Fundamentals of Durable Reinforced Concrete [24, 2 ed.] 9781032199054, 9781032199078, 9781003261414

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Fundamentals of Durable Reinforced Concrete This new edition sets out the fundamental aspects of concrete durability with an emphasis on sustainability and carbon neutrality through performance-based methodologies. Global approaches to managing durability are explained from both a prescriptive and performance viewpoint. Achieving a balance between the interactive factors influencing durability and sustainability is supported by an explanation of the physical and chemical phenomena at play, determination of key performance parameters by mathematical modelling and physical testing, and current guidance for good practice. New chapters and sections examine the holistic approach to durability and significant aspects of traditional and new cementitious systems. The full range of threats to durability are covered in this single volume, including reinforcement corrosion, carbonation, chloride ingress, freeze-thaw effects, sulfate attack, acid and seawater attack, alkali-aggregate reaction, cracking, abrasion, erosion, cavitation, and weathering. The book presents a framework for specification through internationally adopted codes and standards and summarises the background to probabilistic approaches to durability design, providing a state-of-the-art review of mathematical modelling of deterioration mechanisms along with current directions in test methods for performance-based specifications. Fundamentals of Durable Reinforced Concrete is an essential reference on concrete durability for specifiers and researchers, and it is also accessible to undergraduate students.

Mark G. Richardson is Professor Emeritus and formerly Head of the School of Civil Engineering at University College Dublin.

Modern Concrete Technology Series A series of books presenting the state of the art in concrete technology. Series Editor Geert De Schutter Department of Structural Engineering and Building Materials, Faculty of Engineering and Architecture, Ghent University, Ghent, Belgium

14. Diffusion of Chloride in Concrete E. Poulsen and L. Mejlbro 15. Binders for Durable and Sustainable Concrete P. C. Aїtcin 16. Fibre Reinforced Cementitious Composites – 2nd edition A. Bentur and S. Mindess 17. Sustainability of Concrete P. C. Aїtcin and S. Mindess 18. Concrete Surface Engineering B Bissonnette, L Courard and A Garbacz 19. Textile Reinforced Concrete A. Peled, A Bentur and B Mobasher 20. Durability of Concrete: Design and Construction M.G. Alexander, A. Bentur and S Mindess 21. Concrete Permeability and Durability Performance: From Theory to Field Applications Roberto J. Torrent, Rui D. Neves and Kei-ichi Imamoto 22. Shotcrete: Materials, Performance and Use M. Jolin and D.R. Morgan 23. Active Rheology Control of Cementitious Materials Geert De Schutter and Karel Lesage 24. Fundamentals of Durable Reinforced Concrete, Second Edition Mark G. Richardson For more information about this series, please visit: https://www.routledge. com/series-title/book-series/MCT

Fundamentals of Durable Reinforced Concrete Second Edition

Mark G. Richardson

Cover image: Mark G. Richardson Second edition published 2024 by CRC Press 4 Park Square, Milton Park, Abingdon, Oxon, OX14 4RN and by CRC Press 2385 NW Executive Center Drive, Suite 320, Boca Raton, FL 33431 © 2024 Mark G. Richardson First edition published by CRC Press 2002 CRC Press is an imprint of Informa UK Limited The right of Mark G. Richardson to be identified as author of this work has been asserted in accordance with sections 77 and 78 of the Copyright, Designs and Patents Act 1988. All rights reserved. No part of this book may be reprinted or reproduced or utilised in any form or by any electronic, mechanical, or other means, now known or hereafter invented, including photocopying and recording, or in any information storage or retrieval system, without permission in writing from the publishers. For permission to photocopy or use material electronically from this work, access www.copyright.com or contact the Copyright Clearance Center, Inc. (CCC), 222 Rosewood Drive, Danvers, MA 01923, 978-750-8400. For works that are not available on CCC please contact [email protected] Trademark notice: Product or corporate names may be trademarks or registered trademarks, and are used only for identification and explanation without intent to infringe. British Library Cataloguing-in-Publication Data A catalogue record for this book is available from the British Library Library of Congress Cataloging-in-Publication Data Names: Richardson, Mark G., author. Title: Fundamentals of durable reinforced concrete / Mark G. Richardson. Description: Second edition. | Abingdon, Oxon ; Boca Raton, FL : CRC Press, 2024. | Series: Modern concrete technology; vol 24 | Includes bibliographical references and index. Identifiers: LCCN 2023017723 | ISBN 9781032199054 (hardback) | ISBN 9781032199078 (paperback) | ISBN 9781003261414 (ebook) Subjects: LCSH: Reinforced concrete‐‐Deterioration. | Reinforced concrete construction. Classification: LCC TA445 .R475 2024 | DDC 620.1/37‐‐dc23/eng/20230417 LC record available at https://lccn.loc.gov/2023017723 ISBN: 978-1-032-19905-4 (hbk) ISBN: 978-1-032-19907-8 (pbk) ISBN: 978-1-003-26141-4 (ebk) DOI: 10.1201/9781003261414 Typeset in Sabon by MPS Limited, Dehradun

Dedicated to the memory of Tom McCormack (1949–2001)

Disclaimer

Reasonable efforts have been made to publish reliable data and information, but the author and publisher cannot assume responsibility for the validity of all materials or the consequences of their use. In particular, it is neccessary to use codes and standards that are valid in the place of use and current at the time of design, specification and construction.

Contents

Foreword Acknowledgements 1 Durable concrete: the why and the how

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Durable concrete 1.0.1 1 More on ‘why’ and the global impact 3 The big picture of ‘how’ 4 The need to put a timeline on ‘how long’ 11 Summary 12 References 13

2 The prescriptive approach

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Background to the prescriptive approach 15 Significance of permeability and the water/cement ratio 16 The “national durability grade” 18 Classifying the durability threat – first steps 20 Refining the classification of the durability threat 25 Application of the “deemed to satisfy” approach in practice 28 Global consensus and differences in exposure classification 29 Taking stock – is the prescriptive approach fit for purpose? 43 The next step in the prescriptive approach 46 Summary 47 References 48

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3 Durability design and the performance route to specification

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Durability design meets durability specification meets durability verification 51 Durability design – structural design comparison 55 The “design life” concept 60 Probabilistic approach to durability design of major infrastructure 62 Framing the durability specification, including the “ERC” option 70 Verifying that durability performance meets the specification 72 Probability-based durability design and performance specification in action 76 Summary 78 References 79

4 Permeability of traditional and innovative concretes

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Permeability and porosity 83 Overview of permeability significance in the context of concrete durability 84 Influence of water/cement ratio and curing regime 84 Transport processes 89 Modelling transport rates 90 Influence of cement type on permeability of concrete 92 Permeability of innovative concretes 100 Influence of recycled materials 102 Measurement of permeation properties 103 Summary 109 References 110

5 Corrosion of reinforcement in concrete Nature of corrosion damage 113 Electrochemical process 117 Polarisation curves, the “Evans Diagram” 123 Passivity 125 Corrosion mechanism in carbonated concrete 125 Corrosion mechanism in chloride-rich concrete 127

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Influences on corrosion activity 129 Influence of cracking 131 Modelling the rate of corrosion 132 Limit state function for propagation phase time to corrosion induced cracking 136 Monitoring corrosion activity 138 Summary 145 References 145

6 Carbonation

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Carbonation and corrosion 147 Chemistry of carbonation in cementitious systems 148 Primary factors influencing carbonation rate 150 Mathematical modelling of the rate of carbonation 154 Application of models to service life prediction 165 Test methods to detect the carbonation front 168 Standard test methods for carbonation resistance 170 Management of durability in the context of carbonation 175 Summary 179 References 180

7 Chloride ingress

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Chloride ingress and corrosion 185 Critical chloride level for corrosion 189 Primary factors influencing chloride ingress 192 Mathematical modelling of chloride ingress 197 Application of models to service life prediction 205 Detection and expression of chloride levels 209 Standard test methods for chloride resistance 212 Management of durability in the context of chloride ingress 220 Summary 226 References 227

8 Freeze-thaw effects Disruptive forces in frozen concrete 233 Manifestation of distress 234 Process of freezing in porous concrete 236

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Primary factors of influence 237 Developments in modelling freeze-thaw behaviour 242 Test methods for freeze-thaw resistance 246 Testing the freeze-thaw resistance of aggregates 247 Testing of fresh concrete for air content 249 Freeze-thaw resistance tests on hardened concrete 251 Management of the freeze-thaw durability threat 256 Summary 260 References 261

9 Chemical attack: Sulfates

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Introduction 265 Conventional ettringite form of sulfate attack 265 Thaumasite sulfate attack 268 Delayed ettringite formation 269 Factors influencing sulfate attack 270 Tests for sulfate resistance 274 Management of the durability threat from sulfates 275 Summary 282 References 282

10 Chemical attack: leaching, acid, and seawater attack

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Essential facts 285 Physico-chemical aspects 287 Factors influencing attack 290 Mathematical modelling of acid attack 293 Management of chemical attack 295 Specification by performance 297 Summary 297 References 298

11 Alkali-aggregate reaction Forms of reaction 300 Historical and geographical aspects 300 Manifestation of the problem 303 Mechanism of expansion and reaction 305 Primary factors influencing the reaction 307 Additional considerations influencing ASR 313 Modelling and service life prediction 317

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Test methods for detection of ASR and aggregate reactivity 317 Management of the risk of AAR 328 Summary 335 References 336

12 Cracking in reinforced concrete structures

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Context 342 Mechanism of cracking 343 Chronological aspects of cracking 346 Cracking and the design phase 349 Cracking during the construction phase 350 Cracking during the service phase 355 Cracking and corrosion of reinforcement 359 Management of the risk of cracking 361 Summary 365 References 365

13 Abrasion, erosion, and cavitation

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Durability of concrete surfaces 368 Factors of influence 369 Standard test methods 371 Management of abrasion resistance 373 References 375

14 Weathering and efflorescence

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Durability of exposed concrete finishes 377 Weathering 377 Efflorescence 379 References 381

Index

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Foreword

During the past four decades, I have observed researchers, specifiers, and practitioners emphasising the need to move from prescriptive specifications to performance-based specifications in order to ensure that the concrete infrastructure would satisfy the durability requirements. With the recent emphasis on sustainability and carbon neutrality (that is, low carbon infrastructure), specifiers are considering multiple options, but many questions arise. What is the current technological gap between research and practice? How is the current level of research providing convincing evidence that durable concrete infrastructure can still be constructed whilst satisfying the sustainability and low carbon agenda? Are these technologies being adequately dealt with in student education and research projects to ensure that the next generation of concrete technologists and engineers understand the complex nature of interactions between exposure, materials, and performance? In my observation, the journey of performance-based specification was initiated by a team led by the late Dr Tom Harrison on behalf of the British Cement Association. In the early 1990s, he edited a book on performance-based specifications and conducted a number of workshops to bring this message to practicing engineers and concrete technologists. However, even in the revision of British and European Specifications on concrete at that time, only a passing mention was made of performance-based specifications to ensure the durability of concrete infrastructure. So, the first edition of Fundamentals of Durable Reinforced Concrete was a welcome addition to the knowledge domain because it dealt with the topic of durability and performance in detail, and it highlighted the need to emphasise both mechanical properties and durability whilst designing concrete infrastructure. Due to the changes that took place in European specifications and approaches during the late 1990s and early 2000s, the previous edition of the book emphasised the European context, particularly EN206, in relation to durability specifications. For any newcomer to the topic of durability design, the book provided invaluable insight into the

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exposure classes, material-exposure interaction, and performance in service environments. In this revised edition, Professor Emeritus Richardson has incorporated global approaches to specifying and ensuring both the mechanical properties and the durability, whilst retaining the discussion on various mechanisms of deterioration of concrete infrastructure. In this respect, he has carried out an in-depth review of specifications across the globe and highlighted the key features of those standards, benchmarked with EN206. This comparison is useful to construction firms with a global presence, in order to satisfy their design requirements. Furthermore, Professor Emeritus Richardson has tried to address the industrial need for more information from the research community on the fundamental issues underlying each deterioration mechanism, so that the benefits of the new performance-based specification system can be fully harnessed. This global view of the state of the art helps to identify research needs in various regions of the world. A quick comparison of this revised edition with reports published by the American Concrete Institute Technical Committee 201 on Durability of Concrete suggests that those who want to understand the recommendations in ACI 201 documents would benefit tremendously from this revised edition. In this edition, Professor Emeritus Richardson acknowledges that specifying concrete for durability whilst ensuring the sustainability of concrete infrastructure requires a deeper understanding of the interactive factors influencing its performance. This requires academia to bring many practitioners up to speed on the fundamental principles underlying the mathematical models and test methods that underpin the framework. There has been much development in the tools to support this transition over the last 20 years, and this is presented for each mechanism in the new edition of the book. The new edition also allows early-stage researchers an overview of the current state of the art and identification of gaps in their knowledge on modelling and testing concrete infrastructure for determining the durability performance as well as the service life for different exposure regimes. It enables them to take a wider view of the durability and sustainability of concrete infrastructure, and helps them to contribute more effectively in filling in the gaps that still exist in the achievement of fully probabilistic durability designs. This edition is considerably longer than the previous one, because of the greater complexity of achieving multiple sustainable development goals as opposed to the safety and serviceability of RC structures being viewed in isolation from the wider impact of construction on our global village. It was once said “durable and sustainable concrete ─ buy one, get one free!”, but that does not mean that our previous approach to durability automatically translates to sustainable development. The second edition of the book helps us to push the boundaries of existing and emerging technologies in the

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transition from traditional cements to low carbon solutions without compromising our duty to society in respect of durable infrastructure. I have been a contributor to other books on this topic and I can see that both of the editions authored by Mark Richardson were inspired by his journey as an academic with close industrial partnerships. Therefore, the book is pitched at the right level for academic researchers as well as practitioners to get an in-depth knowledge of the complex interactions between materials, exposures, and the resulting performance in service environments. By doing this, he is asking readers to pay attention to both the durability performance and the carbon footprint of concrete infrastructure. He has succeeded in addressing the gap between advanced research and the latest concreting practices. Hence, this is a book that could form a basis for developing modules in undergraduate and continuing professional education, providing the fundamental knowledge of durability specifications and procedures that is required to ensure their compliance in practice. P. A. Muhammed Basheer, CBE PhD, DSc, FREng, FIAE, FICE, FIStructE, FICT, FACI, FIAAM, FIMMM, FRILEM, CEng Chair in Structural Engineering, University of Leeds, Leeds, UK

Acknowledgements

My thanks to Tony Moore, Senior Editor, for encouraging me to embark on the first edition following our discussion at the 25th Anniversary Conference of the Institute of Concrete Technology. This updated version was written in the 50th Anniversary year of the Institute and it is wonderful that the President, Professor P. A. Muhammed Basheer, took time out to write the Foreword to this edition. My writing has been inspired by several decades of collaboration with powerhouses of concrete technology development in academic, regulatory, and commercial partnership. In this regard, I especially acknowledge Colm Bannon, Brendan Lynch, John Newell, and Roger West for their professionalism, advice, and friendship on my journey of learning. The ongoing supportive environment of University College Dublin is acknowledged and I express my gratitude to Associate Professor Amanda Gibney, Head of the UCD School of Civil Engineering. My thanks to Mary Caulfield for assistance with the diagrams and to Aimee Wragg and Kari Budyk, who kept the project on track. Finally, I express my heartfelt thanks to my wife, Rosemary Flynn, and our children, Jo, Niamh, and Andrew, for their invaluable support during the preparation of both editions.

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Chapter 1

Durable concrete The why and the how

DURABLE CONCRETE 1.0.1

Durable ─ why? Firstly, because durable concrete equates to sustainable concrete. The material is a combination of stone, water, and the calcium and silica found in certain rock types. We owe it to ourselves to make the most of these natural resources by building maintenance-free structures with long shelf lives. Secondly, because durable concrete is highly economical. For this reason, concrete is the material of choice for large scale infrastructure projects. Keeping this infrastructure in operation 24/7 over many decades is essential. Unscheduled early maintenance interventions are costly and disruptive to millions of people, impacting on both the quality of life and national economic prosperity. Thirdly, because our clients and the public view concrete as a maintenance-free material that will last a lifetime. As professionals, we must fully meet those expectations to the greatest possible extent. This book outlines the science and engineering that informs current practice in meeting sustainability targets and client expectations.

Durable ─ how? The secret to concrete durability is to ‘get it right the first time’. That entails care in the design, detailing and specification of the material to meet the structural and environmental exposure stresses that it will be subjected to over its lifetime. Equally, quality control is required to monitor compliance with specifications during all phases of construction – material sourcing, mix procedure, transport of fresh concrete, placing, compaction and curing. Quality control is a team effort – designer, specifier, concrete producer, ready-mix truck driver, and especially, the concrete operatives on site. Prescribing the right mix and cover to reinforcement does not necessarily guarantee the achievement of the required level of concrete durability. Poorly cured concrete may have an unacceptably DOI: 10.1201/9781003261414-1

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high level of permeability in the cover zone, despite being made with an appropriately specified concrete of high quality. By better understanding the physio-chemical phenomena that can negatively impact the durability of this two-phase material, especially when it is reinforced, we can ensure that compliance rules make sense to us and get enacted effectively. Furthermore, on innovative structures, we can use our confidence in deep understanding of durability issues to further leverage the guidance and advice available internationally in standards and codes of practice to push the boundaries without compromising the level of acceptable risk. Durable concrete is a product of competence in design and specification in tandem with care in production and construction. This book is a toolbox for the designer and specifier, setting out the science underpinning the engineering and construction practice, also described, so that, together, we can optimise our use of this wonderful material.

Durable ─ for how long? Concrete has the capacity to last for centuries. The Pantheon in Rome is a testament to a concrete structure’s ability to continue in service for 1,900 years – and counting! Thus, today, the concept of contemplating the construction of a concrete structure that is consciously designed to last for a period equal to only 5% of the service life (to date) of the Pantheon may seem laughable. A client is likely to view a specification linked to a defined lifespan as incongruous, given that concrete seems indestructible. The reality, especially in urban environments, is that functional senility has led to replacement of a structure more often than the deterioration of its structural condition. However, unfortunate examples abound, with partial or full replacement of concrete structures due to the unforeseen high maintenance costs of inadequately specified or constructed concrete. As a society we are moving away from the “single use” culture, be it a coffee cup or a building structure. “Refurbish and reuse” is taking over from “demolish and replace”. This pushes concrete durability further up the priority list of designers’ concerns, but enhanced material resistance cannot come with the extravagant use of natural resources. Thus, an increasingly important tool in the designers’ armoury is the concept of a defined lifespan, over which the structure is required to maintain a given service level. This permits the use of a quantifiable notional design life on which to determine the required material resistance. In the “refurbish and reuse” scenario, the clock may be set back to zero on the follow-on project, and the existing concrete condition assessed accordingly. Given the increasing concern over the sustainable use of natural resources, the concrete specifier today must ensure that just the right amount of material is used – not too much, not too little. The key concept is “sufficiency”. This differs from the traditional mindset of “efficiency” as initial cost only. A more helpful approach is to consider “efficiency” in respect of the lifecycle

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cost. That requires our client to indicate a required or expected service life. We may then design an element of sufficient size and adequate material properties to withstand both structural and environmental stresses over a target service life. The end user should define the service life required. Guidance in Chinese Code GB 50010 (MOHURD 2015) is helpful in indicating a typical range. Temporary or industrial structures are considered to have a minimum life of 30 years. Residential buildings, large scale industrial complexes, low- to medium-scale public buildings and infrastructure, such as short to medium span bridges, are expected to serve for at least 50 years. The upper range – 100 years in China and 120 years in some other countries – is considered a minimum for high-cost and prestigious projects. The former would include tall buildings, long-span bridges, and major municipal infrastructure, while the latter would include monumental buildings of symbolic significance. The period that concrete must remain durable is therefore now an accepted quantifiable parameter that we can introduce into the specification process, under both a “deemed-to-satisfy prescriptive” and a “durability design and performance” approach. The mathematical models of durability design described in this book provide a bridge to this concept for practitioners and a springboard for early career researchers to further perfect the art.

MORE ON ‘WHY’ AND THE GLOBAL IMPACT The early decades of the 21st century brought ever-increasing evidence of our intertwined fortunes across the planet, not least through geographically interconnected extreme weather events and a global pandemic. Anticipating the increasing impact of the “global village”, the UN adopted 17 Sustainable Development Goals (SDGs) in 2015 to transform our world. This universal call to action encapsulates the urgency placed on us all to take a more global view of our interdependence on one another, as well as the role of the individual in contributing to sustainability. Those of us associated with the concrete industry can make a significant positive impact by optimising the sustainability of the built environment through durable concrete. Concrete practitioners – researchers, designers, specifiers, material technologists, producers, and construction teams – have a direct influence on many of the UN SDGs through their role in the quality of infrastructure. Consider for a moment the current and future role of concrete in 11 of the UN SDGs. Concrete is the most economic and viable material for the majority of the world’s infrastructure (SDG 9). Basic infrastructure can be a major contributor to economic development (SDG 8), assisting the future wellbeing and resilience of the millions struggling to break free of poverty (SDG 1). Concrete in agricultural infrastructure can aid productivity, helping countries and regions to meet their nutritional needs and end the hunger and malnutrition (SDG 2)

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of an estimated 820 million deprived individuals. Concrete forms the hospitals and clinics needed to bring a more equitable standard of healthcare (SDG 3) to those in greatest need. Educational facilities (SDG 4) need concrete products to provide the infrastructure for access to inclusive and quality education in developed and developing countries. Concrete forms an indispensable element of water supply schemes and wastewater treatment facilities (SDG 6) that bring clean water and adequate standards of sanitation to communities, large and small. Concrete plays its part in the upgrading and expansion of infrastructure to supply clean and efficient energy (SDG 7), especially in developing countries. Urban infrastructure, not least sustainable mass transport systems, requires concrete to meet the growing demands of city dwellers – more than half of the world’s population (SDG 11). High-rise buildings, once the preserve of steel, are now being executed in both concrete and steel/concrete composite construction. Concrete durability is key to ensuring that the built environment endures, providing a high return on investment of natural resources and serving as a major contributor to responsible consumption and production (SDG 12) across our “global village”. Concrete practice has a global effect on climate action (SDG 13). Climate activists have raised awareness of the enormous contribution that our industry can make to climate action. Small actions on each and every concrete building project are producing evermore sustainable solutions by harnessing research findings on durability, innovative cementitious systems, and the process of carbon sequestration. When we scale up these actions on a global scale we can make a huge difference together, given that concrete is the second-most-consumed commodity on earth. While concrete continues to play its crucial and understated role in the planet’s development, the challenge is for it to do this in a more sustainable manner without impairing the rate of progress in achieving many of the highlighted UN SDGs through unforeseen consequences. How do we determine an adequate specification for concrete, project by project, in a manner that will achieve satisfactory performance over the required life of the structure without squandering the earth’s resources and adding an unacceptable amount of carbon dioxide to the atmosphere? Caution that leads to overspecification wastes resources. Underspecification, on the other hand, leads to premature loss of serviceability, often with considerable disruption to third parties and, ultimately, a higher cost in financial and environmental terms. This emphasises the need for concrete that is produced to endure with minimal additional resource consumption during its service life, while being environmentally friendly in its production.

THE BIG PICTURE OF ‘HOW’ The achievement of a durable reinforced concrete structure is a multi-faceted process impacted by decisions made at all stages of a project, from

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Essential elements underpinning a durable reinforced concrete structure

Design and Detailing

Appropriate Specification Route

Designer’s mindset

Designed

Prescribed Competent application of codes

Detailing

Responsible Material Sourcing

Validating Production Quality

Cements, additives, aggregates, mixing water, admixtures and pigments

Certification of personnel and component materials

Standardized Prescribed Performancebased

Quality of Construction

Traing of personnel in placing, compaction and curing

Testing of concrete in the structure Testing of constituents Testing of fresh concrete

Figure 1.1 An overview of the multi-faceted process of achieving a durable concrete structure.

preliminary design to final construction. Five significant stages in the process are presented in Fig. 1.1, from which issues of particular significance are identified and discussed.

Designer’s mindset Decisions made at the preliminary design stage can have significant repercussions on the achievement of a durable concrete structure devoid of wasteful overspecification. The mindset of a designer at the preliminary design stage will prioritise decisions around the risk of collapse rather than the risk of serviceability failure. At first sight, that is a perfectly reasonable starting point. However, let’s reflect on changes in recent decades, which can legitimately lead us to question this mindset. Two changes are particularly relevant. The first is the current imperative to prioritise sustainability. Small refinements in decision making at the preliminary design stage on each and every building project will scale up to significant reductions in carbon footprint at a global level. Prioritising sustainability means prioritising durability. The second change is the ready and cost-effective availability of highly sophisticated structural analysis software for one-off building structures ─ software hitherto justifiable only to designers of prototypes from which many high-value production units would recoup the design costs, such as in the aircraft industry. It can no longer be argued that time spent on durability design is not justifiable in the face of readily available “cheap and cheerful” prescriptive durability guidance rules. That mindset has for

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too long relegated durability considerations to the last stage of the process – concrete specification – often leading to overspecification, where the cement content required for strength and for durability are not aligned. Just imagine the difference it would make globally if the specifier and producer partnered at the preliminary design stage to invest time in selecting a concrete that was “just sufficient” to meet strength and durability requirements over a defined service life, incorporating appropriate material safety factors. This possibility is within our grasp through performance-based specification, but we must work together to make the transition from prescription to performance whenever possible. This book provides a building block for this change in mindset.

Competent application of codes The limit state of serviceability (SLS) is the interface between the desirable state of a structure and a situation whereby it is no longer functioning as intended, short of structural collapse. Incrementally, some international codes are now recognising the “Durability Limit State” (DLS) as a subset of the SLS, and they guide accordingly. The DLS is heavily influenced by decisions made during the preliminary design phase. The probability of a designer delivering a durable concrete structure is greatly enhanced by informed judgements on material properties and cover made as early as possible in the design phase. These judgements are best informed by a fundamental understanding of the philosophy underpinning design codes, and by taking account of the crossover between design codes and material standards at every step of the process. Blind application of a design code and a material standard at different phases of design and construction often leads to compromised decisions that increase either the probability of premature loss of serviceability or the wasteful use of concrete through over-specification. Durability issues, such as the freedom to increase the cover to reinforcement or a more rigorous approach to limiting crack widths in cover concrete, may be ruled out by early and uninformed commitments to geometrical or budgetary constraints. Many practitioners are introduced to locally adopted mandatory design codes as “go-to manuals”, in the form of a set of best practice rules. Many countries have adopted codes produced in other countries that have gained international acceptance. Practitioners across the globe will typically be introduced to the state of the art in design through codes such as ACI 318 (ACI 2019), Chinese Standard GB 50010 (MOHURD 2015), Eurocode 2: EN1992-1-1 (CEN 2004) or Indian Standard IS456 (BIS 2000). Each of these codes represents the agreed distillation of complex theory and local experience to provide guidance that represents safe and efficient practice. However, practitioners today often find themselves tendering for large-scale projects outside their geographical region. In preparation for safely blending use of the locally adopted code with innovative design, there is much to be

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gained by looking to international state-of-the-art thinking reflected in publications by international bodies such as Fédération Internationale du Béton (fib) and the International Standards Organisation (ISO). The researcher and practitioner can thereby gain valuable insights into the bigger picture of design from publications such as the model international code MC2020, in preparation as a successor to the Model Code 2010 (Fédération Internationale du Béton 2013) or specialist topic standards such as that on service life, ISO 16204 (ISO 2012). By studying a range of international codes, the practitioner will become aware that these documents are the work of committees where each sentence will have been carefully crafted and carry significant weight. For example, in the context of durability, one cannot overemphasise the thought that went into the following extract from the introduction to the primary standard for concrete in Europe: “the concrete in the structure is deemed to satisfy the durability requirements … … … provided the appropriate exposure classes were selected” (CEN 2021b). Alas, the designer or specifier who only dips into a local standard or code seeking a particular clause is in danger of missing important collateral guidance. Anecdotal evidence demonstrates that many have failed to grasp the significance of the text in the introductory page of this European standard.

