide section as shown in figure 2 and than characteristics of slide zone in each section subjected to force are analyzed. Development procedure of landslide can be divided into 5 stages, such as creep, squeeze, slide, abrupt slide and relative stability and than characteristics of force and deformation in each stage are analyzed. It is carried out research on variation law from peak strength to residual strength of slide zone soil. It is summed up 5 slide mechanisms such as variation of pore pressure,

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Table 3. Slope structure and slope failure modes. Slope structure

Sub-types

Basic feature

I Homogeneous slope structure

Homogeneous cohesive soil, loess-like soil, accumulated soil and filled soil

Homogeneous rock and soil mass without harmful structural surface on the up part of erosion surface or excavated surface.

II Main bottom type of slope structure

II1 Up part soft and down part hard slope structure (binary structure)

The up part is soil layer or soft rock layer underlain hard rock and rock layer is nearly horizontal or reverse dip.

II2 Up part hard and down part soft slope structure

The up part is sick or very sick hard rock layer underlain soft rock and rock layer is nearly horizontal.

III1 Dip layer-like slope structure

Alternated, interbedding and interstratified bed of hard sandstone and soft mudstone or other rocks, layer dips into the air, dip angle of layer is about 10◦ ∼30◦ .

III2 Reverse-dip layer-like slope structure

Alternated, interbedding and interstratified bed of hard sandstone and soft mudstone or other rocks, rock layer is dip into slope, dip angle of layer is about 10◦ ∼30◦ .

III Layer-like slope structure

III3 Steep dip layer-like slope structure

Profile diagram

Alternated, interbedding and interstratified bed of hard sandstone and soft mudstone or other rocks, rock layer is dip into slope and the air, dip angle of layer is more than 40◦ .

Failure modes Rotational slide and collapse along an arc plane

crack

natural slope line

crack

schist

natural slope line

loess

crack sand and mudstone

reducing load platform

limestone

natural slope line

mudstone limestone mudstone alternated with sandstone

crack

natural slope line sand-mudstone interbedding

crack

natural slope line sand-mudstone interbedding

The up layer or soft rock layer slide or collapse along underlain hard rock Hard rock avalanche, fault fall, soft rock squeeze out

Multi-layer and grade bedding rock landslide

Fault fall, insequent, landslide or avalanche

sandstone mudstone sandstone

crack

natural slope line

natural slope line

Topple, toppled avalanche, toppled landslide, V type joint avalanche or bedding slide avalanche.

(Continued)

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Table 3. (Continued) Slope structure

IV Loose fracture rock slope structure

V Mass slope structure

Sub-types

Basic feature

III4 Incline cross layer-like slope structure

Alternated, interbedding and interstratified bed of hard sandstone and soft mudstone or other rocks, angle between rock layer strike and slope surface strike is more than 30◦ .

IV1 Fault fracture zone slope structure

Profile diagram

Rock mass fractured shape of stone-soil. Secondary structural plane (zone) is rich.

IV2 Large-scale talus, diluvium of gully and ancient landslide accumulation

Fractured stone drift and ancient landslide accumulation

V1 Quasi-layer-like slope structure

Granite, basalt and mass-like rock, link up and layer-like joint surface and leaning suspension

V2 Eyeball-like slope structure

It is formed eyeball-like structure in rock subjected to tectonic action

Failure modes Sliding combined with layer plane and joint plane, V type joint avalanche

fault(joint surface)

layer surface

line strike

loess

secondary fault schist river

Slide and collapse along an arc plane or structural plane

granite-gneiss

crack zone

Avalanche, collapse slide along contacting plane or ancient sliding surface

natural slope line

crack

natural slope line

granite-gneiss

Landslide or avalanche along quasi-layer

gneiss

original ground line # #

existing slope line

strong-medium weathered granite

# #

Fault fall, avalanche, slide

# # # # #

# #

weak weathered granite

#

F 4 fault

V3 General integrity slope structure

General integrity rock, small fault or joint surface dips into the air.

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fall stone

unstable rock

V-type joint avalanche, slide along small fault

Figure 2. Diagram of slide mode of three section type and its stress field.

change of slope stress state, variation of progressive failure and residual strength, squeeze out of soft rock plastic flow and vibration liquefaction. Characteristics of landslide static mechanics, kinematics and landslide transformation are studied. These studies provide a theoretical basis for effectively controlling landslide. 2.4

Investigation and analysis of rock mass structural planes of rock slopes

Geomechanics investigation analysis method of rock mass structural planes of rock slopes is the basic method for determining slope structure. The method from theory expounds the original relationship and distribution law which controls slope deformation failure property and main controlling structural plane of space state. Firstly starting from geomorphology investigation, which preliminarily determines period, order and tectonic framework of the natural slopes subjected to geological structures during their development, the method consequently investigates and recognizes mechanics property of each structural plane, development degree and gonging through degree existing in real rock mass and then carries on systematic analysis to determine distribution law of various structural plane in engineering site and tectonic framework of slope. Finally based on distribution of hard rock and soft rock in slope, slope surface strike and underground water distribution law, slope structure can be precisely determined. Investigation and analysis of geomorphology and structural plane is an essential of this method. It is carried on structural plane investigation around and in the slope and measure group quantity and property of structural plane and it should be pay attention to texture

trace on tow sides of each group structural plane. It is carried on systematic analysis according to structural plane system of rock mass subjected to force and than it is determined how many times in action of texture stress of rock mass in high slope once is subjected to order of texture stress and auxiliary structure plane of each time stress. For controlling rock slope disasters deformation range, dimensions and types of slopes should be determined and the effects caused by engineering action should be forecasted. Then evaluation of slope stability should be given so that it can provide reliable basis for stabilization engineering design. A large and complex rock slope has a complex space structure, which is related with several geological units or several grades, so its failure mode is also complex. Therefore it should be carried on ground investigation and comprehensive survey including remote sensing, physical exploration, boring, excavation exploration, testing and monitoring measures to identify its grade and slice situation. Physical exploration is one of the important measures for slope survey. For two element structure slopes and fractured rock slopes, a concept of ‘‘tendency sliding zone’’ is put foreword after research, it is very useful to guide boring and analyze and determine sliding zone. A new boring technology which drills without water and reverse circulation without pump has been adopted in landslide boring producing core sampling rate of about 85%. This method can be effective for determining potential sliding zone. In order to precisely recognition and control landslides, monitoring of deformation dynamic of landslides is needed. Ground displacement dynamic monitoring includes simple monitoring, monitoring of monitoring network, GPS monitoring, underground water monitoring and monitoring of building subjected to force. Based on analysis of monitoring information it is helpful to determine range, sliding direction, slice and grade of landslide and can provide basis for landslide forecast. Boring hole inclinometer has been widely used in monitoring subsurface displacement of landslides. This method can precisely determine location of landslides’ slip zones. Figure 3 is a case. It can effectively guide construction and has been widely used in slope engineering along railway and expressway (see figure 3). Survey of rock slopes should carry out based on the principle of ‘‘dynamic design and information construction’’ and geological work should participate in construction procedure. Thus a set of scientific and systematic work method has been developed for control of high slopes and landslide. This method includes investigation, analysis, evaluation, engineering geological classification, slope structural classification, failure mode prediction, key disaster survey,

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Figure 3. Diagram of monitoring profile at K28+905 along Changin Expressway.

Figure 4. Evaluation method of landslide and slope stability.

engineering plan optimization, scientific construction, dynamic monitoring and geological whole course tracing. 2.5 Evaluation methods of high slope and landslide stabilities Stability evaluation and variation prediction are a basis for determining slope stabilization and landslide control plane. To evaluate landslide and slope stability Mr. Xu Bangdong puted foreword eight basis methods as shown in Fig. 4 [5][6]. It concludes qualitative evaluation and quantitative evaluation and the former is a main method for evaluating landslide and slope stability. For different types and occurring mechanisms of landslides, 3∼4 of the eights methods are taken to complement and verify each other and result correspondent to actual situation is obtained. Limit equilibrium analysis method is most widely used and

finite element strength reduction method and other numerical analysis methods are also used. Zheng Yingren and Chen Zhuli, academician of the China Academy of Sciences, carried out deeply research on slope and landslide stability calculation, their research achievements have been applied in slope engineering along railway and highway. 2.6

Calculation of landslide thrust force

Landslide thrust force is a basis of anti-slide retaining structure design. In railway and highway system, besides quasi-homogeneous soil slope is calculated using the methods for circular slip surface, for diagonal line slip surface, principal axial section is mainly calculated by transfer parameter method. The formula is: En = KWn Sin αn − Wn Cos αn tgφn − Cn ln + En−1 n

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where, En —landslide thrust force in the profile (kN/m); K—safety parameter adopted in design considered for uncertainty. If a landslide is small in dimension and parameters of its materials are known the small value is adopted and if a landslide is large and dangerous and parameters of its materials are not clear the large value is adopted; For contemporary engineering small value is adopted and for permanent engineering large value is adopted. Small value is adopted 1.05∼1.15 and large value is adopted 1.20∼1.25 or more large. Wk —Weight of slice slide mass (kN/m); αn —Inclined angle of slice slip surface (◦ ); Cn , φn —Cohesion of slide zone soil (kPa) and internal friction angle (◦ ); ln —Length of slice slip surface (m); En−1 —residual slide force transferred from the up slice to calculated slice (kN/m); n —transferring parameter, n = cos(αn−1 − αn ) − sin(αn−1 − αn ) · tgφn . C, φ their values are selected based on results of lab tests, back-calculation and in-situ test, and also considering a slope’s different development period, geometry of slip zone, material and water content state of landslide peak strength, residual strength or a numerical between the tow are separately adopted, it does not use same index in normal circumstance. Northwest Research Institute Co., Ltd. of China Railway Engineering Corporation, Nanjing Hydraulic Research Institute and Hydraulic Research Institute of Songhuajiang and Liaohe Hydraulic Committee introduced a concept of residual strength in earlier time and carried out research on test methods of ring shear apparatus, reversing direct shear apparatus, direct shear apparatus and multi-shear[5] . 2.7 Forecast theory and method of landslide Time forecast for abrupt slide of a landslide is a research subject which draws great attentions of professionals both in China and overseas. The forecast based on creep theory of slide zone soil, created by Seto in Japan, was a precedent. In July of 1994, Northwest Research Institute Co., Ltd. of China Railway Engineering Corporation carried out multi-monitoring measures on Huangci landslide that was 600 × 104 m5 in volumes. On Jan. 27 1995, a critical slide forecast was put foreword ahead of 3 days of its occurrence using deformation power theory. The landslide occurred abruptly at 2:30 on Jan. 30 1995, but no one was injured because of safely evacuating inhabitants after warning issued. The predict time differed 22 hours and 30 minutes from the actual slide time. Methods of multi-measures, systematic monitoring,

progressive analysis, progressive reach, comprehensive decision and total forecast push landslide forecast technique to an advanced world level. 3

DEVELOPMENT OF NEW TECHNIQUES FOR LANDSLIDE AND SLOPE CONTROL

3.1 Principle of landslide and slope control 3.1.1 Principles of landslide control 1. Correctly recognizing landslides 2. Paying great attention in landslide prevention 3. Thoroughly controlling landslides 4. Carrying out landslide design and engineering stage by stage 5. Taking comprehensive engineering measures 6. Implementing engineering measures in landslide deformation’s early stage 7. Considering technical and economic feasibilities and in landslide control 8. Formulating a proper framework in construction of landslide controlling engineering 9. Applying the concept of dynamic design and information construction in landslide control 10. Paying attention to maintenance of anti-slide engineering. The principle of geological select line is a prerequisite for preventing landslide while selecting lines. When railway and highway lines are selected, worse geological section should be avoided (large landslide and landslide, avalanche continuously distributed section, rock layer bedding section thick layer accumulation section and large fault fracture zone). The First Highway Survey & Design Institute of China Communication has carried out systematic research on ancient landslides triggered by various constructions, engineering landslide, unstable high slope and slope deformation occurred along the state highways of No.108, 109, 208, 209, 210, 318 and others and some provincial highways and has put foreword a working principle for preventing landslide disasters after highway complete. The principle is that in the early stage of highway construction it should strength systematic geological work along highway, pay attention to geological select line, strengthen investigation and survey of landslide and potential landslide along lines, and insist on taking prevent as a main measure and integrating prevent with control and controlling thoroughly in once. The principle takes an important part in preventing and mitigating disasters in engineering slope along highways. 3.1.2 Design principle of high slopes 1. Paying attention to analysis of geological conditions. A high slope refers to change a part of geological mass into artificial engineering and

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2.

3.

4.

5.

6.

its stability depends on geological structure of slope mass and artificial reformed degree. Detail geological information is a basis for slope design. Reducing slope total height. If slope excavated in central part is more than 30 m in height it should be considered to compare with an option of tunnel. If fill in embankment central part is more than 20 m in height it should be considered to compare with an option of bridge. It locally adjusts plane location and longitudinal slope rate of line and reduces height of excavation and fill. Principle of integrating slope shape and slope rate with engineering stabilization. Slope below 30 m in height should be designed stable slope shape and slope rate which is suitable for geotechnical strength; if slope is more than 30 m in height, gentle slope will destroy vegetation covered on the slope and increase requisition land and abandon soil and it is disadvantageous for environmental protection. Steep slope rate connected with retaining engineering is adopted to reduce total height of slope. Principle of strengthening toe and middle part of slope. Owing to stress and underground water are concentrated on toe of slope, retaining engineering and drainage should be adopted. In the middle part of slope retaining structure is also used to ensure total and local stability. Principle of strengthening drainage in slope area: it is carried out improving ground and underground drainage system to reduce the effect of water on slope stability. Principle of environmental protection and ecological beautify.

3.1.3 Techniques of pre-stabilization for high slopes Excavation of high slopes will inexorably lead to stress relaxation and deformation of slope mass. ‘‘Pre-stabilization’’ is to control deformation within a certain range and do not make slope occurring failure. Based on geological research, new theory, new calculation method and large-scale model test, and with deeply understanding about deformation and unstable mechanism of slope research on relationship between slope excavation and relaxation deformation gradually tend to quantify. Thus a design concept of ‘‘control deformation’’ for high slope disaster control engineering is put foreword, that is, it should be considered that it may occur deformation type after slope excavated while engineering is designed. Based on a potential deformation type proper retaining engineering measures are used and a certain construction method is adopted to control deformation relaxation of slope in advance so that the slope can be controlled by lowest cost, most optimum design and most optimum construction method.

Techniques of pre-stabilization includes the following methods, i.e. excavation by stages and stabilization by stages method, retain by stages, execute pile first and than excavate afterwards if it is needed to adopt anti-slide pile at toe of slope and pre-stabilization method by anti-slide pile at toe of slope in order to reducing excavation. Technique of pre-stabilization is a new and reasonable design concept in high slope disaster control engineering field. Its specific type adopted is related with specific condition in specific disaster site, therefore the design of pre-stabilization is based on analysis of disaster occurring reason in specific disaster site. 3.1.4 Effect of pre-stabilization techniques on controlling excavated relaxation areas Slope is excavated by layer from the up to the bottom and stabilized while excavated, it is adopted retaining preventing structure to stabilize slope after slope excavated and than the next layer is excavated. The construction method can effectively control relaxation area and deformation caused by excavation. Figure 5 is a construction site composed of three stratum i.e. week weathered grit (W2), strong weathered sand stone and mudstone (W3) and total weathered sand stone with mud stone (W4). Rock in this construction site is characterized by development joint, fractured rock and low strength. After plan comparison it is adopted nail wall to retain and prevent slope. Height of slope is 20 m and is designed as two grades slope, height of each grade is 8 m and slope ratio is 1:0.25. Width of platform between tow grade slope masses is 2 m, thickness of shotcrete is 15 cm, length of anchor rod is 8 m, space is 1 m and dip angle is 14◦ . It is excavated by 10 layers and height of each layer is 2 m. Value of each test point is tested after each layer is excavated and then it is carried out excavation on next layer after procedure of placing wire net, shotcrete and instating anchor rod are finished. When construction method that is excavated without shotcrete prevention is adopted, relaxation area is ABC after excavated, slope surface displacement

Figure 5. area.