Detailing A moment spent in high-quality detailing of a structure can pay dividends in respect of durability. Perhaps the most important point is to consider how rainwater runs off a structure. Details that minimise exposure to water, or that minimise the differential rate of flow across a façade, can be significant positive factors.

Appropriate specification route A good specification must be clear, unambiguous, and not contradictory. It should also state the compliance criteria and the basis of assessment. Many specifications in the past failed to meet these requirements, usually leading to unsustainable waste through knock-on oversupply in the concrete composition, especially cement content. Current practice has been greatly assisted by the clear division of responsibility in the models of good practice that are now available. Basically, four approaches are available within the overarching prescriptive/performance duo: • • • •

designed concrete prescribed concrete standardised prescribed concrete performance-based specification.

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The performance-based specification provides the best opportunity for sustainability and durability through “sufficiency”. The “designed concrete” specification requires the specifier to provide defined basic requirements, primarily strength and the anticipated environmental exposure. It is the responsibility of the concrete producer, using their skill and knowledge of materials, to design and produce concrete that meets all the requirements. The information provided by the specifier about environmental exposure may not have been determined with the precision applied to selecting a strength class. This may force the producer to use a cement content and water/ cement ratio that results in a stronger concrete than required. Although it is a safe and serviceable solution, it is not a sustainable one. Sustainability demands that the designer invest thought into the exposure condition and strength requirements in tandem to produce an optimal solution. It is more likely that such a practice would occur in a durability design framework using performance-based specifications rather than “designed” concrete. The “prescribed concrete” specification route is less-prevalent than the “designed” approach, but is also likely to be overly conservative and unsustainable due to the division of responsibility. It is the approach of choice where a very precise uniformity of finish is required; for example, where exposed concrete finishes are a key design feature. The basis of the specification is solely the quality and proportions of the materials. The specifier is responsible for ensuring that the prescribed recipe will produce concrete that performs satisfactorily to strength and durability requirements. Compliance tests cannot, therefore, include strength testing. Responsibility for meeting strength and durability requirements rests solely with the specifier, who is likely to act with an abundance of caution. The solution is likely to be less sustainable than a performance-based specification. The “standardised prescribed concrete” approach is never used where durability requirements are onerous. These concretes are “off the shelf” low-strength mixes available from certified concrete producers in situations where the complexities of “designed” or “prescribed” specifications are unnecessary. The benefits of simplicity are counterbalanced by mix proportion recipes that are conservative, using more cement than the likely minimum required for the project. Nevertheless, there is a distinct and valid role for these standardised concretes in low-concrete-strength elements, outside the scope of this book, where quality control on site is nonexistant. The “performance-based” specification provides an exacting approach to meeting durability targets over the service life of a structure. Basically, the method involves the specifier using mathematical models of concrete durability to evaluate, in precise quantifiable terms, the performance required for a particular lifespan. An example of performance would be a target value of chloride diffusion resistance. It would then be up to the concrete producer to supply a concrete that could demonstrate compliance with the required chloride resistance in standard pre-qualification tests. The contractor would then follow on with acceptance testing of the finished

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product as cast. The performance-based concept holds the promise of a more sustainable concrete than heretofore: match the supplied capacity to the required demand with a specific probability of failure. More data from field experience is required to perfect this probabilistic approach, but great progress has been made.

Responsible material sourcing The durability of concrete is underpinned by the fact that all constituent components are covered by standards. In this brief review of the overall picture on durability, it is worth mentioning the potentially unforeseen cumulative effect of making decisions on concrete mix at different stages of the design and construction process. For example, certain aggregates, mixing water and admixtures, might individually be certified as meeting the requirements of a published standard, but used in combination might create a high chloride level, potentially injurious to prestressing wire. The need for caution, especially when making hurried decisions due to some problem or opportunity on-site, is required from beginning to end.

Validation of production quality Durability failures have occurred in situations where defective materials have found their way into the production process, notwithstanding certification of components. Aggregate properties in particular may vary as a quarry moves into a fresh seam. Variations in colour, texture, or working properties of sitebatched concrete may signal a change. Training and certification of personnel is an important step in nurturing a highly effective quality culture. Potential durability failures can be arrested at source by challenging the validity of material quality and invoking the testing of a component or product. The aviation industry has taught us that the culture of “challenge” can become an accepted and healthy norm ─ distinct from the culture of confrontation ─ in a professional environment. Examples exist where the cyclical nature of the construction industry has discouraged challenge in stressful and busy times, and significant durability failures have become apparent within a decade or two.

Quality of construction No matter how good the paperwork, it counts for nothing if the quality of execution of the concrete element falls below expectation. One cannot overstate the importance of competent workmanship in all stages of handling fresh concrete, and the arrangements for adequate curing. The quality of the communication link between the design and construction teams has been variable over the decades, resulting in a wide range of

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Fundamentals of Durable Reinforced Concrete

outcomes. These have spanned from universally acclaimed and awardwinning buildings and structures to those which have failed to deliver a fraction of their service life. Two significant developments are helping to improve the consistency of high-quality construction. These are the adoption of internationally recognised standards documentation regarding execution of concrete structures, and increased investment in the training of site personnel. International standards such as ISO 22966 (ISO 2009) and EN 13670 (CEN 2009c) are very helpful in scoping out the common requirements to be mutually appreciated at the design/construction interface. These documents provide a set of standardised technical requirements and also help to eliminate the need for contractors to fill in the gaps in situations where the designer failed to communicate information, as was an accepted part of the craft industry in the “master builder” era. Increased opportunities and investment in formal training of site personnel is to be warmly welcomed in respect of concrete durability. The concrete operative has long suffered from the dichotomy of being expected to learn a craft, while at the same time not being accorded the status of a recognised craftsperson. This has had two unfortunate outcomes. The first is that “learning by doing” on live building projects has often involved a transfer of knowledge (“how do we?”) without the parallel development of an understanding of the reasons (“why do we?”) underpinning good concrete practice. As can be imagined, this level of skill is limited to satisfactorily handling predictable situations. It does not empower the individual to make wise judgements when troubleshooting. A common example is the uncontrolled addition of water to concrete when consistence – what we previously referred to as ‘workability’ – is proving troublesome. Technological developments in concrete as a material increases the need to formally recognise concrete site practice as a craft, with attendant structured and certified training. Some durability failures are caused by the unfortunate coincidence of an error in the supply chain combining with inadvertent perpetuation of poor site practice. The level of investment in training at various technical levels is, thankfully, increasing. Examples include the Irish Concrete Society certified “Concrete Ticket”, valid for the five-year period following a course on current best practice in receiving, transporting, placing, and finishing concrete. Another example is the Concrete Field Testing Technician Course, jointly accredited and administered by the American Concrete Institute and the Institute of Concrete Technology. The course is being extended to other international partners, based on regionally relevant test standards. Concrete durability and sustainability are dependent on combining technical developments in the increasingly complex suite of available materials with greater emphasis on formal skillset development of those responsible for concrete on-site.

Durable Concrete

11

THE NEED TO PUT A TIMELINE ON ‘HOW LONG’ The limit state of serviceability is reached much more often than the limit state of collapse. What is a reasonable expectation for the service life of concrete? Concrete based on Joseph Aspdin’s Portland cement patent of 1824 was in use for well over a century before anyone questioned its “shelf life”. The physical similarity with rock engendered an assumption that concrete had an infinite service life, but this of course did not automatically extend to reinforced concrete. Even so, why now worry about designing to a notional life expectancy? The answer comes back to the concept of “sufficiency” and the best use of material and energy resources to achieve the client’s objectives for the building or infrastructure. Constructing each reinforced concrete element involves earth, water, fire, wood and metal: the five elements at the heart of Chinese philosophy regarding interactions in the universe. We play a “zerosum game” when we exploit the earth’s materials and energy to transform natural resources into the built environment. Designing to a client’s defined expectation of the lifespan of a structure – the “service life” – can better serve society’s investment in its future wellbeing and sustainability. Accepting that life is inherently uncertain, we need to introduce the concept of probability of failure when translating the “service life” into a design value – the “design working life”. The definition of “design working life” in Eurocode EN1990 is the: “assumed period for which a structure or part of it is to be used for its intended purpose with anticipated maintenance but without major repairs being necessary” (CEN 2010). This is a notional value determined by the designer as a function of the user-defined service life requirement, increased by a factor of safety based on the probability of failure. The serviceability requirement may be specified in a qualitative or quantitative way. A qualitative example would be the time to onset of spalling due to corrosion. A related quantitative example would be the time at which a stated percentage of a surface area is in a deteriorated condition due to spalling. Probability theory is implicit in the design process for structural resistance to load (dead, imposed, wind). Such an approach can also be usefully applied in design for material resistance to environmental load (carbonation, chloride ingress, etc.). A risk analysis can then be performed by balancing load (environmental factors causing deterioration) against resistance (material properties and any planned maintenance interventions). The permissible allowance for significant maintenance interventions in achieving the design life would depend on access, aesthetic, or economic considerations. For example, foundations are relatively inaccessible; occasional patch repairs are unacceptable in visual concrete; functional obsolescence might render it uneconomic to invest in maintaining structures beyond a certain age.

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Fundamentals of Durable Reinforced Concrete

These key concepts underpin developments in the future production of specifications for concrete that can provide a better level of reliability than that experienced heretofore. The risk analysis approach involves a change from prescriptive to performance-based specifications for the design team. The philosophy and operational aspects of durability design, however, are identical to the approach already adopted by the design team in determining structural resistance. The client will get what they pay for, but will have the assurance that the investment is matched to an expected service life. The further extension of this risk-analysis approach to embrace life cycle costing will be another step on the journey to greater sustainability, allowing the costs and benefits of any trade-offs to be assessed on an evidence-based analysis.

SUMMARY A quality product is one that meets predetermined expectations. In the case of concrete, the expectation of the client is that it will perform maintenance-free over its required service life; the expectation of society is that it will contribute to sustainable development. Concrete durability is central to meeting these expectations. We have failed to always meet these expectations in the past. Given the state of the art, the risk of concrete failing to perform satisfactorily over its service life should be low, yet in many cases, the risk is still unquantified. In the context of sustainability, durability of concrete is essential from both an economic and an environmental viewpoint. Unforeseen expenditure on the repair or replacement of concrete elements during the anticipated service life of a building reduces the competitiveness of the industry. A matter of increasing importance is that loss of concrete durability inevitably involves wasteful use of the earth’s resources in both materials and energy. Demolition due to unserviceability involves production of construction waste, which the planet has a dwindling capacity to absorb. Future generations may favour retaining existing durable reinforced concrete frames when refurbishing buildings through several functional incarnations. The ‘why?’ of durable concrete is not up for discussion. In the past, less was known about the physical and chemical processes that affect reinforced concrete in even moderate exposure conditions. This led to a misguided expectation that structures would remain serviceable until deemed redundant for other reasons, such as a change in use. However, even reasonable expectations of a maintenance-free multi-decade service life were too often not met. Armed with research findings, particularly over the last 50 years, combined with material developments and communication of best site practice, we can now deliver enhanced durability for a defined service life with less use of scarce resources. The ‘how?’ of durable and sustainable concrete is already available. We must educate ourselves to harness the

Durable Concrete

13

current knowledge now and encourage our most talented researchers to direct their talents and interest in advancing the sustainability of reinforced concrete even further to match the growth in society’s demand for more and better infrastructure in less-developed regions. Change to business practice has always been resisted to a lesser or greater extent in every industry. The “master builder” craft industry origins of the construction sector has been perpetuated through the centuries to good effect. However, some lateral thinking is now urgently required to save the planet. Given the scale of the concrete industry as a force for good in this regard, we must challenge certain mindsets. Firstly, we must forge greater partnership between the design, production, and construction teams at the earliest point possible in each project, to jointly seek reductions in wasteful overspecification of concrete. Evidence from satisfied clients’ experience with the “early contractor involvement” (ECI) process on projects prompts the desire to see this trickle down to other levels at the concept phase of design and construction. Secondly, we must educate the client to embrace the concept of “sufficiency”, translated into the notional concept of each concrete element having a predefined “shelf life”. The “for how long?” concept of defining a concrete structure’s service life must become an integral part of the design and specification process. The chapters that follow present the state of the art in prescriptive and performance-based specifications; our understanding of the physics and chemistry underpinning each durability threat; examples of approaches to the mathematical modelling of these threats; and test methods commonly used to research or specify performance. Empowered with this knowledge and understanding, we can advance the sustainability of today’s infrastructure, while students and early-career researchers begin their journey of enhancing the next generation of sustainable reinforced concrete structures that will deliver a better quality of life, in whatever global circumstances we may face.

REFERENCES ACI Committee 318. 2019. Building code requirements for structural concrete (ACI 318-19) and commentary on building code requirements for structural concrete (ACI 318R-19). Michigan: American Concrete Institute. BIS. 2000. IS 456: Plain and reinforced concrete - code of practice. New Delhi: Bureau of Indian Standards. CEN. 2004. EN 1992-1-1. Eurocode 2: Design of concrete structures - Part 1-1: General rules and rules for buildings. Brussels: Comité Européen de Normalisation. CEN. 2009c. EN 13670. Execution of concrete structures. Brussels: Comité Européen de Normalisation.

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CEN. 2010. EN 1990:2002+A1:2005(E). Eurocode – Basis of structural design, incorporating corrigenda of December 2008 and April 2010. Brussels: Comité Européen de Normalisation. CEN. 2021b. EN 206:2013 +A2:2021. Concrete – Specification, performance, production and conformity. Brussels: Comité Européen de Normalisation. Fédération Internationale du Béton. 2013. fib model code for concrete structures 2010. New Jersey: Ernst & Sohn. ISO. 2009. ISO 22966, Execution of concrete structures. Geneva: International Standards Organisation. ISO. 2012. ISO 16204, Durability - Service life design of concrete structures. Geneva: International Standards Organisation. MOHURD - Ministry of Housing and Urban-Rural Development. 2015. GB 50010-2010 updated 2015, Standard for design of concrete structures. Beijing: General Administration of Quality Supervision, Inspection and Quarantine of the People’s Republic of China (AQSIQ).

Chapter 2

The prescriptive approach

BACKGROUND TO THE PRESCRIPTIVE APPROACH During the latter decades of the twentieth century, a significant number of reinforced concrete structures exhibited signs of distress at a relatively early point after their entry into service. A typical case was the spalling of concrete from bridge soffits, exposing corroded reinforcement, as illustrated in Fig. 2.1. The scale of the problem and additional cost to the taxpayer was highlighted by many researchers (National Research Council 1997; Hobbs 2001). These studies indicated a cost burden in the order of $200/capita in the United States (US) and £10/capita in the United Kingdom (UK), just to maintain existing concrete infrastructure. Surprisingly, these instances were in contrast with older reinforced concrete structures built earlier in the century that were providing maintenance-free exemplary service. How did this change in pattern happen? It happened because, over the course of that century, we moved ahead of our technical understanding of the range of stresses that can be induced in concrete. There was a concentration on design for structural stresses to the virtual exclusion of consideration of environmental stresses. Technical strides in concrete technology at the time were concentrated on measures that progressively increased the maximum achievable compressive strength. Clients pushed for more load on a given span or spanning further for a given load. The industry responded with higher-strength concretes. This was also reflected in a corresponding emphasis on strength as the sole measure of quality. Thus, the developed world’s highway infrastructure at the end of the twentieth century did not represent sustainable development. Let’s explore what went insidiously wrong, and how the problem was rectified by an increasingly sophisticated set of prescriptive rules on the composition of concrete: the “deemed to satisfy” approach to specification. The prescriptive approach is falling somewhat out of favour for major projects but is likely to remain appropriate for the majority of day-to-day works for some time to come. It is also instructive to review the drivers of DOI: 10.1201/9781003261414-2

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Fundamentals of Durable Reinforced Concrete

Figure 2.1 Evidence of corrosion and spalling in a concrete bridge soffit.

successful change in recent decades, and how this may assist researchers and future policy makers in persuading a somewhat conservative industry to accelerate adoption of innovative, if complex, solutions to enhancing overall durability performance.

SIGNIFICANCE OF PERMEABILITY AND THE WATER/ CEMENT RATIO A key factor in durable concrete is impermeability of the cover zone material. The topic of concrete permeability is examined in detail in Chapter 4, but it will suffice for the moment to note that the higher the water/cement ratio the more permeable the concrete. Achievement of durable structural concrete was traditionally linked to cement content. During the 1950s the cement content required to achieve specified strength was generally high, resulting in an acceptably low water/ cement ratio and, consequently, an adequate level of impermeability. The coarseness of the cement grains led to ongoing hydration over a prolonged period, further reducing permeability through the ongoing growth of hydration products in the pore structure. Over time, the investment in grinding cement finer was justified by the improved performance of concrete in respect of early age and 28-day compressive strength. Perversely, the historical decline in durability performance referred to earlier lay partly in the development of these higher-strength cements. This insidious sequence is illustrated in Fig. 2.2.

The prescriptive approach

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Figure 2.2 Chronological relationship between mix parameters and durability.

The demand for higher-strength concretes was initially met by an increase in cement content, which collaterally enhanced durability. Later the standard production cements on the market became progressively stronger. However, as the strength of the cements themselves increased, a lower cement content could be used to attain a given 28-day concrete strength level. The cement contents in producers’ mix design recipe books were progressively decreased for each benchmark of 28-day concrete compressive strength. This enhanced the economic advantage of concrete, but unwittingly reduced its durability through increased water/cement ratios.

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Fundamentals of Durable Reinforced Concrete

Further developments in fresh concrete manufacture included improvements in the quality of plant and machinery, allowing refined tolerances in mix design and batching. Further efficiency was achieved through greater employment of statistical methods in quality control. The net result was that concrete strength grades and workability requirements were maintained with progressively lower cement contents. As cement contents decreased, the water/cement ratios correspondingly increased. This led to more permeable, and inherently less durable, concretes. Concern for the implications regarding durability were increasingly recognised in the 1970s, at a time when structural design codes were being updated. This opportunity was used to introduce empirical advice to practitioners in codes of practice on limiting water/cement ratios for various severities of exposure conditions. Advice was expressed through both minimum cement contents and maximum water/cement ratios, thus introducing the concept of a prescriptive approach to durable concrete. If the specifier adhered to the prescriptive guidance in the design code, the concrete was “deemed to satisfy” the code requirements in respect of durability. It was a somewhat “light touch” regulation, but it helped to raise awareness among practitioners of the link between water/cement ratio and the key durability parameter of permeability. A perceived weakness in the quality chain was that specifiers could prescribe mix parameters that were “deemed to satisfy” the durability criteria, but how would compliance be assured? There was no simple site test at the time for measuring the actual cement content, neither was there one for determining the water/cement ratio of fresh concrete. Regular testing of batches for workability/consistence could detect significant outliers relative to a norm, but the consistence test is not applicable as an absolute measure of quality. The test is used to highlight significant departures from a norm, through comparing relative values. Compliance with limiting values of concrete composition needed to take account of the fact that the most commonly specified and tested parameter was still the 28-day concrete compressive strength. This led to an exploration of the relationship between the compressive strength and both the cement content and water/cement ratio. Although the absolute relationship was highly dependent on the materials used in the mix, leading to significant regional variations, it was recognised that valid relationships could be established for population subsets, such as in a particular country. This led to the concept of the “national durability grade”.

THE “NATIONAL DURABILITY GRADE” The introduction of the “national durability grade” was proposed in the UK by Deacon and Dewar (1982). Their hypothesis was that durability could be assured for a specific exposure class if a nationally established

The prescriptive approach

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Figure 2.3 Durability grade concept, illustrating that concrete strength could be used as a proxy for cement content and water/cement ratio in the assessment of potential durability.

minimum 28-day concrete strength, defined in terms of a concrete “grade”, was adopted as the “deemed to satisfy” code value and was achieved in practice. The basis of the concept is illustrated in Fig. 2.3, which shows the relationship between durability and mix composition being routed through concrete strength, with strength used as a surrogate for the combination of cement content and water/cement ratio. Establishment of the benchmark strengths (“grades”) for local exposure conditions was dependent on surveying a large sample of mix compositions that were typical of local practice. In this way, it was argued, control of the minimum cement content and maximum water/cement ratio could be monitored through the 28-day concrete strength results. Achievement of a specified minimum benchmark – the “national durability grade” ─ would thereby provide assurance of mix composition compliance. The concept has been adopted by several countries, including some in Europe that have perpetuated its use at a national level following the transition to the larger regional framework of a European standard for concrete production and conformity. The advantage of the “national durability grade” concept was that it established a link between specification for durability and a measure of its potential attainment. The 28-day concrete compressive strength acts as a measure of strength compliance and as an indicator of compliance with the specified limits on cement and water content. Compressive strength is proportional to the total cement content; compressive strength is inversely proportional to total water content. Thus, the 28-day compressive strength should also allow assurance of meeting particular targets in respect of maximum water/cement ratio and minimum cement content, if the relationship between strength and each of the other two parameters is established. Such an approach had the advantage of being easily bolted onto national practice at the time, as the 28-day strength test was well-accepted by designers, specifiers and producers as the common currency of concrete quality measurement.

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Fundamentals of Durable Reinforced Concrete

Equally important, it encouraged greater investment in quality control by rewarding producers who reduced the variability of their input parameters ─ notably material sourcing, testing, batching accuracy, and trend monitoring. Producers with strong quality control could typically manage to achieve a given characteristic cube strength (i.e. mean strength minus 1.64 standard deviations) with the minimum specified cement content, and thus make careful and economical use of scarce natural resources. Producers with lower standards of quality control ─ and therefore higher values of standard deviation ─ needed to aim for higher mean strengths through excess cement content in order to achieve the same characteristic cube strength target. The relationship between concrete strength and its mix composition is a variable across localities because of the uniqueness of locally-sourced materials. Concrete is made from naturally-occurring materials, typically with aggregates sourced in the same region as the concrete producer’s plant. Also typically, although not exclusively, the cement is sourced in the same region. Therefore, it is unsurprising that the relationship between concrete strength and the cement content and, separately, with the water/cement ratio, is a variable from producer to producer within a region. Thus a survey of practice in a region at a particular time will not show a precise relationship. It will instead produce a spread of results from which an envelope of compliant practice by certified concrete producers may be determined, represented by the shaded areas in the graphs of Fig. 2.4. These show the results of a survey in Ireland analysed by West and Keating (1999). It may be seen that within the boundaries of a nationwide industry, the practice in respect of materials sourcing and quality control level can lead to significant variation. Such a survey does, however, allow for the establishment of a maximum water/cement ratio and a minimum cement content that would be statistically expected to yield a minimum characteristic compressive strength. Equally, one could deduce statistical values, such as those likely to yield a satisfactory probability of compliance at the 95% confidence level. Establishing these relationships in a region at a particular point in time was a cornerstone of implementing the “national durability grade” concept.

CLASSIFYING THE DURABILITY THREAT – FIRST STEPS The concept of applying limits to two important mix parameters was a significant step on the journey to assuring concrete durability. The next task was to provide the specifier with a suite of exposure conditions. The first iteration of a classification system was based on a simplified qualitative description of the exposure condition. This ranked the severity of the durability threat using a five-point scale from “mild” to “extreme”. Each of the five exposure classes covered a range of durability threats, some of which might be found in combination and others which were mutually

The prescriptive approach

21

Figure 2.4 Example of relationship between mix parameters and characteristic cube strengths, showing the wide range of values identified in a survey of national practice in one country.

exclusive. For example, a combination of regular exposure to seawater spray and many freeze/thaw cycles would be unusual, but both were covered in the “very severe” category, described in UK guidance as concrete exposed to seawater spray, de-icing salts (direct or indirect), corrosive fumes or severe freezing conditions whilst wet (BSI 1997a). Each of the five exposure classes had an associated “national durability grade” and minimum cover-to-concrete value, calibrated against previous experience in the place of use, to form a set of “deemed to satisfy” combinations. A trade-off between increased concrete cover and reduced concrete strength formed an integral part of the guidance. There was no explicit differentiation, however, between Portland cement concretes and those incorporating blended cements – any implied differentiation would have been accounted for by reference to their relative 28-day strengths. In the case

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Fundamentals of Durable Reinforced Concrete

of concrete structures in chloride environments, this was counterintuitive. As discussed in a later chapter, cements incorporating ground granulated blast furnace slag enhance chloride resistance but have lower strengths at 28 days than Portland cement concretes. This all-encompassing methodology relied entirely on the judicious exercise of engineering judgement. Guidance was provided on interpreting the applicable exposure classification, but it could not fully account for any subjective bias or inexperience of the user. The specifier’s choice of exposure class was somewhat subjective. Increasingly, the knowledge gained from surveys of deteriorated structures and research findings raised doubts about the adequacy of this all-encompassing exposure classification system. Particular concern was expressed regarding the adequacy of the advice to cope with Portland cement concretes in chloride-laden environments (Browne 1986; Bamforth 1994). However, the designer/specifier was presented with codes of concrete practice that were inconsistent in respect of the balance between advice on meeting structural stresses and advice for environmental stresses. The dilemma faced by the specifier may be considered by an analogy with structural design. Imagine that design calculations were based on qualitative descriptions of the demand side of the equation. A designer could not size a reinforced concrete beam with an acceptable level of structural resistance reliability if the information on span and loading was not quantified, but was qualitatively classified as “moderate” and “severe”, respectively. Such an imaginary and absurd scenario is presented in Fig. 2.5, contrasted with actual practice in the twentieth century in respect of choosing parameters for durability resistance. It is readily apparent that such a system would lead to an unknown factor of safety, if any, or the wasteful use of excessive material in cases of unwitting overdesign. Unfortunately, many extant structures were built during this period by designers who did not fully appreciate the significance of complex-durability micro-environments when selecting a “once size fits all” exposure classification for the entire project. Some of these structures are now deteriorating at an unacceptable rate. The format and imbalance of the early code guidance encouraged a routine whereby designer/specifiers considered strength and durability requirements as separate issues to be dealt with at the beginning (strength) and end (durability) of the design process. Designers typically locked in a concrete strength class based on structural considerations early in the process. Much later on in the design process, they specified limits on concrete composition for durability considerations based on the requirements in the codes of practice. This often led to an incompatible set of constraints from the viewpoint of economical production of concrete. Specifications with redundant parameters forced the producer to make inefficient use of materials in order to meet the most onerous requirement of the specification. Worse still, the specifications sometimes included mixes that satisfied the designer’s intention regarding strength, but not the durability requirement.

The prescriptive approach

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Figure 2.5 Analogy showing the dilemma posed by a “broad brush” qualitative description of design parameters.

The end use of the fresh concrete was not always known to the producer, who was therefore powerless to advise on situations which today would be covered by “fit for purpose” legal obligations. Because the prevailing prescriptive approach prior to the year 2000 was not entirely objective, it could not be used to optimise economic considerations in a statistical way because it did not take account of the required service life or the accepted level of reliability. The method was also cumbersome in adapting to the benefits of emerging technologies in the form of new materials and construction techniques. A code based solely on the all-encompassing approach could not easily harness the potential benefits of new technologies. The specifier, for example, would find it difficult to do a cost-benefit analysis on the use of controlled permeability formwork or corrosion inhibitors, because their use would not specifically change the prescriptive requirements of the code. Shortcomings remained, despite the advances in concrete durability through the “national durability grade” concept. Three aspects are worthy of critical comment. The first issue was the validity of an approach to controlling a complex set of independent or interacting concrete degradation mechanisms based on simplified qualitative exposure classifications. Reliance on simplified empirical relationships derived through previous satisfactory experience was total.