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Deformation curve of slope surface in relaxation

is cd and fc. it can be seen that the displacement is too much in central part of slope. If construction method that is excavated with shotcrete prevention is adopted, relaxation area reduces a little after excavated and becomes A, B, C. Slope surface displacement is reduced and becomes ab and ec. Therefore, the construction method that is adopted shotcrete prevention while excavated can reduce disturbed slope excavated and increases stability after slope excavated. The effect of this construction method is obviously. 3.2

Construction methods of pre-stabilization

In view of characteristics of slope engineering, besides ensuring structural stability it is considered how to construct on the structure in design of stabilization (retaining structure) engineering structure. In design of stabilization (retaining structure) engineering structure type of stabilization (retaining structure) engineering structure is adopted according to engineering geological property and construction condition. Different construction technology and stabilization method are used in view of different geological condition and height of slope. At the present the effective pre-stabilization construction method and technology that are often adopted are as the following: 1. Stabilization by grades and pre-stabilization at toe by anchor pile; 2. Excavation by grades and anchor by grades; 3. Excavation by grades Stabilization by grades and pre-stabilization at toe. By adopting the above mentioned construction technology and method in slope engineering, it can meet the needs of mechanized construction, raise construction work efficiency, reduce effective range subjected to excavation, control large deformation companied with large excavation and

a-total buried pile

e-h-like pile

ensure safety during construction for cut slope and long term stability of slope. 3.3 Research and application of new 3.3.1 Stabilization measures As far as the landslide control engineering is concerned it can be divided into two categories according to its action. One is drainage engineering to improve water contain state in slide mass and slide zone and the other is reducing weight and retaining engineering to recovery equilibrium of slide mass. 3.3.2 Underground water drainage engineering Since the 1050’s, retaining trench drain, intercepor basin, drainage tunnel vertical drainage boring hole group have been used to drain underground water. At the end of the 1970’s horizontal hole drainage technique started to be used. It has been proved in practice that these cut off and drainage measures that are suitable for different conditions have played an important role in landslide control and slope stabilization. 3.3.3 Retaining engineering 1. Large section gravity anti-slide retaining wall, excavated hole reinforced concrete anti-slide pile used in the 1950’s and pre-stressed anchor wire antislide pile used in the 1950’s of 20 century can save 30% cost compared with the common cantilever pile. Figure 6 is structural types anti-slide pile used at present. 2. In the 1990’s century engineering measures of prestressed anchor wire ground beam and framework for stabilizing slide mass were introduced. Owing to the advantages of simple structure, mechanized

b-cantilever pile

f-frame pile

c-buried pile

g-rigid frame pile

Figure 6. Basic forms of anti-slide pile.

2071

d-abutment pile

h-anchor wire pile

construction, safety construction, small disturbing for slope, rapidly stabilizing slope and reducing slope deformation, these measures are widely used in railway and highway construction in resent years. 3. In the resent years, relying on the principle of ‘‘strengthening middle part of slope and stabilizing toe of slope’’ and concept of ‘‘environmental protection road’’, a optimum comprehensive technique which is a combination of retaining stabilization, drainage and dewatering and greening slope is put foreword. ‘‘Strengthening middle part of slope and stabilizing toe of slope’’ is an important principle in slope design. By practice it gradually finds a set of combination structure type i.e. ‘‘pile-anchor’’ structure, ‘‘wall-anchor’’ structure, ‘‘pile-pile’’ structure and ‘‘reducing weight-anchor (pile)’’ structure which are suitable for stabilizing slope during construction and operation. These combination structures are suitable for controlling engineering of complex and fracture rock high slope. 4. Application of combination structure of mini-pile. The structure of micro-pile is simple and feasible for construction. It is suitable for emergency engineering and can freely form many types according to geographical and geological condition. In recent years it is successfully used in slope stabilization and landslide control. But engineering application is ahead of its theoretical study, in China mini-pile is designed based on experience or only shearing action on shaft is considered. Design and calculation theory of micro-pile is not maturity and needs to further study and improvement. Relying on disaster control engineering of landslide group in Qiandiaoqiao-Longdanxi section along Sichuan-Tibet Highway, ‘‘Research on Landslide Mechanism and Control Technique of Landslide in Qiandiaoqiao-Longdanxi Section along Sichuan-Tibet Highway’’, which was a research project in the western authorized by Communication Ministry, has been conducted. In this section topography is danger and rainfall and snowfall is rich. Owing to the effect of multi-period structure, stratum is fracture and landslides distribute in groups. In 1997 it caught rainstorm in this area, ground surface water seepage downward. Three landslides occurred along the section of 3 km long. Through fully proof it wais decided to control landslides along the highway by stages based on detail investigation and monitoring of 4 years on landslide. 1 landslide was the largest and most serious among the 3 landslides (highway surface settlement is about 1.6 m, effective range is about 350 m and volume is about 400 × 104 m3 ). Rock of landslide is fracture and the landslide is characterized by multi-grades, multi-slices and multi-layers. Key point of research is to carry out research on formation mechanism of

landslide and its control technique. Thus, a large quantity of testing elements and devices were installed in engineering structures and occurring mechanism of landslide and corresponding control techniques were deeply studied by deep displacement monitoring, laboratory model test, finite element numerical calculation and geological log of pile pits. The research taked an important role in control engineering design, construction and engineering safety and accumulates experiences and data for controlling similar large and complex landslide along highway in China. In principal part of the engineering prestressed anchor wire framework, common anti-slide pile and prestressed anti-slide pile are adopted. Length of anchor wire executed in practice is about 73 m and length of anti-slide pile is about 67 m, control effect is well. Neither complex degree of slide nor engineering construction difficulty is ever seen in highway engineering in China. In addition, ‘‘Research on Retaining Structure in Mountainous Area’’ and the other 13 items related to landslide and slope control techniques, which are scientific research items in the western authorized by Communication Ministry, will be completed one after another or completed soon. These scientific research achievements will lay foundation for prevention, stabilization and stability evaluation of slide-susceptible slope mass along highway.

4

SLOPE PREVENTION MEASURES FOR DISASTERS MITIGATION

Disasters caused by slide-susceptible slopes are not only related to slope type, geological topography and artificial engineering but also related to weather and natural factors. Most disasters caused by slidesusceptible slopes are resulted from many factors. Normally prevention of natural slopes is based on the understanding of slide-susceptible slopes, many recognized slide-susceptible slopes have been forecasted for minimizing the loss of disaster before slope failures occurred. Although understanding level on the effects of human activities and natural factors on slope stability has be constantly enhanced, but such understanding is far from enough. Thus, the effective method of reducing disaster loss caused by engineering slopes is to carry out monitoring and forecast on the unstable slopes. With constantly development of express railway and expressway construction, unstable slopes in a small range will result in large disaster accidents. These post great risks for construction and safety operation, disaster indemnity may be a heavy burden for the owner. ‘‘Research on Techniques of Monitoring and Forecast of Geological Disasters along Highway in the Western Area of China’’, a research project authorized

2072

by Communication Ministry, has carried out systematic research on slide-susceptible slopes. Six hundred thousand Yuan (RMB) were invested on real time monitoring on slide-susceptible slopes for several years and satisfied research achievements have been obtained.

5

SLOPE ENGINEERING COMBINED WITH ENVIRONMENTAL PROTECTION

Environmental protection is an important issue in expressway construction. Considering with geological conditions and unstable failure types of slopes several combined engineering structures are generally used in slope engineering, such as retaining structure, drainage and slope green, combination of anti-slide pile and prestressed anchor wire and earth beam (framework), anchor rod framework and prestressed anchor wire framework and etc. Tree and grass can be plant among earth beam (framework) to built green slopes. This not only controls high slope but also has an effect of natural scenery. Figure 7 is a diagram showing a slope stabilization using combined engineering structures. Xi’an to Hanzhong expressway, about 258.65 km in total length, was

the first expressway designed according to design concept of ecological protection and environmental protection, is. The expressway goes through the Qinling mountains in which slopes are high and valleys are deep. The construction was very difficulty in the area with complex topographical condition, more bridges and tunnels and high requirement of engineering techniques were needed. Total investment of the construction was 13.45 billion Yuan (RMB) and construction began in 2002 and planed to complete for operation in 2006. The expressway starts a prelude of new concept for constructing expressway with green scenery and environmental protection. In addition, in view of ecological and environmental protection of expressway many research projects authorized by Communication Ministry were carried out, such as ‘‘Research on Technique of Ecological Engineering in highway Area’’, ‘‘Research on Protection and Environmental Protection of Rocky Slope along highway’’ and etc. Results of these research projects have played an important role in constructing expressway with green and environmental ecological protection.

6

anchor wire

(a)

Slope engineering is not only a field in engineering practice, but also a scientific field involving many academic subjects. Combination of theoretical research and engineering practice is hence an effective way to solve engineering problems of complex high rock slope. In engineering practice in the future it should pay attention to analysis of geological processing mechanism, combination of both geology and engineering, and combination of both qualitative and quantitative evaluation. Specifically, these problems need to be paid more attention: (1) relaxation of excavated high slope with different slope structures; (2) reasonable engineering structures for high slope stabilization; (3) quantitative evaluation of underground drainage engineering in slope stability; (4) light-type construction, mechanization and rapidly construction in slope stabilization and etc.

anti-slide pile horizontal drainage hole (b)

anchor rod

(c)

PROBLEMS NEED TO BE SOLVED IN THE FUTURE IN SLOPE ENGINEERING

(d)

anchor wire

REFERENCES (e)

(f)

Figure 7. Diagram of stabilized types for slope (a) stabilization with prestressed anchor wire; (b) stabilization with pile and anchor wire; (c) stabilization with two role piles; (d) stabilization with anchor wire and rod by layer; (e) stabilization with reducing load.

Zhang Xianggang, Li Guanghui, Geological choice of BaojiChengdu Railway line and its main engineering geological problems, Journal of Railway Engineering Society, 2005.12 Beijing. Shi Wenhui, Railway geological disaster in China and its control, Journal of Railway Engineering Society, 2005.12 Beijing.

2073

Wang Gongxian, Xu Junling, Liu Guangdai, Li Chuanzhu, Landlidology and Landslide Control Technique, China Railway Publishing House, 2004, Beijing. Ma Huimin, Wang Gongxian, Zhou Depei, Control cases of high slope disasters along mountainous road, the People’s Communication Press, 2006, Beijing.

Northwest Research Institute of China Academy of Sciences of Communication Ministry, Landslide and its Control, China Railway Publishing House, 1977. Xu Bangdong, Landslide Analysis and its Control, China Railway Publishing House, 2001.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Mining slope engineering in China Sijing Wang Institute of Geology and Geophysics of Chinese Academy of Sciences, Beijing, China

Qian Gao University of Science and Technology, Beijing, China

Shiguo Sun North China University of Technology, Beijing, China

ABSTRACT: Slope stability becomes one of the key issues in opencast mining, particularly for slopes with increasing gradient and height as a result of deep excavation. This paper is to provide a critical review of mining slope engineering and design principles, including challenges confronted and countermeasures taken. Some suggestions are discussed in the study on meta-system of complex mining slopes for safe and economic operation and environmental protection. 1

INTRODUCTION

Slope, as one of the most important components of opencast mines, may directly impact on safety of production and economic benefits of mines. Open-pit mine slopes are characterized by their frequent change of slope configuration, limited service period, nonselectivity of site, influence on surrounding buildings and so on. The design and construction of mining slope should take all these factors into consideration. And this is the particularity of mining slope engineering to be different from civil slope engineering of infrastructures. This paper attempts to give an overview of the slope engineering in open-pit mines of China, including a summary of recent progress and main challenges confronted. The development of design procedure and achievements in scientific and technological research are introduced. A series of methodological topics are discussed for future development of mining slope engineering in China.

2

coal mines have rapidly increased in recent years (Lu Shizong, 1999). The distribution of China main open-pit mines is shown in Figure 2. Most of them are distributed in Northern China. Coal and iron mines are of large scale and excavation depth. In the 50’s of last century, most of the open-pit mines in China are of side-hill or shallow excavation with exploitation depth normally less than 100 m. But at present there are a number of large open-pit mines with a slope height of over 300 m, some even up to 600 m. For example, the slope of Shuichang iron mine will reach 700 m high in the near future (Table 1). The mining slopes keep on increasing in height as the exploitation keeps on going. And the stability of high and steep slopes is becoming more and more practical risk in many mines.

DEVELOPMENT OF OPENCAST MINING AND PROBLEMS

China is one of the few countries which have many opencast workings. Approximately, 90% of iron ore, 50% of nonferrous metals, 70% of chemical materials and almost 100% of building materials are extracted by opencast mining, as shown in Figure 1 (Tong Guangxu, 1995). Among them the number and scale of open-pit

Figure 1. The proportion of opencast vs underground mining in China.

2075

N Harbin Urumqi

Shenyang

Sinkiang

Hohhot

Beijing Tianjin

Yinchuan Xining

Jinan Lanzhou

Tibet

Xi'an

Zhengzhou Shanghai

Lhasa

Legend open-pit mine capital Beijing manpower waterway river

Figure 2.

Chengdu Chongqing Wuhan

Taibei

Kunming

Hongkong Haikou

Distribution of main open-pit mine un China.

Since 1970’s of last century, the slope slides have increased both in number and scale because of change of many mines from side-hill excavation into deep excavation.

A study of mining slope slides has shown that the frequency of failure increases obviously when the slope height exceeds 100∼150 m. Many serious failures were recorded in major open pit mines, such as Fushun, Dagushan, Baiying, Jinchuan, Nanshan, Haizhou, Baguanhe mines, and etc. For example, more than 70 slope slides happened in the Fushun west mine during its operation over 70 years. In order to increase the slope stability, totally 100 million m3 of rock and earth have been unloaded from the slope. In 1959, a major slope slide occurred along a weak layer of tuff and destroyed the main upgrade belt line, stopping the coal production for a period of 3 years. 14 slope slides were recorded in the north wall in 1978 because of the extended excavation along the northern slope. In 1979, another large slide happed in the west slope, burying the up-grade belt line and once again stopping the production. In 1984, intensive coal mining in the north side l induced a lot of slope failures and severe ground deformations, consequently resulting in a output reduction of 3 million tons of oil shale and 3.89 million tons of coal. Local subsidence and slides have been observed in the east

Table 1. List of high slopes of open-pit mines in China. No.