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Fundamentals of Durable Reinforced Concrete

The second issue was the specifier’s accuracy in identifying the appropriate exposure classification based on descriptive clauses. The nature of the descriptions, and the nomenclature, prompted a mental image of weather-related exposure conditions (mild, severe, etc.) rather than physio-chemical phenomena (permeability, carbonation, etc.), especially in the mind of a less-experienced designer consumed by concerns regarding the limit state of collapse, or immediate serviceability issues such as excessive deflection. Experienced concrete technologists will recognise that the description of the exposure classes in the superseded codes of the 1980s were sufficiently comprehensive to ensure good practice by designers predisposed to concrete durability concerns. However, many young designers learned about durability the hard way: through their mistakes in misinterpreting the seriousness of the concisely presented durability clauses in voluminous codes of practice overwhelmingly devoted to structural design. The likelihood of over- or underspecification was significant. The third issue is that national codes assumed a relationship between concrete grade and current industry practice in regard to minimum cement content and maximum water/cement ratio. This is related to the fact that the concrete durability grades quoted in codes were based on industry norms in a country as surveyed at a particular time. Changes to practice – such as preferred cement type – can significantly alter the relationships over time. It is impractical to expect codes of practice to rapidly adapt to subtle market changes. Although codes and standards lag technological developments by several years, these same codes provide some legal protection to designers and specifiers, who must take a conservative approach to risk. An example in recent decades was the slow pace of adopting new durability guidelines for cements other than Ordinary Portland Cement, CEM I, in chloride-rich environments. Thus a significant link in the chain changes with time, but this chain is not readily amenable to regular scrutiny and updating in published codes and standards. Recognition that the all-encompassing qualitative prescriptive approach had failed in too many cases in the past resulted in more emphasis being placed on specifying low water/cement ratios. There is a limit, however, on how far one can push down the water/cement ratio without compromising the amount of mixing water needed for placing and compaction, which directly influences the permeability of the finished product. This is especially the case in labour-intensive projects across developing countries, where high-technology construction plant may not be available. The learning curve, informed by an excess of failures in practice in the 1960s and 1970s ─ together with extensive research findings in the 1980s and 1990s ─ yielded an enhanced deterioration-specific prescriptive approach while maintaining the framework of the “deemed to satisfy” route. The prescriptive approach was comprehensively modified to move the spotlight from exposure condition to the durability-threatening phenomena themselves.

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REFINING THE CLASSIFICATION OF THE DURABILITY THREAT Achieving change to national standards is generally a slow and difficult journey. Happily, around this period of the 1970s and 1980s, other fora were gathering to pool knowledge and recommend a new European standard for concrete. This provided the impetus for a step change in enhanced guidance to concrete practitioners in respect of concrete durability. Starting from a clean sheet of paper is often an easier route to radical change, rather than negotiating major revision of an existing document. The context was the Construction Products Directive (European Union 1989) – the predecessor of the current Construction Products Regulations (European Union 2011), implemented since 2013. The Directive sought to eliminate or reduce barriers to trade by harmonising standards and uniformity in technical approvals. Initially, two committees of the Comité Européen de Normalisation (CEN) worked in parallel to produce separate documents for concrete specification and ready-mixed concrete practice respectively. Eventually, they merged to pursue agreement on a non-harmonised European concrete standard to which national bodies could append National Application Documents if local conditions required variations for particular circumstances in the place of use. It proved to be a protracted but highly successful endeavour. Significant milestones on the road to a European standard included publication of two proposed standards in the 1980s. These were prEN 206 ‘Concrete: Properties, production, processing and quality attestation’ and prEN 199 ‘Ready-mixed concrete – production and delivery’. The reason for two documents lay in significant previous work carried out in the 1970s by the European Ready Mixed Concrete Organisation (ERMCO) on a pan-European code of good practice for ready-mixed concrete. The code was seen as a natural precursor to a proposed European standard for the sector – hence prEN 199. In the event the two pre-standards failed to get the required support for publication as EN documents, a decision was taken to merge the work of the two CEN technical committees in pursuit of a single document. The resulting draft document achieved European pre-standard status in 1989, and further development in the 1990s brought the document to full European standard status in 2000, as EN 206-1 ‘Concrete – Part 1: Specification, performance, production and conformity’ (CEN 2000a). Development of pre-standard prEN 206 provided an opportunity to take a more rational approach to specification and design for durability than that in the various national standards it would ultimately replace. This standard for concrete production, taken together with advice on cover to reinforcement published in the design code EN1992, ‘Eurocode 2, Design of Concrete Structures’ (CEN 2004), provided an opportunity to take a more rational approach to specification and design for durability. Crucially, the standard required that the intended working life of the structure be

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Fundamentals of Durable Reinforced Concrete

considered, and it allowed for anticipated maintenance. Reflecting the state of the art, each relevant durability-threatening physical or chemical phenomenon was explicitly identified in the standard, with guidance to specifiers on each topic. The standard continues to evolve over time. A new version was adopted in 2013 and, to date, two amended versions (A1 and A2) have followed, resulting in the current version, EN206+A2 (CEN 2021b). The standard is accompanied by national annexes in the respective countries, and the annexes themselves evolve over time. A major contribution to European practice was the identification of 18 distinct exposure conditions in the context of concrete durability, and this has informed the redrafting of national standards globally over the last two decades, notwithstanding differences in regional climatic conditions. The European suite of exposure sub-classes currently serves as a benchmark from which regional variations in other parts of the world are often developed. The 18 European environments are grouped into six distinct durability threats: no risk (X0), carbonation (XC), chlorides (XD and XS), freeze/ thaw (XF), and chemical attack (XA). A separate exposure class for abrasion (XM) has also been included in some national complimentary codes, and is discussed in a later section of this chapter. The 18 sub-classes, designated by the alphanumeric codes presented in Fig. 2.6, readily lead the specifier to consider both the nature of the threat and its intensity. Further assistance is provided by the description of the threatening environments, as presented in Table 2.1. The potential shortcomings of the contemporaneous “deemed to satisfy” rules with broad-brush qualitative descriptions of exposure severity were addressed, in part, by this enhanced deterioration-specific prescriptive

Figure 2.6 Exposure classification system in European practice.

The prescriptive approach

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Table 2.1 Summary of classification by degradation phenomena and environments in European practice (adapted from information published in EN206 ( CEN 2021b)) Degradation phenomenon

Sub-class

Environment

No risk of corrosion or attack

X0

Corrosion induced by carbonation

XC1 XC2

Unreinforced concrete other than that in conditions at risk from freeze-thaw damage, abrasion and/or chemical attack Reinforced concrete in very dry conditions Permanently wet or in dry conditions Wet environment where concrete is rarely dry Moderate humidity Wetting and drying cycles Moderate humidity Wet chloride environment where concrete is rarely dry Wetting and drying cycles Exposure to airborne salt Permanently submerged in seawater Tidal, splash and spray zones Moderate water saturation but without de-icing agents Moderate water saturation with de-icing agents High water saturation but without de-icing agent High water saturation with de-icing agent or sea water Slightly aggressive (eg. groundwater SO42− < 600 ppm) Moderately aggressive Highly aggressive (eg. groundwater SO42− > 3000 ppm)

Corrosion induced by chlorides other than from seawater Corrosion induced by chlorides from seawater Freeze/thaw attack

XC3 XC4 XD1 XD2 XD3 XS1 XS2 XS3 XF1 XF2 XF3 XF4

Chemical attack

XA1 XA2 XA3

approach. The specifier is required to consider a range of specifically categorised mechanisms and subclasses of environmental conditions. The determination of limiting values of concrete composition is then based on the most onerous condition identified from consideration of each relevant durability threat. The advice on concrete composition in the standard, code or accompanying national annex is typically benchmarked to a service life of 50 years. The specifier can then relax or increase the limiting factors on mix composition parameters or cover to reinforcement if the required service life is lower or higher.

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Fundamentals of Durable Reinforced Concrete

APPLICATION OF THE “DEEMED TO SATISFY” APPROACH IN PRACTICE The actual values of the limiting compositions and conditions vary significantly from country to country, reflecting local materials and experience. The limiting values on concrete composition are set nationally and published in national standards or codes of practice. Designers and specifiers working across national boundaries are required to consult locally published guidance, irrespective of the transnational nature of some standards, to ensure that local experience and provisions valid in the place of use are fully harnessed in decision making. Even across Europe, for example, it did not prove possible to frame a uniform set of concrete composition values in EN 206 for use across all countries that have adopted this CEN code. This non-harmonised standard permits national standards bodies to publish provisions valid in the place of use, and the relevant limiting values through national application documents. The consequent divergence across 31 countries using EN206 has been surveyed at intervals since 2009, with the latest report, CEN/TR 15868 (CEN 2018a), highlighting the persistence of some surprising differences in provisions valid in the place of use. Helland (2013) suggests that this reflects the divergence of opinion among national standards bodies as to the accepted norm for a structure’s satisfactory service life in their countries, rather than any scientific arguments regarding material resistance. Although the “deemed to satisfy” approach is practiced universally, there is variation in the parameters used to specify and verify conformity with durability criteria. The requirements in any one country are drawn from subsets of the following parameters: • maximum water/cement ratio • minimum cement content, sometimes with limitations on permissible cements, variations in minimum content with use of blended cements in selected exposure classes or requirements to use particular binders, such as sulfate-resisting cement • minimum compressive strength at 28 days, generally calibrated to local concrete producers’ experience of characteristic strengths yielded by particular combinations of cement content and water/cement ratio in a ‘durability grade’ concept approach • minimum compressive strength at 56 days • minimum air content and/or freeze-thaw resisting aggregates • curing regime • crack width limitations • minimum cover, often including a trade-off with reduced concrete cover permitted with more onerous limiting values of concrete composition

The prescriptive approach

29

• waterproof covering • maximum water soluble chloride ion level • maximum chloride ion penetrability test level. The practitioner may find that all of these parameters are covered somewhere in the national standards applicable in any one country, other than the chloride ion penetrability test, which is not as common. What varies is the location of the guidance within a wealth of standards and codes related to construction. Sometimes excellent guidance on durability using these parameters is distributed across design, material, production, conformity, and execution codes in a country. Thus, the composite situation is satisfactory, but from the practitioner’s viewpoint the process of specifying durable concrete is hampered by the distribution of the information across the standards. This may also inadvertently lead to an implied distribution of responsibility. How many early career structural designers find themselves cast into the role of specifier for concrete while unaware of the existence of codes on the execution of structures which might, or might not, contain essential information on concrete practice in the relevant jurisdiction? From the client’s viewpoint, if “everyone is responsible for concrete durability” the process may get distributed to the extent that, in reality, no one takes responsibility for concrete durability on a particular project. Unfortunately, when things go wrong, many of the parties involved can demonstrate that their part of the process was in accordance with their contractual obligations, which is of little comfort to the client and diminishes the reputation of the concrete industry.

GLOBAL CONSENSUS AND DIFFERENCES IN EXPOSURE CLASSIFICATION Following the publication of European Standard EN206-1 at the beginning of the new millennium, there was broad consensus across the globe on the distinct physio-chemical phenomena to be considered when framing guidance in respect of concrete durability through the “deemed to satisfy” route. The universally agreed physio-chemical phenomena were corrosion due to carbonation or chloride ingress, scaling due to freeze-thaw action or salt crystallisation, chemical attack (sulfates, acids, or seawater), and abrasion. As may be seen, these are phenomena primarily influenced by aggressive agents in the external environment. Issues such as alkali-aggregate reaction and delayed ettringite formation are generally considered in codes and standards as separate issues from chemical attack in this context. This is because they relate more to internal factors in the concrete than the external environment. It is interesting to review practice across the globe on how the same physio-chemical degradation mechanisms are translated into environmental

30

Fundamentals of Durable Reinforced Concrete

Table 2.2 Summary of principal categories and environmental sub-categories in general Chinese practice for design of concrete structures ( MOHURD 2015) Category 1 2a

2b

3a

3b

4 5

Subcategory Indoor and dry Permanently wet Indoor, humid Outdoor, not in extreme cold or cold area Outdoor, not in extreme cold or cold area, exposed to non-erosive soil and water In extreme cold and cold area (below frozen line), exposed to non-erosive soil and water Wetting and drying cycles Frequent fluctuation of water level Outdoor, in extreme cold and cold area In extreme cold and cold area (above frozen line), exposed to non-erosive soil and water In extreme cold and cold with frequent fluctuation of water level Exposed to de-icing salt Exposed to sea breeze Saline soil Direct contact with de-icing salt Seashore Seawater Exposed to man-made or natural corrosive substance

or exposure categories by different national standards committees in drafting guidance to specifiers. The following tables, Tables 2.2 to 2.9, demonstrate little evidence of unanimity in terminology, nor even in the classification of the intensity of a universal chemical phenomenon such as sulfate attack. Despite the overall consensus of the degradation process at play, there are significant differences in classifying these threats in codes and standards across the globe. Nevertheless, this review of selected standards in China, the American continents, Africa, Asia, and Australia reveal sufficient common ground when exceptional circumstances are excluded ─ such as extremes of temperature or exposure to industrial scale acidic environments – that a future consensus around terminology might emerge. It is clear from a review of Tables 2.3 to 2.9 that a global standard of exposure condition terminology could be built around about 20 subclasses if there were the will to do so. Local conditions in respect of materials, microclimates, and experience in the place of use would, however, preclude any suggestion of global unity on guidance such as minimum cement contents.

Class

I

II

III

IV

Degradation phenomenon

Corrosion induced by carbonation

Freeze/thaw attack

Corrosion induced by chlorides from seawater

Corrosion induced by chlorides other than from seawater

III-C Medium III-D Serious III-E Very serious III-F Extremely serious IV-C Medium

II-E Very serious

II-D Serious

I-C Medium II-C Medium

I-B Mild

I-A Slight

Environmental action intensity

(Continued)

Indoor and dry with relative humidity 60% Outdoor, sheltered from wetting and drying Outdoor, exposed to wetting and drying Moderately saturated vertical surfaces, highly saturated horizontal surfaces Moderately saturated vertical surfaces in combination with de-icing salts, highly saturated horizontal surfaces in combination with de-icing salts Highly saturated horizontal surfaces in combination with de-icing salts Immersion in seawater Light airborne salt Heavy airborne salt fog Tidal, splash and spray zones, hot climate Light de-icing salt spray Immersion in chloride-laden groundwater Groundwater with low chloride level combined with wetting and drying cycles

Example of environment

Table 2.3 Summary of exposure classes and intensity of environments in Chinese practice (adapted from information published in MOHURD 2019b)

The prescriptive approach 31

Chemical attack

Degradation phenomenon

V

Class

V-E Very serious

V-D Serious

V-C Medium

IV-E Very serious

IV-D Serious

Environmental action intensity

Moderate de-icing salt spray Groundwater with medium chloride level combined with wetting and drying cycles Direct contact with deicing salts Heavy de-icing salt spray Groundwater with high chloride level combined with wetting and drying cycles Moderately aggressive chemicals in water, soil or atmospheric pollution Severely aggressive chemicals in water, soil or atmospheric pollution Very severely aggressive chemicals in water, soil or atmospheric pollution

Example of environment

Table 2.3 (Continued) Summary of exposure classes and intensity of environments in Chinese practice (adapted from information published in MOHURD 2019b)

32 Fundamentals of Durable Reinforced Concrete

The prescriptive approach

33

Table 2.4 Summary of classification by degradation phenomena and environments in Indian concrete practice (adapted from information in BIS 2000) Degradation phenomenon

Class

No risk of corrosion or attack

Mild

Corrosion induced by carbonation

Moderate

Severe Corrosion induced by chlorides from seawater

Severe

Freeze/thaw attack

Very severe Extreme Severe

Chemical attack

Very severe Very severe Extreme

Environmental exposure Structures located away from the coast and aggressive conditions, with surfaces protected from weather Surfaces sheltered from severe rain and saturated salt air; concrete in areas of condensation; concrete immersed in water other than seawater; concrete in contact with or buried under nonaggressive soils or groundwater; no risk of freeze/thaw while wet Exposed to severe rain; alternate wetting and drying Permanently submerged in seawater or located in a coastal environment Seawater spray Tidal zones Alternate wetting and drying; occasional freezing Severe freezing conditions Corrosive fumes; contact with aggressive subsoils or groundwater Direct contact with aggressive liquid or solid chemicals

Table 2.5 Summary of exposure classes and environments in Canadian practice (adapted from information published in CSA 2019a) Degradation phenomenon Chloride with or without freeze/thaw

Sub-class

Environment

C-XL

Reinforced concrete exposed to chlorides or other severe environments, with or without freeze/thaw, requiring a higher durability performance than in other sub-classes Reinforced concrete exposed to tidal, splash and seawater spray zones, salt water pools, with or without freeze/thaw Non-structurally reinforced concrete exposed to chlorides and freeze/thaw

C-1

C-2

(Continued)

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Fundamentals of Durable Reinforced Concrete

Table 2.5 (Continued) Summary of exposure classes and environments in Canadian practice (adapted from information published in CSA 2019a) Degradation phenomenon

Sub-class C-3 C-4

Freeze/thaw attack without chlorides

F-1 F-2

Chemical attack (aggressive liquids and gases)

A-XL

A-1

A-2

A-3

A-4

Chemical attack (sulfates)

No risk of corrosion due to chlorides or freeze/thaw damage

S-1 S-2 S-3 N N-CF

Environment Continuously submerged concrete exposed to chlorides but not freeze/thaw Non-structurally reinforced concrete exposed to chlorides but not freeze/thaw Saturated concrete subject to freeze/thaw climate Unsaturated concrete subject to freeze/thaw climate Reinforced concrete exposed to manure or silage gases or both, requiring a higher durability performance than in other sub-classes Severe exposure to aggressive gases in agricultural environment or hydrogen sulfide vapour atmosphere in municipal or industrial effluent, with or without freeze/thaw Moderate to severe exposure to aggressive gases or liquids in agricultural environment, with or without freeze/thaw Moderate to severe exposure of continuously submerged concrete to aggressive gases or liquids in agricultural environment or continuously submerged concrete in municipal or industrial effluent, with or without freeze/thaw Non-structurally reinforced concrete with moderate exposure to aggressive gases or liquids in agricultural environment, without freeze/thaw Very severe exposure to sulfates Severe exposure to sulfates Moderate exposure to sulfates Interior elements or buried in nonaggressive soils Interior concrete floors with a steel trowel finish (Continued)

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Table 2.5 (Continued) Summary of exposure classes and environments in Canadian practice (adapted from information published in CSA 2019a) Degradation phenomenon

Sub-class

Residential concrete not subject to freeze/thaw and de-icing salts

R-1 R-2 R-3

Environment Foundations Walls, grade beams, etc. Interior slabs except garage floor

Table 2.6 Summary of exposure classes and environments in American practice (adapted from information published in ACI 2019) Degradation phenomenon

Sub-class

Environment

No risk of corrosion or attack

C0

Corrosion induced by carbonation Corrosion induced by chlorides

C1

Benign ─ dry or protected from moisture (but be aware of possible threat from chlorides in the materials used to make the concrete) Exposed to moisture but not to an external source of chlorides Exposed to moisture and an external source of chlorides from deicing chemicals, salt, brackish water, seawater or spray from these sources Benign ─ not exposed to freeze/thaw cycles Freeze/thaw with limited exposure to water Freeze/thaw with frequent exposure to water Freeze/thaw with frequent exposure to water and exposure to deicing chemicals Benign (eg. groundwater SO42− < 150 ppm) Mildly aggressive (eg. groundwater SO42− 150 – 1,500 ppm) Highly aggressive (eg. groundwater SO42− 1,500 – 10,000 ppm) Extremely aggressive (eg. groundwater SO42− > 10,000 ppm) Benign ─ dry in service Contact with water where low permeability is not required Contact with water where low permeability is required

C2

Freeze/thaw attack

F0 F1 F2 F3

Chemical attack (Sulfates)

S0 S1 S2 S3

Contact with water

W0 W1 W2

Beginning our review in the Peoples Republic of China, the national codes GB 50010 (MOHURD 2015) and GB/T 50476 (MOHURD 2019b) apply across a geographical area of almost 10 million square kilometres, more than twice that of the European Union, where even national

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Fundamentals of Durable Reinforced Concrete

Table 2.7 Summary of exposure classes and environments in Brazilian practice (adapted from information published in ABNT 2014) Degradation phenomenon

Class

Aggressiveness Risk of deterioration

Example of environment

No risk of corrosion or attack Corrosion induced by carbonation

I

Weak

Insignificant

Rural or submerged

II

Moderate

Low

Corrosion induced by chlorides

III

Strong

High

IV IV

Very strong Very strong

Very high Very high

Urban, can drop down one class if surface is always dry or protected from moisture Coastal, can dropdown one class if rain is infrequent Industrial, can drop a class below if surface is protected from moisture Sea spray Industrial with aggressive chemical atmosphere

Chemical attack

Table 2.8 Summary of exposure classes and environments in South African practice (adapted from information published in SABS 2014) Degradation phenomenon

Sub-class

No risk of corrosion or attack

Moderate (Mild)

Corrosion induced by carbonation

Moderate

Environment and examples of relevant surfaces Above ground ─ dry or protected from moisture. The “mild” class is a subset of the wider “moderate” class and covers situations where reinforced concrete of 20 MPa compressive strength would endure for the intended service life. Sheltered or submerged exposure conditions where surfaces are protected against wet and dry conditions caused by water, rain and sea water. This would include enclosed surfaces and those with a waterproof cover. The category envisions reinforced concrete of minimum 20 MPa compressive strength and cover to concrete typical of day-to-day practice may be used in a trade off with strength. (Continued)

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37

Table 2.8 (Continued) Summary of exposure classes and environments in South African practice (adapted from information published in SABS 2014) Degradation phenomenon

Sub-class

Contact with water

Severe

Corrosion induced by chlorides

Very severe (Saline)

Chemical attack (Water)

Very severe (Aggressive water)

Chemical attack (Soil)

Very severe (Aggressive soils)

Abrasive action of liquids

Extreme

Freeze/thaw attack

(No specific exposure class)

Environment and examples of relevant surfaces All exposed surfaces subjected to hard rain, contact with soil or permanently under running water and where alternatively wet and dry conditions may occur. Exposed to sea water or a saline atmosphere up to 30 km from the coast. The category envisions onerous requirements on both minimum compressive strength and a generous depth of cover. Exposed to river water polluted by industries. Similar to onerous requirements for concrete in saline conditions. Surfaces of cast insitu piles exposed to aggressive ground conditions. Similar to onerous requirements for concrete in saline conditions but with a much higher value of characteristic cover due to the uneven nature of the surface. Surfaces exposed to sea water or very aggressive waters in industrially polluted or marshy conditions. Similar to onerous requirements for concrete in saline conditions but with a slightly higher value of characteristic cover. Not listed as a separate exposure class because of the mild climate in South Africa. Nevertheless there are requirements for air entrainment where freeze/thaw conditions exist, such as exposed surfaces of cold liquid storage facilities of freezer stores.

“deemed to satisfy” guidance among the member states is quite divergent. The guidance to specifiers across the various microclimates of China, therefore, emphasises specific environmental conditions in the general code for structures (GB 50010) backed up by the European style of classification based on degradation phenomena in the standard for concrete durability (GB 50476). In the former, seven main categories

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Fundamentals of Durable Reinforced Concrete

Table 2.9 Summary of exposure classes and environments in Australian practice (adapted from information published in Standards Australia 2018) Degradation phenomenon

Sub-class

No risk of corrosion or attack

A1

Corrosion induced by carbonation

A1

A2

B1

Corrosion induced by chlorides other than from seawater

A2 B1 B2

Corrosion induced by chlorides from seawater

B1 B2

C1 C2 Chemical attack

A1

A2

Environment In contact with ground and protected by damp proof membrane; residential footings in non-aggressive soils Interior environment of fully enclosed residential buildings; Exterior surfaces of buildings inland, more than 50 km from the coast, in non-industrial setting and arid climate Interior environment of fully enclosed nonresidential buildings; Exterior surfaces of buildings inland, more than 50 km from the coast, in nonindustrial setting and temperate climate Interior environment, members subject to repeated wetting and drying in industrial buildings; Exterior surfaces of buildings inland, more than 50 km from the coast, in nonindustrial setting and tropical climate Members in contact with salt-rich soils of low electrical conductivity Members in contact with salt-rich soils of moderate electrical conductivity Members in contact with salt-rich soils of high electrical conductivity Exterior surfaces of buildings in regions from 1 km to 50 km from the coast Exterior surfaces of buildings in regions within 1 km of the shoreline or further inland where strong prevailing winds create a saline atmosphere. A higher level than B2 is advisable in such regions impacted by vigorous surf Surfaces in the spray zone, 1 m above the wave crest level Vertical surfaces in the tidal/splash zone and all horizontal soffits above seawater Members in low permeability, mildly aggressive soils (eg. groundwater SO42− < 1,000 ppm) In contact with ground of non-aggressive soils, for elements other than residential footings; Members in highly permeable, mildly aggressive soils; (Continued)

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Table 2.9 (Continued) Summary of exposure classes and environments in Australian practice (adapted from information published in Standards Australia 2018) Degradation phenomenon

Sub-class

B1

B2

C2 U

Environment Members in low permeability, moderately aggressive soils (eg. groundwater SO42− 1,000 – 3,000 ppm) Exterior surfaces of buildings inland, more than 50 km from the coast but within 3 km of an industrial setting yielding atmospheric pollutants; Members in highly permeable, moderately aggressive soils; Members in low permeability, highly aggressive soils (eg. groundwater SO42− 3,000 – 10,000 ppm); Surfaces exposed to freshwater which is neither soft nor running Surfaces permanently submerged in seawater; Members in highly permeable, highly aggressive soils; Members in low permeability, extremely aggressive soils (eg. groundwater SO42− > 10,000 ppm) Members in highly permeable, extremely aggressive soils Surfaces exposed to soft or running freshwater; Members in aggressive soils where the magnesium content is equal to or greater than 1 g/L; Other environmental situations not listed above

are denoted, within which a total of eighteen distinct exposure conditions and microclimates are differentiated. These are presented in Table 2.2. In the latter, the degradation phenomena were classified into five main categories as illustrated in Fig. 2.7. Sixteen subcategories are differentiated by the intensity of the environmental load that threatens the durability, coded from the mildest “A” to the most severe “F”, across the main categories. These are presented in Table 2.3. The intensity levels provide an indication of the relative severity of the threat, in that the “slight” (A) and “mild” (B) level codes apply only in the case of atmospheric carbonation, whereas the “extremely serious” (F) level code applies only to one of the marine environments. Li et al. (2008) caution that the intensity codes are relative within each of the five categories, independent of their use in other categories. That is to say that a code “C” in respect of the

40

Fundamentals of Durable Reinforced Concrete

I-A

III-C

Atmospheric Environment I-B

Marine chloride Environment III-D III-E

I-C

III-F

V-C

II-C

Freeze-thaw Environment II-D

II-E

Chloride (excluding cryohydrate) Environment IV-C IV-D IV-E Chemical Environment V-D

V-E

Figure 2.7 Exposure classification system in Chinese practice for concrete structure durability design, adapted from GB/T 50476 ( MOHURD 2019b).

most onerous case of carbonation does not equate to a code “C” in respect of the least severe marine chloride environment. The specifier in China is required to comply with cover depth and crack control criteria, as commonly found in other national and international codes and standards. The cover to reinforcement is quoted in the codes for a 50-year service life. The requirements for a 100-year service life are met by increasing the cover by a factor of 1.4. Additionally, there are specified curing requirements. Regarding another example in Asia, the classification of exposure in India is based on generic descriptions of intensity rather than specific degradation phenomena. The relevant standard, IS 456 (BIS 2000), was updated in the year 2000, using five classes: mild, moderate, severe, very severe, and extreme. This represented an increase in the number of classes. There have been five amendments since then, but without a move to change the form of classification or to introduce subclasses. This reflects the fact that many structures are located in regions of low risk to concrete durability if the quality of construction adheres to national standards. A correlation between degradation phenomena, class and environmental exposure conditions is presented in Table 2.4. Turning our attention to another large land mass, microclimates are also a factor in guidance on the vast North American continent. Although there is strong alignment between Canada and the US on the degradation mechanisms that drive concrete durability guidance, the categorisation of relevant exposure conditions differs significantly. This reflects differences in climatic extremes and national divergence in the sectoral importance of concrete to specific uses, with a particular focus in Canada on the agriculture industry. The emphasis in Canada, reflected in standard CSA A23.1 (CSA 2019a), is on corrosion due to chlorides, freeze-thaw action, and chemical attack due to naturally occurring sulfates or aggressive compounds in municipal infrastructure and industry, especially the agriculture industry. This is

The prescriptive approach Chloride Environment C-XL

C-1

C-2

C-3

Agricultural (manure and silage) Environment A-XL A-1 A-2 A-3

Freeze-thaw Environment F-1 F-2

C-4

Sulfate Environment A-4

S-1

Residential Environment R-1

R-2

41

S-2

S-3

Not exposed to chlorides or freeze/thaw N N-CF

R-3

Figure 2.8 Exposure classification system in Canadian practice, adapted from CSA A23.1 ( CSA 2019a).

categorised through six main classes yielding 20 subclasses, as indicated in Fig. 2.8 and Table 2.5. It may be observed that within each category the intensity of the environmental load decreases with increasing number in the series, unlike the sequence in other codes and standards internationally. The approach in the US, reflected in the building code ACI 318 (ACI 2019), is broadly similar to European practice, whereby specific durability loads are the drivers in defining four main categories, illustrated in Fig. 2.9. These categories yield 14 subclasses, differentiated by the severity of the environment, presented in Table 2.6. In practice, four of the categories are benign, denoted by the zero “0” suffix. Overall, it may be seen that the primary difference from practice elsewhere relates to a distinct classification (W) related to contact with water. This was introduced to highlight the dangers of alkali reactive aggregates in wet conditions – a topic generally treated separately from exposure classifications in other jurisdictions. Additionally, there are wide differences in the values of groundwater and soil sulfate levels that define European exposure classes (XA1, XA2, XA3) compared with American practice (S1, S2, S3). Brazil provides one example of South American practice. The guidance in standard NBR 6118 (ABNT 2014) provides limits on maximum water/ cement ratio, minimum compressive strength and minimum cover for a suite of four exposure classes. The related descriptions of the environment are quite broad. Therefore, selection of the appropriate exposure class from those outlined in Table 2.7 provides significant scope for the specifier to Water penetration W0

F0

W1

Freeze-thaw Environment F1 F2

Corrosion W2

F3

C0

S0

C1

C2

Chemical Environment (Sulfates) S1 S2

Figure 2.9 Exposure classification system in American practice.