Mines

Location

Type of mines

Heights of slope (m)

States of mining

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30

Shuichang Changba Nanfen Daye Pingzhuangxi Fushun west Lanjian Dexing Haizhou Xincheng Yanqianshan Antaibao Jinchuan Jinduicheng Baiyun Tuanjie gou Dagushan Qidashan Cangshang Axi Xiaolongtan Heiwang Baoguo Yunfu Huangmailing Donganshang Shirengou Anjailing Gushan Qinshuigou

Hebei Gansu Liaoning Hubei Inner Mongolia Liaoning Sichuan Jiangxi Liaoning Heilongjiang Liaoning Shanxi Gansu Shanxi Inner Mongolia Heilongjiang Liaoning Liaoning Shandong Xinjiang Yunnan Shandong Liaoning Guangdong Hubei Liaoning Hebei Shanxi Anhui Yunnan

Iron lead an zinc Iron Iron Coal Coal Coal Copper Coal Gold Iron Coal Nickel Molybdenum Iron Gold Iron Iron Gold Gold Coal Iron Iron Sulfur iron phosphorite Iron Iron Coal Iron Phosphorite

700 600 500 428 400 400 400 380 350 330 300 300 300 290 200 200 200 200 200 200 200 190 180 180 150 120 120 120 100 100

Exploitation Changing into underground Exploitation Changing into underground Exploitation Changing into underground Exploitation Exploitation Closed in 2005 Changing into underground Changing into underground Exploitation Closed in 1990 Exploitation Exploitation Closed in 2001 Exploitation Exploitation Exploitation Exploitation Exploitation Exploitation Changed into underground Exploitation Exploitation Exploitation Changing into underground Exploitation Exploitation Exploitation

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and west side walls in the Dagushan open pit iron mine after its turning into deep exploitation. 19 slides have occurred in the Daye iron mine since 1972, involving a total volume of 1.4 million m3 of sliding mass. The largest slide, which occurred in 1979, has a volume of 20 × 103 m3 , covering a section of 890 m in length, accounting for 19.8% of the total slope length. Two slides occurred in the Baguanhe limestone quarry. The first one was slide of 10 million tons of limestone, and the second was 8 million tons, resulting in serious disruption of mining production. 10 slides or collapses are observed in a period of 20 years in one slope of the Baiying mine involving a total volume of 650 thousand m3 of sliding mass. During the period from 1975 to 1982, about 1.48 million m3 rock and soil debris were disposed

from the instabilities in the Jinchuan open pit. And in March 1990 a large slope slide occurred with a volume about 4 × 105 m3 , destroying the El.1514 m bench, eventually resulting in closing of the mine ahead of schedule in 1990. 21 slides happened in the Nanshan iron mine during 1979∼1990, and one of the largest occurred in 1980 with a volume of 2 × 103 m3 . The slope has reached up to 240 m in height in the Haizhou open pit since the mine was put into operation in 1952, and 73 slope slides were recorded in a period of 40 years, including three large ones. One slide happened in 1977 with a volume about 3 million m3 , burying 700 thousand m3 of coal and stopping the coal production for more than 1 month. Another one occurred in 1986 with a volume about 300 thousand m3 and the third one happened in 1989, involving 6 sidesteps in the south side. Table 2

Table 2. Geo-hazards of slope failure in open-pit mines. No.

Mines

Minerals

Year of failure

Scale and features

Consequences

1

Fushun west

Coal

1959

Sliding along tuff

1964

Large sliding

1979

Large sliding

Destruction of main upgrade belt line and recovery lasting 3 years Destruction of machinery-electric overhaul plant Burying of upgrade belt line and stop of production

1977

Sliding mass 168 × 103 m3 Sliding mass 20 × 103 m3 Sliding mass 250 × 103 m3 Rock fall

2

Daye

Iron

1979 1996 3

Jinchuan

Nickel

1979 1990

4

Baiyin

5

Haizhou

Copper

1983 1977 1986 1989

6

Baguan he

Limestone

1981

7

Sandaoling

8

Qianhe

Coal

1964

9

Pingzhuangxi

Coal

1983

10

Lvhe

Coal

1977

11

Hongqiling

1982

1986

Sliding mass 400 × 104 m3 Sliding mass 1 million m3 Sliding mass 3 million m3 Sliding mass 30 × 104 m3 Sliding of 6 benches Sliding mass 5 million m3 Sliding mass 21.63 million m3 Sliding mass 3 million m3 Sliding mass 47 × 104 m3 Sliding mass 10 × 104 m3 Sliding mass 63 × 104 m3

2077

Instability section length 890 m, account 9.8% of total slope Deformation and failure of benches, and safety of production threatened by minor rock falls Compelled close ahead of schedule in 1990 Forecast successfully ahead of three days Stopping production for a month

Excavation length reduced Transport line was compelled to move for 40 m Coal pit was buried Sidesteps destruction No. 1 winch line deformation, bunker line sliding down 5–10 m Sliding mass threatened operation of ropeway station

shows the severe consequence recorded in some open pit mines due to slope failure. By threatening the mining safety and economics benefits, the instability of high and steep slopes, has become the largest obstacle of mining production. Along with more and more opencast excavation being replaced by underground mining, more safety and technical problems are arising, to which more attentions should be paid by mining engineers and geologists in this field (Wang Jinxue et al., 2005, Cai Meifeng 2003). On the other hand, the slope angles of opencast mines are generally over 45 degrees in overseas countries, while under 40∼45 degrees in China. Therefore, to increase the slope gradient and at the same time to keep slope stability is much concerned by the community of mining engineers and geologists. However, due to the complexity and uncertainty of slope geology, especially the complexity and levity of transition stage from open-pit turning into underground excavation, there are still confronted a number of difficult problems in stability maintenance and failure forecast for high and steep mining slopes. 3

OPENCAST MINE SLOPE DESIGN AND STABILITY CONTROL IN CHINA

3.1 Main factors to be considered in mine slope design From the engineering point of view the mining slopes are of its particular characteristics being different from other engineered slopes of infrastructures. In study and design of open mine slope, some important factors should be taken into consideration, which can be summarized as follows: (1) The slope of open cast mines is changing time to time in the excavation and operation period. Open pit mine slopes are always in the process of excavation. In the initial phase of excavation, they are just provisional slopes. Only in the middle and final excavation phase do they become the final pit slopes. (2) Excavations can only be carried out under preexisting engineering geological conditions Mineral resources are the outcome of natural geological process, and occur in a particular geologic environment and structure. The location of mines cannot be chosen. Therefore, the mining excavation should adapt any geological conditions which are accompanying with the type of minerals. (3) Slope excavation is closely related to the economic benefits of mines Mining slopes are formed in the process of mine excavation. In order to reduce overburden excavation

in the initial stage, the mines have to be excavated not only by steps, but also by steeper slope gradients. When open pit mine excavation turns to underground mining, the underground mining generally starts from the toe of open pit slopes where the geo-stress is, however, concentrated. This kind of excavation model is unfavorable to slope stability and may result in a lot of slope safety problems. (4) Open pit mine slope engineering is a combination of complex structures For an open pit, there are involved various kinds of engineering structures, including of production platforms (such as haulage platform, working platform and so on), drainage channels and ramps. Therefore, the design of slope engineering usually has to be coordinated with the whole mining complex. (5) The criteria for assessment of slope engineering safety are flexible Since open pit slope belongs to a kind of provisional structure, so the slope stability just needs to meet the safety requirements for mining production. The service time for a mining project is relatively short, and on the other hand, the requirement for safety evaluation is also different under different surrounding conditions. The strength and deformation criteria should be accepted in different circumstances. (6) The objective of mine construction is to look for maximum economic benefits The eventual goal for open pit mine slope engineering is to seek an optimal point between slope safety and economic benefits of the mine. In most cases, an appropriate slope design is not that the slope can still keep stable after closure of the mine, but that it may be in a critical condition or fail eventually. In general, the principle in designing of mining slope is to find a best solution for guaranteed safety, economic benefits and environment protection, so-called ‘‘three in one’’. 3.2 Main codes for opencast mine slope design in China Along with the development of opencast mining technologies and improvement of related equipments, the reasonable opencast excavation depth has been increasing and the mining techniques have been improving. The maximum excavated depths have reached around 300 m for some closed mines such as Jinchuan nickel mine, Baiyin open pit mine and Daye iron mine. The design maximum excavation depth has reached up to 500∼700 m for some large ongoing mines such as Nanfen iron mine, Qi Dashan iron mine, Shuichang iron mine and Dexing copper mine.

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Figure 3. Reduction of overburden excavation in volume vs slope angle and depth of open-pit.

For a large scale opencast mine, one-degree increase in slope angle will significantly reduce the amount of the stripped rock materials, as shown in Figure 3 (Wang Min 1998). For example, for an opencast slope ranging between 30◦ ∼60◦ in angle without overburdens, one-degree increase in slope angle will reduce the stripped rock

materials 3.43%∼3.91%. For a opencast mine with a depth of 500 m, if the slope angle increases from 40 degrees to 45 degrees, the stripped rock material will be reduced by 48 million m3 in volume for a width of 1000 m. The rationality and reliability of a slope gradient mainly depend on study of the optimum balancing point between safety and benefit, as well as environmental requirements, which involves slope safety and mining maintenance. The China Safety regulations for metal and nonmetal opencast mines (GB 16423–1996) gives detail regulations on slope engineering, including the slope height, angle, maximum number of combined segments, water control and drainage for each working phase, and the blast requirements for the final slope. In addition, it provides requirements on regular site inspection and observations for mining slopes and it also requires establishment of a complete system for slope management and inspection for large opencast mine slopes or mines with large potential hazards. Compared with the former Design code for opencast coal mine engineering (GB50197–94), the latest Design code for opencast mines of coal industry (GB50197–2005, implemented from October 1, 2005) gives a special description on slope stability in a whole chapter, and provides detailed requirements on design of excavation site, landfill site, water proof and drainage. The Design code for opencast coal mines of chemical industry (HG/T 22810–1997) and the technical Code for opencast mines of rock building materials (to be issued soon) require the final excavation slope should be determined by designers according to the rock slope stability and relevant technical conditions. The Regulation of slope stability inspection for opencast mines (1992, compiled by the Supervision Bureau of Mine Safety under China Ministry of Labor) defines the range of slope angles according to the excavation depth and rock hardness factor f (Table 3). In addition, it provides specification on slope stability inspection. The inspection works are carried out according to the Regulations on slope stability inspection for township opencast mines issued by the Ministry of Labor. The slope inspection process

Table 3. Hardness factor of open pit excavation vs slope angle and depth. Depth (m) hardness factor (f)

Final pit slope angle (◦ ) ≦90 m

≦180 m

≦240 m

≦300 m

Bench angle (◦ )

15∼20 8∼14 3∼7 1∼2 0.6∼0.8

60∼80 50∼60 43∼50 30∼43 21∼28

57∼65 48∼57 41∼48 28∼41

53∼60 45∼53 39∼45 26∼39

48∼54 42∼48 36∼42 24∼36

70∼76 65∼70 60∼65 48∼60 48

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Table 4. Development of slope design method in China. Stage

No.1

No.2

No.3

No.4

Period Main design methods

1950s to 1970s Engineering analogy

1970s to 1980s Engineering analogy combined with limited equilibrium method

1980s to 2000 Complete quantitative analysis design, including numerical modeling

2001-present Slope system synthesis analysis and dynamic design

can be summarized as follows: collection of basic data of mines, processing of the basic data, analysis of the inspection data, conclusions and proposals, submission of inspection reports. Due to frequent safety accidents in non-coal mines, slope is clearly listed as an object for safety assessment in the Guidelines of safety assessment for non-coal mines, issued in 2003 by the State Administration of Production Safety Supervision. Case studies are made on slope failure types and specific methods are proposed for slope design in the Technical code for underground and opencast mining and the manual of implementation of safety assessment published by the Tianjin Electro-power Press. As well-known to all, an opencast slope without any risk is unreasonable. On contrary, a slope designed with too much risk is also unadvisable. In fact, an ideal slope design is to increase the slope gradient to the largest extent, reduce excavated waste rocks, and avoid risks through proper slope maintenance. The development of opencast mine slope design in China can be divided into four phases as shown in Table 4 (Wang Min, 1998). Slope design in 1950’s was carried out in China according to the excavated slope angles in former USSR. Through more than 50 years’ development, a dynamic (revisable or flexible) design method based on comprehensive analysis has been widely used now for opencast mine slope design. The slope design has experienced a transfer from engineering analogy based on experiences to dynamic design through quantitative and comprehensive analysis. 4

ADVANCEMENT OF OPENCAST MINE SLOPE RESEARCH IN CHINA

4.1 Slope optimization and steep side excavation techniques of large opencast mines An analysis of the cost of over 300 open pit mines in the world indicates that the cost for stripped rocks accounts for about 22% of the total production cost. Therefore, reduction of the stripped rocks by increasing slope angle can not only reduce mining cost, but also the area for dump, which is important in the environmental protection of mines. Compared with that in overseas mines, the slope angles of the opencast mines in China are relatively low

(Jiang Ke 2000), which is attributed by two aspects: one is that too much attention is paid to the safety of mine slopes while the economic benefits are not duly considered; another is that works are not done enough in slope stability analysis, site test, risk prediction and slope maintenance, as well as slope reinforcement techniques. Along with the development of slope analysis and monitoring techniques in China, attentions have been paid to the optimization of mine slopes in recent decade. For example, the original designed slope angle of the Dagushan iron mine has been increased by 2◦ ∼9◦ through slope optimization research, and the slope has been steepened through application of steep side excavation techniques, consequently producing additional 5.95 million tons of iron ores, reducing 25.86 million tons of stripped rocks and increasing largely the economic benefits. The original designed slope angle of the Daye Iron Mine East Opencast Mine was increased by 1◦ ∼4◦ through special-themed research, including engineering geological study, determination of mechanical parameters and theoretical analysis, as a result, producing additional 0.95 million tons of iron ores with a significant economic benefit (Sun Zainan 1999). Three principal problems need to be solved for slope optimization and steep side mining: 1. Zoning of the opencast mine slopes according to geological conditions identified by geo-mapping; 2. Establishment of engineering geological models for different zones, so as to provide a basis for slope optimization; 3. Slope gradient optimization for each zone through detailed stability analysis. During 2001–2005, a special research program was carried out for the Shuichang iron mine. In that research, the studied area (i.e., the Northern slope) was divided into five zones as I—V, based on the characteristics of rock formations, rock mass structures, mining designs and slope orientations (Figure 4). An analysis of the rock mass structures was made for each zone and five geological failure models were established, including plane sliding, circular sliding, wedge sliding, toppling and falling (Li Changhong et al., 2004). Comprehensive stability analysis was carried out for the excavation process including 3D

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Figure 4. Slope engineering zoning of Shuichang iron mine.

Table 5. Recommended slope angles in Shuichang open-pit mine. Subarea Profile Position

Dividing Original Gradient line scheme (◦ ) level/m (◦ )

I I I I I I II II II II II II III IV IV IV IV IV V

50 50 43 45 43 49 45 47 45 48 48 50 48 41 44 48 46 48 46

I-1 I-1 I-2 I-2 I-3 I-3 II-1 II-1 II-2 II-2 II-3 II-3 III-1 IV-1 IV-2 IV-2 IV-3 IV-3 V-1

top Lower part top Lower part top Lower part top Lower part top Lower part top Lower part whole whole top Lower part top Lower part whole

+10 +10 +50 +50 +10 +10 −50 −50 −111 −111 −112 −112 – – +10 +0 −50 −50 –

each zone, as a result, reducing 30∼50 million tons of stripped overburdens with enormous economic benefits (Cai Meifeng et al., 2004; Feng Jinyan et al., 2005). 4.2

46 46 41 41 41 41 46 46 46 46 46 46 42 40 42 42 42 42 42

Slope stability for the phase from opencast mining to underground mining

Up to now, many large mines in China such as Daye iron mine, Shirengou iron mine, Baoguo iron mine, Heiwang iron mine, Changba lead and zinc mine, Fushun west strip mine, and Xincheng gold mine have turned from opencast excavation to underground mining. The extensive opencast excavation has resulted in significant stress disturbs around the pit and stress concentration at the toe of the the slopes. In this condition, underground excavation will cause larger scale stress disturbs, consequently inducing slope failure, which may create great difficulties and potential risks for the excavation changing from opencast to underground (Li Siji 1999 & Fu Helin 2002). The principal problems for the change from open pit excavation to underground mining include:

solid-fluid coupling analysis and GIS analysis (Cai Meifeng et al., 2006). Accordingly, the slope gradient optimization was carried out (Table 5) and the overall designed slope angle was increased by 1◦ ∼6◦ for

1. The impacts of hanging-wall and toe mining on the slope stability in the transition period; 2. Selection of the optimum underground excavation method on the basis of analyzing the impacts of the different excavation methods on the slope stress redistribution;

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3. Necessary slope deformation control when the transportation system of underground excavation is located on the slope. Hanging-wall excavation will further disturb the stress-concentrated slope toe, leading to higher risk of slope failure. Competent ores can be left as retaining walls for slope deformation control. Limit equilibrium analysis and numerical modeling can be used for determining rational excavation method. When slope deformation control is required, necessary strengthening measures are needed in addition to optimization of mining method and real-time displacement monitoring. After more than 100 years of excavation, a steep slope of 428 m in height has formed in the Daye iron mine when excavation has to change into underground. The hanging ore on the opencast slope is the first to be mined, which will be bound to further weaken the stability of the toe part of the high slopes. Due to the short period of excavation at this part, extensive treatment is economically unreasonable. Two alternative mining schemes have been compared for excavation of the hanging ores at the northern side of the Xiangbishan district (Figure 5, Li Hui 2005),: Scheme 1: excavation to elevation +18 m, all the iron ore is mined without retaining wall; Scheme 2: excavation to elevation +18 m, leave an iron ore wall of 4–6 m in thickness as retaining wall at the slope toe. The results of numerical simulation indicate that the displacement of scheme 2 is 10 times less than that of scheme 1 and its plastic zone is also much smaller, then the retaining walls can be accepted for deformation control and maintaining slope stability. The Fushun opencast coalmine is one of the largest slope engineering in China. In transition from opencast to underground mining, a compound vector effect special study has been proposed to study on

Qh intense weathering diorite

scheme 1

5

scheme 2

4 48

retaining wall 36 Fe 18

3 2 1

F25

diorite

Figure 5. Mining scheme and control points of Daye opencast slope.

the combined open-underground mining techniques (Sun Shiguo et al., 2000). The features and causes of the subsidence and ground cracking at the Fushun No. 1 petroleum plant close located to the slope were studied (Wu Jieqing & Han Guogang, 1990). Special-themed researched have also been carried out on the effects of the transition mining from opencast to underground on the slope stabilities and other relevant problems for other mines, including the Changba lead and zinc mine (Wang Gongming et al., 2003), Xinchen gold mine, Shi rengou iron mine, Baoguo iron mine and Heiwang iron mine (Jiang Yanmei et al., 2005, Huang Ruiquan 2002, Wang Zhixue & Xi Wangfang, 2004, Li Ming et al., 2001). In the recent years, studies have been made on slope stabilities related to high bench exploitation, slopes without benches and pre-blast design for opencast slopes (Luo Yuanfan 2005, Zhang Rongjin et al., 1997, Wu Chaoyang 2002). 4.3 Slope deformation induced by combined opencast and underground mining The large coalfields in China are mainly located in the loess plateau region of west and northwest China and most of the coal seams are gently inclined. Up to now many of them have begun to use combined opencast and underground mining techniques to increase production and reduce waste, which raised many new challenges for opencast slope stability research (Wang Zhenwei 2006). In space, the opencast slope is located in the influencing region of underground excavation. As a result, the two influencing areas induced respectively by the opencast and underground mining are overlapped partly, which makes the deformation mechanism of slope rock mass in this part more complex, and the deformation and the its influencing area much larger. The Fushun coalmine is one of the typical examples of this. The Fushun coalmine is located in a transition belt between of Songliao plain and Liaodong peninsula hills. The original landform is gentle with a height difference of about 100 m and the average altitude is around +80 m. After more than 90 years of excavation, a pit of 6.6 km (length) × 2.2 km (width)×400 m (depth) has been formed with a slope angle of about 31 degrees. Due to the gentle inclination of the coal seams, the combined opencast and underground mining technique was adopted for further excavation in the depth, i.e. the shallow part was excavated by opencast while the deep part was excavated through the Shenli coal shaft. Mining in the Shenli mine began as early as 1907 and was mainly located under the level −100 m at the early stage, and between −400 m and −650 m

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at the later stage (the level between −400∼−425 m was kept as a barrier pillar between the Fushun opencast mine and the Shenli mine). Severe deformation was induced at the industry establishments around the mining area by the Shenli mine. The mine was closed in 1979 for the safety of the industrial establishments above. The mining location of the Fushun opencast coalmine and the Shenli mine is shown in Figure 6. The influencing area is located between W1400 and E2000 determined according to projection of the influencing angle of underground mining, which indicates that the north slope of the Fushun opencast mine is all located within the influencing area of the underground mining. Figure 6 shows the location of the slope of profile W200 and the mining area at the east section of the deep shaft. The deformation modes induced by the sequence of underground mining followed by opencast excavation are studied (Sun Shiguo et al., 2000) and summarized in Figure 7. 4.4

Comprehensive treatment of opencast mine slopes

Comprehensive treatment refers to the slope maintenance by multi measures that can be summarized as follows: 1. Treatment and drainage of surface water and groundwater

The treatment for surface and ground water in the mining area includes the prevention of surface water from entering the fissures in the slopes and drainage of groundwater to reduce water pressure along sliding planes. . 2. Controlling blast The controlling blast includes reducing the charge of every differential blast to the minimum extent; using pre-blast for the final slope surface; using cushion blast between pre-blast and normal production blast. 3. Slope reinforcement Reinforcement can not only stabilize the potential unstable slopes, but also increase the slope angle to produce more benefits. Reinforcement measures mainly include anti-slide piles, rock bolts, anchor cables, pressure grouting, shotcrete and seepage control grouting. 4. Deformation monitoring and observations. The prediction of slope deformation and adoption of corresponding reinforcement measures are made according to the magnitude and rate of measured displacement; On the other hand, workers and equipments can be evacuated in time by warning before slope failure. In recent years some progresses have been made in study of the integrated treatment of opencast mine slopes, including reinforcement schemes, deformation monitoring and design optimization.

Figure 6. Locations of underground excavation under open pit slope.

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Figure 7. Ground deformation due to slope and underground mining along profileW200 (1–6 are lateral deformation curves for different period, 1′ -4′ are lateral deformation curves for different period; 1,1′ : 1959-12∼1987-10; 2,2′ : 1959-12∼1987-03; 3,3′ : 1959-12∼1966-12; 4,4′ : 1959-12∼1964-12; 5: 1959-12∼1962-12; 6: 1959-12∼1961-12.

In late 1990s, a integrated concept in slope treatment was applied to the east open pit of the Daye iron mine, including retaining, dewater, anti-sliding, anchorage and protection measures, which has provide important experiences of slope management of deep open pits (Sun Zainan & Chen Yinfang, 1993). In order to explore both theoretically and practically the possibility of reinforcement of gentle bedded coal mine slopes by pre-stressed anchorage cables, studies have been carried out on pre-stress anchorage reinforcement design and stability analysis for coalmine slopes of gentle beddings (Gao Qian & He Bengui, Gao Qian et al., 2004). Opencast slope monitoring techniques have developed very fast in recent years. The research in this aspect has focused on dynamic analysis of slope deformations and slide failure forecast based on observation information. The application of GPS monitoring technique in the Jinchuan (Zheng Guozhong et al., 1998), Shuichang and Changba opencast mines has greatly improved deformation monitoring techniques of opencast mine slopes in China and real-time automonitoring techniques have come into use in slope rock mass (Xu Wancai et al., 2005; Zhao Jingbo et al., 2005; Yu Qingzai 1997). 4.5

Environmental protection of opencast mine

Exploitation of mine resources is one of the most serious destructions to environment imposed by human

activities. The environmental problems of open cast mines include slope excavation and solid mine dumps. Environmental protection for opencast mines includes two aspects. One is reduction and control of mine dumps; another one is disposal and recycling use of the mine dumps, vegetation recovery in the opencast mined-out areas, and comprehensive rehabilitation of mine slopes and dump sites. The approaches for further industrialization of land reclamation in mined areas include (Wang Jianjun et al., 2006): 1. Speed up legislation for land reclamation in mined areas and perfect relevant law and policy systems. 2. Introduce competition mechanism through market and promote the market development of mine land reclamation. 3. Formulate reasonable benefit distribution mechanism. 4. Improve the capacities of workers in this sector. 5. Strengthen international cooperation and introduce advanced concepts and experiences in mine land reclamation. A lot of mine companies, such as the Sichuan Huidong Mankuang Corp. (Yuan Shilu 2004), the Pan zhihua Steel and Iron Corp. (Li Yuchang 2000; Luo Dejun & Liu Hanchao, 2003), Fujian Zijinshan gold mine company (Li Feng et al., 2005) and others, offered large investment for mine environmental

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protection and management, and achieved significant results. 5

CHALLENGES AND PROSPECTS OF OPENCAST MINE SLOPE ENGINEERING

In the previous paragraph some advancements in the study of opencast mine slopes are introduced, however, the problems confronted are far from appropriate solution, and the challenges still exist in connection with the rapid development of increasing mining in China. Along with the continued exploitation of mine resources, the mining conditions become more and more complex for many opencast mines, as the steep excavation, deep-pit excavation, extended side excavation are used. Comprehensive studies have given a relatively satisfactory result (Wang Sijing, 1997). The transition from opencast to underground mining and extensive use of combined opencastunderground excavation, as this happens in Daye iron mine, the Shirengou iron mine, the Baoguo iron mine, the Heiwang iron mine, the Changbei lead and zinc mine, the Fushun coalmine, and the Xincheng gold mine and so on, are two unsolved problems. The complexity of the problems has brought about new challenges in its solution. Much more difficulties in guarantee of mine safety, economy and environment management appear for mining design and operation. The interaction between underground and open pit excavation will show an integrated effect to the complexity of stress distribution and deformation pattern, therefore, the subject to be studied becomes a complex meta-system and the problem will develop from single slope stability to a system engineering issue. A meta-system composes of systems and subsystems which are interactive between various factors and levels. This interaction causes the meta-system to be of dynamicity and non-linearity. Therefore, a metasynthesis methodology is to be developed to solve the complex mining slope engineering problems. Understanding and quantitative characterization of the engineering features of system elements and influencing factors is the basis for meta-system integration. In the evaluation of element properties a number of new techniques, including geophysical and in-situ testing methods, can be used, however, the nondeterministic behavior of geo-systems becomes a most important factor to be concerned. The authors of this paper have proposed the research thinking of engineering geology, studied the interactions of human engineering activities and geological settings and their impacts on the environment and, based on the fact that the existing researches have mainly focused on unilateral impacts of the relevant

factors instead of the coupling effects of internal and external factors, pointed out that the geosphere dynamics is the basis of geologic environment, geological disaster and engineering researches (Wang Sijing, 1997a; 2004). The coupling theory of earth internal and external dynamics has been explored and the genesis of major geological disasters has been analyzed (Wang Sijing 2002). The stability status of an opencast slope engineering system reflects also the coupled effects of internal and external dynamic processes. Therefore the coupling theory of internal and external dynamics of opencast slope engineering will become a new approach for opencast slope engineering researches. For solution of connective processes of meta-systems of mining slopes the THMC coupling analysis method is proposed to be used (Wang Sijing et al., 2003; 2005). Currently, 3S (GPS, GIS, RS) techniques have been widely used in acquisition, management and analysis of mine and environmental information, which may significantly improved the economic and social benefits of mines. The 3S techniques are very helpful in the meta-system management and in the integrated evaluation of the whole system behavior, as well as in the environment management. At present, GPS has been already applied for slope deformation monitoring in several opencast mines in China. GIS has found application in slope engineering of the Shuichang opencast iron mine. Increasing attentions have been paid to the application of ‘‘3S’’ techniques in mines (Xie Mowen & Cai Meifeng, 2005; Luo Deren, 2001; Luo Zhouquan 2002; Cheng Jiehai and Hao Limin, 2004). In the time of knowledge globalization, any research can not be accomplished independently by a single country or region. Chinese researchers are looking forward to the cooperation with overseas counterparts in opencast mine slope researches for further development and mining prosperity.

ACKNOWLEDGEMENT The authors are grateful to Professors Cai Meifeng, Peng Suping, He Manchao, Liu Quansheng for kind cooperation in mining research programs. Many thanks are to give to Mr Zhang Xiaoping, Hu Bo, Liu Shungui for help.

REFERENCES Cai Meifeng 2003. Main issues metallic mines now are facing and solutions of the problems. Mining Engineering 1(1): 40–43. Cai Meifeng, Feng Jinyan & Wang Jinan 2006. Threedimensional hydraulic coupled stability of a high steep

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open pit slope. Journal of University of Science and Technology Beijing. 28(1): 6–11. Cai Meifeng, Qiao Lan, Li Changhong & Wang Jinan 2004. Stability Analysis and Design Optimization of High and Steep Slope in Shuichang Open-pit Mine. Journal of University of Science and Technology Beijing. 26(5): 465–470. Cheng Jiehai & Hao Limin 2004. Application of ‘‘3S’’ Technology in Coal Mine. Express Information of Mining Industry. 17(6): 52–54. Feng Jinyan, Wang Jinan & Cai Meifeng 2005. Study on the Optimization of the Slope Angle and Ability of High Deep Open Pit Slope. China Mining Magazine. 14(4): 45–48. Fu Helin 2002. Slope Stability Analysis During the Period of Underground Excaving. Hunan Nonferrous Metals. 18(4): 4–6. Gao Qian & He Bengui 2004. Stability analysis and reinforcement approach research of the jointed and bedding rock slope. China Mining Magazine 13(5): 34–36. Gao Qian, Zhao Jingbo & Wu Xuemin 2004. The Application and Stability analysis of Prestressed Anchor Rope Reinforce Slope. Mining Engineering 2(3): 9–12. Huang Ruiquan 2002. The Characteristic of Baoguo open pit iron mine turn into underground excavation. Express Information of Mining Industry (8): 5–6. Jiang Ke 2000. The approach to improve surface mine’s economic benefits. China Mining Magazine. 9(4): 109–111. Jiang Yanmei, Zhao Jianjun & Song Aidong 2005. The Discussion of the Problems during the Process of Open Pit turn into Underground Exploitation in Shi Rengou. Express Information of Mining Industry (10): 30–31. Li Changhong, Cai Meifeng & Li Juncai 2004. Analysis of the Rock Mass Structure and Destruction Mode of Shuichang Open-pit Mine Slope. China Mining Magazine. 13(2): 48–51. Li Feng, Wei Aimin & Zhu Ruigeng 2005. Research of Green Ecological Materials and Mine Recovery. Journal of Anyang Institute of Technology. (6): 1–3. Li Hui 2005. Control Measure and its Numerical Analysis for Slope Satety during Hanging-wall Ore Mining below a Landslide. Wuhan: Institute of Rock & Soil Mechanics Chinese Academy of Sciences. Li Ming Zheng Huaichang, Zhang Pengbuo & Ju Fengyu 2001. Application of Combined Mining of Boundary and Corner Orobocly at Heiwang Iron Mine. Metal Mine (2): 20–22. Li Siji 1999. Study on transition from open pit to underground mining without stopping of operation, Design and Construction of metallic mines [J]. 31(5):3–8. Li Yuchang 2000. Present situation and protection of geological environment in mines in Pan Zhihua city. Acta Geologica Sichuan 20(2): 125–129. Lu Shizong 1999. Basic Conditions and Prospect of China’s Mine Slope Research. Metal Mine (9): 6–10. Luo Dejun & Liu Hanchao 2003. Geological environment character & evaluation of mainly reveal mine in Pan Zhihua city. Journal of Geological Hazards and Environment Preservation. 14(4): 25–29. Luo Deren 2001. Review on Application of 3S techniques in Mining, Hunan Metal [J]. 17(6):52–54. Luo Yuanfan 2005. The Research of application of Open-air High Stage of Liancheng Mn-Mine in Fujian Province. China’s Manganese Industry 23(4): 40–42.