S3

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Fundamentals of Durable Reinforced Concrete

exercise their judgement. This degree of freedom puts a significant onus on the specifier to inform themselves of the risks to durability posed by the environment. Turning to the African continent, the unified practice in the partners of the East African Community may be placed in contrast with that of South Africa to give an overview of the direction of travel in respect of the “deemed to satisfy” approach across a significant area of the continent. Standards harmonisation in East Africa is being pursued by representatives of the national standards bodies of Burundi, Kenya, Rwanda, South Sudan, Tanzania, and Uganda. Working under the auspices of the East African Standards (EAS) Committee of the East African Community (EUC) secretariat, guidance on the approach to concrete durability was published within the standard for concrete EAS 131-1 (East African Community Secretariat 2008). The six national bodies have published the standard prefixed by the national code (NB, KS, RS, etc.) in the form that applies to their country, similar to the situation in Europe, where CEN codes are prefixed by the national codes (ÖNORM, NBN, CSN, etc.). The scale of work in developing hundreds of harmonised standards and getting each of these adopted is daunting, and the pace of success to date has fallen behind schedule. Happily, the concrete standard EAS 131-1 is already in its second edition. The prescriptive approach to concrete durability mirrors that of European practice EN206, with the same suite of exposure classes. However, adoption of this significant change in specification for durability is proving difficult to apply in all projects. For example, contractors working to the Kenyan Standard Specifications for Road and Bridge Construction (Ministry of Transport and Communication 1986) will often find that the tender documents include clauses giving them scope to agree concrete mix designs with the engineer. Contractual requirements may specify strength, maximum aggregate size, water/cement ratio, and slump without any reference to the exposure conditions. This situation is also replicated in many other countries across the world, with state bodies such as highway agencies having in-house specifications that do not necessarily align with state-of-the-art guidance in updated core national standards on concrete durability design and production. In South Africa, the classification of conditions of exposure in standard SANS 10010-2 (SABS 2014) uses internationally familiar terms, but these are not readily interchangeable with other countries’ use of similar terminology. That is because the classification in South Africa combines exposure conditions with specific structural elements, such as insitu piles. The exposure conditions presented in Table 2.8 are presented in ascending rank order of requirements in respect of a combination of concrete strength and characteristic minimum cover. Interestingly, there is sole reliance on strength as an indicator of impermeability. There are no specific constraints on concrete composition but the designer and specifier is referred to a

The prescriptive approach

43

bibliography which includes state-of-the-art knowledge on sustainable and durable concrete compositions. Not least among these references is Fulton’s Concrete Technology, the 10th edition edited by Alexander (2021) now serving an international audience as faithfully as previous editions have done for South African practitioners over six decades. Another example of practice in the southern hemisphere is that of Australia, presented in standard AS 3600 (Standards Australia 2018). Some similarity is apparent with the South African approach, whereby the specifier is guided by a wealth of examples of exposure conditions combined with specific concrete elements, from which the relevant exposure class may be readily discerned. However, the Australian suite of examples is more extensive, with the intersection of five environments (contact with ground, interior, exterior, in water, in seawater) and six categories of increasing intensity of exposure (A1, A2, B1, B2, C1, C2) together with a special category (U), which captures any number of other environments. The vast expanse of Australia introduces a further layer of subdivision across three microclimates. These take account of exposure differences caused by significant variations in seasonal rainfall across the arid climate of the west and interior; the tropical conditions north of 20oS latitude, and the temperate conditions of the southwest and east. These combinations are summarised in Table 2.9. The standard specifies minimum early age and characteristic strengths of concrete for each of the intensities of exposure. Additionally, a continuous curing requirement of three days is specified for categories A1 and A2, with seven days required for B1, B2, C1, and C2. A trade-off is available whereby the minimum cover can be reduced with increased concrete strength.

TAKING STOCK – IS THE PRESCRIPTIVE APPROACH FIT FOR PURPOSE? Notwithstanding their inability to capture every nuanced environmental condition that a concrete member might be exposed to, the many subclasses of exposure classification in current codes and standards across the world represent a major advance on practice prior to the year 2000. Despite the somewhat conservative approach underlying the “deemed to satisfy” approach, national standards bodies throughout the world are to be congratulated on the dynamic approach they are demonstrating in frequently updating guidance. The topic of concrete durability is being heavily researched and it is gratifying to see the fruits of these labours are being harnessed by national authorities in collaboration with industry on a continual basis, rather than with the protracted step-change approach of the twentieth century. The time period between updates of standards is much shorter than was traditionally the case, and this helps to take account

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Fundamentals of Durable Reinforced Concrete

of the move away from Portland cement to blended cements as the norm rather than the exception. The range of exposure classifications presented in the tables of this chapter are testament to the great advances in our understanding of degradation mechanisms and the herculean efforts of those involved in standards drafting committees to translate this new knowledge into a practical form for the designer and specifier. We must recognise that construction is a craft industry whereby previous satisfactory experience in a place of use is still a powerful argument against change. However, the plethora of different exposure classes across the world are a double-edged sword. Yes, the deterioration-specific classes (XS3, III-D, C-2, B1, etc.) represent something better than that before (mild, moderate, severe, etc.). But on the other hand, the fact that the same chemical and physical phenomena that affect concrete across the world lead to the sort of divergence apparent in Tables 2.3 to 2.9 is an indictment of our failure to date to bring more certainty to the client that their concrete will be durable in every instance. The climatic conditions and precise chemical composition of common cements may vary across the planet, but our current understanding of the same chemistry and physics at play in these concretes is well known and does not respect international boundaries. Minimum cement content requirements in Europe, for example, vary significantly across the continent. Differences of up to 20% in minimum cement content for chloride environments are evident between national jurisdictions. One may ask whether this reflects the influence of microclimates, local materials, or the risk-averse nature of some national standards committees. Probably all of these, but a variation of 20% is certainly not based on science. Imagine the benefit for carbon reduction if we could narrow the gap to even 10% by revising down the highest values. The differences across Europe also reflect the lag between changes in practice by concrete producers and revision to guidance in standards. The availability of a greater range of cement types on national markets is being fully utilised by concrete producers to offer appropriate solutions to clients seeking to optimise the cost-strength-durability-carbon footprint matrix. It is therefore difficult for national standards bodies to capture average or typical practice across producers, from which to assess when revisions to “deemed to satisfy” guidance is warranted. This has exacerbated a situation that existed even when the range of cement types in use was low. The relationships between cement content, water/cement ratio, and characteristic strength are determined through an industry-wide survey of practice in a particular region at a particular time. The resulting relationships are critically dependent on the characteristics of local materials, most-common regional cement type, and industry-norm quality control standards. These characteristics change with time, as illustrated in Fig. 2.10, showing temporal changes in one region and also differences between neighbouring regions at the same time step. Many researchers are critical of the current prescriptive approach. Kulkarni (2009) laments the continued use in some countries of generic

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Figure 2.10 Variation over time and between neighbouring regions of the relationship between minimum cement content and concrete “grade” (characteristic strength) even when the readily available cement types were less prevalent than today.

classifications of environment by threat level (“mild”, “moderate”, etc.). He argues that specifiers’ awareness of risk is heightened if durability is presented in the form of degradation exposure classes and subclasses, as in EN206. Even with numerous classes and subclasses, Beushausen et al. (2021) point out that it leads specifiers to consider individual degradation mechanisms in isolation, whereas combinations can be more deleterious. For example, carbonation releases bound chlorides; chloride penetration rate is more pronounced in tandem with sulfate attack. Equally there are shortcomings in helping the specifier to recognise particular threats: the influence of temperature on the rate of chloride ingress is not sufficiently addressed globally; carbonation exposure classes fail to take account of the regular situations where elevated carbon dioxide levels arise in daily use of buildings; other than Australia, few countries take account of saline atmospheres some distance inland and caused by prevailing winds. Ali et al. (2021) argue that the prescriptive approach is detrimental to durability. There is some measure of truth in that statement. By presenting the designers, specifiers and clients with such a complex web of exposure classes, are we providing a false level of comfort that the concrete has a probability of failure in the service limit state that is as low as that in the limit state of collapse? Is it acceptable to base our specifications on prescriptive rules with origins in a craft industry when the tools are readily available to engineer scientifically based solutions? It may be argued that the prescriptive approach represents the engineering link between the precision of the research laboratory and the variables of the construction site, where conservative tolerances are added to control risk. An impressive engineering link it is, and a greatly improved one compared with decades ago. But the priority of the concrete industry now must be sustainable development ─ literally saving the planet. This demands that we use the best available technology in every step

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of the design and construction process. Currently, we know that the best available technology is the performance-based approach described in Chapter 3. The prescriptive approach has its place in the construction process, and further refinement will continue to see its effectiveness improve. However, the community of researchers, designers, specifiers, and building contractors needs to work together to reduce our reliance on it as the “go to” method in national codes and standards.

THE NEXT STEP IN THE PRESCRIPTIVE APPROACH Accepting that a global overnight move to the performance-based approach is both impractical and undesirable in an industry that best uses incremental change to innovate responsibly, there remains work to be done to improve the existing prescriptive approach in practice. Two examples of incremental change in Europe are the rationalisation of concrete durability guidance into one document and the bringing of abrasive resistance into the exposure class fold. A parallel development is the introduction of performance-related concretes to minimise risk of corrosion using a system that will overlap with the now-familiar exposure classification system in the prescriptive approach, through the “Exposure Resistance Class” described in Chapter 3. Taking the first example, one slightly awkward aspect of Europe’s approach to specifying concrete durability was the division of guidance across two documents. The designer following the prescriptive “deemed to satisfy” approach needed to consult the concrete materials standard (EN206) for information on the exposure classes and limits on the concrete composition and, separately, the concrete design code (EC2) for correlation with different structure types, limits on minimum cover to reinforcement, and maximum crack widths. The intention is to rationalise the guidance on exposure classes, structure types, limitations on concrete composition, and cover to reinforcement as part of the design code only. The other example relates to abrasion. A difficulty for standards committees worldwide has been to find a logical home for guidance on abrasion resistance. A starting point in the drafting of exposure classes was to limit the list to degradation phenomena that are directly related to the naturally occurring external environment, such as carbonation or chemical attack. Internally generated degradation mechanisms such as alkali silica reaction or delayed ettringite formation are thereby excluded from the list and dealt with in separate sections of the code or standard. But what of degradation phenomena that are directly related to the man-made external environment? Abrasion of concrete surfaces, for example by forklift traffic on industrial floors or flowing liquids, is neither exclusively a naturally occurring phenomenon nor an internally generated durability failure! However, there is emerging consensus in several countries that this mode of durability failure should be flagged in the overall exposure classification system with attendant guidance on concrete

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composition. The proposal in Europe is to reference it as Exposure Class XM, ‘wear resistance of concrete’, in the international standard, as has for many years been the practice in some national application documents, such as the German regional application document DIN 1045-2 since 2001, now in its third edition (Deutsches Institut fur Normung 2022), and the Czech complementary standard CSN P73 2404, also recently updated (Czech Standards Institute 2021). A bridge between the prescriptive and performance-based approaches is being pursued. This recognises the desirability of a “one stop shop” unifying concrete requirements in European practice, whereby a concrete strength determined at the preliminary design stage for use in structural design calculations will also satisfy the anticipated durability requirements. This durability design approach, referred to as the ‘Exposure Resistance Class’ (ERC), builds on the existing “deemed to satisfy” exposure class system of the prescriptive approach and has strong parallels with the “strength class” approach to durability that was used in the “durability grade” approach. The topic is discussed in more detail in Chapter 3.

SUMMARY The craft industry of construction continues to evolve best practice in respect of specifying durable concrete. Initial steps involved highlighting a small number of different exposure conditions ranging from “mild” to “very severe” allied to guidance on minimum cover to reinforcement. Allowance was made for reducing the minimum cover through a trade off against increased concrete strength. Durable concrete is impermeable concrete, and early reliance on strength was replaced by recognition that minimum cement content and maximum water/cement ratio were better guarantors of impermeability. This led to the tying together of guidance on concrete durability through exposure classes and limiting values on concrete composition, in respect of minimum cement content and maximum water/cement ratio, sometimes calibrated against national concrete “durability grades”. Since the turn of the millennium this “deemed to satisfy” prescriptive approach has evolved through a more finely tuned and greatly increased suite of environmental categories and exposure subclasses. This refinement has raised awareness amongst designers and specifiers of the processes involved in degradation of concrete structures. It also introduces the concept of a defined, if notional, service life standardised around a period of 50 years, with more relaxed or onerous requirements for shorter or longer periods, respectively. However, the prescriptive approach is still seen as falling short of what is possible in reducing the probability of serviceability failure. The imperative of making concrete both durable and sustainable demands that we work smarter with the component materials. The pathway to a more sustainable

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concrete is that of durability assured through a performance-based specification approach described in Chapter 3. The approach itself will therefore continue to evolve. An example of this is the “Exposure Resistance Class” method of specifying concrete. It is hoped that other innovations will see more international unity on the terminology used to categorise environments and exposure conditions, notwithstanding the freedom to determine national guidance in respect of concrete compositions. Equally, it would be helpful if there were global agreement on the criteria to use in respect of prescriptive guidance. Some countries rely on limiting compositions through minimum cement contents and maximum water/cement ratios, with no reference to minimum compressive strength. Others rely heavily on minimum compressive strength as an indicator of impermeability. Some prescribe specific curing regimes as part of the durability criteria, while others treat this as a separate issue to be dealt with under the execution of structures standard, again with an emphasis on compressive strength rather than impermeability. For example, a principle whereby the curing period is long enough such that 50% of the characteristic strength is reached. Although this is clearly a durability-related measure, the optics are wrong from the educational and attitudinal perspective if we constantly refer to strength rather than impermeability. This disconnect between a global consensus on the degradation mechanisms and the guidance on dealing with these threats is disappointing. We need to globally coalesce around the suite of criteria to be used in guidance to specifiers, though not the actual values of mix composition which, of course, will vary from region to region. Nevertheless, the role of the prescriptive approach remains valid for many projects, given the reality that our built environment will continue to be created by an industry with its roots in the craft and art of the master builder of previous millennia.

REFERENCES ABNT. 2014. Reinforced concrete structures – procedures. Rio de Janeiro: Associaçäo Brasileira de Normas Técnicas. ACI Committee 318. 2019. Building code requirements for structural concrete (ACI 318-19) and commentary on building code requirements for structural concrete (ACI 318R-19). Michigan: American Concrete Institute. Alexander, M. – editor. 2021. Fulton’s concrete technology. 10th edition. Johannesburg: Cement and Concrete SA. Ali, S., Naganathan, S. and B. Mahalingam. 2021. State of the art review on prescriptive and performance based approaches for concrete durability. International Journal of Sustainable Construction and Technology 12, 2: 80–88. Bamforth, P. 1994. Admitting that chlorides are admitted. Concrete 28, 6: 18–21.

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Beushausen, H., Ndawula, J., Helland, S., Papworth, F. and L. Linger. 2021. Developments in defining exposure classes for durability design and specification. Structural Concrete 22, 5: 2539–2555. BIS. 2000. IS 456: Plain and reinforced concrete - code of practice. New Delhi: Bureau of Indian Standards. Browne, R. 1986. Building deteriorology – the study and prediction of building life and performance. Chemistry and Industry 15: 837–844. BSI. 1997a. BS 8110. Part 1, Structural use of concrete: Code of practice for design and construction. London: British Standards Institution. CEN. 2000a. EN 206-1. Concrete - Part 1: Specification. performance. production and conformity. Brussels: Comité Européen de Normalisation. CEN. 2004. EN 1992-1-1. Eurocode 2: Design of concrete structures - Part 1-1: General rules and rules for buildings. Brussels: Comité Européen de Normalisation. CEN. 2018a. CEN/TR 15868. Survey on provisions valid in the place of use used in conjunction with the European concrete standard and developing practice. Brussels: Comité Européen de Normalisation. CEN. 2021b. EN 206:2013 +A2:2021. Concrete – Specification, performance, production and conformity. Brussels: Comité Européen de Normalisation. CSA. 2019a. A23.1-19/CSA A23.2-2019, Concrete materials and methods of concrete construction, test methods and standard practices for concrete. Mississauga: Canadian Standards Association. Czech Standards Institute. 2021. ČSN P 73 2404, Concrete – specifications, properties, production and conformity - additional information. Prague: Czech Standards Institute. Deacon, C. and J. Dewar. 1982. Concrete durability - specifying more simply and surely by strength. Concrete 16, 2: 19–21. Deutsches Institut fur Normung. 2022. DIN 1045-2, Concrete, reinforced and prestressed concrete structures - Part 2: Concrete. Berlin: Deutsches Institut fur Normung. East African Community Secretariat. 2008. EAS 131-1, Concrete – specification. Arusha: East African Community Secretariat. European Union. 1989. Council Directive 89/106/EEC of 21 December 1988 on the approximation of laws, regulations and administrative provisions of the Member States relating to construction products OJ No L 40 of 11 February 1989. Luxembourg: Publications Office of the European Union. European Union. 2011. Regulation (EU) No 305/2011 of the European Parliament and of the Council of 9 March 2011 laying down harmonised conditions for the marketing of construction products and repealing Council Directive 89/106/EEC OJ L 88 of 4 April 2011. Luxembourg. Publications Office of the European Union. Helland, S. 2013. Design for service life: Implementation of fib model code 2010 rules in the operational code ISO 16204. Structural Concrete 14, 1: 10–18. Hobbs, D. 2001. Concrete deterioration: Causes, diagnosis, and minimising risk. International Materials Review 46, 3: 117–144. Kulkarni, V. 2009. Exposure classes for designing durable concrete. The Indian Concrete Journal 83, 3: 23–43. Li, K., Zhou, C. and Z. Chen. 2008. Chinese code for durability design of concrete structures: A state-of-art report. In ICDCS 2008, First International Conference on Durability of Concrete Structures. Hangzhou: Zhejiang University.

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Ministry of Transport and Communication. 1986. Standard specifications for road and bridge construction, 1986 edition. Nairobi: Ministry of Transport and Communication. MOHURD - Ministry of Housing and Urban-Rural Development. 2015. GB 50010-2010 updated 2015, Standard for design of concrete structures. Beijing: General Administration of Quality Supervision, Inspection and Quarantine of the People’s Republic of China (AQSIQ). MOHURD - Ministry of Housing and Urban-Rural Development. 2019b. GB/T 50476-2019, Standard for design of concrete structure durability. Beijing: General Administration of Quality Supervision, Inspection and Quarantine of the People’s Republic of China (AQSIQ). National Research Council. 1997. Nonconvential concrete technologies: Renewal of the highway infrastructure. Washington: The National Academies Press. SABS. 2014. SANS 10100-2, The structural use of concrete. Part 2: Materials and execution of work. Pretoria: South African Bureau of Standards. Standards Australia. 2018. AS 3600:2018, Concrete structures. Sydney: Standards Australia. West, R. and D.-M. Keating. 1999. Report on industry survey of cement content and free water content for various concrete grades. Report for the Irish Concrete Society. Dublin: University of Dublin. Trinity College.

Chapter 3

Durability design and the performance route to specification

DURABILITY DESIGN MEETS DURABILITY SPECIFICATION MEETS DURABILITY VERIFICATION The prescriptive approach relies on controlling a subset of variables – concrete composition and cover to reinforcement – while assuming that all other variables in construction are to a high standard, not least the curing of fresh concrete. The alternative approach is durability design leading to specified performance for resistance to all the pertaining deterioration mechanisms. The concrete producer may prove the adequacy of a proposed mix by reference to relevant approval tests, demonstrating the potential of the proposed mix to meet the defined performance level. The finished work on-site can then be tested to confirm that the potential performance has been reached in practice. The producer’s approval test might be conducted specifically for a major contract. Alternatively, the concrete mix might be accepted where adequate performance has been established by previous tests on concretes of similar materials and mix composition. Control testing could then be used to monitor the key performance parameter. The performance approach reduces or eliminates many of the uncontrolled variables in the prescriptive approach. Performance-based specification was extensively discussed by Spekkink (2005) in the wider context of performance-based design, following extensive consultation with international stakeholder groups. The final report reflected on the writings of Vitruvius in the first century BC and the centrality of performance – firmitatis, utilitas, venustatis – in the mindset of the designer then and now. The core of these in any performance specification for hardened concrete is the definition of the functional requirements – utilitas. Performance-based design may be seen as the translation of the clients’ and the users’ needs into performance specifications, combined with assessment methods from which to affirm success. However, Spekkink (2005) expressed concern that barriers have built up over the centuries in the building process that are impeding the readoption of Vitruvius’s emphasis on performance at DOI: 10.1201/9781003261414-3

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the heart of what we do as designers. The barriers include the traditional culture in which the ability to improvise is still seen as a major merit; a prevalent conviction of most designers that the most important quality aspects of buildings cannot possibly be translated into performance specifications; the fragmentation of design engineering and construction; and the guilds mentality in the industry. The move from prescriptive to performance specifications in the wider industry may therefore be more challenging than it first appears. The self-evident weaknesses of the prescriptive approach in respect of concrete durability should, however, provide a strong impetus for change in this sector of the market. Overcoming the inherent weakness of the prescriptive approach to durability means replacing specified limitations on concrete composition ─ intended to yield unspecified desirable qualities ─ by a direct connection from the specification document to the desired performance. Thereby, the unspecified desirable quality of a product becomes a specified mandatory and determinable property. The difference between these approaches has been elegantly summarised by Bickley et al. (2006). They state that, in the prescriptive approach the specified requirements are a means to an end, and that it is the means that are verified. In the performance approach, the end is both specified and verified. Among ten identified reasons given by Spekking (2005) in support of performance-based design was the opportunity to make better use of the knowledge and expertise of contractors and suppliers, allowing them to come up with innovative cost-effective solutions. The application of this to the production of durable concrete is self-evident. Looking back to the time of Vetruvius again, it might be noted that the craft industry of construction derives from a time when a central commanding figure (“archi”) was expected to be the master of all the technology involved. Sadly, this unrealistic expectation pervades the division of roles and responsibilities in the building process under the prescriptive approach to specification. To date, this has reduced the instances in practice of aligning expertise with responsibility in the production of durable concrete. Tradition in the allocation of responsibilities in the construction industry has often prevented those with the most relevant knowledge, technology, and experience from playing their optimal role. Expecting designers to be experts in the elastic-plastic response of reinforced concrete sections to structural loading, while also being masters of the art of concrete production in a dynamic materials market, is both unrealistic and ineffective. The concrete supplier knows best, and the imperative of sustainable concrete is to harness the best that each party can bring to the table. This translates into an incontrovertible argument that performance should replace prescription in the specification of concrete where possible, with designers changing focus to durability design through mathematical models of degradation rate, as naturally as they have always done for structural design. Having thereby determined the required performance over a defined

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service life, the designer should express this as a minimum material performance target. It should then be left to the expert concrete producer to design a mix that can be tested to verify durability performance. We replace the “master builder” with the “master building team”. Integral in this team in current practice is certification of personnel as part of a quality management system overseeing the performance-based approach in practice. Current practice, reflected in international codes and standards, is not yet at the tipping point for change to a more scientifically informed approach. Ideally, the most accurate mathematical models of degradation would be applied to determine the key performance parameter that would then be specified. In the case of a major infrastructure project it would be applied in a fully probabilistic way. The impediment to greater use of the durability design method at present is the availability of a sufficient number of databases from which the statistical distribution of key parameters can be determined. This constraint will be progressively overcome in time. Nevertheless, the road to this enhanced method of ensuring both durable and sustainable concrete is already paved by three forms of specification ─ built on the performance concept ─ that represent, to a greater or lesser extent, the ideal. A useful summary of these approaches is presented in Table 3.1, adapted from an investigation of international specification practice by PIARC (2003). All three represent a significant step forward from reliance on the prescriptive approach. The obvious questions arise: if the performance approach to specification is so much better than the prescriptive, why is it not the default route in international codes and standards? If researchers had already published mathematical models of degradation that received international acceptance decades ago, why did European concrete practice in 2000, exemplified by EN206-1, relegate performance specifications to an “informative annex”? More than 20 years into the new millennium, why does the updated

Table 3.1 The essential attributes of three forms of specification built on performance (adapted for the case of concrete structures from guidance on road work projects in PIARC 2003) Form of Specification Performance specification Performance-based specification Performance-related specification

Essential Attributes Statement of the required performance of concrete in a project over its service life Statement of the engineering properties of concrete to ensure satisfactory performance over its service life Statement of the minimum acceptable values of specific material properties of the concrete, which have been correlated with the engineering properties of concrete to ensure satisfactory performance over its service life

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European standard EN206, in common with practice in many countries across the world, permit performance-related specifications but fail to set out the methodology? There are at least three possible reasons. The first is mindset, the second is inertia, and the third is an incomplete toolbox. Regarding mindset, we are unwittingly setting up our designers, specifiers, producers, and contractors to operate in two different leagues: the specialists and the rest! In introducing a discussion document on performance-based durability design (Knights 2019), a view is advanced that performance-based specification is the reserve of specialists and likely only to be applied on large projects in aggressive environments where long-term durability is desired or where an element is critical. Why are we hesitant to trust the industry to universally rise to the challenge of using concrete more sustainably through durability by design? Regarding inertia, the performance route would involve a fundamental shift in the roles and responsibilities of the parties involved. The performance route to achieving durable concrete involves a four-stage process: designing the concrete resistance to meet the durability load; specifying the resistance in a form that allows a concrete producer to design a mix that will meet the minimum level required; testing the proposed concrete mix before construction begins, to validate its potential compliance with the specified resistance; and testing the finished construction to verify that the potential resistance has been achieved in practice by the contractor. The responsibility for durable concrete – “fit for purpose” concrete – now becomes a chain with clear boundaries between each link, to which a particular party would clearly be identified as the lead. Long-established roles and responsibilities would change. As we know, the construction industry has the weight of millennia of craft history on its shoulders, and this can be an impediment to change. Finally, our toolbox needs updating. To date, there is a failure to internationally agree on the test methods to be used to benchmark an acceptable level of performance. This chapter reviews the “how to?” of implementing performance specifications. It introduces the probabilistic approach to durability design through a comparison with the well-established basis of structural design underpinning international design codes. It explains the concept of “design life” as a notional extension of “service life”. It demonstrates that mathematical formulations can be configured in different ways to suit the designer’s chosen approach to determining the performance requirement. It takes a more detailed look at the practical application of the three specification routes, including the “ERC” concept. It presents a range of options for verification of performance, using absolute values of a material property or durability indicators: “durability indices”. The final section highlights case studies demonstrating the power of the performance route to satisfy a range of stakeholders on high-value infrastructure projects: designer, client, contractor, and insurance underwriters.

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DURABILITY DESIGN – STRUCTURAL DESIGN COMPARISON Ideally, the performance-based specification derived through an analysis of the durability load and service life requirement would be achieved through a probabilistic approach. The process of designing the concrete resistance to meet the durability load is analogous to that of designing a concrete member’s resistance to the structural load. Achieving harmony in the design approach to both “loads” has been the goal of dedicated researchers in recent decades. In 1998, the European Commission funded a research network under the “Duranet” title to support the concept of limit state design for durability, building on the vision of a group of European experts. This important work has continued under the umbrella of the International Federation for Structural Concrete (fib), with a dedicated international team seeking to provide the rational tools to accompany the concept of designing to a defined “end of service life”. The International Standards Organisation recommended through ISO 13823-1 (ISO 2008) that the limit state approach be universally adopted for durability, following in the footsteps of the ultimate limit states and limit states of serviceability first set out in 1986 through ISO 2394, now in its fourth edition (ISO 2015). Thus, the limit state of durability would be the extent of damage beyond which an element or structure no longer meets its required condition for a certain degree of reliability. Examples of “damage” could be a critical chloride level, a ratio of carbonation depth to cover, the appearance of cracking following corrosion activation, etc. Introducing the probabilistic approach to durability design into international practice will involve agreement of safety factors, mean values, design values, distributions, performance functions, and capacity functions. The fundamental principles that underlie the application of probability analysis to durability design may best be introduced by reviewing a familiar example from structural design. The example chosen for demonstrating the principles is that of providing adequate resistance to bending in the design of a reinforced concrete beam. Consider the case of a singly reinforced concrete beam, simply supported at each end and carrying a uniformly distributed load, as illustrated in Fig. 3.1. The beam may fail in a number of ways but, for the purpose of this example, only the phenomenon of bending will be considered. The design of the beam is based on the assumption that it will fail if the maximum demand bending moment, occurring at mid-span, exceeds the ultimate moment of resistance of the section. Thus we may impose a constraint: R

S

0

where R = “resistance” (ultimate moment of resistance of section) and S = “load” (maximum bending moment).

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Figure 3.1 Conditions assumed in structural design comparison example.

To convert this relationship into a form that yields design values for the section, we formulate relationships describing the resistance of the section based on the concrete failing first; alternatively, resistance based on the reinforcement yielding first; and for the load in terms of the maximum demand bending moment. These relationships are functions of the following parameters:

R = f (fcomp, b , d) R = f (fsteel , As , z) S = f (W , L) where fcomp = compressive strength of the concrete b = width of beam d = effective depth of beam fsteel = tensile strength of the steel As = area of steel reinforcement z = the lever arm W = the total load L = the span. It may be further noted that z is a function of d and that d is a function of the total depth h. The design problem becomes one of selecting values and either validating that the constraint of the formula is met or solving the formula for an unknown value. The issue of uncertainty now enters. The intended values of width and total depth are generally assumed to remain constant during the service life and are readily achievable within specified tolerances during construction. What of the assumed concrete strength and load, however? The actual concrete strength that will be achieved in the structure cannot be known with certainty at the design stage. It will depend on the materials selected by

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the producer, the degree of compaction achieved by the operatives, and the curing conditions, which will be influenced in part by the weather at the time of construction. The load is variable also. The imposed load component in particular will fluctuate on a daily basis to a degree that is primarily dependent on the function of the building, but what of the unexpected loads that may arise over the lifetime of a building? Design based on the worst case ─ weakest possible concrete and potentially heaviest load, together with an allowance for design and construction blunders ─ would unnecessarily result in ungainly structural elements with a high moment of resistance (Rmax). These would be uneconomic, would severely limit the technical advancement of span, and would fail to meet the requirements of sustainable development. On the other hand, design based on the most optimistic case ─ concrete achieving its full potential strength, no unexpected load combinations, and no allowance for blunders ─ would result in elements with low moments of resistance (Rmin) which would have a high probability of failure. This is illustrated in Fig. 3.2, which charts a theoretical profile of the bending moment values (Sactual) resulting from load combinations that vary with time, as may be expected in reality. It is a question of balancing economy and safety in an acceptable way. This is achieved by the use of characteristic values of strength and load based on probabilistic considerations, and by the application of safety factors. The characteristic values of strength and load are determined by consideration of the mean values encountered in practice, their variability and the application of statistical parameters. The characteristic strength is determined by reducing the mean strength by an amount based on a chosen multiple of the standard deviation, while the characteristic load is based on values above an anticipated mean. These values are modified further to produce design values through the application of partial safety factors (Fig. 3.3).

Figure 3.2 Profile of resistance and applied bending moments showing possibility of wasteful overdesign or catastrophic underdesign if probabilistic methods are not used to describe variables.

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Figure 3.3 Concept of design strength and design load based on statistical considerations.

The design problem may be solved in a deterministic way, despite the fact that the values of the moment of resistance (R) and design bending moment (S) are calculated in an approach that incorporates allowances for the variability encountered in practice. The underlying probabilistic nature of the problem is further reinforced by the fact that the magnitude of both the factors of safety and the allowance for variability in the design codes has been selected to yield an acceptable probability of failure. The concept is illustrated in Fig. 3.4.

Figure 3.4 Incorporation of probabilistic concepts into an essentially deterministic model showing that, although the resistance moment has been designed to be greater than the applied moment, the effect of variability of these values leads to a chance of failure.

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Thus, while it appears that the problem being solved is of the form: R

S

0

it is set in the wider context of ensuring that the probability of failing to satisfy this condition is less than a maximum allowable failure probability (Pf max). This may be stated as follows: P {R

S < 0}

Pf

max

Hence a reliability analysis is being conducted to assess the probability of failure in the context of a defined limit state condition. Building on existing practice in design codes for the limit state of collapse, guidance on the serviceability limit state related to durability may be expressed through the reliability index (β) rather than the probability of failure. The relationship between the failure probability and the reliability index is illustrated in Fig. 3.5. The terms are related by the following formula: Pf = y (

)

where Pf = probability of failure y = normal distribution function β = reliability index For example, if −β has a numerical value of two then the area outside that multiple of standard deviations from the mean, on one tail of the curve, is about 3%, giving a failure probability of the order of 10−2. Bamforth (1999) made a study of existing and draft codes of practice to assess a typical international value of the order of magnitude of the

Figure 3.5 Relationship between acceptable level of failure probability and reliability index.

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probability of durability failure. He noted that the reliability index for the ultimate limit state of collapse was 3.8 for a 50-year life, indicating a probability of failure of the order 10−4. The reliability indices for the limit states of serviceability were approximately 1.5, indicating a probability of failure of the order 10−2. This is slightly more conservative than the fib Model Code for Service Life Design (Fédération Internationale du Béton 2006) which countenanced reliability indices of approximately 1.3 ─ a probability of the order 10−1. The order of magnitude associated with reinforced concrete could be less stringent than that for prestressed concrete in corrosive environments. Formulae for structural design, both precise and simplified in codes of practice, are now well-established, based on the key parameters. Equally, in respect of durability design, researchers have now clearly identified the dominant mechanisms and key parameters controlling the rate of concrete deterioration. Mathematical models of the deterioration mechanisms are very well established for the initiation phases of carbonation and chloride induced corrosion. Further refinement of these models is continuing. Modelling of the other phenomena that compromise durability are in various stages of advancement. However, at this point our analogy breaks down slightly. In structural design we assume that the strength, representing the resistance, does not diminish over time. In durability design we must allow for the fact that the conditions are incrementally changing due to the cumulative effect of the applied load (carbonation front advance, chloride buildup, etc.) and therefore, time is an important factor. Thus durability resistance is integrally associated with time, leading to a finite service life, which thereby leads to the concept of a notional “design life”. This is discussed in the next section.

THE “DESIGN LIFE” CONCEPT An integral concept of specification for durability in structures is that the material properties should meet the performance requirements over a defined life. This concept has been in the background in standards and codes of practice for many decades, but was not specifically exploited in practice. For example, the United Kingdom Code of Practice CP3, Chapter IX, Durability (Council for Codes of Practice for Buildings 1950) included definitions of what was termed the “designed life” and the “satisfactory life”. It was in use until superseded in 1992 by BS 7543, and was most recently updated in 2015 (BSI 2015). The British standard provides guidance on the application of life cycle design in parallel with the published and developing parts of ISO 15686, first published in 1998 and Part 1 of which was most recently updated in 2011 (ISO 2011). This international framework for service life planning derived from the deliberations of a joint RILEM/CIB group (Masters and Brandt 1989),

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Figure 3.6 Concept of safe durability design whereby the service life is defined by statistical methods to be less than the likely mean time to failure.

which itself built on earlier work from Sweden (Sentler 1983) and in the Netherlands (Siemes et al. 1985). This was further studied and a clearer focus on the way forward was provided by the publication of RILEM Report 14 (Sarja and Vesikarri 1996) and CEB Bulletin 238 (Schiessel et al. 1997). Meanwhile, a significant impetus to the development of the tools necessary to apply the service life principles was provided through the European Union Brite EuRam IIIa project “DuraCrete”, concerning probability performance-based design of concrete structures, (COWI 2000). An overview of the concepts of design life and service life is illustrated in Fig. 3.6 in a development of work by Tuutti (1982) and Sarja and Vesikarri (1996). This shows that the structure reaches its design life when the maximum tolerable level of damage is reached. An acceptable value of the service life may then be evaluated, for example by statistical considerations. This introduces the concept, but in practice, the degradation mechanism can itself be modelled stochastically and the risk of reaching the limit of damage is considered probabilistically. This is explained in later sections. A simple concept of design life is to define it as the period of use anticipated when designing the structure (Concrete Society 1996). Previously, the concept was introduced by Rostam (1984) as the combination of possible technical life with economic considerations. Somerville (1986) had recognised the reality that some maintenance could be expected, but saw design life as the minimum period that one could expect the structure to perform its designated function without significant loss of utility and without significant maintenance intervention. A similar description was incorporated into EN 206 (CEN 2021b), as a working life period without major repair. Viewing it from the performance perspective, the design life may be regarded as the period during which the performance requirements are exceeded (ISO 2011).

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Design life may thus be seen in the context of supporting decisions relating to specification of material properties, and it could therefore be extended to life cycle cost analysis. However, it must be seen as a notional concept ─ the structure will not necessarily reach the failure criterion at the appointed year. Serviceability limit states need to be debated and agreed. A wide range of possibilities exists even, for example, in the common case of crack development in cover concrete from corroding reinforcement. In some structures, the serviceability limit state would be reached at the appearance of the first crack, whereas in others a certain level of spalling would be tolerable. Circumstances will vary ─ the risk of concrete spalling is less acceptable in the case of a bridge over a motorway than for a bridge over a minor river. Furthermore, there is typically variable corrosion activity and damage within a single structure, which supports the drive to adopt a probabilistic approach to the problem.

PROBABILISTIC APPROACH TO DURABILITY DESIGN OF MAJOR INFRASTRUCTURE Durability design is informed by mathematical models that correlate the rate at which concrete can resist a durability threat with the service life it is expected to deliver. Harnessing the design life concept in this context allows the determination of appropriate material properties on a rational basis. Researchers continually strive to perfect models of degradation in performance with time-dependent functions in this context to be used in one of a number of overarching approaches to durability design. Several overarching approaches have been advanced by, for example, Siemes et al. (1985), Sarja and Vesikari (1996), Siemes and Rostam (1996). Progress is being made on controlled allowance for uncertainties by adopting a probabilistic approach to durability design as an increasing volume of published field data becomes available. The use of statistical methods is highly appropriate in durability design because the requirement is to minimise the risk of failure in the face of even more uncertainties around the variables than in structural design. Whatever the “known unknowns” ─ such as introducing innovative cementitious materials but with long-established chemistry ─ what about the “unknown unknowns”? Amongst the latter, Alexander (2018) mentions global warming. To what extent will atmospheric temperature rise increase the rate of degradation reaction in concrete over the service life of existing structures and those currently under construction? In the interim, a partial factor method may be used, whereby the yet-to-be-defined statistical data for resistance and durability load in the full probabilistic approach is substituted by characteristic values and partial coefficients. This is a semi-probabilistic approach to deterministic calculations, with variability taken into account by the partial factors.

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Irrespective of the overarching approach, durability design allows consideration of the type of structure, material properties, microclimate, required life, quality of site practice, and the probability of failure or the level of reliability in a manner whereby optimal solutions may be determined in the context of sustainable development. Appropriate material parameters may be determined based on a balance of the economics of initial and life cycle cost. Durability design is thereby conducted in the context of a risk that is quantifiable. Besides “risk of failure”, one may equally express a defined level of reliability. This is significant to the designer, client, and insurance underwriter of high-cost infrastructure projects such hospitals, tunnels, and bridges. There are several overarching approaches to formulating the expressions used for durability design. The applied load to resistance may be expressed as a limiting ratio or an inequality through a factor of safety in several ways. Thus, the mathematical solutions to the problem of design for durability are not quite as straightforward as those presented in the structural design analogy. Allowance for the multitude of variables and their distributions in durability design problems leads to several approaches, each of which may be more readily applicable in any given case, depending on the deterioration mechanisms involved. The application in each case, however, involves aspects that are familiar from the principles of the approach to structural design. Although the terminology in the literature varies for similar approaches, these may be broadly described as the “Lifetime Safety Factor Method”, the “Intended Service Period Design”, and the “Lifetime Design”. The first of these is essentially deterministic, but the others are stochastic design methods.

Lifetime safety factor method Firstly, the concept of the Lifetime Safety Factor Method involves consideration of: • a function ( R or R(t)) that describes the resistance of the structure; it may or may not be time-dependent; it may be based on mean values of input parameters • a function (S or S(t)) that describes the “load” (for example, the chloride level) on the structure; it may or may not be time-dependent; it may be based on mean values of input parameters • the safety margin (R(t) − S(t)) • the mean service life (tmean). The resistance of the structure may reduce with time until failure is reached when it equals a pre-determined value of the “load”. In such cases, R(t) is described as a “performance model”. Equally, the resistance of the structure may be a constant but the load may increase with time until failure is

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reached when it equals a pre-determined value of the resistance. In such cases, S(t) is termed a “degradation model”. Note that the requirement may be phrased in a limiting manner: for example, the ability of a concrete element to keep a given parameter below a certain level. Degradation models are common in durability design problems. In the case of the carbonation phenomenon, for example, the depth of carbonation increases with time and may eventually reach the reinforcement. In the case of chloride ingress, the level of chlorides at the reinforcement builds up over time and may reach a critical corrosion threshold level. Consider the case of a typical degradation problem. The two conditions that prevail during the mean service life (Fig. 3.7) are such that: R R

S (t) > 0 t < tmean S (tmean) = 0

It may be seen from Fig. 3.7 that, due to the distribution of values of S(t), the possibility of the condition {R − S(t) = 0} being met occurs in advance of tmean and that the probability of this occurring increases with time. Thus, the maximum allowable failure probability must be considered (Pf max), leading to a design constraint that: P {R

S (t) < 0}T

Pf

max

T = tmean

Figure 3.7 Profile of a “degradation model” for durability design showing its variation over time, inherent variability and interaction with a resistance level assumed to be constant.

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and this leads to a target service life (tg) that is the time at which Pf max occurs. The challenge for the specifier is to meet the target service life (tg) that represents the client’s expectation of service life. To achieve this duration of satisfactory performance it is necessary to specify based on the anticipated mean service life (tmean), which becomes the design service life (td). This introduces the concept of the “Lifetime Safety Factor”, which is described by the relationship: td =

t tg

where td = design service life γt = Lifetime Safety Factor tg = target service life. Introduction of probability to durability design in a similar way to that of structural design for the traditional form of “loading” requires a more simplified route than that of calculating probabilities. It may be shown (Sarja and Vesikari 1996) that, once values for Lifetime Safety Factors have been calibrated, design would involve the following steps: • agree the target service life of the structure • determine the design service life based on the Lifetime Safety Factor and the target service life • apply the relevant degradation model or performance model using the design service life and select appropriate material properties, sections sizes, and/or protective measures • check that the reduction in the safety margin, for example, R − S(t), from time t = 0 to time t = tg is less than an allowable value • if the reduction in safety margin is too great, redesign using higher performance materials, or larger sections, or introduce additional protective measures • if the redesign fails to produce a satisfactory value of the minimum safety margin, consider a shorter target service life, agreed with the client. The Lifetime Safety Factor method provides a good introduction to the concept of extending structural design principles to durability design. The next development is to extend consideration of the probability of failure as a significant criterion. Two other methods, which are essentially both sides of the same coin, have been described (Siemes and Rostam 1996) as the “Intended Service Period Design”, and the “Lifetime Design”. They approach the same problem from different perspectives, but yield the same outcome. The choice of method depends on the information available at the design stage.

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Intended service period design method The principles underlying the Intended Service Period Design and the Lifetime Design methods are similar to the Lifetime Safety Factor method; however, the methods extend the use of probabilistic analytical tools. In relation to the Intended Service Period Design, consider: • • • • •

the function (R or R(t)) that describes the resistance of the structure the function (S or S(t)) that describes the load on the structure the condition that represents a technical failure (R(t) − S(t) < 0) the maximum allowable failure probability (Pfmax) the target service period (tg) during which it is expected that the probability of technical failure will not exceed the maximum allowable level.

The probability of the difference between resistance and capacity becoming negative at least once during the target service period may be calculated at any value of time t: P {R (t )

S (t) < 0}T where T = tg

This value may then be compared with the maximum allowable value of the failure probability to calculate the probability of the former being less than the latter. This may be stated as follows: Pf , T = P {R (t)

S (t) < 0}T

Pfmax

where T = tg

Thus it is possible to chart a series of values of Pf,T for different values of t and check whether the required level of reliability is achievable during the target service life. The effect of changing parameters in the resistance or load functions may also be examined by plotting the probability distribution curves. Typical curves are illustrated in Fig. 3.8. The shape of the curves will be familiar from the traditional quality control procedure of assessing by statistical means the acceptability of a batch of goods by testing only a sample. Thus, design would involve the following steps: • agree the target service life of the structure by consultation with the client • apply the relevant degradation and/or performance model using the mean values of the input parameters together with their distributions • check the probability of the resistance falling below the required level during the service life or, equally, check the probability of the load exceeding the available level of resistance

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Figure 3.8 Probability distribution curves showing the effect of different load parameters.

• chart the failure probability through consideration of the acceptable maximum level of failure • if the failure probability is unsatisfactory, redesign the mean values of the input parameters by specifying higher performance materials or larger sections and/or reduce the distributions by specifying stricter quality control measures if this is feasible • if the redesign fails to produce a satisfactory level of failure, consider whether the client could accept a shorter target service life or a higher risk of failure. The acceptable maximum level of failure would depend on the consequences. The greater the negative consequences of exceeding the limit state, the lower the acceptable probability of failure. For example, a higher figure might be more acceptable for carbonation-induced depassivation leading to a low rate of generalised corrosion than chloride-induced depassivation leading to pernicious pitting. The other issue, discussed in later sections, is the definition of end of service life in a corroding structure. Conservatively, it might be assumed to occur at the end of the initiation phase when the steel depassivates. Alternatively, it could include an additional period allowing for the years to elapse in the propagation phase before damage leads to serviceability failure, even if collapse is not yet a risk. Much remains to be agreed on these fundamental questions. For example, the time to reach a 50% probability as a level of failure for initiation only would equate only to 35% if a decade of the propagation period were included (Helland 2013). This exemplifies the opportunities to be grasped in the quest for the most sustainable solutions to durable concrete. Designing for the time to reach a 50% probability for both the initiation and part of the propogation phase would lead to a significant global saving in material resources.

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Fundamentals of Durable Reinforced Concrete

Lifetime design method The principles of the Intended Service Period Design may also be used in the Lifetime Design method, but the problem is approached in a different way. For example, in the former, one might consider the probability of the level of chloride at the reinforcement reaching 0.4% by weight of binder during the target service life. In the latter, one considers the probability of the service life being less than the target service life for the condition whereby the chloride level is 0.4% by weight of binder at the reinforcement. The first method is used where the distributions of the performance and load are known. The Lifetime Design method is used where the distribution of the service life is known or may be assumed to follow a certain profile. The Lifetime Design method uses the functions R(t) and S(t) to formulate a relationship in terms of the life of the structure (L): L = f (R (t), S (t))

Knowing, or assuming, the appropriate model of the service life distribution and the maximum permissible value of the probability of failure, it is possible to evaluate the target service life. The probability of the calculated life of the structure being less than the target service life may be determined:

P L

tg < 0

The concept is illustrated in Fig. 3.9 for a deterioration process in which time-dependency promotes a higher risk of failure in earlier years.

Figure 3.9 Lifetime design: service life distribution.

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69

Figure 3.10 Probability distribution curves showing the effect of different resistance parameters.

A further calculation may be made of the probability of failure. This is the probability of the difference between the expected life and the target service life being less than an acceptable value: Pf = P L

tg < 0

Pacceptable

As before, the effect of changing parameters in the resistance or load functions may be examined by plotting the probability distribution curves as illustrated in Fig. 3.10. Thus, design would involve the following steps: • agree the target service life of the structure by consultation with the client • apply the relevant degradation and/or performance model using the mean values of the input parameters, together with their distributions, to determine the calculated life of the structure • check the probability of the calculated life falling below the target service life • chart the failure probability by considering the acceptable maximum level of failure • if the failure probability is unsatisfactory, redesign the mean values of the input parameters by specifying higher-performance materials or larger sections and/or reduce the distributions by specifying stricter quality control measures, if this is feasible • if the redesign fails to produce a satisfactory level of failure, consider whether the client could accept a shorter target service life or a higher risk of failure. Using the Lifetime Design method involves knowing the service life distribution. The distribution will depend on the deterioration process under consideration.

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The distributions will not necessarily be normal. Some phenomena lead to a situation in which the values are not evenly distributed about the mean, with damage occurring more frequently in earlier years than later ones. This would be typical in the case of deterioration models involving the square root of time. These models include the significant cases of corrosion initiation by carbonation and chloride ingress. The service life distribution for such phenomena may be modelled by, for example, the log-normal distribution.

FRAMING THE DURABILITY SPECIFICATION, INCLUDING THE “ERC” OPTION By using mathematical models of degradation integrated into one of the overarching approaches to durability design, the designer will have determined the required level of performance to be specified. In framing the durability specification, the specifier needs to tie down the level of performance that is acceptable without unnecessarily restricting the concrete producer from exploiting innovative materials and expert knowledge of local materials in their design of an optimal mix. Such a collaboration should not only produce a concrete of known performance, but also one that is economical and an example of sustainable development, harnessing as it does the producer’s expert knowledge of materials, available technological advances, and innovative combinations. An interesting overview of the process was presented in a nomogram by Dhir, Jones, and Ahmed (1991) in respect of chloride resistance. The nomogram relates the intended service life, environmental chloride content, cover, chloride threshold value for corrosion, and the required coefficient of chloride diffusion. The format of the nomogram is illustrated in Fig. 3.11. The nomogram is based on Fick’s second law of diffusion. It relates diffusion coefficient (D) and the period required to reach a specific chloride concentration at the reinforcement (t1). The initial relationship is based on an assumption of a 50 mm cover, a surface chloride level of 0.5 M, and an initial chloride content of zero. The effect of different values of cover on the period may be assessed by use of the southeastern quadrant of the nomogram to determine the alternative period (t2). Should the effect of a different surface chloride level be of interest, the southwestern quadrant can be used to determine the relevant period (t3). The northwestern quadrant includes a graph to yield a multiplication factor to be applied if the initial chloride content differs from the assumed value of zero. A nomogram was also employed by Bamforth (1994) to highlight the significant difference in performance of PFA and slag concretes in comparison with Portland cement concretes. International leadership in the drafting of performance-based codes and standards in concrete has been studied by Tanesi et al. (2010). In common with other researchers, they note the seminal place of the Nordic Model and

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71

Figure 3.11 Format of Dhir et al. (1991) nomogram for determination of required chloride resistance.

its influence on the drafting of the international standards that have followed. The Nordic Model applied to concrete calls for the performance requirements to derive from the functional needs that derive directly from the users’ needs. Translation of this into general concrete practice is likely to be on the basis of hybrid approaches. These will fall somewhere between prescription at some point in the process and performance at the production point. The need for performance testing of the finished product, to close the quality control loop, remains a matter for debate. An interesting European development is the introduction of the Exposure Resistance Class (ERC), which opens the door to a more performance-directed mindset among all stakeholders by incrementally introducing a link between a scientifically derived target and the already familiar prescriptive exposure and strength classes. The ERC concept is initially directed to the initiation of reinforcement corrosion through depassivation by carbonation or chloride ingress. These phenomena have been successfully modelled by many researchers (Chapters 6 and 7). The models of carbonation were studied by Von Greve-Dierfeld and Gehlen (2016a, 2016b, 2016c) as the basis for a partial safety factor approach to performance-based durability design from which to move on from the deemed-to-satisfy rules while building on the internationally accepted exposure

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classification system. The exposure class listing in Europe, set out in Chapter 2, and heretofore solely associated with deemed-to-satisfy limiting compositions, is the framework for the alternative ERC performance approach. The ERC concept provides a roadmap for designers and specifiers of major projects who previously shunned the performance approach due to a lack of codified guidance. The imperative of using innovative material combinations for which there is little or no “experience in the place of use” acts as a further driver in pursuit of sustainable development. An example of the ERC concept is illustrated in Table 3.2 in the case of corrosion risk from carbonation. Essentially, a level of performance is determined through durability design in respect corrosion risk from carbonation or chloride ingress. This performance level is specified in respect of a maximum rate of carbonation (mm/year0.5) or chloride diffusion coefficient (10−13 m2/s). Based on these benchmarks, the ERC is selected – the “XRC” series or “XRDS” series for carbonation and chloride ingress, respectively. The designer is guided through Eurocode 2, underpinned by limit state principles, to specify the relevant ERC together with a minimum cover to reinforcement. Following specification, the next step is the production of a concrete which meets the test value criteria. The producer may already have determined an acceptable concrete mix for each ERC through initial type testing. The key to the concept is European agreement on the test methods. A systematic review of all European test methods for concrete is regularly carried out. About half of the 19 methods in the EN 12390 series of tests on hardened concrete relate to durability, and each one is referred to in more detail in later chapters as appropriate. Although the ERC concept has been welcomed by Geiker et al. (2021), they have expressed some concern that its application may require design based on conservative assumptions that might limit innovation. Also, the associated limit state function may not be favourable in the context of sustainability.