Luo Zhouquan 2002. GIS, RS, GPS and their application in mining. China Molybdenum Industry. 26(5): 47–53. Sun Shiguo Yang Suzhen & Qin Siqing 2000. Rock Mass’s Deformation Research of composite mining. Beijing: Seismology press. Sun Zainan & Cheng Yinfang, 1993. Stabilization of landslide No.1 in Daye Iron Mine, Proceedings of Daye iron mine. Sun Zainan 1999. Study on potential increasing of slope gradient in Daye east open pit mine, Proceedings of Daye iron mine. Tong Guangxu 1995. Mining in Hard Rock, Metallurgy Publisher, Beijing. Wang Gongming & Ren Fengyu 2003. Group mining destroy and the problem and practice of open pit mine turn into underground excavation. Mining Technology 3(3): 1–8. Wang Jianjun Hu Ke & Wang Yonghui 2006. The realization ways for the industrialization of mine reclamation. Resources & Industries. 8(1): 55–57. Wang Jinxue Wang Jiachen, Dong Weijun & Ji Zhaoning 2005. Study of Deep Mining Technology of Large-scale Opencast Metal Mine. Metal Mine (7): 14–16. Wang Min 1998. Review of Metallurgy Mine Slope Design of China. China Mining Magazine 7(3): 21–24. Wang Sijing, Xiaoli Liu & Enzhi Wang 2005. Hydrogeodynamic Analysis and Application, Proceedings Geoproc2006, Nanjing, Hehai University, 14–30. Wang Sijing 1997. Interaction between human engineering activity and geoenvironment and its environmental effects. Journal of Geological Hazards and Environment Preservation 8(1): 19–26. Wang Sijing 1997a. Scientific Thinking of Engineering Geology. Journal of Engineering Geology. 5(4): 289–291. Wang Sijing 2002. Coupling of endogenic and exogenic geodynamic processes and origins of serious geological disasters. Journal of Engineering Geology. 10(2): 115–117. Wang Sijing 2004. Geosphere dynamics—foundation for study in geoenvironmental, geohazard and geoengineering. Journal of Engineering Geology. 12(2):113–117. Wang Sijing & Wang Enzhi, 2003. Recent Study on Coupled Processes in Geotechnical and Geo-environmental Fields in China, Proceedings Geoproc2003, Stockholm, 66–76. Wang Zhenwei 2006. Study on the slope stability of loess floor earth-disposal site which has been affected by well mine. Fushun: China Coal Research Institute in Fushun. Wang Zhixue & Xing Wanfang 2004. Preliminary discussion on the Change of mining technique from open pit mining to underground mining in Xincheng Gold Mine. Gold 25(10): 22–25. Wu Chaoyang 2002. Designing of Presplitting Blasting in Opencast Mining. Mining and Metallurgical Engineering 22(2): 45–46. Wu Jieqing & Han Guogang 1990. Ground Cracking and Collapse in Site of Fushun Petroleum Factory, Chinese Journal of Rock Mechanics and Engineering, 9(4): 286–299. Xie Mowen & Cai Meifeng 2005. Theory and Practice of Information Slope Engineering, Beijing: Science Press[M]. Xu Wancai, Ouyang Zhenhua & Liu Zhiqiang 2005. Application of GPS technology in slope monitoring of Shuichang iron mine. Opencast Coal Mining Technology (1): 12–15.

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Zhao Jingbo, Zhang Liguo, Yu Yalun, Yang Shuo & Gao Qian 2005. Slope stability prediction analysis based on GPS monitoring. Nonferrous Metals (Mining Section) 57(1): 29–31. Zheng Guozhong, Xu Jiamo, Ma Fengshan & Cao Jing & Xia Yi & Zheng Shilin 1998. GPS Monitoring of Slope Deformation in JinChuan Open pit Mine. Journal of Engineering Geology 6(3): 282–288.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Structure and failure patterns of engineered slopes at the Three Gorges reservoir Y.P. Yin China Geologic Survey, Beijing, China

ABSTRACT: A large number of engineered slopes have been created in the city, town and road reconstruction of the Three Gorges reservoir area. According to statistic data, the high slopes are more than 3000, and 90% are unstable that severely threat the resettlement plans. The reclamation work is distinguish and remarkable, compared to other similar projects in China. The stability-based slope utilization is considered, and the houses can be constructed at the upper, bottom, and even in the middle part of the slope. Meanwhile, the landscape should be protected mostly for tourism purposes. This paper summaries the slope patterns and studies on deformation and failure mechanics. Case-studies on slope stabilization and utilization are also presented. 1

INTRODUCTION

The Yangtze River Three-Gorges Project is the largest hydropower-complex under construction in the world and the dam of the Three-Gorges Project is located on Sandouping, Yichang City, Hubei Province (refer to Figure 1). It is well known that the resettlement project of the Three-Gorges Reservoir Areas is of great significance. However, there always exist hazards such as debris flow, landslides and collapse among the twelve relocation counties (cities) and 116 towns where relocation is necessary. Landslide prevention has been a prime concern as these areas suffered severely from this disasters. The following are some typical examples. • On June 12, 1985, a slope with a height of 800 m and a total volume of 20×106 m3 dumped into

Figure 1. Overview of the resettlement and reconstruction plan of Three Georges reservoir area.

the Yangtze River at Xintan, creating a surge of 80 m. 1371 people evacuated 24 hours prior to the landslide. • The Wu Long landslide is an example that brought disastrous event happened on 1st May 2001 (Figure 2). 79 people were buried and by 16,000 m3 rock fall. • The Qian Jiang Ping landslide, happened on 13 July, 2003, is believed to be partly attributed to the impoundment of the Three-Gorges Reservoir whose water level rose from 100 m to about 135 m. 24 people were killed and 1100 people homelessness. Figure 3 gives an overview of Qian Jiangping failure. The cross section before and after the failure are shown in Figure 4 and Figure 5 respectively.

Figure 2.

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Overview of Wu Long Landslide.

2

Figure 3. Overview of Qian Jiangping landslide.

Figure 4. Cross section before the failure of Qian Jiangping.

Figure 5. Cross section after the failure of Qian Jiangping.

With the implementation of the project of resettlement of emigrants from the reservoir region of the Three Gorges, a lot of human-cutting slopes had been formed as a result of construction of cities, roads, and bridges. To make full use of limited land resources, resettlement engineering projects are usually in the form of reinforcement in the front and excavation in the back. The exposure of these slopes had changed the natural seepage field and stress field and the spatial configuration and stress field of the slope rock mass had been modified. Generally, the stability of slopes decreased significantly, which is dangerous to the nearby roads and buildings. The problem of human-cutting slopes is a very crucial in the reservoir region. In the reservoir region of the Three Gorges, the top, bottom, and even middle part of the slope would all be used for building, which is quite different from other regions. So, the utilization of slopes should be considered in the process of controlling slopes. Furthermore, planting and tourism should also be accounted for.

BEDDING SLOPE

The bedding slope, which is mostly distributed in such counties as Xingshan, Zigui, Badong, Wushan, Fengjie and Yunyang, is the most widespread slope around the reservoir. It mainly occurs in areas with formations of Triassic and Jurassic, which are composed of interbedded sandstone and mudrock, or limestone and marl, with a dip value ranges from 20 to 70 degrees. The deformation of the slope is mainly controlled by the bedding planes. The process of the slope destruction can be divided into three stages: natural slope, cutting slope, and destroying and slipping (Figure 6). In the natural condition, the stability of slopes is usually high as a result of no turbulence. Buckling failure is the most common mechanism leading to deformation and unstability of slopes, which is related to unloading, weathering, and hydrostatic pressure and uplift pressure caused by groundwater. Dip slopes are widely occurred around the reservoir, and the resultant damage is huge. Based on observation of the Wufengshan Slope in the old city of Yunyang, the slope body is composed of layered sandstone interbeded with mud-stripe or mudrock of Suining Formation. The mud-stripe and musrock form the potential sliding surface, with a thickness of 4–6 m. The slope has a dip of about 40◦ ∼45◦ , a height of about 150 m, and a length of about 230 m. All these factors have led the failure of the rock mass similar to that of sheet crack rock mass. The widely distributed pine trees in the form of Sabre trees indicate that creep deformation has been developing for a long time, especially when rain lasts a long time, the decrease in pore pressure would reduce the stability of slopes. The critical length of sheet crack rock mass without engineering measures can be calculated by the following equation which was originally proposed by Sun:  2π 2 Eh2 lcr = 3 (1) 3γ sin α where, lcr—The critical length of sheet crack rock mass; E—Deformation modulus of rock mass; H—thickness of sheet crack rock mass;

Figure 6.

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Same-layer Slope structure and failure pattern.

α—dip of sheet crack rock mass; γ —specific weight of sheet crack rock mass. Accordingly, the safety coefficient, K, can be obtained by: lcr (2) l where, l is the actual length of sheet crack rock mass. Assume that γ is 26 kN/m3 , E is 1 000 000 kN/m2 , and lcr is 242 m, then the corresponding safety coefficient, K is 1.04. This case shows that in general conditions, the stability of the Wefengshan layered crack rock mass is below normal. If the rock mass contains several sets of discontinuities, and rain lasts and creep develops for a long time, then its critical slope length would decrease considerably. According to the deformation and failure characteristics of the new landslide at Wufengshan, if the safety coefficient, Ks is recommended to be 2.5, then the recommended slope length should be 96.8 m and the corresponding vertical spacing should be 62.2 m to keep the Wefengshan layered crack rock mass stable. When reservoir area inhabitants build new houses and excavate the slope bottom, the support to the dip slope would lose, and the slope would be unstable and fail, leading to either shallow deformation or bedding landslide, both of which are harmful to the resettlement of inhabitants. The stability of slopes is determined by the ratio of the interior angle of friction of the bedding plane to the dip of rock formation. When several sets of discontinuity is present, the ratio is usually below 1, shallow deformation would happen. Due to the absence of major discontinuities, the dip slope at a road in Xinshan is high in stability, even though its dip is as high as 40◦ ∼50◦ and the ratio is below 1. The high slope to the west of the Baiyangou Bridge in Badong County is characterized by slope topplingsliding caused by excavation of slope bottom. Here the natural dip is around 25∼30◦ , and the lithology is massive marly limestone interbedded with limestone and marl of the Third Member of Middle Triassic Badong Group (T2b3), with a trend of 350◦ and a dip of 32◦ . Around the slope, the strata are mainly monoclinal. Three sets of joints occurs: ① strike 260∼275◦ , trend SSE∼S, dip 60∼85◦ ; ② strike 330∼345◦ , trend SWW, dip nearly 90◦ ; ③ strike 285∼290◦ , trend SW, dip 70∼85◦ . Among them, the second set is the most prominent, and its extension is as long as more than 10 m (Figure 7). As a result of the reconstruction of urban roads, the road near the west of Baiyangou Bridge had been shifted innerly to the rock mass for about 25 m, and the excavation height is as high as 25 m. In the rain season of 1998, large scale sliding failure developed, covering an area of 1.6×104 m2 , a width of 160 m at the front part and a length of 200 m. The sliding K=

Figure 7.

Bayangou Bridge slope structure.

extent is as large as more than 20 m, and about an area of 300,000 m3 rocks were made loosen. Even more, three long and big troughs and numerous short troughs developed, the biggest of which is as wide as 19 m. The slope failure had caused the damage of road for about 160 meters, and several houses at the outer side of the road had been influenced. The discrete element method had been utilized to model the process of slope excavation and deformation. In the initial condition, the length and height of the domain is 240 m and 132.8 m, respectively. The dip of the bedding plane is 27◦ ; the thickness of layer is 3.5 m; the dip of joints is 88◦ ; The spacing of joints is 16 m. The size of cells is set to be 35 m; The Mohr-Coulomb model is chosen; The left, right and bottom boundaries are all set to be fixed boundary; the initial stress is considered to be geostatic, and the ratio of vertical stress to horizontal stress is 2. As shown in the modeling results (Figure 8), three aspects about the deformation and failure of bedding marly limestone slope can be summed up: (1) free-faces would develop as a result of excavation without support; (2) the intercated weak layers in the rock mass are potential slip surfaces; (3) the widely occurred fractures contributed to the infiltration of surface water, which reduced the anti-shearing intensity considerably, especially in the rain season. Generally, such measures as anti-slide-pile, and drainage of both surface water and groundwater should be taken. A lot of experience had been gained in the prevention and controlling of dip slope. Due to the shortage of land for construction, controlled slopes should be used as land for construction. In the slope at Forestry Administration of Wushan City, bedding slipping happened after excavation in 2001(Figure 9), and it spread backward in the heavy rain in May, 2003. Such measures as adopting Anti-Sliding Piles of reinforced concrete in the back, Anchor Spray Net in the front, and Load-Bearing and Anti-Sliding Piles in the middle part had been proposed by the author and others to keep the slope safe. These measures had been carried out successfully and are good examples of utilizing slopes for construction.

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Figure 8. UDES result of Bayangou Bridge slope (step = 4300) (a) Resultant displacement (b) Shear displacement.

Figure 9. Prevention-based slope utilization of houses at Wushan.

3

‘‘SANDWICH’’ - SHAPED SLOPE

‘‘Sandwich’’ - shaped slope is composed of interbedded sandstone and shale of Jurassic Shaximiao Group and massive sandstone and shale of Jurassic Suining Group. Based on investigations at a profile in Wanzhou, 17 units of sandstone and mudrock, which constitutes 8 cycles, had been identified. This kind of slope occurs in Yunyang, Wanzhou, Zhongxian, and so on. The rock formations contain two sets of joints, which reduces the stability greatly after digging (Figure 10).

Figure 10. ‘‘Sandwich’’-shaped slope structure and failure (a) Natural slope (b) Manmade slope (c) failed slope.

The main failure mechanism of such slope is toppling. For a natural slope, after excavation, the underlain soft rock would be further weathered and denuded, and the overlain sandstone would deform resiliently, which led to the outward movement of the center of gravity of the deformation body. The stability of rock mass can be evaluated by the ratio of anti-rollover bending moment to toppling moment. When the ratio is below 1, toppling would occur. It is clear that during heavy rain the hydrostatic pressure caused by infiltration of water would increase toppling moment, and the groundwater would soften the underlain mudrock, which might lead to uplift pressure. All these factors can make the rock mass unstable. When the

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dip of formations is large, bedding-slip mechanism should be considered. In the reservoir area, as a result of interbedding, a characteristic of multi-order toppling-sliding deformation and failure is displayed. The human cutting high slope in the Prosecuting Bureau of Yunyang is used as an example. It is composed of sub-horizontal layered sandstone and mudrock. The sandstone is in the upper part, and the mudrock is in the lower part. The strike and dip of rock formation are 210◦ and 15◦ , respectively. A set of joints (200◦ ∠58◦ ) is the most prominent. The height of slope is about 20 m, and dip angle is about 80◦ . Owing to the difference in anti-weathering ability of sandstone and mudrock, research on the deformation and failure characteristic of ‘‘Sandwich’’ - shaped slope is of practical importance. The available results show that, although the red bed mudrock is weak in dilatability, most of the mudrock is durable. Field investigations also reveal that, the strong weathering of bare mudrock in the bottom of slope would lead to development of cavities. The formation of cavities and redistribution of stress would cause the failure of the overlain sandstone. Based on these, the discrete element method is used to analyze the evolution of the deformation and failure of the slope before and after controlling.