VERIFYING THAT DURABILITY PERFORMANCE MEETS THE SPECIFICATION Verification that the concrete as constructed will satisfactorily meet the specified performance criteria is typically a three-stage process. The first requirement is that the concrete be produced at a plant certified to have an adequate quality control process in place. The second issue would be provision of test data showing that the proposed concrete mix will meet the performance requirement through pre-qualification testing. The third aspect would be verification that the contractor can place, compact, and cure the fresh concrete delivered to the site such that the specified performance is met in the finished product on-site. This acceptance testing

0.6 1.2

2.4

3.6

4.8

6.0

7.2

8.4

XRC2

XRC3

XRC4

XRC5

XRC6

XRC7

mm/year

0.5

Characteristic carbonation rate

XRC0.5 XRC1

ERC

50

100

50

XC2

100

Design life (years) 50

XC3

100

50

XC4

100

Minimal level of cover is acceptable for all 4 classes and both design life periods Minimal level of cover acceptable Minimal level of cover for 50 years and very low level for 100 years Minimal level of cover for 50 years and very low level for Very low level of cover for 50 years, increased by 10 mm 100 years for 100 years Minimal to low level of cover for 50 years, increased by 5 mm Low level of cover for 50 years, increased by 10 mm for for 100 years 100 years Minimal to low level of cover for 50 years, increased by 10 mm Moderate level of cover Moderate level of cover for 100 years for 50 years, increased for 50 years, increased by 10 mm for 100 years by 15 mm for 100 years Low level of cover for Moderate level of cover Intermediate level of Very low level of cover for 50 years, increased by for 50 years, increased cover for 50 years, 50 years, increased by 10 mm for 100 years 10 mm for 100 years by 20 mm for 100 years increased by 15 mm for 100 years Very low level of cover for Moderate level of cover for High level of cover required for 100 years but for 50 years 50 years, increased by 50 years, increased by it may be reduced by 20 mm for XC3 and 15 mm 10 mm for 100 years 10 mm for 100 years for XC4 Very low level of cover for Moderate level of cover for Very high level of cover required for 100 years but for 50 50 years, increased by 50 years, increased by years it may be reduced by 20 mm for XC3 and 15 mm 15 mm for 100 years 15 mm for 100 years for XC4

XC1

Exposure class

Minimum cover to reinforcement

Table 3.2 Draft guidance on cover to reinforcement for a given service life in combination with a concrete of known minimum performance in respect of carbonation resistance using the ERC concept (adapted from CEN/TC 104 2023)

Durability design and the performance route to specification 73

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would require generation of field data from elements over a short to medium period after casting. Given the greater degree of variability for concrete cast on site, the criteria for acceptance testing of the finished product may be somewhat less restrictive than pre-qualification testing conducted in the more controlled environment of the producer’s plant. It may again be noticed that a not-so-subtle aspect of the three-stage process is the division of responsibility. The specifier is responsible for ensuring that the selected level of performance is adequate. The concrete producer is responsible for designing and manufacturing a fresh concrete mix that has the potential to meet the specified performance. The contractor is responsible for ensuring that the concrete members, as cast, achieve the specified performance level. Therefore, the specifier is seeking from the producer a certificate of compliance with the performance requirement, supported by evidence from test results, rather than a proposed set of mix constituents and proportions for the designer to sign-off on. Ideally, the designer would also receive certification based on test results from a sample of the constructed concrete to verify that the potential of the producer’s concrete had been achieved by the building contractor. However, that may be more difficult to achieve in practice because of the long-term nature of monitoring concrete in service before definitive trends are verifiable. Both fundamental and applied research has identified the core material factors influencing concrete durability. The former, based on thermodynamics and physio-chemical materials science, has accurately described the transport and reactivity of potentially deleterious ions, gases, and fluids through the permeable and porous network of a cementitious matrix. Applied research has characterised the apparent and semi-empirical coefficients that contribute to the rate at which specific ions, gases, and fluids penetrate concrete under a range of service conditions. Data from combinations of fundamental and applied science, laboratory and field experiments, has allowed the engineering of measurable parameters from which performance specifications may be based. The “engineering” of these material properties is the bridge between highly accurate but impractical-to-measure parameters and sufficiently accurate semi-empirical durability indicators that have the practicality to gain candidacy for inclusion in international codes and standards. The concept of using a semi-empirical measure of carbonation resistance in durability design and specification – a “k-factor” value for each combination of member type, exposure subclass, surface finish and orientation – was proposed in the 1980s by Richardson (1988). He proposed a factorial approach with quantification of the range of k-factors to be resisted in different environments and surface finishes from which the cover to reinforcement could be specified, depending on the required service life. For example, using typical concrete practice of the time (characteristic compressive strength of 30–35 MPa), he determined that cover to reinforcement would need to withstand a minimum rate of 0.8 to a

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75

maximum rate of 6.5 mm/year0.5 over a range of exposure conditions. This wide span of rates extended from saturated slabs to walls in moderately humid indoor conditions caused by a temperate climate subject to frequent Atlantic weather fronts. The direction of travel has taken two parallel roads since then, targetting both the concrete resistance property required for a given cover and service life rather than Richardson’s approach of making cover the variable with a fixed material property of characteristic strength. One road has been to perfect test methods for determining quickly and accurately a fundamental material resistance property that could be used in specification. A good example in the case of marine exposure conditions is the test method for chloride ion penetration, ASTM C1202 (ASTM 2022d). The other road has been based on readily obtainable and highly relevant material parameters that can serve as a durability index in place of an elusive material property that is difficult to measure. Enduring research on this as a proposed approach has been notably conducted by Alexander (Alexander and Ballim 1993; Alexander et al. 1999a and 1999b; Alexander et al. 2008; Muigai et al. 2009); Andrade (Andrade 1993; Bjegovic et al. 2016); Torrent (1992) and Baroghel-Bouny (Baroghel-Bouny 2004; Thiery et al. 2006). The durability index is based on two pragmatic concepts. The first is the replacement of a material property that is most relevant but difficult to measure by a parameter that is easily measured at a young age, and which provides a soundly based prediction of performance during the service life. The second aspect is that a single durability index can also be used to embrace the multifaceted interactions that dictate the rate of degradation, rather than a separate series of individual material properties. Practice in South Africa, for example, is built around three tests on pre-conditioned disc specimens that are 75 mm in diameter and 30 mm long. Ballim (1991) correlated an oxygen permeability index (OPI) with the carbonation rate, and the 28-day chloride conductivity index (CCI) with chloride diffusivity. The third test yields a water sorptivity index (WSI). Further development included a very rapid chloride conduction test (Streicher and Alexander 1995), modifications to the chloride test rig (Otieno and Alexander 2015) and refined correlation of the oxygen permeability index with the carbonation coefficient and carbon dioxide diffusion coefficient in a revised model of carbonation (Salvoldi et al. 2015). This has formed the basis for the South African SANS 3001-CO3 (testing concrete in structures) series reported in Chapters 4, 6, and 7 (SABS 2015, 2021, 2022), with the WSI test method pending adoption. The employment of electrical resistivity as an index of the rate of chloride-induced corrosion has been demonstrated by Andrade et al. (1993). A method of measuring the air permeability of concrete cover (kt), developed by Torrent has ready applicability in performance-based specifications (Torrent et al. 2012) and has been incorporated as an annex to Swiss Standard SN 505 - SIA 262/1 (SNV 2013). The maximum acceptable coefficient of air permeability of the cover concrete (kTmax)

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can be determined by durability design and thereby become the specified performance target. The producer can then propose a mix, based on prior results of kT values in their database of production concrete. The contractor can verify that the specification has been met through kT values determined from site testing. Following the lead of South Africa (oxygen permeability, chloride conductivity, water sorptivity) and of practice in Switzerland (coefficient of air permeability), other countries providing codified guidance on acceptable performance-related tests include Australia, Canada, France, Netherlands, New Zealand, Norway, Spain, and the US. Identified rapid and reliable tests centre around chloride permeability, chloride bulk diffusion, electrical resistivity, sorptivity, and air void systems. Some of the national codes have introduced specification by performance in a hybrid form. Also, Alexander (2018) has emphasised that some of these codes represent only a partial step on the road to performance-based and performance-related specifications in that the performance of the as-built concrete member is not always verified. Equally, the need to refine standard test methods for innovative mixes in laboratory and site conditions may also be a feature of the performance specification methodology for some time until large databases are built up, from which trends can be reliably extrapolated. For example, Harrison (2019) found that carbonation performance tests could underestimate the durability potential of slowly reacting cements where the replacement level of Portland cement clinker exceeded 55%. This is because the relative humidity levels of the carbonation tests may be too low compared with in-service levels, prematurely terminating longer-term hydration.

PROBABILITY-BASED DURABILITY DESIGN AND PERFORMANCE SPECIFICATION IN ACTION An early case study of durability design is the use of the DuraCrete methodology reported by Breitenbucher et al. (1999). This example pertains to the design of the concrete lining for the Western Scheldt Tunnel in the Netherlands. The methodology was used to fulfil a requirement that the contractor document the achievement of an anticipated life in excess of 100 years. The tunnel was to be constructed in chloride-contaminated soil. The tunnel segments were to be jointed together with interlocking nibs and recesses. It was recognised that in time the joints may leak, thus exposing the joint surfaces to chloride ingress. The cover to reinforcement at the joints was identified as being critical ─ a compromise was required between minimising the cover for structural reasons and providing adequate cover for durability reasons over a service life in excess of 100 years.

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The serviceability state chosen for analysis was that of the onset of corrosion. The acceptable probability of failure was derived by a study of the reliability indices in a Dutch code of practice and in a draft Eurocode. In the absence of better data a requirement was set that the minimum reliability index would be in the range of 1.5 to 1.8, equating to an acceptable failure probability of about 10−2. Chloride ingress was modelled by a formula based on a solution to Fick’s second law of diffusion. Further detail on the background to this formula is presented in Chapter 7. The form of the equations used was as follows: x (t )

= 2C(Crit) kt DRCM,0 ke kc

kt DRCM,0 = D0 C(Crit)

= erf

1

(1

CCrit CSN

( )t t0 n t

)

where x = concrete cover C(Crit) = critical chloride content CSN = surface chloride level DRCM,0 = chloride migration coefficient measured at time t0 D0 = effective chloride diffusion coefficient at time t0 kt, ke, kc = constants to take account of method of test, environment and curing on the value of D0 n = age exponent erf −1 = inverse of error function t0 = reference period t = exposure period. The condition tested was for a cover equal to or in excess of 35 mm. Quality control procedures for achieving the specified cover were anticipated to be good, and so a mean value of 37 mm was adopted with a standard deviation of 2 mm, exponentially distributed. The chloride migration coefficient was determined by a chloride migration method to be 4.75 × 10−12 m2/s (mean value) with a standard deviation of 0.71, normally distributed. The critical chloride content was determined from a literature review, taking account of the anticipated humidity, and was taken to be 0.70% by weight of binder with a standard deviation of 0.10%, normally distributed. The surface chloride level was taken as 4.00% by weight of binder with a standard deviation of 0.50%, normally distributed. Mean values and standard deviations were assigned for the coefficients and age exponent. The value of the reference period, 0.0767 years (28 days) was, of course, deterministic. An analysis of the use of a 37 mm mean cover yielded a reliability index of 1.5 at 100 years, and was therefore acceptable.

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Other interesting case studies have been collated by Alexander (2016). The use of the chloride permeability test to ASTM C1202 has been described by Heath (2016) in the case of a significant marina in Abu Dhabi where initial field test results indicated a need for further investigation of predicted performance. The use of the ASTM International test has also been outlined by Andrade et al. (2016) in the design and specification of concrete in the new Panama Canal structures. Durability design for a 120-year design life combined with real-time durability assessment was reported in respect of the Hong Kong to Zhuhai and Macau bridge and tunnel complex by Li et al. (2016).

SUMMARY Application of the durability design and performance method to widespread practice will require three developments. Firstly, consensus on a set of mathematical models of deterioration that can be easily applied in practice. Secondly, development and acceptance of relatively easy to use on-site tests for properties of concrete from which the specified performance or material properties in the finished construction can be verified. Thirdly, the publication of comprehensively supported performance-based specification methodologies in national codes and standards, side-by-side with the existing guidance on the prescriptive approach. The relegation of durability design and performance-based or performance-related specification to annexes in such documents has already acted as a barrier to their widespread introduction to practice. The concepts involved are neither new nor complex. Models used as design tools need to be relatively straightforward so that they can be used on an everyday basis. Complex formulations based on parameters that cannot be readily specified or tested are unlikely to find favour. The refinement, or to put it more correctly, the coarsening, of the models will involve replacement of terms which are difficult to test routinely. Practitioners need to specify on the basis of material characteristics that can easily and inexpensively be checked for compliance. Replacement of terms through conservative assumptions may involve some loss of exactitude, but this will be compensated for by an increase in reliability in service. However, change to traditional practice is always challenging, and the prescriptive approach will always be an effective alternative in low-risk, low-value, repetitive projects. Use of the performance approach is already commonplace in high-value infrastructure projects. Further adoption of the full probabilistic approach to durability design, with internationally agreed values of the acceptable probability of failure or reliability index, will require refinement of the statistical modelling of variables based on “big data” from field studies. The most appropriate type of distribution appropriate to each input parameter will need to be studied. Meanwhile, a partial factor method may be used whereby the yet-to-be-defined statistical data is substituted by characteristic values and partial coefficients.

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Chapters 4 to 11 present mathematical models of transport processes and the parameters that influence the rate of individual degradation mechanisms. Use of these in drafting codes and standards for specification by performance is already being harnessed in some countries for concretes exposed to corrosion risk from carbonation or chloride ingress. Models of the initiation and propagation phases of corrosion are available, although sole reliance on the former is likely to be used by most designers. Researchers have also proposed approaches for modelling freeze-thaw, chemical attack, and alkali-silica reaction. Further refinement of these models may be expected, leading to their eventual use in specification. Specification by performance using a fully probabilistic approach will require further determination of the statistical distributions to be associated with a wide canvas of variables. Another step will be the broadening of models to allow inclusion of the chemical and physical properties of innovative binders. The research journey will be lengthy, but the effort will be rewarded through the ability to specify innovative and sustainable low carbon concretes with a quantifiably minimal risk of durability failure. Some commentators suggest that progress be focused on the partial factor method, using the semi-probabilistic approach of deterministic calculations with variability taken into account by the partial factors. However, calibrating the factors presents a significant challenge, so striving for full probabilistic design remains a worthwhile endeavour for all in the industry.

REFERENCES Alexander, M. – editor. 2016. Marine concrete structures, design, durability and performance. Sawston: Woodhead Publishing. Alexander, M. 2018. Service life design and modelling of concrete structures. Revista ALCONPAT 8, 3: 224–245. Alexander, M. and Y. Ballim. 1993. Experiences with durability testing of concrete: A suggested framework incorporating parameters and results from accelerated durability tests. In Proceedings, Third Canadian Symposium on Cement and Concrete, 248-263. Ottawa: National Research Council. Alexander, M., Mackechnie, J. and Y. Ballim. 1999a. Guide to the use of durability indexes for achieving durability in concrete structures. Research Monograph No. 2. Cape Town: Department of Civil Engineering, University of Cape Town. Alexander, M., Ballim, Y. and J. Mackechnie. 1999b. Concrete durability index testing manual. Research Monograph No. 4. Cape Town: Department of Civil Engineering, University of Cape Town. Alexander. M., Ballim, Y. and K. Stanish. 2008. A framework for use of durability indexes in performance-based design and specifications for reinforced concrete structures. Materials and Structures 41, 5: 921–936. Andrade, C. 1993. Calculation of chloride diffusion coefficients in concrete from ionic migration measurements. Cement and Concrete Research 23, 3: 724–742.

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Andrade, C., Alonso, C. and S. Goni. 1993. Possibilities for electrical resistivity to universally characterize mass transport processes in concrete. In Concrete 2000: Economic and durable construction through excellence, ed. R. Dhir and M. Jones, 1639–1652. Cambridge: E and FN Spon. Andrade, C., Rebolledo, N., Tavares, F., Pérez, R. and M. Baz. 2016. Concrete durability of the new Panama Canal: Background and aspects of testing. In Marine concrete structures. Design, durability and performance, ed. M. Alexander, 429–458. Sawston: Woodhead Publishing. ASTM. 2022d. C1202-22e1, Standard test method for electrical indication of concrete’s ability to resist chloride ion penetration. West Conshohocken: ASTM International. Ballim, Y. 1991. A low cost falling head permeameter for measuring concrete gas permeability. Concrete/Beton, Journal of the Concrete Society of Southern Africa 61: 13–18. Bamforth, P. 1994. Admitting that chlorides are admitted. Concrete 28, 6: 18–21. Bamforth, P. 1999. Double standards in design. Concrete 33, 3: 33–35. Baroghel-Bouny, V. 2004. Durability indicators: A basic tool for performance-based evaluation and prediction of RC durability. In Proceedings of the International Seminar on Durability and Life Cycle Evaluation of Concrete. ed. R. Sato, Y. Fujimoto, and T. Dohi, 13–22. Higashi-Hiroshima: Research Center of Hiroshima University. Bickley, J., Hooton, R. and K. Hover, K. 2006. Performance specifications for durable concrete, current practice and limitations. Concrete International 28, 9: 51–57. Bjegovic, D., Serdar, M., Oslakovic, I., Jacobs, F., Beushausen, H., Andrade, C., Monteiro, A., Paulini, P. and S. Nanukuttan. 2016. Test methods for concrete durability indicators. In Performance-based specifications and control of concrete durability, ed. H. Beushausen and L. Fernandez Luco, 51–105. RILEM State-of-the-Art reports. Vol. 18. Dordrecht: Springer. Breitenbucher, R., Gehlan, C., Schiessl. P., van den Hoonaard, J. and T. Siemes, 1999. Service life design for the Western Scheldt Tunnel. In Proceedings of the 8th International Conference, Durability of Building Materials and Components, ed. M. Lacasse and D. Vanier, 3–15. Ottawa: NRC Research Press. BSI. 2015. BS 7543. A guide to the durability of buildings and building elements, products and components. London: British Standards Institution. CEN. 2021b. EN 206:2013 +A2:2021. Concrete – Specification, performance, production and conformity. Brussels: Comité Européen de Normalisation. CEN TC104. 2023. Background of exposure resistance classes (ERC) concept in EN 206-1. (Report in development). Brussels: Comité Européen de Normalisation. Concrete Society. 1996. Developments in durability design and performance-based specification of concrete. Special Publication CS109. London: Concrete Society. Council for Codes of Practice for Buildings. 1950. British standard code of practice, CP3 Chapter IX – Durability. London: British Standards Institution. COWI. 2000. General guidelines for durability design and redesign: DuraCrete, probabilistic performance based durability design of concrete structures. Gouda: CUR Civieltechnisch Centrum Uitvoering Research en Regelgeving. Dhir, R., Jones, M. and E. Ahmed. 1991. Concrete durability: Estimate of chloride concentration during design life. Magazine of Concrete Research 43, 154: 37–44.

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Fédération Internationale du Béton. 2006. fib Model code for service life design. fib Bulletin 34. Lausanne: Fédération Internationale du Béton. Geiker, M., Hendriks, M. and B. Elsener. 2021. Durability-based design: The European perspective. Sustainable and Resilient Infrastructure 8, 2: 169–184. Harrison, T. 2019. Specifying resistance against carbonation-induced corrosion by performance. Magazine of Concrete Research 71, 7: 341–348. Heath, K. 2016. Marinas in the Arabian Gulf region. In Marine concrete structures. Design, durability and performance, ed. M. Alexander, 215–240. Sawston: Woodhead Publishing. Helland, S. 2013. Design for service life: Implementation of fib model code 2010 rules in the operational code ISO 16204. Structural Concrete 14, 1: 10–18. ISO. 2008. ISO 13823-1, General principles on the design of structures for durability. Geneva: International Standards Organisation. ISO. 2011. ISO 15686-1, Buildings and constructed assets – Service life planning – Part 1: General principles and framework. Geneva: International Standards Organisation. ISO. 2015. ISO 2394, General principles on reliability for structures. Geneva: International Standards Organisation. Knights, J. 2019. Performance-based durability design for concrete. Concrete Society Report CS176. Camberley: Concrete Society. Li, K., Li, Q. and Z. Fan. 2016. Hong Kong-Zhuhai-Macau sea link project, China. In Marine concrete structures. Design, durability and performance, ed. M. Alexander, 339–370. Sawston: Woodhead Publishing. Masters, L. and E. Brandt. – editors. 1989. Systematic methodology for service life prediction of building materials and components. Materials and Structures 22, 5: 385–392. Muigai, R., Moyo, P., Alexander, M. and H. Beushausen. 2009. Application of durability indexes in probabilistic modelling of chloride ingress in RC members. In Second International RILEM Workshop on Concrete Durability and Service Life Planning – ConcreteLife’09, ed. K. Kovler, 408–415. Paris: RILEM Publications. Otieno, M. and M. Alexander. 2015. Chloride conductivity testing of concrete – past and recent developments. Journal of the South African Institution of Civil Engineers 57, 4: 55–64. PIARC. 2003. A fact finding review of performance specifications in 2002. Technical Committee 7/8 Road Pavements. Paris: Permanent International Association of Road Congresses report. Richardson, M. 1988. Carbonation of reinforced concrete: Its causes and management. Dublin: CITIS Ltd. 118–121. Rostam, S. 1984. Durability of concrete structures. Bulletin No.152. 415–432. Lausanne: Comite Euro-International du Beton. SABS. 2015. SANS 3001-CO3-1, Civil Engineering test methods: Concrete durability testing index – Preparation of test specimens. Pretoria: South African Bureau of Standards. SABS. 2021. SANS 3001-CO3-3, Civil Engineering test methods: Concrete durability testing index – Chloride conductivity test. Pretoria: South African Bureau of Standards. SABS. 2022. SANS 3001-CO3-2, Civil Engineering test methods: Concrete durability testing index – Oxygen permeability test. Pretoria: South African Bureau of Standards.

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Salvoldi, B., Beushausen, H. and M. Alexander. 2015. Oxygen permeability of concrete and its relation to carbonation. Construction and Building Materials 85: 30–37. Sarja, A. and E. Vesikari. 1996. Durability design of concrete structures. RILEM Report 14. London: CRC Press. Schiessl. P., Hergenroder, M., Moller, J., Nilsson, L-O. and A. Siemes. 1997. Durability design – performance and reliability based design of concrete structures. Bulletin No.238. Lausanne: Comite Euro-International du Beton. Sentler, L. 1983. Service life prediction of concrete structures. Report TVBK-3018. Lund: Department of Structural Engineering. Lund University of Technology. Siemes, A. and S. Rostam. 1996. Durability, safety and serviceability – a performance based design format. In Proceedings IABSE Colloquium. Basis of Design and Actions on Structures – Background and Application of Eurocode 1. Delft. IABSE Report 74: 41-50. Zurich: IABSE. Siemes, A., Vrouwenvelder, A. and A. van den Beukel. 1985. Durability of buildings: A reliability analysis. Heron 30, 3: 3–48. SNV, Swiss Association for Standardisation. 2013. SN 505 262/1 – SIA 262/1, Concrete structures – supplementary specifications. Zurich: Schweizerischer Ingenieur- und Architektenverein. Somerville, G. 1986. The design life of concrete structures. The Structural Engineer 64A, 2: 60–71. Spekkink, D. 2005. Performance based design of buildings. PeBBu Domain 3 Final Report. Rotterdam: CIB (PeBBu). Streicher, P. and M. Alexander. 1995. A chloride conduction test for concrete. Cement and Concrete Research 25, 6: 1284–1294. Tanesi, J., Da Silva, M., Gomes, V. and G. Camarini. 2010. From prescription to performance: international trends in concrete specifications and the Brazilian perspective. IBRACON Structures and Materials Journal 3, 4: 420–431. Thiery, M., Baroghel-Bouny, V., Villain, G. and P. Dangla. 2006. Numerical modeling of concrete carbonation based on durability indicators. In Proceedings of 7th CANMET/ACI Conference on Durability of Concrete, ed. V. Malhotra, 765–780. ACI SP-234-48. Farmington Hills: American Concrete Institute. Torrent, R. 1992. A two-chamber vacuum cell for measuring the coefficient of permeability to air of the concrete cover on site. Materials and Structures 25, 6: 358–365. Torrent, R., Denarié, E., Jacobs, F., Leemann, A. and T. Teruzzi. 2012. Specification and site control of the permeability of the concrete cover: The Swiss approach. Materials and Corrosion 63, 12: 1127–1133. Tuutti, K. 1982. Corrosion of steel in concrete. Stockholm: Swedish Cement and Concrete Research Institute. von Greve-Dierfeld, S. and C. Gehlen. 2016a. Performance-based durability design, carbonation part 1 – benchmarking of European present design rules. Structural Concrete 17, 3: 309–328. von Greve-Dierfeld, S. and C. Gehlen. 2016b. Performance-based durability design, carbonation part 2 – classification of concrete. Structural Concrete 17, 4: 523–532. von Greve-Dierfeld, S. and C. Gehlen. 2016c. Performance-based durability design, carbonation part 3 – partial safety factor (PSF) approach and a proposal for revision of deemed-to-satisfy rules. Structural Concrete 17, 5: 718–728.

Chapter 4

Permeability of traditional and innovative concretes

PERMEABILITY AND POROSITY The durability of concrete is essentially influenced by processes that involve the passage of ions, liquids, and gases into or through the material. The service life will depend on the rate at which aggressive species may move through the concrete. The passage of these potentially aggressive agencies is primarily influenced by the permeability of the concrete. Permeability to a given agent is a function of the pore size distribution, the degree of interconnection of the pore structure, and the moisture content of the permeable pore structure. The diameters of most ions and gas molecules are smaller than the pores in concrete, so even the highest-quality concrete will be permeable to some extent. Although permeability is intrinsically related to the pore structure, there is a world of a difference between “permeability” and “porosity”. In respect of durable concrete, impermeability is the important factor, irrespective of the level of porosity. Porosity is the ratio of the volume of voids to the total volume. Permeability may be defined as the ease with which an ion, molecule or fluid may move through the concrete. A permeable material must always include a pore network and permeability is a function of the tortuosity of the interconnected pore structure. However, permeability is not directly proportional to porosity. For example, a large cube of material with just one small-diameter hole drilled all the way through it has a low porosity but is very highly permeable. The moisture state of the pores is a very important factor and can either facilitate or reduce the rate of passage. Pores that are water-filled reduce the permeability to gases, such as carbon dioxide and oxygen, but may facilitate higher rates of ionic diffusion, such as in chloride environments. The processes involved in fluid and ion migration include the distinct mechanisms of capillary attraction, flow under a pressure gradient and flow under a concentration gradient. These mechanisms are characterised by the material properties of sorptivity, permeability and diffusivity, respectively. The term “concrete permeability” has often been popularly used, incorrectly, in an DOI: 10.1201/9781003261414-4

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all-embracing manner to refer to material properties that influence the rate of ingress of ions, gases and liquids, with or without a pressure differential. Probably the most important distinction for concrete practitioners is to distinguish “diffusivity” from “permeability”. We do not want to annoy our scientific colleagues by being as imprecise in our terminology as many journalists, who annoy us in their writing about “cement buildings”!

OVERVIEW OF PERMEABILITY SIGNIFICANCE IN THE CONTEXT OF CONCRETE DURABILITY The relationship between permeability and concrete durability is apparent from a brief consideration of the phenomena that negatively impact service life. The most common problem is corrosion of reinforcement. Corrosion is preceded by depassivation of the reinforcement. This may be caused by carbonation or chloride ingress. Carbonation rates are a function of both physical and chemical phenomena, but clearly the ease of ingress of carbon dioxide is a key feature. Transfer of ions through the concrete is also a rate-controlling feature. Depassivation due to chloride ingress is caused by the buildup of chlorides to a critical level, which is related to the ease of ingress of chloride ions from external sources. Once initiated, the rate of corrosion propagation is related to ease of ingress of moisture and oxygen. Sulfate attack is caused by the ingress of sulfate ions, typically from groundwater. Deterioration by freeze-thaw behaviour is a function of the number of freeze-thaw cycles and is related to flow of water and its distribution within the pore structure of the concrete. An alkali-silica reaction (ASR) can occur in many concretes but it becomes deleterious ASR, and thereby a durability failure, only when sufficient moisture can be imbibed from the permeable structure to cause the production of gel in amounts which cause damaging expansion.

INFLUENCE OF WATER/CEMENT RATIO AND CURING REGIME The extent to which the surface and cover zone of a concrete member is permeable derives from a number of factors. The most important of these, assuming well-proportioned materials and good compaction, are the water/ cement ratio and the degree of hydration. The permeability of a concrete will be predominantly influenced by the permeability of the cement paste fraction, especially in the cover concrete and at the interface with aggregate particles – the interfacial transition zone (ITZ). It is the reaction between the unhydrated cement grains and the water that predominantly influences the permeability, through the production of a gel which is particulate in nature and has its own pore structure. The aggregate is usually

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impermeable or, importantly in the case of porous aggregate, is surrounded by a layer of gel that is sufficiently impermeable to isolate the aggregate from the permeable network. Hardened concrete may be considered as a three-phase material consisting of solid matter, water and air. The relative proportions of each phase depend on the age of the concrete and the nature of the environment to which the element is exposed. The water-filled spaces found in fresh cement paste become filled, partially or totally, by hydration products. The extent to which the pores are filled depends on the total water content relative to that needed for full hydration of the cementitious material. During hydration only a proportion of the water becomes chemically bound, leaving a quantity of free water within the concrete that creates the canal-like network. Some of this water may evaporate after the concrete has hardened, leaving a pore structure that is both interconnected and accessible from the atmosphere. The production of sufficiently impermeable concrete is made possible by the fact that the gel formed during hydration has a greater bulk volume than the parent cement grains and thereby has a blocking effect in reducing the diameter of pores over time. Early concrete practice involved coarse cement grains which hydrated for several months and helped to reduce permeability on an on-going basis. Modern cements are ground more finely to provide early strength through a higher surface area for reaction sites. The hydration reaction is thereby quicker, but the reactive components exhaust themselves within the first fortnight or so, at which point the enduring permeable pore structure is well established. An excellent insight into the process of hydration may be gained from the classic work of Powers (1958). The unhydrated cement grains may be assumed, in a slight simplification, to be formed of silicates, aluminates, and aluminoferrites. The dominant compounds are tricalcium silicates (3CaO.SiO2) and dicalcium silicates (2CaO.SiO2). These silicates are predominantly granular in nature. The silicates, on hydration, produce calcium silicate hydrates (for example 3CaO.2SiO2.3H2O) and calcium hydroxide. The calcium silicate hydrates have varying physical properties but it is now thought that most of them are fibrous in nature with straight edges and lengths up to ten times their width. The calcium hydroxide is more clearly crystalline. The calcium silicate hydrate crystals interlock and form both physical and chemical bonds. It is thought that the physical bonds are more significant in giving concrete its structural properties. The calcium silicate hydrate crystals are so small that the product is regarded as a gel. Crosslinking of fibres leads to a particulate network with interstitial spaces. The clusters of gel particles will have spaces within them known as gel pores. Gel pores, sometimes characterised as “micropores”, exist as interlayer spaces between the calcium silicate hydrate sheets. The gel pores occupy about one third of the gel volume. Larger spaces are formed by the boundaries of the clusters, and these are known as the capillary pores (Fig. 4.1). The capillary pore structure is particularly significant in the context of durability. Capillary

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Figure 4.1 Representation of the pore structure in concrete.

pores may be described as the space originally occupied by the mix water. Hydration of cement grains can continue in the water-filled capillary pores. The resulting gel will be porous, but it forms at the expense of the capillary pore volume and, therefore, the overall effect is one of reduction in pore volume. This is because hydrated grains occupy about twice the space of unhydrated grains. Clearly the state of the capillary pore structure at the end of the hydration stage is critically dependent on the water/cement ratio and the quality of curing. Fig. 4.2 indicates schematically the influence of water/cement ratio and curing. The graph shows the growth in gel at the expense of water during hydration. It also shows that at higher water/cement ratios the remaining free water creates a large capillary space. In the case of poor curing, the excessive amount of mixing water that evaporates, especially in the cover zone, deprives the unhydrated cement grains of the ability to react and therefore the volume of gel produced is reduced.

Figure 4.2 Influence of water/cement ratio and quality of curing on the pore structure of concrete.

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Poorly cured, high water/cement ratio concretes are susceptible to durability failure due to the resulting high porosity and high permeability. The production of durable concrete therefore requires: • the water/cement ratio to be low enough to limit the initial volume of the capillary pore network produced by the mix water • the water/cement ratio to be high enough to provide a water-filled capillary pore network with an initial volume at least twice that of the unhydrated cement • the curing regime being such that the capillary network remains water-filled long enough to ensure that the hydration process is not stopped through lack of fuel for the reaction. In addition to gel pores and capillary pores, voids caused by entrapped or entrained air will be found in hardened cement paste. The relative size of these pores is represented graphically in Fig. 4.3 based on studies by researchers including Setzer (Comite Euro-International du Beton 1989) and Mehta (1993). Gel pore diameters are of the order of nanometres, and range from 0.001 μm to 0.008 μm. The permeability of such pore diameters to potentially aggressive species is not significant. In theory, all capillary voids could be filled in a concrete with a very low water/cement ratio of 0.38. In practice, full hydration is rarely achieved and, in any case, workability requirements demand higher water/cement ratios. Thus it is inevitable that a capillary network will exist. The challenge is to produce a concrete that has the least capillary network for a given workability. The achievement of capillary pore diameters in the range 0.01 μm to 5 μm is possible in well-hydrated, low water/cement ratio concretes, but values can reach 50 μm in lesser-quality concretes. Air voids are formed by entrapped or entrained air and pore diameters range from 100 μm (entrained) to 2 mm (entrapped). The size and distribution of entrained air depends on the use of air-entraining admixtures. Entrapped air should be released during compaction, but clearly, some entrapped air remains, particularly on inclined shuttered surfaces.

Figure 4.3 Classification of the pore structure in concrete by pore radius.

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Voids may also be found in aggregates. Aggregate voids contribute to the porosity of hardened concrete. Normal weight aggregates have porosities in the typical range of zero to 20%. Lightweight aggregate voids may total 50%. The aggregate voids are, however, isolated from the permeability network if a satisfactorily cured and compacted cement paste surrounds each aggregate particle. The capillary pores, initially filled with water, accommodate the hydration products of the chemical reaction. These products occupy double the space of the original solid constituents being hydrated. Clearly, therefore, water/ cement ratio is a critical parameter in the production of low-permeability concrete. In theory, all capillary voids could be filled in a concrete with a very low water/cement ratio. In practice, full hydration is rarely achieved and in any case consistence requirements demand realistic water/cement ratios. The challenge is to produce a concrete that has the least capillary network for a given consistence. The significance of the water/cement ratio is illustrated by the trend shown in Fig. 4.4. It may be noted that exceeding a ratio of 0.6 can lead to an exponential rise in permeability. Optimising total water content for hydration and ease of placing fresh concrete leads to a water/cement ratio in the range 0.4 to 0.5. Regarding a lower limit, Neville (1995) demonstrates that below a water/cement ratio of 0.38 the capillary pore volume would be insufficient to allow complete hydration of the cement grains. As the water/ cement ratio increases above 0.38, the space available becomes progressively greater than that required. However, the production of impermeable concrete does not demand that the capillary network become completely filled by gel during hydration – it is sufficient if the gel partially fills the network in a manner which makes the capillary pore network discontinuous or tortuous. At the upper end of the range, the use of a water/cement ratio greater than 0.65 is unlikely to produce a pore network of acceptable impermeability for durable concrete, especially reinforced concrete.

Figure 4.4 Influence of water/cement ratio on permeability.

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Moving from pore structure to permeability, the capillary pore space will have a fluctuating moisture level in service dependent on the environment, and this will also influence permeability and the rate of ingress of aggressive species. The relative amounts of water and air remaining in the pore structure of hardened concrete depend on the temperature and relative humidity of the surrounding atmosphere. If the atmospheric conditions are cyclical so too will be the permeability of the concrete.

TRANSPORT PROCESSES The transport processes involved in the passage of potentially harmful agencies through concrete are: • • • • •

gaseous diffusion (oxygen, carbon dioxide) vapour diffusion (moisture movement) ionic diffusion (chlorides, sulfates) absorption and capillary rise (chlorides dissolved in water) pressure-induced flow (aggressive groundwater, freeze-thaw).

In certain mechanisms, for example, carbonation and sulfate attack, the rate of ingress is also influenced by an associated chemical reaction. The phenomenon of adsorption should also be mentioned. It is a phenomenon in which molecules from one material may adhere to the surface of another. Adsorption is relevant to concrete technology, for example in the case of gas molecules adhering to the solid surface of the pore walls in concrete by Van der Waals forces or chemical bonds. This is due to the surface energy: the molecules at the pore walls have excess energy because they are not surrounded. The surface area of a given pore volume increases as the pore radius decreases leading to a consequent increase in water adsorption. Thus the relative humidity of concrete can be higher than the surrounding environment. The relative humidity of the pore structure can be a significant factor in the rate of transport of other species in concrete. In most concretes, gaseous and ionic diffusion will be the critical mechanisms. Gaseous diffusion is rapidly accommodated in unsaturated concrete and ionic diffusion occurs in saturated conditions. Therefore, concretes subject to cyclical environments are often the ones that deteriorate most rapidly. Examples include the tidal zones of coastal structures, and structures subject to wetting and drying. The former allows continual ingress of chlorides by ionic diffusion and by absorption, leading to rapid depassivation of reinforcement. Concurrently, access for moisture and oxygen promotes corrosion following depassivation. Structures subject to wetting and drying will carbonate as the relative humidity falls and, following depassivation, will experience reinforcement corrosion as the relative humidity rises.

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MODELLING TRANSPORT RATES The principal transport processes in concrete are gaseous diffusion, ionic diffusion, absorption, and liquid flow under pressure. It is useful to review the equations governing the rate of transport for each of these processes before seeing how they form the starting point on which mathematical models are built of the durability degradation mechanisms detailed in the chapters that follow.

Gaseous diffusion Gaseous diffusion is a transport process by which a flow of matter is observed to occur under the influence of a concentration difference. The actual transport mechanism is by random molecular motion, with diffusing molecules moving independently of one another. These molecules constantly undergo collisions with the molecules of the material through which they are diffusing and, although the motion is random, an average fraction of diffusing molecules move past a given section at any one time. A higher number of molecules move from an area of high concentration to an area of low concentration than vice versa. The deterioration models for concrete where gaseous diffusion is involved generally include Fick’s first law of diffusion. This law may be used to describe the rate of diffusion of a gas into a uniformly permeable material:

J=

D

dc dx

where J = mass transport rate (g/m2s) D = diffusion coefficient (m2/s) dc/dx = concentration gradient (g/m4) x = distance (m) The moisture content of concrete greatly influences the rate of diffusion. The diffusion rate of gases such as oxygen and carbon dioxide is 104 to 105 times greater in the gas phase than in solution.

Ionic diffusion Ionic diffusion requires a concentration difference and saturated conditions. Ions, such as chlorides and sulfates, exploit free water as a passage for ionic diffusion. Ionic diffusion may be represented through Fick’s second law of diffusion and, in certain conditions, by the error function solution of Crank (1975):

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2C C =D 2 t x Cx x = 1 erf C0 2(D t)0.5

where Cx = concentration of species at depth x (x in metres) C0 = concentration of species at the surface D = diffusion coefficient (m2/s) t = time of exposure (s) erf = error function A common use of the formula is in chloride profile analysis, in which instance the concentrations are often expressed as percentage chloride by mass of cement. The error function is available in references such as Crank (1975) or Lawrence (1981). It may also be found in commercial spreadsheet packages. One problem in using spreadsheet packages is that some find difficulty in handling problems with parameters of very diverse magnitudes. The effective diffusion coefficient may typically be of the order 10–12 m2/s, whereas time, for example, could be 109 seconds (approximately 30 years). This problem may be overcome by first evaluating the coefficient using the value in years and then converting to the required units.

Absorption Absorption is a process in which fluid is taken into the spaces within a material. It involves the ingress of a fluid, such as water, by capillary action. It is related to the pore structure of the sample, but not to its permeability. Absorption may be significant in concrete subject to freeze-thaw cycles or in situations where potentially harmful agents, such as chlorides, are dissolved in the water that has cyclical access to the member. The transport of liquid by capillary rise is caused by the pressure differential across the meniscus. Capillary rise is characterised by the equation of Washburn (1921), but in concrete technology, it is often adequate to determine the “sorptivity” of a sample. Sorptivity may be defined by the following relationship:

V = St 0.5 A where V = volume of material absorbed in time t (mm3) A = cross sectional area of sample in contact with water (mm2) S = “sorptivity” (mm/min0.5) t = time (min)

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Liquid flow under pressure Pressure induced flow is often used in test methods to characterise the permeability of a material. Its relevance to concrete in service is limited to cases such as that where the element is under a hydrostatic head, such as groundwater flow. Liquid ingress under a pressure head may be modelled empirically by D’Arcy’s Law and the D’Arcy coefficient may be converted to the intrinsic permeability by a conversion factor. The following relationship results: Q = A

k

g

h L

where Q = flow rate (m3/s) A = cross-sectional area (m2) k = intrinsic permeability (m2) ρ = density of fluid (kg/m3) g = acceleration due to gravity (m/s2) μ = viscosity of fluid (Ns/m2) Δh = head loss across the sample (m) L = thickness of sample (m) Intrinsic permeability is a property of the material independent of the fluid concerned. In the case of concrete, the coefficient of permeability based on water flow (at room temperature) is often quoted as a measure of its permeability:

Q = A

K

h L

where K = coefficient of water permeability according to D’Arcy (m/s)

INFLUENCE OF CEMENT TYPE ON PERMEABILITY OF CONCRETE The development of new cement types with reduced clinker content fit nicely into the “sufficiency” concept described in Chapter 1. Must every structural element forming the internal and external elements of a building structure be made of the same concrete? A review of the exposure classification tables in Chapter 2 would suggest not. Must every structural element in a building structure be specified to reach its characteristic strength at an age roughly equivalent to the lunar cycle? Not if we decide

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otherwise, even if it means leaving formwork and props in place during construction for longer periods. Taking these two points together prompts us in the direction of reducing the carbon footprint of our cements by matching durability and strength capacity to the exposure and structural demands of each element in service, taking account of construction constraints. In this way we can supply just what is needed – no less, no more. This approach is greatly facilitated by the wide range of certified cements already available and also in development. The approach to clinker reduction by substitution has not only reduced the carbon footprint, but has also contributed to reducing permeability through the incorporation of constituents that promote ongoing hydration and/or are of nano scale. The shift in concrete technology focus from strength alone to also including durability and sustainability has led to diversification in certified cement types. Many of these cements confer durability enhancements over Portland cement concretes. Currently, there are more than 30 distinct cements recognised in Europe alone, but the use of Portland cement concrete will continue in many parts of the world for some time to come, due to economic and supply chain issues. Meanwhile, the current drive to continually reduce clinker contents will plateau and then reverse when clinker becomes an attractive low-carbon option. This will come about as a result of clinker production becoming less carbon intensive, combined with a reduction in the availability of secondary cementitious materials. The supply of pulverised fuel ash, for example, will decrease with the enforced closure of coal-burning power stations while supply of slag will reduce as steel plants decarbonise primary production. The readily controllable aspects that can have the greatest influence on permeability of Portland cement concretes are mix design, compaction, and curing. Irrespective of cement type, concrete mix design should aim to produce the lowest possible water demand consistent with durability and workability requirements. Dewar (1985) has shown that consideration of mean particle size, voids ratio diagrams, and particle interference effects can be used in mix design to determine the minimum water demand required. Software is now available to assist in this process. The water/cement ratio should be as low as possible, consistent with consistence requirements. Some cement types aid the rheology of fresh concrete, and there is significant experience in the use of water-reducing admixtures. Excessive permeability resulting from poor site practice cannot be compensated for by choice of cement type. Proper compaction is essential in the production of impermeable concrete. Entrapped air bubbles may have relatively large diameters and would therefore provide linkage between otherwise discrete capillary pores. Poor curing is probably the greatest single cause of permeable concrete. The quality of the cover zone is as important as its depth in terms of the defence mechanism against durability threats. The quality of the cover zone is critically influenced by the curing regime. Inadequate curing prohibits the required blocking of

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capillary pores by the products of hydration. Vigilance on-site is a given when casting concrete in extremes of hot and cold weather, but must not be ignored even in temperate climates, despite the forgiving nature of the material. Mr. Bill Fleeton, an expert in concrete repair, reflecting on many years of practice, once commented, “We never repaired poor concrete. We repaired poor workmanship”! The dominant range of cement compositions used globally are presented in Tables 4.1 to 4.4 based on a selection of European and American standards: EN197-1 (CEN 2011b), EN197-5 (CEN 2021a), C150 (ASTM 2022a), C595 (ASTM 2021b), and C989 (ASTM 2022c). The primary constituents are denoted as follows: clinker (K), silica fume (D), limestone, total organic carbon content < 0.50% by mass (L) and < 0.20% by mass (LL), natural

Table 4.1 Examples of Portland cement compositions Descriptor Portland cement

Designator and permitted composition EN 197-1: CEM I, 95–100% K ASTM C150 Type I, 95–100% K, ≤ 5% L, ≤ 5% inorganic, ≤ 1% organic

Table 4.2 Examples of Portland-Slag, -Fly Ash, -Pozzolans, -Limestone and –Silica Fume cement compositions Descriptor Portland-Slag

Portland-Fly Ash

Portland-Pozzolana

Portland-Limestone

Portland-Silica Fume

Designator and permitted composition EN 197-1: CEM II/A-S, 80–94% K, 6–20% S EN 197-1: CEM II/B-S, 65–79% K, 21–35% S ASTM C595, IS(x), 5–94% K, x = 6–95% S EN 197-1: CEM II/A-V, 80–94% K, 6–20% V EN 197-1: CEM II/A-W, 80–94% K, 6–20% W EN 197-1: CEM II/B-V, 65–79% K, 21–35% V EN 197-1: CEM II/B-W, 65–79% K, 21–35% W EN 197-1: CEM II/A-P, 80–94% K, 6–20% P EN 197-1: CEM II/A-Q, 80–94% K, 6–20% Q ASTM C595 IP(x), 60–94% K, x = 6–40% P EN 197-1: CEM II/B-P, 65–79% K, 21–35% P EN 197-1: CEM II/B-Q, 65–79% K, 21–35% Q EN 197-1: CEM II/A-L, 80–94% K, 6–20% L EN 197-1: CEM II/A-LL, 80–94% K, 6–20% LL ASTM C595, IL(x), 85–94% K, x = 6–15% L EN 197-1: CEM II/B-L, 65–79% K, 21–35% L EN 197-1: CEM II/B-LL, 65–79% K, 21–35% LL EN 197-1: CEM II/A-D, 90–94% K, 6–10% D ASTM C595 IP(x), approximately 85–94% K, since typically x = 6–15% P

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Table 4.3 Examples of Portland-composite cement compositions Descriptor Portland-Composite

Designator and permitted composition EN 197-1: CEM II/A-M (x-y-z), 80–88% K and a 12–20% combination of other constituents (x, y, z) drawn from D (max. 10%), L or LL, S, V, W, P, Q, and T EN 197-1: CEM II/B-M (x-y-z), 65–79% K and a 21–35% combination of other constituents (x, y, z) drawn from D (max. 10%), L or LL, S, V, W, P, Q, and T ASTM C595 IT (Lx)(Py), 70–88% K, and a 12–30% combination of other constituents drawn from L (x = 6–15%) and P (y = 6–15%) where x ≥ y ASTM C595 IT (Lx)(Sy), 70–88% K, and a 12–30% combination of other constituents drawn from L (x = 6–15%) and S (y = 6–15%) where x ≥ y ASTM C595 IT (Px)(Ly), 45–87% K, and a 13–55% combination of other constituents drawn from P (x = 7–40%) and L (y = 6–15%) where x > y ASTM C595 IT (Px)(Py), 60–87% K, and a 13–40% combination of other constituents drawn from two pozzolans P1 (x = 7–34%) and P2 (y = 6–19%) where x > y and x+y ≤ 40% ASTM C595 IT (Px)(Sy), 30–87% K, and a 13–70% combination of other constituents drawn from P (x = 7–40%) and S (y = 6–30%) where x ≥ y ASTM C595 IT (Sx)(Ly), 30–87% K, and a 13–70% combination of other constituents drawn from S (x = 7–64%) and L (y = 6–15%) where x > y ASTM C595 IT (Sx)(Py), 30–87% K, and a 13–70% combination of other constituents drawn from S (x = 7–36%) and P (y = 6–34%) where x > y EN 197-5: CEM II/C-M (x-y), 50–64% K and a 36–50% combination of two other constituents (x, y) drawn from D (6–10%), L or LL (6–20%), S, V, W, P, Q, and T

pozzolana (P), natural calcined pozzolana (Q), slag (S), burnt shale (T), siliceous fly ash (V), and calcareous fly ash (W). The total percentage of components for cements to EN197-1 and EN 197-5 includes allowance for minor additional constituents to a maximum of 5%. As may be noted, the list is long because cement has always been tailored to local manufacturing conditions and industry requirements. Over time, the dominance of Portland cement has given way to blended and slag cements, which are manufactured from several main constituents other than Portland cement clinker. Most popular among these are limestone cements, fly ash cements and ground-granulated blastfurnace slag (ggbs) cements. Together, these blends currently represent about 60% of the global market, twice that of Portland cement. However, “necessity is the mother of invention”, and new ternary cements are being patented to further enhance the sustainability of concrete by reducing the clinker content to 35%.

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Table 4.4 Examples of Blastfurnace, Pozzolanic and Composite cement compositions Descriptor Blast furnace cement

Pozzolanic cement

Composite cement

Designator and permitted composition EN 197-1: CEM III/A, 35–64% K, 36–65% S EN 197-1: CEM III/B, 20–34% K, 66–80% S EN 197-1: CEM III/C, 5–19% K, 81–95% S A comparable cement is ASTM C595, IS(x) with high levels of ASTM C989 slag cement EN 197-1: CEM IV/A, 65–89% K and 11–35% combination of other constituents drawn from D (max 10%), P, Q, V, W EN 197-1: CEM IV/B, 45–64% K and 36–55% combination of other constituents drawn from D (max 10%), P, Q, V, W EN 197-5: CEM VI (S-P), 35–49% K, 31–59% S, 6–20% P EN 197-5: CEM VI (S-V), 35–49% K, 31–59% S, 6–20% V EN 197-5: CEM VI (S-L), 35–49% K, 31–59% S, 6–20% L EN 197-5: CEM VI (S-LL), 35–49% K, 31–59% S, 6–20% LL EN 197-1: CEM V/A, 40–64% K, 18–30% S and 18–30% combination of other constituents drawn from P, Q, V EN 197-1: CEM V/B, 20–38% K, 31–49% S and 31–49% combination of other constituents drawn from P, Q, V

Portland cement concrete The permeability of Portland-cement concrete is at the mercy of the standard of workmanship. The hydration of Portland cement produces an abundance of calcium hydroxide in addition to the calcium silicate hydrates. Thereby inherently permeable, the material requires a low water/cement ratio and careful curing to combine a sufficiently low level of initial porosity with sufficient production of hydration products after early hardening to block and disconnect the permeable pore network. Unfortunately, the uncontrolled addition of water on-site to boost shortterm consistence is a natural inclination of those unaware of its detrimental impact on long-term durability. The humorous phrase “what did posterity ever do for me?” comes to mind! Curing of concrete in temperate climates is often a matter of chance. Curing is conducted with due care and attention in conditions of extreme heat or cold, but complacency may arise in temperate zones where the atmospheric temperature and relative humidity will often combine to successfully cure a forgiving concrete. However, for our clients, “often” is not enough, and this will inevitably lead to an occasional durability failure if conditions are not benign. For example, even a moderate breeze blowing over an unprotected slab can rob the cover zone of the water needed for full hydration, irrespective of a prevailing moist atmosphere.

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Portland-slag and blastfurnace cement concretes The introduction to practice of blended cements provided a win-win-win opportunity to address the inherent impermeability of Portland cement concrete while decreasing the carbon footprint of the construction industry and recycle a waste material. One of the most impressive developments in this regard has been the incorporation of blastfurnace slag in concrete. The ASTM International standard specification for slag is C989 (ASTM 2022c), and in Europe it is EN 15167-1 (CEN 2006a). Blastfurnace slag is a by-product of the steel industry. The blastfurnace process produces iron and an accumulation of waste materials in the form of a molten slag. The molten slag is quenched by high pressure cold water spray to produce a granulated material (gbs), somewhat like sand. Suitability for use in cement occurs for slags with two-thirds by mass of a combination of calcium oxide (CaO), magnesium oxide (MgO), and silicon dioxide (SiO2), where the sum of the mass of CaO and MgO exceeds that of SiO2. This material is finely ground to produce ground granulated blastfurnace slag (ggbs). The finely ground material is a naturally cementitious hydraulic material. It has the benefit of continuing to produce hydration products over a more extended period than current Portland cements. This continuing process, long after the early hardening of concrete, can significantly act in a pore blocking capacity to greatly reduce permeability. In particular, the ingress of chloride ions is hampered, which delays or eliminates the onset of corrosion initiation. If corrosion occurs, its rate is lowered by reduced permeability to oxygen and moisture from the atmosphere. An important consideration, however, is the quality of curing in the first weeks after casting. It must be of a high standard to ensure ongoing hydration. A threat to the ongoing supply of the material is that of decommissioning blastfurnace foundries on environmental grounds. One way or another, even the current supply of this waste stream from the steel industry can supply only 10% of the global need (O’Riain 2022), and a wide range of existing and future alternatives needs to be employed. One step is the grinding of gbs to such a fine particle size that it behaves as an activator to further encourage hydration.

Portland-fly ash cement concretes The majority of the world’s electricity generating stations are coal-burning. Fly ash is a by-product of pulverized coal burnt in these plants. The flue gases contain ash of varying coarseness. The finest particles are harnessed through stages of mechanical separation and electrostatic precipitation as pulverised fuel ash (PFA), also known as fly ash. Two ash types are distinguished through standard C618 (ASTM 2022b), based on calcium

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oxide content. Class F is made from burning anthracite and/or bituminous coal, resulting in a low CaO content of less than 10% and a silicon dioxide level of about 50%. Class C is produced from burning lignite or sub-bituminous coal, resulting in a CaO content of 10–30% and a silicon dioxide level of about 35%, which confers on it self-cementing properties in addition to pozzolanicity. This practice is mirrored in Europe through standard EN 450 (CEN 2012), which uses the 10% CaO benchmark to distinguish siliceous ash (V) from calcareous (W), with five subclasses based on fineness and loss on ignition. The physical and chemical characteristics of the material qualify it as a useful manmade or artificial pozzolan for use as a cement replacement in concrete. The PFA particles are spherical, and this physical characteristic aids the consistence of fresh concrete, making it easier to place and compact. This characteristic can be used to reduce the water required to make the fresh concrete workable and hence lower the water/cement ratio for a given consistence. Carefully cured fly ash concretes can thereby yield enhanced impermeability compared with Portland cement concrete, through reduced water demand. Chemically, its pozzolanic characteristics are such that it reacts with calcium hydroxide to produce calcium silicate hydrates and calcium aluminates. Porosity is reduced by supplying less excess water over and above that required for full hydration, and permeability is reduced by the conversion of calcium hydroxide to hydrates that act as pore blockers. The current and future supply of certified fly ash to the market is more uncertain than other waste materials. Coal is currently sourced for each power station from a variety of geographical locations, depending on prevailing market conditions. Some burnt coals require greater processing of the ash than others to yield a material suitable for use in concrete. The energy required to assess, process, and certify inferior ash may negate the planned environmental and economic benefits. Looking to the future, the use of PFA cannot meet long term supply chain requirements because the global number of coal-fired power stations will significantly reduce as part of the drive to net-zero carbon emissions by 2050, even if the target of a 90% reduction in coal use is not fully achieved.

Portland-pozzolana and pozzolanic cement concretes Natural pozzolans are sourced from mineral and volcanic deposits. Artificial or manmade pozzolans are produced by heating clay or shale to incorporate oxygen into the finished product – calcined clays or calcined shales – which are further processed to produce fine powders. Metakaolin, for example, is a type of calcined clay derived from very high temperature processing of kaolin. The chemical composition is essentially silicon dioxide

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(SiO2) and aluminium oxide (Al2O3), with at least 25% by mass of the former. When used as fine powders in the production of concrete the pozzolanic materials react with the calcium hydroxide fraction to produce calcium silicate hydrates and other cementitious compounds, which leads to a reduction in permeability. Silica fume, also referred to as “microsilica”, is a by-product of elemental silicon metal or ferrosilicon alloy production in electric arc furnaces. It is captured through filtering of the emitted fumes at the plant. Chemically, the material certified for use in concrete is at least 85% silicon dioxide. Physically, it is non-crystalline and its specific surface is 50─80 times finer than Portland cement. The diameter of most particles is less than 1 μm, one hundred times smaller than the average cement particle. The ASTM International standard specification for silica fume is C1240 (ASTM 2020c), and in Europe is EN 13263 (CEN 2009b). These amorphous characteristics make it highly reactive. It is therefore effective in producing a dense impermeable concrete by altering the microstructure through pozzolanic reaction with free calcium hydroxide in the paste to produce additional hydrates, and by physically filling in the tiny voids between cement grains. Generally, the percentage used in a concrete mix is low, partly due to expense, but also because of the increased water demand of the fine particles, which must be offset by superplasticisers. The trend of reducing clinker content has introduced CEM II/C-M to European practice with clinker contents of 50 – 64%. The balance is made up of two components drawn from those that already have approval for inclusion in cement. Two popular combinations in practice are likely to be slag with limestone and fly ash with limestone, recognising the global limited supply of slag and fly ash. The slight change in mix proportions of these components does not significantly alter the permeability landscape, but strict control of a low water/cement ratio and extended curing to ensure extensive generation of pore-blocking hydration products is absolutely essential for durability.