Figure 11.

Cutting slope displace vector (step = 5000).

Figure 12.

Cutting slope displace vector (step = 30000).

3.1 The evolution of slope before controlling (1) establishment of the model and selection of parameters A model using discrete element method based on field measurements is established. The influence of bedding planes and joints has been accounted for. The spacings between bedding planes and between joint planes are both 1.0 m. The ranges of joints in the sandstone and the mudrock are 4 m and 2∼3 m, respectively. The Mohr-Coulomb model is chosen; The left, right and bottom boundaries are all set to be fixed boundary; The initial stress is considered to be geostatic. (2) results and analysis Results from numerical analysis show that, as the excavation goes on, the stress field redistributes, and the shift of slope body rotates to the free-surface. However, due to the difference of rock in strength, the cumulated displacement of the mudrock in the lower part is much larger than that of the sandstone in the upper part. When the iteration reaches 5000 steps (Figure 11), considerable deformation has happened in the mudrock, which led to the increase in deformation of the sandstone, and caused local pull cracks. When the iteration reaches 30000 steps (Figure 12), the aforementioned deformation enhanced greatly. It is clear that with the increase in iteration times, the distribution of stress differs: stress concentrates in the mudrock, and tension stress occurs in the sandstone.

The evolution line of tracking body displacement indicates that the deformation of slope is still continuing. As a result of the low strength of mudrock, as well as the softening effect of groundwater, the deformation of mudrock toward the free-surface could lead to tensile and bending yielding deformation of the upper sandstone, and the impact range in around 10 m. 3.2 Analysis on deformation under shotcrete (1) Analysis on deformation under shotcrete with Cable and Wire mesh To protect the slope from deformation and failure, 4 12∼15 m-long cables were employed. Each cable was pulled sequentially to 35T, and the slope surface is

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4

SLIDE DEPOSIT SLOPE

Slide deposit slope is also a common kind of slopes in the Three Gorges Reservoir Area. It is widely occurred in Badong, Wushan, Fengjie, Yunyang, Wanzhou, etc. The deformation and failure of Slide deposit slope usually follow the existing sliding surface (Figure 14). The deformation and failure process of the Slide deposit slope in Qingganhe Bridge, Zigui, Hubei Province had been modeled using discrete element method. The slide front edge has an elevation of 135 m which is 44 m higher than the river bed and the slide trailing edge has an elevation of 330∼335 m. The slope looks like a long stripe, and has a trend of NW∼SE. From bottom to top, the dip changes from 25∼30 to 15∼20, then from 25∼30 to 15∼20 again. The slope is bounded by a mountain to the east, and by Dachonggou to the west, covering an area of 67×104 m2 . The thickness of sliding plane ranges from 0 to 1.5 m, and Figure 13.

Anchored slope displace vector (step = 30000).

covered by wire mesh. In addition, wells for drainage were drilled. Based on these conditions, the aforementioned model was modified to account for the cables and wire mesh. The parameters are chosen as follows: the thickness of mesh is 0.1 m; the deformation modulus of cables is 100 GPa; Pretensioning is 3.5×105 Pa; Strength standard value is 10 GPa; Cement Mortar strength is 30 GPa. All other parameters are the same as those in the former model. The Mohr-Coulomb model is chosen; The left, right and bottom boundaries are all set to be fixed boundary. Results show that, after shotcrete, no obvious deformation can be found in the mudrock. The stress field changed considerably. As to the evolution line of tracking body displacement, the displacement reaches a contant. So, the controlling work is effective. (2) Analysis on deformation under shotcrete with Anchor Rod and Wire mesh The aforementioned Cable is substituted with Anchor Rod. the deformation modulus of anchor rod is 200 GPa; Strength standard value is 310 Mpa. All other parameters are the same as those in the former model. The Mohr-Coulomb model is chosen; The left, right and bottom boundaries are all set to be fixed boundary. Results also show that, after shotcrete, no obvious deformation can be found in the mudrock. The stress field changed considerably. As to the evolution line of tracking body displacement, the displacement reach a contant. So, the controlling work is effective. The outcome of this method is almost the same as the former method with cable. However, in this method, the deformation displacement is a little larger.

Figure 14. Slide deposit slope structure and failure (a) Natural slope (b) Manmade slope (c) failed slope.

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Table 1. Compared stability index with pre- and cutting of the Yanchanghou slope.

Before excavation After excavation Decrease rate (%)

a=0 C = 20kPa φ = 20◦

a=0 C = 20kPa φ = 18◦

a=0 C = 20kPa φ = 17◦

a = 0.1g C = 20kPa φ = 20◦

a = 0.1g C = 20kPa φ = 18◦

a = 0.1g C = 20kPa φ = 17◦

1.24 1.121 9.6

1.147 1.037 9.6

1.102 0.996 9.6

1.062 0.96 9.6

0.982 0.888 9.6

0.943 0.853 9.54

Figure 15. Rockfall deposit slope structure and failure (a) Natural slope (b) Manmade slope (c) failed slope.

mostly from 0.2 to 0.4 m. Its lihology is gravel soil, in a state of stiff-plastic. Discrete element method is used to model the deformation of slope after excavation without any protecting measures. Results show that the slope above the sliding plane would slide downward. As the failure of slope continues, sections with reduced stress and decreased stress introduced along the sliding plane. At the shear

outlets in the front edge, the shear displacement is the largest, and the stress approximates to 0. The modeling results show that, the natural slope is stable before excavation, and was cut in the middle part after excavation. The upper sliding body would slide downward along the sliding plane as a result of losing support, and the back soil would be pulled as a result of the towing force. The secondary sliding happens at an elevation of 360 m. The landslide behind the Wushan Cigarette Factory had been used as an example to study the failure of deposit slope caused by excavation in the front edge. The deposit slope is located at the slope between Pinghu Road and Jixian Road, which has a trend of 115◦ and an average dip of 34◦ . Excavation in the front edge led to the development of a slope with height of 20 m, and the sliding plane was exposed along the bottom of the slope. Under the influence of gravity and infiltrated water, the slope would slide along the weak layer or sliding plane within the deposit slope, leading to secondary retrogressive landslide. By using the Fellenius method, before excavation, the safety coefficient is calculated to be 1.102∼1.240, and after excavation, the safety coefficient is calculated to be 0.996∼1.121, which has a decrease of 9.6% (Table 1). It can be concluded that for a slide deposit slope, excavation in the front edge would decrease the stability considerably, and the decrease rate in safety coefficient can be as high as 10%. Furthermore, before excavation, the dip of the front edge of landslide is usually low, which would stop the sliding of landslide. However, after excavation, the landslide force can increase for as high as several times. The deformation and failure of avalanche deposit slopes is significantly different from that of slide deposit slopes. The deformation and failure mainly happen at the front edge, and new sliding planes develop within the deposits, which has little to do with the boundary between deposits and the underlain bedrock (Figure 15). For example, many large scale landslides in the city of Wanzhou are caused by regression of avalanche deposits. However, no integrated sliding plane develops. As a result, large deformation and sliding can seldom take place.

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5

Figure 16. Karst breccia slope structure and failure (a) Natural slope (b) Manmade slope (c) failed slope.

This kind of slope is composed of karst breccia which was formed as a result of karstification and gravity on the layered limestone and marlstone of Middle Triassic badong Group. It is mainly distributed in such counties as Badong, Wushan and Fengjie. Due to the excavation for reconstruction, the stability of karst breccia reduced considerably and the slopes fails in several places (Figure 16). The configuration of karst breccia slopes is similar to that of avalanche deposit slopes. But the former has a much higher strength as a result of calcareous cementation or the process of ‘‘falling—cementation—dissolution—falling’’. However, due to karstification, the strength of such slope decrease significantly as a result of infiltration of water. Thus, such slopes fail easily. Similar to avalanche deposit slopes, new sliding planes usually develop within the deposits, which has little to do with the boundary between deposits and the underlain bedrock. According to investigations on the karst breccia slope at Erlang Temple in Wushan, the shallow layer consists of soil with rock agglomerations, whose size can be as large as 3 m × 4 m, and has a loose texture. After excavation, karst breccia in the form of blocks in different sizes is exposed. The rock mass is cemented with calcium and is compact. Caviies can be found in the rock mass. The dip of the human-cutting slope is 18–75, and the height is 35 m. The stability of the slope is extremely low, and collapses happened in several places, which is harmful to the project and the Xiaoshanxia Dock below the slope. It is found out that although the stability of karst breccia is low, its distribution is regular. In gullies, the deposits are thick. On the contrary, in ridges, the deposits are thin, and bedrock can be exposed easily. So slopes in ridges have a high stability. Anchor Net Supporting Technique had been utilized to protect this slope, and has been proved to be very effective (Figure 17).

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Figure 17.

Prevented section of Karst breccia slope.

KARST BRECCIA SLOPE

FRACTURED LAYER SLOPE

This type of slope, which is mainly composed of mid-thick layer of sandstones of the Upper Triassic Xujiahe Group and partly composed of layered mudrocks and shales, usually contains multiple sets of joints (Figure 18). As a result of gravity of the slope as well as human cutting, landslide failure happens in these slopes. They are widely occurred in the reservoir region, such as Zigui, Xingshan, Badong, Wushan, Fengjie, Wulong and so on. The stability of layered fractured-rock slope mainly depends on the strike of formation. For anti-dip formations, slope failure are controlled by the cutting of fractured joints.

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Figure 18. Fractured layer slope structure and failure (a) Natural slope (b) Manmade slope (c) failed slope.

Figure 19. A 9-storey building destroied and 79 died since Slope failure.

In the Three Gorges Reservoir area, the ‘‘5.1’’ landslide in Wulong, Chongqing is a disaster caused by failure of layered fractured-rock slope. The slope has a vertical height of 46.8 meters, a front edge width of 55.2 meters and a back edge width of 25–30 meters. On May 1, 2001, a landslide with a volume of about 16,000 cubic meters occurred, causing the burying of a nine-storey building and 79 deaths as well as 7 injuries (Figure 19).This slope, which is composed of sandstones interbedded with mudrock of the Upper Triassic Xujiahe Group, is a tangential slope with low dip angles. The strike of the formation is 130 ◦ ∠11◦ . As a result of the cutting by two sets of joints, one with a strike of N50◦ W/220◦ ∠88◦ and the other with a strike of N30◦ E/310◦ with a strike of 88◦ , the rock mass belongs to layered fractured rock mass. Separate blocks of different size were formed, the largest of which has a volume of 10 m3 . Since 1989, it suffered two times of serious excavation in the bottom. The first time was in 1989, when the 319 National Road (Xiamen - Chengdu) was to be built. A steep slope of about 20 m in height and an angle range from 60◦ to 80◦ , without additional support, was formed as a result of the excavation along the Wujiang River valley slopes. The second time was in 1997, when this area was levied by the government for construction. A channel space with a width of 160 m, a depth of 15.5 m, and a height of 46.6 m, and a back edge dip of 70–80◦ , was formed without effective support. In situ investigation showed that the rock mass in the area of nine-storey building is fractured and the stability of the back edge of the slope is poor. The supporting measures by pulp stone block wall’s is found to be not effective. There were no waterproofing anti-filter layers and draining holes. The Dahegou Slope in the new city of Fengjie is a typical fractured layered slope. The lower part of the formation is composed of purple interclated with Kelly calcarious shale of T2b2; the loose rock mass develops in the gray marly limestone interbeded with kelly shale; On the top, there is a layer of reddle caused by karstification. Three sides of the loose rock are empty, to the south is the Yangtze River, to the west is a big gully, to the east is another gully. The terrain conditions and the T2b2 red soft-rock at the bottom provided good conditions for the development of loose rock. The loose belt lies 70 m inner of the Yangtze River bank, and has a width of 60 m and a visible thickness of 24 m. The loose rock displacement is along a set of nearly vertical NW trending joints. The joints disappears when it come cross shale. In the west of the loose belt is a big discontinuity with an aperture of 0.8 m. A zone filled with fractures occurs in the east, whose aperture is 0.3 m. The blocks rotated and were overhead, causing the combinational contact changes into point contact, and the rock becomes even steeper. The loose also happens along a set of fractures parallel

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to the gully, which can be seen on the Dahegou Slope and the top of the ridge. In the process of loosening, karstification had played a very significant role. In the cracks, there are ditches, troughs, and small stalactites. After the excavation of the slope, a slope of about 35 m formed, and the slope dip is greater than 75◦ . The rock mass of the slope is thick and fractured, with a poor stability. Because the slope belongs to cutting slope, large scale failure will not happen, and only small scale collapse occur in the shallow part.

7

OTHER TYPES SLOPE

According to the occurrence or rock material composition of formations, there are three other types of slopes in the Three Gorges reservoir area. 1. Reverse Rock Slope: the control interface of slope deformation andfailure is the anti-trend fracture. The weathering and spalling of mudstones and the crashes of sandstone blocks are the main forms of slope deformation and failure. In general, unless large discontinuities (joint, fracture) trending off the slope are developed, large-scale landslides seldom occur (Figure 20).The stability of the reverse slope depends on the overall strength of the rock and characteristics of discontinuities. In order to simulate the deformation and failure of this type of slope, two sets of discontinuities, one of which is parallel to the bedding plane and the other of which is parallel to the trend of slope, are considered. The spacing between bedding planes is 2 m, the spacing between fractures is also 2 m, the length of fracture is 8 m. All fractures distribute continuously. The results (Figure 21) indicate that after the excavation of slope, as a result of the large slope dip angle, the stress field of the slope body changes and the stress field of the shallow rock redistributes. The displacement of the shallow slope rotates toward the free-surface, and the displacement in the slope bottom is large while that of the slope top is small. In addition, the displacement curve on x and y direction indicate that the block displacement increases rapidly after the slope excavation, and continues. So, in the absence of supporting measures, the area influenced by rock deformation is large, especially in the area 20 meters inside the slope surface. The displacement increase gradually toward the slope top, and shearing takes place between adjacent layers. As the iterative step increases, the magnitude and range of rock deformation increases gradually, and the interlayer shearing are enhanced. The results also show that the slope deformation and failure is dependent on time, the deformation

Figure 20. Counter-layer slope structure and failure (a) Natural slope (b) Manmade slope (c) failed slope.

evolves from the slope bottom to the slope top, and from shallow to the deep. Therefore, this type of slope should be reinforced as soon as the excavation finished. 2. Inclined cutting-slope: Inclined cutting-slope refers to those slopes whose strike is bevel with the strike of formation. This kind of slopes are usually high and the terrain is usually steep. The deformation and failure of slope happen in the form of local collapse. If the rock body has a high content of weak rocks, the weathering and erosion of soft rock becomes the main form of deformation and failure. In the Three Gorges area, the mudstone of the Middle Triassic Badong group is a weak rock with high particle density, medium size, the middle porosity,

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no significant directional arrangement and middle roundness. As a result of physical weathering, the rock particles changes, which leads to reduction in integrity of rock mass and engineering properties, and then leads to the soil sliding along the surface of bedrock or formation of arc-shaped sliding surface (Figure 22). 3. Residual soil slope: the layer thickness of residual slope is usually less than 10 m, and the structure is usually in the form of superposition of rock and soil. The ‘‘Wushan loess’’, which is widely occurred in Wushan and was formed by transport of water flow, mixed with other weathering residuals and formed brown silty clay, which is high in Fe, Mn, low in Ca, high in montmorillonite (as high as 12.75 %), and high specific surface area. All of these factors led to a special type of disasterous soil. In the slopes in the new city of Wushan and along the reservoir, ‘‘Wushan loess’’ can easily lead to landslides. Figure 21. 50000).

Conter-layer slope displace vector (step =

8

SUMMARY

In this paper, the classification of slope structure in the Three Gorges Reservoir Region is proposed, and the deformation and failure processes of representative slopes before and after excavation are studied. In the context of resettlement project, the measures for protection of slopes are introduced. ACKNOWLEDGEMENT The author wishes to express his gratitude to the cooperation and guidance fromSong Yuansheng, Duan Zhide, Liu Di, Luo Yuanhua, Zhou Wei, Liu Daosheng, Feng Jinghua, Sun Kaiwu and Zhou Yanping. The author also wants to thank Qu Yongxing, who did the laboratory measurements and analysis. REFERENCES

Figure 22. Remaining clay slope structure and failure (a) Natural slope (b) Manmade slope (c) failed slope.

Chen Jianping, Tang Huiming, Li Xuedong. Problems on rock slope anchoring. Earth Science, 2001, 26 (4): 357∼360. (In Chinese). Cui Zhengquan, LiNin. Slope Engineering—New advance in Theory and Practice. Beijing: Science and Tech Press of China. (In Chinese). Jin Renxiang, Yang Liangche, Ren Guangming, et al. Analysis of Deformation Feature of Bedding Slope in Three Gorges Reservoir Region [J]. Journal of Geological Hazards and Environment Preservation. 2003, 14 (22): 39∼43. (In Chinese). Miao Jichuan. Expansive soil engineering feature and effect on cutting slope. Engineering Survey, 1994, (4): 18∼22. (In Chinese).

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Sun Guangzhong. StructureMechanics of Rockmass. Beijing: Science Press, 1990. (In Chinese). Yin Yueping. Regulation of Design and Construction of landslide Prevention at the Three Gorges. Beijing: Geologic Press, 2001. (In Chinese). Yin Yuep ing. Major geologic hazard and prevention on the relocation site at the Three Gorges. Beijing: Geologic Press, 2004. (In Chinese).

Zhang Xuenian, Sheng Zhuping, Sun Guangzhong, et al. Studies of Bedding Bank Slopes in Three Gorges Reservoir Region [M]. Beijing: Seismological Press. 1993. (In Chinese).

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

Slope engineering in hydropower projects in China J.P. Zhou & G.F. Chen China Hydropower Engineering Consulting Group Co., Beijing, China

ABSTRACT: High steep slopes of hydropower projects in China are characterized with complex geological conditions and large scale. Their stability plays an important role in decision-making, construction and operation of the project, and is one of key technical issues for most hydropower projects in China. On the basis of analysis of and a review on reconnaissance, design, construction and operation experiences of the slope engineering, authors summarized the effects of geological conditions on slope stability and slope treatment, clarified the principle for determining the safety criteria for slope stability, concluded the common failure patterns of slope and the commonly used numerical approaches for slope stability analysis, put forward practical requirements for slope excavation and slope treatment construction, pointed out that the integrity of reconnaissance, design, construction and management of slope was a preferential issue at present and would be the development trend in the future. It hopes that the viewpoints and suggestions presented in this paper can provide useful information for study, analysis, design and construction of slopes in China.

1

INTRODUCTION

There are many large slopes with great height and very complex geological and hydrological conditions in China’s water resources and hydropower projects. Slope stability problems greatly influence the feasibility study, investment decision-making, construction and safe operation of these projects. In the southwest and northwest regions, slope problems have become the most important issues for project development. Slope failures have caused serious damage to human lives and properties. On January 7, 1989, a catastrophic landslide happened during excavation of the left abutment of Manwan concrete dam when the elevation went down from 1017 m to 911 m. This case caused delay of the online schedule of a 1500 MW power plant for one year. Figure 1 shows the pictures before and after the failure. Figure 2 captures a landslide that took place on July 27, 2001at the outlet of the Diversion Tunnel No.2 of the Zipingpu Water Resources Project. The excavation of the slope started at the elevation 830 m until 779 m where the slip was triggered. The landslide material consisted of Quaternary deposit that overlays the bedrock with a thick clay seams. The Xiaowan Hydropower Project, currently under construction, includes a 292 m high arch dam that necessitates a large scale excavation at the left abutment in order to accommodate the arch and an access road. Long, continuous cracks developed in December, 2003 when the excavation went from elevation 1645 m to 1245 m. The slope movement rate

was measured 1 mm/day. Huge amount of reinforcing work was applied to rescue as will be described in Section 5. Slope stability is normally influenced by many factors such as the properties of the slope rock structure and its structural planes, surface water, ground water, undertaken actions, slope height, slope ratio and the pertinence of the adopted treatment measures, as well as the sequence, procedure, measures and details of slope construction. The above factors are difficult to be truly simulated by numerical approaches. The evaluation of the slope stability by an approximate model is only the reference for the slope design. Thus, the slope can not be designed without project comparisons, experiences summarization and engineer’s judgement. During construction of a slope, the geological mechanical model should be improved according to the actual geological conditions, the stability of the slope should be reevaluated by using measured data, and the design parameters and the construction scheme is optimized when necessary. And then, with the development of geological reconnaissance technologies, numerical analysis method, computation technique, internet techniques and information techniques, the design and construction of the slopes are being quickened up for the informational style, cooperating with multiple disciplines, three-dimensional visualization, dynamical design and construction, detailed computation and quick feedback make it possible for the integrity of reconnaissance, design, construction and managements for the slopes. The design and construction for slopes of hydropower projects has become

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(a) before failure

Figure 1.

(b) after failure

The landslide happened during the exaction of the left abutment of Manwan concrete dam.

Figure 2. The landslide happened during the exaction of outlet of the No. 2 diversion tunnel of the Zipingpu Project.

Figure 3. Cracks developed during excavation of the left abutment of Xiaowan arch dam.

the most active issue in the field of research on and application for intercross and cooperation with multiple disciplines, and modern computation techniques. A number of large-scale engineered slopes have been successfully built in China’s water resources and hydropower development, using the advanced

Figure 4.

The intake structure slope of Xiaolangdi Dam.

slope analysis and designed methods and modern construction technology. The Xiaolangdi Multi-purposes Water Resources project consists of a 154 m high embankment dam which requires that all the water discharge facilities must be arranged in the left abutment. This has created a 127 m high intake structure slope and an outlet slope that contains long persistent bedding planes with thick seams, refer to Figures 4 and 5. The slopes kept stable throughout their construction and operation periods thanks to a correct design and the reinforcing approaches that include prestressed cables, piles and drainage tunnels. The Three Gorges project has a two-lane and fivestage ship lock (Fig. 6) that creates a 174 m high rock slope in granites, which involves the removal of 22 × 106 m3 rocks. Serious stability concerns over its stability have been raised due to the relaxation of rock mass. This giant slope project has beenproved successful since its operation in 2003. In the areas of hydropower slope engineering techniques have made continual achievements, and have already established a complete set of technical

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Figure 5. The thick clay seams developed at the outlet structure slope of Xiaolangdi Dam. Figure 7. The left abutment of the dam, which involves a 310 m high vertical cut.

Figure 6. The ship lock slope of the Three-Gorges Project.

standard including reconnaissance, design, construction, instrumentation and management. The principles for design and construction of the hydropower slopes are safe, applicable, economic reasonable, advance in techniques, assuring the quality. By summarizing the experiences in executing the above principles, the following basic considerations during design and construction of hydropower slopes are, adopting comprehensive reconnaissance techniques, deepening the recognition of the slope geological conditions, constituting and improving the geological mechanical model; analyzing the slope deformation and stability condition, and evaluating the safety and the key factors with the help of rock mass classification and slope engineering classification; selecting an appropriate slope type and strengthening measures by engineering comparisons and engineer’s judgments; adopting controlled dynamiting techniques for excavation, alleviating the damage for slope rock to the most extend, supporting and strengthening the slopes rock in time; making use of the excavating geological condition and the observation data, conducting real-time dynamical feedback analysis, optimizing the

Figure 8. The group of intake structure slopes built on limestone inter-beddings.

slope design parameters and construction scheme for better pertinence and adaptability. For illustration, let us review the 180 m high Hongjiadu concreted faced rockfill dam that consists of two high slopes at the left bank: (1) the left abutment of the dam, which involves a 310 m high vertical cut as shown in Figure 7; and (2) a group of intake structure slopes built on limestone inter-beddings dipping into the river as shown in Figure 8. Modern numerical methods include the finite element and threedimensional analysis. The large scale piles and cables helped the successful performance of these slopes. 2

ENGINEERING GEOLOGICAL CONDITIONS OF SLOPE

The complex slope projects have many geological elements such as rock properties, structural planes,

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weathering, unloading, geo-stress and ground water, topography and physiognomy, as well as density, porosity ratio, water content, permeability, strength and temperature of rock mass, these factors all may influence the slope stable stability. Since the slope treatment is generally dependent upon the geological information obtained, it is necessary to find out the engineering geological and hydrological condition, to obtain the physical and mechanical parameters of slope mass, structural planes and sliding planes, and to analyze and forecast the geological problems of slope by adopting geological mapping, reconnaissance techniques. In recent 20 years, the reconnaissance equipment has been improving and updating considerably, the precision and reliability of reconnaissance results are improved. However, because of the limited working time and money at the preparation stage of project, reconnaissance of the slope engineering is mainly earth surface mapping and investigation, reconnoitering measures such as investigating cave and drills are only used as verifying measures; physical and mechanical tests for slope mass are mainly preformed in laboratory. A few in situ tests are only performed fro specific structural planes or sliding planes. The analysis and evaluation results may be different from one to another due to engineer’s experiences. Presently, a code entitled ‘Geological Reconnaissance Technical Specifications for Slope of Hydropower and Water Conservancy Project DL 5337’ was issued in China to clarify the content, method, technical requirements and evaluation standards of engineering geological reconnaissance for slopes in water resources and hydropower projects. The reconnaissance tasks are to find out the geological conditions, analyze the engineering factors contributed to the stability of slope, evaluate the current stable status, forecast the deformation development trend, provide geological data for design and construction, participate in the design and research of slope treatment scheme and the check of the treatment effect. On the basis of slope rock mass quality grading, stable classification of slope engineering is an important method for slope stability evaluation. According to the geological and hydrological condition, and considering the formation cause, material composition, rock layer property, rock mass structure, stable state, failure mode and characteristics, the water resources and hydropower slopes are classified as stable slope, potentially stable slope, doformed slope, unstable slope and landslide. There are 6 kinds of slope failure mode, that is, falling, sliding, toppling, bulking, cracking and debris flowing, in which, the sliding is the mostly occured. According to the statistical results of the slope failure in water resources and hydropower projects in China, the number of slopes failured by sliding is more than 80% of the total failed slopes.

It is suggested by authors that the structure of slope rock mass is an extremly important governing factor that influences slope stability. For different type of rock structure, the stability changes greatly, especially the developing extent, properties and combination of structural planes, which basically determine the slope stability. During the stability analysis and treatment design, the objective of slope geological reconnaissance is to depict and simulate the distribution and the property of the structural planes. As for slope rock structures, the geometry boundary controlling the slope stability is the bottom sliding plane that inclines outward or lateral sliding plane, the connect ratio, inclination angle, plane mass content and its physical mechanical behavior is the key aspect, after finding out the inclination angle, with or without mud and the underground water, essential judgment can be made for the slope stability.

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SAFETY CRITERIA FOR SLOPE DESIGN

The safety of a structure is jointly decided by the safety of structure itself, its foundation and its surrounding slopes. The following principle, safety criteria and other specification for design, construction, operation and maintenance of structure are also important for the slope stability evaluation. In view of economical reasonableness, the safety of the foundation and the slopes should be basically consistent with the structure, and is suited with the structure classification. During the service period, hydraulic structure, its foundation and slopes must have enough safety and meet the functional requirements for normal service. According to Specifications of Classification and Design Safety Standard of Hydropower Projects DL5180-003, the safety classification is usually closely relative to the corresponding structure; if slope failure only affects the normal operation structure, but does no harm to the safety of structure and human life, the slope class can be lowered by one grade under the precondition that the rehabilitation is cheaper than the slope strengthening treatment; for the slopes near the dam in the upper reaches, if its stability state changes can be forecasted by observation data, and effective measures against the possible damage caused by the slope failure is assured, the slope class can be lowered by one grade. In fact, the safe class of the slope is jointly decided by the importance of the influenced structure, the damage to the structure and the severity of the possible sequent disaster caused by the slope failure. According to Specification DL5180-2003, the minimum anti-slide safety factor for slope of different safety classification is listed in Table 1.

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Table 1. The minimum anti-slide safety factor for slope design. Load Combination∗

Slope Class

Basic Case

Special Case I

Special Case II

Class 1 Class 2 Class 3

1.30∼1.25 1.25∼1.15 1.15∼1.05

1.20∼1.15 1.15∼1.05 1.10∼1.05

1.10∼1.05 1.05 1.00

∗ During

the stable analysis, different action combination or operation status should be differentiated. The listed safety factor is only suitable for the lower limit method of Limit Equilibrium Analysis Approach.

The safety factors listed in Table 1 have a range, for specific engineering design, the relation between slope and structures, slope scale, stable status, the certainty of calculating parameters and boundary condition should be accounted for and be determined after analysis. For slopes with high failure risk, or more uncertainty in stable analysis, large safety factor should be adopted, contrarily, small factors can be used. For the very important slopes or slopes with deformation limitation, the safety factor should be determined after risk analysis and verification, and are usually higher than that listed in Table 1. The stable status of slope is not constant, but continuously variable. As time goes by, the changes of undertaken actions, rock properties and the effects of strengthening measures all caused the change of slope rock stable status, and these changes generally go against the slope safety. Therefore, it is necessary to leave a proper safety margin in the reconnaissance, design and construction of slopes. Besides the safety factor analysis, appendix D of DL/T5337-2006 listed the reliability design method for slope stability, thereby, the effect of the uncertainty, variability of influencing factors on reliability of slope stability can be included.

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STABILITY ANALYSIS FOR SLOPE

The engineering geological analysis is usually qualitative analysis. Because more correct boundary condition and slope rock mechanical parameters are needed, thus, quantitative analysis is only conducted for the important slope, possibly unstable slopes and deformational slopes, and mainly focused on the treatment scheme which influences the operation safety. Only by finding out the slope rock structure, sliding model, basic parameters, boundary condition, adopting proper analysis model, can correctly analyze and judge the slope stable status. Due to the differences between the assumption of numerical model and actual

object, each method has its own limitation. Specification DL5108 and DL 5353 specify that, slope stable analysis should be developed according to the slope type and sliding mechanism , and through selecting proper numerical model, geotechnical parameters and numerical method; meanwhile, the lower limit methods of Rigid Body Limited Equilibrium method is the basic method for safety factor analysis of slope. For class 1 and class 2 slopes, more than one kind of methods should be adopted to analyze the stability and comprehensively evaluate its stability. The limit equilibrium methods include MorgensternPrice’s method, Chen-Morgenstern’s method, Spencer’s method, Swedish method, Janbu’s method, Simplified Bishop Method, Transfer of unbalanced push forces Method. The upper-bound limit analysis methods include EMU Method, Sarma’s method and Pan’s Method. The common numerical methods include finite element method, discrete Element method and block movement method. The limit equilibrium methods recommended by Code DL 5353 are Simplified Bishop’s Method, Morgenstern-Price’s method, Transfer of unbalanced push forces Method and EMU method. For many cases, a three-dimensional analysis plays unique and important roles. For instance, structural plane is terminated in limit range and lateral resistance of the separate body should be taken into account for stability analysis; contraction joints were placed in the gravity, and the dam foundation is continuous, and evaluate the stability of a certain dam segment along a deep structural plane or weak layer; the space incision of structural planes can not be reflected by two-dimensional procedures; when analyzing the slope stability of tunnel outlet, the assumption of infinite long excavation plane for 2D are much different from the truth. In addition, 3D method has advantages when solving the 3D resistant capacity and analyzing the abutment stability of arch dam. The common 3D methods are Rigid DEM, DDA, NMM, FLAC, but FLAC assumes that the medial is continuous. For numerical analysis based on complex model and complicated constitutive relationship, it is not necessary to go in for too fine calculation because of the uncertainty of numerical model and the difficulty of acquiring accurate mechanical parameters, and the detailed computation can not remedy the error caused by the uncertainty of calculation conditions. However, for numerical analysis, the complexity of boundary, media mechanical properties, structural figure and the actual process of construction and operation can be taken in account; the effect of different structural planes can be evaluated; the coupling effect of seepage and temperature can be considered; optimal design for reinforcing structures can be conducted with the help of comprehensive strengthening treatments.

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5

REQUIREMENTS FOR SLOPE CONSTRUCTION

The improper construction procedures may lead to the failure of a stable slope; even cause a severe construction accident. For the slope under critical stable condition, it is more necessary to adopt rational construction process, techniques and measurements, thus the construction safety can be assured. So, for slope design, not only scheme comparison, calculation and evaluation, but also the controlling standard and inspection criteria for adopted material, construction procedure, quality, operation and maintenance should be specified, the above requirements should accord with the special specification of the codes for reconnaissance, design, construction, operation and maintenance. Reduction of the effects of dynamite vibration on slope rock mass to the most extent is an important process to assure the slope stability. Positive effects of shock absorption can be achieved by optimizing the dynamite parameters, adopting the techniques of pre-crack dynamite, glaze dynamite, reduction of vibration dynamite, constituting rational safety criteria and controlling standard for dynamite vibration. Anchor bar, anchor cable, anchor piles, reinforcement grid and shot concrete are common measures for strengthening the rock slope, and are widely applied in hydropower projects. Selection of an appropriate supporting time is vitally important for anchor-shoring; in addition, the sequence of excavation, anchoring and supporting are different for slope with different stable capacity. For slopes on the verge of failure, a careful excavation or strict measures controlling dynamite, and strengthening measures should be applied before excavation. For slopes with poor geological conditions and local unstable areas, the strengthening measures should be applied in time in the process of excavation. For slope with good geological condition can stabilize itself for a certain long time, usually select the supporting chance with consideration of construction condition, construction schedule and slope rock stability. For high slope, usually excavated from top to bottom, and support along with excavation, the support should not be later than the excavation by more than two excavation steps. For slow slope with good geological conditions, can stabilize itself for a long period, no support or random support is enough, and the support chance can be selected more freely. The left abutment of 304 m high Jingping arch dam as shown in Figure 9, which is between Elevation 2110 m and Elevation 1580 m, has a slope ratio of 0.5:1∼0.3:1. Systematic pre-stressed anchors with 2000 kN have been installed to stabilize the left abutment slope. Figure 10 shows a drilling technique that used in the Xiaoman deposit slope, which enables the protection

Figure 9. An overview of the 540 m high left abutment slope of Jingping arch dam.

Figure 10. The group of intake structure slopes built on limestone inter-beddings.

of the bore holes in alluviums by a steel pipe. This technique allows more 1000 cables to be installed for rescuing the slope. The change of hydrogeology condition will influence the stability of slope. Rainfall, reservoir leakage, seepage around the abutment, water from production and life, leakage of water pool located at high elevation all can cause the change of seepage field of slope rock mass, and deteriorate the hydrology condition of the slope. Seepage control and drainage is one of important measures for slope treatment, sometime, even can not be replaced by other measures. For slopes in a poor stable state, shallow drains should be constructed in advance at the top of slope surface, even drainage galleries and drainage curtains should be constructed before slope construction. Slopes should not be excavated in rainy seasons, and stabilizing measures should better be done in one time. Temporary enclosing measures should be adopted for construction in rainy seasons. The drainage holes and

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Figure 11. Ground water surface of the Three—Gorge ship lock slope.

Figure 13.

The large scale piles in the Hongliadu Project.

During the slope treatment, many other new strengthening measures are developed, such as antislide cantilever piles, anti-slide anchoring cable piles, anti-shear tunnel, anchoring tunnel, many types of retaining wall and concrete sash beam. The above concrete strengthening structures have their own capability, and should be selected or combined according to the geological conditions and the construction condition of the actual slope. For weathered fragmentized slope rock mass, pre-stressed anchor cables are difficult to work, concrete structures such as anti-slide piles can be adopted. Figure 12 shows the cantilever piles used in the Xiaowan slide (refer to Fig. 3). Figure 13 shows the layout of the pile group, each having a size of 20 m × 5 m in the Hongjiadu Project (refer to Figure 8). When the rock of the slope toe is weak, and is prone to be weathered or eroded by water, retaining structure or anti-scouring protection should be adopted to assure the stability of the slope toe. Figure 12. The large scale piles that reinforce the intake structure slope.

drainage galleriess may cause the local increase of seepage gradient in the slope rock mass, which goes against the seepage stability of fragmentized slope rock mass, anti-filtration protection should be adopted at certain location of drainage holes and drainage galleries at this moment. Shallow and deep drainages, have been widely applied in hydropower practice in China, have been proved to be successful. Figure 11 shows the field measured ground water table of the Three Gorge ship lock slope that clearly shows the dry zone outside the 7 levels of drainage tunnels.

6

INFORMATIONAL DESIGN FOR SLOPE

With the advent of computers, slope engineering design has been greatly benefited from the information gathered from past experience and digital data. 3D geological visual digital model has been widely adopted in engineering geological condition evaluation. The information includes physical mechanical properties of slope media, construction disturbance, weathering and unloading effect, surface water and ground water, construction process. The results of the primary design are only applied for the basic judgment of the slope global stability, more actual geological information is needed to forecast and solve the problems met during

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(a)

Figure 14.

(b)

The large scale piles in the Hongliadu Project.

Table 2. The statistical table according to the slope property. Classification Number Total

Excavation slope 42

Reservoir slope 46 117

River bank slope 29

construction, and have to pay a special attention to the change of engineering geological conditions and mechanical parameters of the slope rock. It is very important to develop geological analysis and deformation observation during construction. Designers have to conduct slope deformation analysis and stability feedback analysis; evaluate the effects of excavation, support and strengthen treatment, forecast the trend of slope rock behavior, deformation, and stability for the slope. Based on the economical reasonable principle, a possible adjustment of design parameters and construction schemes is also determined by using expert information system, putting forward and executing the most appropriate treatment scheme. In the past several years, China’s water resources and hydropower workers had recorded 117 slopes with complete technical information, and published these documentations in the web as shown in Figure 14. These slopes can be divided into three kinds: excavation slopes, reservoir slopes and the river bank slopes. The number of each kind slope registered in Table 2. 7

projects is much complicated. The high steep slopes in hydropower projects are characterized with complex geological conditions, large scale and great influence on structures. These features are seldom encountered in slopes of other engineering practice. Many experiences and lessons have been learnt from the reconnaissance, design, construction, operation and maintenance of slopes of hydropower projects in China, and a careful study and summarization of these lessons are necessary. According to author’s working experience and knowledge learnt from involved hydropower projects, the following was made: • the geological conditions and stability evaluation methods for the slope are summarized; • the basic principle of safety criteria for the slope design is clarified; • the common slope failure and deformation model and the commonly used numerical approaches for stability analysis are presented; • the requirements for slope construction are put forward; • the importance of informational design for the slope is emphasized. It hopes that the viewpoints and suggestions presented in this paper can provide a useful reference for the study, analysis, design and construction of high slopes of hydropower projects in China.

REFERENCES

CONCLUSIONS

Compared with slopes of highway, railway, mine and civil engineering, the slope involved in hydropower

Chen, Z.Y., Wang, X.G., Yang, J., Jia, Z.X. & Wang, Y.J. 2005. Principle, method and program for rock slope stability analysis. Beijing: China Hydropower Press.

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Design Specification for Slope of Hydropower and Water Conservancy Project DL 5353. 2007. Beijing: China Electricity Power Press. Geological Reconnaissance Technical Specifications for Slope of Hydropower and Water Conservancy Project DL 5337. 2006. Beijing: China Electricity Power Press. Li, Z. 2004. Analysis and studies on project and major technical issues of super-high arch dams. Beijing: China Electricity Power Press.

Pan, J.Z. 1980. Anit-sliding Stability Analysis for Buildings and Sliding Slope Analysis. Beijing: China Hydropower Press. Specifications of Classification and Design Safety Standard of Hydropower Projects DL 5180. 2003. Beijing: China electricity power press. Zhou, W.Y. 2005. Numerical calculation method for rock mechanics. Beijing: China Electricity power Press. Zou, L.C. 2006. Theory and engineering practice of treatment to complex high slope. Beijing: China Hydropower Press.

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Landslides and Engineered Slopes – Chen et al. (eds) © 2008 Taylor & Francis Group, London, ISBN 978-0-415-41196-7

A thunder at the beginning of the 21st century – The giant Yigong Landslide Zhihua Wang China Aero Geophysical Survey and Remote Sensing Center for Land use and Resources-AGRS

ABSTRACT: On April 9, 2000, a huge landslide occurred in Yigong of Tibet, China, drawn world wide attention because it has revealed a large extent of geological process of gravitational erosion in a very short time and brought heavy economic loss and ecology disaster. Afterwards, many professional researchers have been there to investigate the landslide and over 40 technical papers published. In this paper, a method of RS+GIS combining with site survey was used to interpreate the landslide features and characteristics of each part of the Zhamunong Gorge where the landslide occurred. Based on the investigation results, the location, shape and dimensions of the pre-sliding mass were determined, and it is demonstrated that this landslide has flied through a 3700 m long narrow canyon at high speed. It has found much evidence of the high-speed sliding mass crashing the left slope of the gorge mouth, being disintegrated and forming high-speed debris-flow. The deposit structures of the debris flow and the size of each part of the deposit were obtained in this paper. Based on the previous study results, it is considered that the Yigong Landslide is a type of high-speed landslide-debris-flow. The motion model of this kind of landslide was established based on its movement characteristics. By comparison of the structure of landslide and pre-sliding mass with the dimension of the landslide-debris-flow deposit, it is calculated that the Yigong Landslide is 3150 meter high with a volume of 91.1 ×106 m3 , which is slightly different from the previous estimations. 1

INTRODUCTION

High speed landslide-debris-flow, which is compromised of collapse, landslide and debris flow, and moves at high speed with a huge dimension and potential energy, is a special mass erosion disaster carried terrible destroying power. The Yigong Landslide (as shown in Fig.1), a giant landslide occurred in Yigong district of Tibet China

Figure 1. An aerial view of Yigong Landslide.

on April 9, 2000, has attracted world wide attention because it has expressed the specific geological process of huge mass gravitational erosion and caused heavy economic loss and ecology disaster in large range and very short time. So many professional researchers went there for investigating the natural disaster phenomena. Since the Yigong Landslide was occurred, satellite remote sensing (RS) has been used to monitor the change of water level of the Barrier Lake caused by the Yigong Landslide, to estimate the flood discharge if the dam of the Barrier Lake fails, and to identify disaster type and range in the lowerstream areas. (Liu, 2000, Zhou Gangyan, 2000, Wan, 2000). A plane view of the Yigong Landslide and the Barrier lake formed are shown in Figure 2. It concludes by summarizing all site investigation observations that: a collapse was occurred initially at the head of the Zhamunong Gorge, the collapse body crashed and activated the clastic deposit on the bottom of the gorge, then a high speed landslide, blocking the Yigong River, with a volume of about 3.0 × 108 m3 was formed. Recently, more comprehensive study by using RS+GIS technique has been carried out and significantly different results from former common understanding on the Yigong Landslide have been obtained.

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5520m

Zh

am

Barrie

on

g

Go

rge

r Lak e

Yigong Tea Factory

un

2264m

Deposit Depositarea area 2340m

2165m

Figure 2. A plane view of the Yigong Landslide and the Barrier Lake formed.

2

The scenery of Yigong Landslide.

Figure 4.

The location of the Yigong Landslide.

BRIEF INTRODUCTION OF THE YIGONG LANDSLIDE

Since the Yigong Landslide was occurred far away the populated area, there are only few people who may really saw it, a person named Cheng Keshan, a staff of Yigong Tea factory, described this event as: ‘‘In the evening on April 9, 2000, it was fine raining. I was watching TV at home that is about 300 m from the front edge of the the Zhamunong Gorge and about 10 km from the head of the gorge, and felt the floor was shocked suddenly at 20:05, then I heard an oppressive sound. I walked to the door fens about 10 seconds later I heard another outburst sound, house and fence shook violently, I couldn’t even stand stable. Meanwhile, I saw a thick smoke rising from the Zhamunong Gorge higher than snow line, then it rushed out from the gorge. The bottom of the smoke was thick black, light in the middle, and light grey in the upper. The thick bottom smoke rushed rolling towards the Yigong River from the mouth of the Zhamunong Gorge and flied over the valley followed by terrible sound, it last for nearly one minute’’.

3

Figure 3.

LOCATION AND GEOLOGICAL CONDITIONS OF YIGONG LANDSLIDE

The Zhamunong Gorge (the center is in E 94◦ 58′ 03′′ , N 30◦ 12′ 11′′ ), where the Yigong Landslide was occurred, is located in the north bank of the lower reach of the Yigong River, about 59 km away the great Yaluzangbu River bend, as shown in Figure 3. In addition, the Zhamunong George is located in an area where two huge faults intersect. The Cainozoic era Jiang-gu-la Granite and Pre-Sinian Period Gangdisi Group (AnZgd) metamorphic rock distribute in the gorge, as shown in Figure 4.

4

METHODOLOGY

‘‘Digital Landslide Technique’’ adopted in this study include: taking multi-temporal and multi-resolution satellite data as the information resource; taking DEM as the geographical control information resource; obtaining base image for interpretation by image processing; taking Men-PC interact interpretation to obtain basic information on the Yigong Landslide and its developed environment, which is stored and managed by GIS; then the information and geo-principle of landslide were integrated by means of GIS space analysis for studying landslide features and the process of the high speed landslide-debris-flow. 4.1 Information resources 1987-12-12TM (1987TM here after), 1999-0923ETM, 2000-12-30 ETM, 2000-09-23 IKONOS are adopted as the RS information resources for understanding conditions before and after 2000-04-09 Yigong Landslide. Digital 1:100,000 topographic maps are used as the geographic coordinates control for before the Yigong Landslide. 107 on site points are measured by GPS (by Etrex sumit GPS meter, horizontal precision