Portland-limestone cement concrete About 15% of the earth’s crust is limestone, and finely milled limestone may be used as a partial replacement for cement clinker. The ASTM International standard specification for limestone is C1797 (ASTM 2017c). The requirement in Europe is for a calcium carbonate (CaCO3) content of at least 75% by mass. A European standard test method EN 13639 (CEN 2017) distinguishes limestone’s purity in respect of total organic content, which must not exceed 0.50% for type “L” or 0.20% for type “LL”. The powder is primarily used for economic and green credential reasons rather than inferring enhanced properties compared with Portland cement concrete. Nevertheless, its physical and chemical attributes can marginally enhance impermeability if used at modest replacement levels of 5–15%.

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Physically, the fine particles improve the pore size distribution by reducing porosity. This is effective up to a replacement level of about 10%. Chemically, the main contribution is that the fine particles act as nucleation sites. Together with the cement grains, this increases the surfaces available for the precipitation of hydrates, thus enabling pore blocking and disconnection of pores. At replacement levels above 10%, the law of diminishing returns applies, due to the low chemical reactivity of the limestone compared with the Portland cement clinker. By the 15% replacement level, the Portland limestone cement can match the performance of Portland cement if the former has a Blaine fineness about 100 m2/kg above the latter. Above 15% replacement, there is a tradeoff between carbon footprint reduction and performance (Thomas and Hooton 2010, Tennis et al. 2011). As an aside, although limestone is abundant, it might also be possible to manufacture synthetic limestone as part of a process to sequester carbon dioxide. It, therefore, has a significant role to play in future research and development of durable low-carbon construction.

Composite cement concrete Pushing the boundaries of low-carbon cements within the current framework of common cements has seen the introduction of Type CEM VI cements in Europe. Lowering the clinker content to 35–49% is achieved through ternary combinations using slag with natural pozzalana, slag with fly ash, or slag with limestone. The durability properties of concrete made with two such cements were studied by Król et al. (2020) at the 45% clinker level. They found that slag combined with either fly ash or limestone was sufficiently resilient if the water/ cement ratio was low and the curing period was lengthy. This emphasises the basics of impermeable concrete: keep the porosity to a minimum by using as little mixing water as possible, and then cure carefully to let the hydration products grow into the initially interconnected pore structure, transforming it into a tortuous set of narrow laneways and cul-de-sacs.

PERMEABILITY OF INNOVATIVE CONCRETES The pace of innovation in construction is ever-increasing and embracing new technologies, and this has extended to concrete practice. A notable aspect of materials technology has been the exploitation of additives to achieve properties over and above those obtainable with certified cements alone, and exploration of nanotechnology. Several of these are durability-enhancing through decreased permeability. Although most specifiers rely on certified cements, there is no legal impediment to the use of new cements, additions, or minor components in the production of concrete in advance of certification and inclusion in a national standard.

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Reches (2018) reviewed the use of nanoparticles in concrete practice. The addition of nano-SiO2 or nano-clays in the 40–100 nm size range has been particularly effective in reducing permeability to ions and gases. The combined effects of strong pozzolanic reactivity and high surface area has encouraged more complete hydration than in traditional particle size concrete. Additionally, the fineness of the particles has facilitated density of packing. Disadvantages to date include difficulty in overcoming agglomeration to better disperse the particles and high cost. Graphene, a single layer of carbon atoms, can be added as an additive in concrete. Pure crystalline graphene is available as graphene oxide sheets that have the thickness of a single atom. Following on from the pioneering science of Geim and Novoselov (2007), Dimov et al. (2018) researched the application of this technology at an industrial scale by infusing graphene as flakes into the mixing water for concrete. Graphene acts as a surface catalyst to enhance bonding in the hydration reaction between water and cement. Although its primary advantage is to counteract the tensile weakness of concrete, graphene also reduces permeability to moisture by a factor of four. As will be noted in later chapters, this has positive implications in controlling phenomena such as rate of reinforcement corrosion or swelling of alkali silica reaction gel. The term “cementene” has been registered as a trademark (Ricketts 2022). The role of alkali-activated materials, such as slags and ash, has already been described in the context of supplementary cementitious materials in blended Portland cements. However their use as alkali-activated pozzolanic cements in innovative low-carbon concretes is also an option. The term “geopolymer” is often associated with this technology due to its long chains or networks of inorganic molecules. Also, the term “earth-friendly concrete” (EFC) has been registered as a trademark in this field. Specification PAS 8820 (BSI 2016) for the production of an alkali-activated cementitious material (AACM) limits Portland cement to less than 5% by mass of the binder, and subsidiary constituents to less than 25% of the mass of the cementitious material. The silicon and aluminium components are activated by an alkaline solution. Although most durability research attention has focused on their role in eliminating material from concrete that is subject to acid attack or abrasion, it may be envisaged that their application as durable concretes in aggressive chloride environments will be further enhanced by research on chloride ion impermeability (Wong 2022). Calvo et al. (2019) reported the effectiveness of crystalline hydrophilic additives in reducing the water permeability of concrete by half. The hydration reactions in the presence of water produce solid amorphous phases in the concrete bulk with a pore blocking affect which reduces permeability. It can also act as an aid to self-healing in the case of microcracks. A bonus aspect is that the additive allows a slight reduction in the water/cement ratio for similar consistence. The American Concrete Institute Committee Report 212 3R-16 (ACI 2016) includes advice on permeability reducing admixtures for hydrostatic and non-hydrostatic conditions.

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Regarding self-healing, Seymour et al. (2023) postulated that “hot mixing” could be used to encapsulate aggregate-sized lumps of lime in concrete to aid durability. These could then be called upon as sources of reactive calcium to progressively block pores and fill cracks (self-healing) over a highly extended service life. Their hypothesis was based on an examination of the source, chemistry, and behaviour of relict lime clasts in concrete from the ancient walls of Privernum, built during the Roman Empire, which have survived for millenia. “Hot mixing” is a technique whereby quicklime, CaO, is used in the production of fresh concrete instead of slaked lime, Ca(OH)2. The reaction can lead to local hotspots of about 200oC, forming calcium-rich clasts. These may later react with water flowing through the permeable network to form calcium alumino-silicate hydrates, leading to local sealing of cracks and pores and reducing permeability. The technique was demonstrated with a modern fly ash mix.

INFLUENCE OF RECYCLED MATERIALS It may be expected that the production of sustainable concretes will increasingly involve the incorporation of recycled materials as part of the innovation process. Primary among these will be the use of recycled aggregate. The irregular nature of recycled aggregate could adversely impact the cohesiveness of the cement paste/aggregate interface, leading to increased concrete permeability. Even with natural aggregate, the significance of the interfacial transition zone (ITZ) in increasing permeability was demonstrated by Bustos et al. (2015), who maximised cement paste at the expense of aggregate content through use of natural pozzolans. The ITZ was represented by surface density, and it was found that lower aggregate content yielded lower permeability. Extrapolating this finding to jagged recycled aggregate would suggest that the ITZ would make an even more significant contribution to adverse permeability. Evidence to support this was reported by Kubissa et al. (2016), who found that water absorption, sorptivity, and chloride migration increased when recycled aggregate was used as the coarse fraction. However, the negative impact could be mitigated through the use of supplementary cementitious materials such as silica fume and metakaolin. Inevitably, many recycled aggregates will not meet the criteria applicable to higher-quality natural aggregates for use in concrete. Nevertheless, Paine and Dhir (2010) demonstrated that a performance-related approach could produce recycled aggregate concrete that was satisfactory. Direct replacement of natural aggregates by recycled aggregates normally yields a more permeable concrete, but this in itself should not be a reason for dismissing its role in sustainable concrete. Sufficiently durable concrete is as successful as highly durable concrrete if the structure remains fully serviceable over the target service life. This again emphasises the importance of the

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durability design and performance specification route in opening up opportunities for greater sustainability in the construction industry. An overview of self-compacting concrete manufactured with recycled concrete aggregate, by Revilla-Cuesta et al. (2020), indicated that satisfactory durability performance relative to traditional concrete could be achieved up to a certain level of replacement. Kapoor et al. (2016) specifically studied the durability of self-compacting concrete made with recycled aggregate and mineral admixtures. There was a loss of durability at the 100% recycled aggregate level that could not be mitigated by addition of mineral admixtures. However, the addition of silica fume or, especially, metakaolin at 10% by weight of Portland cement was effective at the 50% recycled aggregate replacement level in restoring the durability performance in respect of rapid chloride penetrability, initial surface absorption, water penetration and capillary suction tests.

MEASUREMENT OF PERMEATION PROPERTIES Many test methods are available, some of which are a direct measure of permeability, while others detect surface absorption or porosity and may be used to rank a concrete’s vulnerability to permeation in relative terms. Among many possible techniques, the following are noteworthy either as standard test methods, quality control options, or investigative and research tools, either in the laboratory or on-site: • • • • • •

depth of water penetration under pressure surface absorption test water absorption test sorptivity test air and water pressure permeability tests detection of high-risk air voids.

Permeability and air voidage may also be inferred from standard tests for carbonation rate, chloride ingress, and freeze-thaw resistance, presented separately in Chapters 6, 7, and 8 respectively. Additionally, a comprehensive Concrete Society report on permeability testing of site concrete is now in its second edition (Concrete Society 2008a).

Depth of water penetration under pressure The foremost European standard test for concrete permeability is a water penetration test, EN 12390-8 (CEN 2019). The test is useful for comparative purposes but does not yield an absolute value of water permeability. The test surface of a cubic, cylindrical, or prismatic specimen is exposed to water under a pressure of 500 ± 50kPa for a period of 72 ± 2 hours. The test is conducted on the untrowelled, surface-roughened face. The maximum depth

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of water penetration is measured by splitting the concrete face open to reveal the penetration front. Some difficulty in detecting the front can occur in darker concrete faces. Newlands et al. (2022) reported overcoming the problem by spiking the water reservoir with ulta-violet active dye and subsequently examining the split surface under ultra-violet-light illumination.

Surface absorption test Surface absorption is commonly measured by the Initial Surface Absorption Test (ISAT). The observed flow rate may be used in comparative trials of different concretes and to characterise the performance of cast concrete against benchmarks of high and low absorption rates. The test results provide information on the quality of the surface layer and cover concrete, which may differ from the characteristics of the inner body material if curing is substandard. The ISAT has the advantage of being covered by a British Standard, BS 1881: Part 208 (BSI 1996). The BS 1881 suite of standards covers tests methods in use in the United Kingdom which are supplementary to European ‘EN’ standard tests on concrete. The method involves sealing a cap onto the concrete surface and monitoring the rate of seepage of water into the surface and near-surface pores (Fig. 4.5). Clearly, the moisture content of the surface has an influence on the results. On-site testing must therefore take account of current and recent weather patterns that would significantly influence the surface condition.

Water absorption test Bulk concrete absorption testing determines the amount of water absorbed by an unsaturated specimen in a prescribed period, thus providing a relative measure

Figure 4.5 Initial surface absorption test.

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of the permeable pore structure. It does not measure permeability directly, but provides an indication of the volume of interconnected voids from which the relative quality of the concrete in respect of permeability may be inferred. Testing may form part of the process of determining the density of concrete. One such method is set out as Part 122 of the BS 1881 suite of standards in the UK (BSI 2020a). Typically, this test is carried out on cores drilled from concrete members, providing access to the pore structure through the cylindrical surfaces of the element. Other valid specimens are cast cylinders and prisms where no point in the specimen is more than 50 mm from a free surface. Lower absorption through surfaces cast against the smooth mould faces may not yield representative values of the same concrete mix in service. The test procedure involves immersing oven-dried cores in water to detect the uptake of water in a specified time (Fig. 4.6). Concrete should be tested at an age of 28─32 days. The specimens are oven-dried for a period of 72 ± 2 hours, cooled in an airtight container, and immersed in water for 30 ± 0.5 minutes. The percentage mass increase is determined. The result is expressed as an equivalent value to that of a notional 75 mm long cylinder of 75 mm diameter, through the actual surface to volume ratio. A related test in the United States, used to determine the density of concrete, ASTM C642 (ASTM 2021c), yields data that also allows calculation of the permeable voids percentage. Samples are oven-dried for successive 24-hour periods and cooled, preferably in a desiccator, until the difference between successive oven-dry mass readings is less than 0.05% of the lower value. The dry specimens are immersed in water until saturated, determined by successive 24-hour immersion period surface dry mass readings being less than 0.05% of the higher value. The saturated specimens are boiled in tap water for five hours and their mass determined on natural cooling to between 20oC and 25oC over at least a 14-hour period. The specimens are then suspended in water to determine their apparent mass in water. The percentage of permeable voids may be calculated as:

V = 100

C C

A D

Figure 4.6 Water absorption test.

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where V = percentage of permeable pore space volume (g) A = oven dry mass in air (g) C = mass in air after immersion and boiling (g) D = apparent mass in water after immersion and boiling (g) The value is dependent on the extent to which water is driven out of the pore structure on drying, and forced in under immersion and boiling. Vacuum techniques and pressure would assist the determination of more accurate values of voidage where very small pores form a significant part of the pore size distribution.

Sorptivity test Test methods for determining sorptivity are based on exposing the flat surface of an unsaturated cylindrical specimen to a source of water from which it may gain mass solely through capillary action. The rate of water ingress is monitored through the mass of water absorbed over time. A graph is plotted of mass absorbed against the square root of time. Sorptivity may be computed from the slope of the best fit line (Fig. 4.7). The methodology has been formalised as ASTM C1585 (ASTM 2020e) and further extended in the Water Sorptivity Index (WSI) test method of the Durability Index Testing Procedural Manual (Alexander et al. 2018). The specimen in ASTM C1585 is a 50 mm slice from a cylinder, cured for 28 days and then conditioned for 3 days in an environment of 50oC and 80% relative humidity. The slice is stored for 15 days in a sealed container and epoxy coated on all but the flat top surface to be exposed to water absorption. The mass gain is monitored over time.

Figure 4.7 Sorption test.

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The Water Sorptivity Index (WSI) forms one leg of the three durability indices of South Africa’s performance-based approach to durability (Ballim and Alexander 2018). The WSI method integrally associates determination of sorptivity with measurement of porosity to produce a value that is more sensitive to interpretation (Moore et al. 2021). The index is the rate of mass increase from water uptake by absorption divided by the water-penetrable porosity of the specimen. Sorptivity is determined through a 30 mm long, 70 mm diameter concrete disc, recovered from cubes or a concrete element on site. The specimen is pre-conditioned by oven-drying at 50 ± 2oC for 7─8 days and cooling for 2─4 hours in a desiccator. The curved surfaces are epoxy coated or covered with packaging tape. The first part of the WSI test involves determining the dry mass and the increasing mass at eight intervals over 25 minutes following first contact with water. The rate of increase may be expressed by graphing the test results and determining the slope of the best fit line. The vacuum saturated mass is determined by vacuum saturation and subsequent soaking in calcium hydroxide saturated water. This is achieved through a vacuum tank at -75 to -80 kPa in which the specimen is stored for 3 ± 0.25 hours before the water is introduced and maintained at the vacuum level for a further 3 ± 0.25 hours. The vacuum is released and storage continues for 18 ± 1 hours, a period determined to be adequate for precision of porosity determination (Moore et al. 2020). Finally, the vacuum saturated mass is measured. The sorptivity is calculated as follows: S=

F d MSV MS0

where S = sorptivity (mm/hour0.5) F = slope of the best fit line (g/hour0.5) d = specimen thickness (mm) MSV = vacuum saturated mass (g) MS0 = dry mass (g)

Air and water pressure permeability tests Commercially available kits have been developed to monitor the air and water permeability of concrete through flow under a pressure difference. Examples include the “Figg-type” tests (Fig. 4.8) and the “Autoclam” test (Fig. 4.9). These are designed for ease of testing the quality of concrete in existing structures, but could equally be used for research in laboratory conditions. A significant feature of both tests is detection of the permeability of concrete within the all-important cover zone. Based on the research of Figg (1973), the Figg method involves drilling a 10 mm diameter hole to a depth of 40 mm in the cover concrete and sealing

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Figure 4.8 Figg-type permeability test.

Figure 4.9 Autoclam permeability test.

the top 20 mm to create a test chamber. The device achieves a pressure differential between the hole and the surrounding concrete by reducing the chamber to -55 kPa through a vacuum pump. The time taken to increase pressure to -50 kPa from the flow of air through the permeable structure is monitored. Water permeability can also be estimated by forcing water into the test chamber. Both tests yield a number unique to this test method. The Autoclam Permeability Test measures the air permeability, unsaturated water permeability, and sorptivity of concrete. It was developed at the Queen’s University of Belfast for use both in the laboratory and on-site to measure air and water permeability (Basheer et al. 1993). A base ring is bonded to the concrete surface to isolate the test area. The main body of the apparatus is then screwed to the ring. A cylinder and piston are used to pressurise the test area. The change in pressure over time is monitored.

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Detection of high-risk air voids Air voids forming part of the permeable pore network are a particular concern if they reside at the interface of reinforcement and concrete. These sites are points of accelerated corrosion rate if passivity is compromised. A non-destructive method has been developed to measure the volume of entrapped air voids at the interface (Davies and Buenfeld 2021). The sensor consists of a latex sleeve over a perforated stainless-steel tube slotted over a brass rod. The assembly is cast into the concrete mimicking the reinforcement. The latex sleeve is inflated by water under pressure of nitrogen gas to fill the interface voids, from which the void volume can be estimated. The sensor was found to accurately determine air voidage in the range of 1–6%. Further development for routine laboratory quality control testing is foreseeable, although use for measurements on-site would require resolution of practical issues. SUMMARY Durability is a function of the ease with which aggressive species may move through an element. This is characterised by the permeability of concrete, particularly in the cover zone. The need for mixing water in a greater amount than that required solely for hydration of the cementitious components leads to the creation of voids that may interconnect to form a network of channels. Defence from the ingress of aggressive agents in the environment is achieved by making this network tortuous and narrow. Key factors are a low water/cement ratio to minimise the initial porosity, and curing to allow hydration products to form before and especially after the concrete has achieved its initial set. Secondary cementitious materials such as slag and pozzolans have physical and chemical properties that may enhance impermeability, but only if the fundamentals of low water/cement ratio and a high standard of curing are followed. Equally, low-carbon cements using ternary blends with components such as limestone can be used to produce adequately impermeable concrete if the fundamental rules are followed. There are many ways to characterise the relative permeability of concrete. Most commonly employed are water absorption, sorptivity, air and water pressure indices, and ionic diffusion coefficients. From the viewpoint of the durability of most concrete structures, the diffusion coefficients are most relevant – and are discussed more in Chapters 6 and 7 regarding carbonation and chloride ingress – but permeability testing is more amenable to pressure or capillary suction methodologies if rapid results are required, and for testing on site. The results of these tests provide strong correlation with service life prediction and are readily amenable for use in performance-based specifications.

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REFERENCES ACI Committee 212. 2016. Report on Chemical Admixtures for Concrete, ACI 212.3R-16, 43–48. Michigan: American Concrete Institute. Alexander. M., Ballim, Y. and J. Mackechnie. 2018. Concrete durability index testing manual. Version 4.5.1. Cape Town: University of Cape Town and University of Witwatersrand, Johannesburg. ASTM. 2017c. C1797, Specification for ground calcium carbonate and aggregate mineral fillers for use in hydraulic cement concrete. West Conshohocken: ASTM International. ASTM. 2020c. C1240, Standard specification for silica fume used in cementitious mixtures. West Conshohocken: ASTM International. ASTM. 2020e. C1585, Standard test method for measurement of rate of absorption of water by hydraulic-cement concretes. West Conshohocken: ASTM International. ASTM. 2021b. C595, Standard specification for blended hydraulic cement. West Conshohocken: ASTM International. ASTM. 2021c. C642, Standard test method for density. absorption. and voids in hardened concrete. West Conshohocken: ASTM International. ASTM. 2022a. C150, Standard specification for Portland cement. West Conshohocken: ASTM International. ASTM. 2022b. C618, Standard specification for coal fly ash and raw or calcined natural pozzolan for use in concrete. West Conshohocken: ASTM International. ASTM. 2022c. C989, Standard specification for slag cement for use in concrete and mortars. West Conshohocken: ASTM International. Ballim, Y. and M. Alexander. 2018. Guiding principles in developing the South African approach to durability index testing of concrete. In Proceedings, Sixth international conference on durability of concrete structures, ICDCS2018 ed.P.A.M. Basheer, 36–45. Caithness: Whittles Publishing. Basheer. P. A. M., Montgomery, F. R. and A.E. Long. 1993. The Autoclam permeability system for measuring the in-situ permeation properties of concrete. In Proceedings, NDT in Civil Engineering Conference. 14-16 April 1993. Liverpool University, 235–259. Northampton: British Institute of Non-Destructive Testing. BSI. 1996. BS 1881: Part 208, Testing concrete. Recommendations for the determination of the initial surface absorption of concrete. London: British Standards Institution. BSI. 2016. PAS 8820, Construction materials. Alkali activated cementitious material and concrete. Specification. London: British Standards Institution. BSI. 2020a. BS 1881: Part 122: 2011 +A1:2020, Testing concrete - Method for determination of water absorption. London: British Standards Institution. Bustos, F., Martinez. P., Videla. C. and M. Lopez. 2015. Reducing concrete permeability by using natural pozzolans and reduced aggregate-to-paste ratio. Journal of Civil Engineering and Management. 21, 2: 165–176. Calvo, J., Moreno, M., Carbelossa, P., Pedrosa F. and F. Tavares. 2019. Improvement of the concrete permeability by using hydrophilic blended additive. Materials 12, 15. doi: 10.3390/ma12152384.

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CEN. 2006a. EN 15167-1. Ground granulated blast furnace slag for use in concrete, mortar and grout - Part 1: Definitions. specifications and conformity criteria. Brussels: Comité Européen de Normalisation. CEN. 2009b. EN 13263-1:2005+A1:2009. Silica fume for concrete - Part 1: Definitions, requirements and conformity criteria. Brussels: Comité Européen de Normalisation. CEN. 2011b. EN 197-1. Cement - Part 1: Composition, specifications and conformity criteria for common cements. Brussels: Comité Européen de Normalisation. CEN. 2012. EN 450-1. Fly ash for concrete - Part 1: Definition, specifications and conformity criteria. Brussels: Comité Européen de Normalisation. CEN. 2017. EN 13639. Determination of total organic carbon in limestone. Brussels: Comité Européen de Normalisation. CEN. 2019. EN 12390-8. Testing hardened concrete - Part 8: Depth of penetration of water under pressure. Brussels: Comité Européen de Normalisation. CEN. 2021a. EN 197-5. Cement - Part 5: Portland-composite cement CEM II/C-M and composite cement CEM VI. Brussels: Comité Européen de Normalisation. Comite Euro-International du Beton. 1989. Durable concrete structures design guide. London: Thomas Telford. 112pp. Concrete Society. 2008a. Permeability testing of site concrete, a review of methods and expansion. Technical Report TR31. Camberley: Concrete Society. Crank. J. 1975. The Mathematics of Diffusion. Oxford: Clarendon Press. Davies, R. and N. Buenfeld. 2021. Non-destructive measurement of air voids at the reinforcing steel-concrete interface. Concrete 55, 8: 46–48. Dewar, J. 1985. The structure of fresh concrete - a new solution to an old problem. First Sir Frederick Lea Memorial Lecture. Beaconsfield: Institute of Concrete Technology. Dimov, D., Amit, I., Gorrie, O., Barnes, M., Townsend, N., Neves, A., Withers, F., Russo, S. and M. Craciun. 2018. Ultrahigh performance nanoengineered graphene–concrete composites for multifunctional applications. Advanced Functional Materials 28, 23. doi: 10.1002/adfm.201705183. Figg, J. 1973. Methods of measuring the air and water permeability of concrete. Magazine of Concrete Research 25, 85: 213–219. Geim, A. and K. Novoselov. 2007. The rise of graphene. Nature Materials 6: 183–191. Kapoor, K., Singh, S. and S. Singh. 2016. Durability of self-compacting concrete made with recycled concrete aggregates and mineral admixtures. Construction and Building Materials 128: 67–76. Król, A., Giergiczny, Z. and J. Kuterasińska-Warwas. 2020. Properties of concrete made with low-emission cements CEM II/C-M and CEM VI. Materials 13, 10: 1–17. Kubissa, W., Jaskulski, R. and M. Brodňan (2016) Influence of SCM on the permeability of concrete with recycled aggregate. Periodica Polytechnica Civil Engineering 60, 4: 583–590. Lawrence, C. 1981. Durability of concrete: molecular transport processes and test methods. Technical Report No. 544. Slough. Cement and Concrete Association. Mehta, P. 1993. Concrete structures, properties and materials. New Jersey: Prentice-Hall Publication.

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Moore, A., Bakera, A. and M. Alexander. 2020. Water sorptivity and porosity testing of concrete. Concrete Beton 162: 13–16. Moore, A., Bakera, A. and M. Alexander. 2021. A critical review of the Water Sorptivity Index (WSI) parameter for potential durability assessment: can durability be considered in isolation of porosity. Journal of the South African Institution of Civil Engineers 63, 2: 27–34. Neville, A. 1995. Properties of Concrete. 4th Edition. Harlow: Longman. Newlands, M., Jones, R., McCarthy, M., Dyer, T., Csetenyi, L. and M. Jozwik. 2022. Low-carbon concrete at Dundee. Concrete 56, 9: 13–15. O’Riain, C. 2022. The ggbs advantage. Concrete 56, 8: 38. Paine, K. and R. Dhir. 2010. Recycled aggregates in concrete: a performance related approach. Magazine of Concrete Research 62, 7: 519–530. Powers, T. 1958. Structure and physical properties of hardened Portland cement paste. Journal of the American Ceramic Society 41, 1: 1–6. Reches, Y. 2018. Nanoparticles as concrete additives: review and perspectives. Construction and Building Materials 175: 483–495. Revilla-Cuesta, V., Skaf, M., Faleschini, F., Manso, J. and V. Ortega-López. 2020. Selfcompacting concrete manufactured with recycled concrete aggregate: an overview. Journal of Cleaner Production 262. doi: 10.1016/j.jclepro.2020.121362. Ricketts, N. 2022. Cementene: a sustainable innovation for the construction industry. Concrete Engineering International 26, 2: 38–39. Seymour, L., Maragh, J., Sabatini, P., Di Tommaso, M., Weaver, J. and A. Masic. 2023. Hot mixing: mechanistic insights into the durability of ancient Roman concrete. Science Advances 9, 1. doi: 10.1126/sciadv.add1602 Tennis, P., Thomas, M. and W. Weiss. 2011. State-of-the-art report on use of limestone in cements at levels of up to 15%. SN3148: 34-35. Skokie: Portland Cement Association. Thomas, M. and R. Hooton. 2010. The durability of concrete produced with. Portland-limestone cement: Canadian studies. SN3142: 23-25. Skokie. Portland Cement Association. Washburn, E. 1921. The dynamics of capillary flow. Physical Review Series 2, 21: 273–283. Wong, L. 2022. Durability performance of geopolymer concrete: a review. Polymers 14, 5: 868.

Chapter 5

Corrosion of reinforcement in concrete

NATURE OF CORROSION DAMAGE Corrosion of reinforcement is the most significant cause of durability failure in the world’s concrete infrastructure. Although much attention is focused on factors that initiate corrosion, such as carbonation and chloride ingress, it is important that the designer understand the fundamental aspects of an active corrosion cell in reinforced concrete. Understanding the challenge can influence decision-making that mitigates the risk. It may even prove possible to extend the notional life of a structure in durability design by a decade or more through modelling the propagation phase of corrosion damage in addition to the initiation period. At the very least, an understanding of corrosion in reinforced concrete structures will explain why localised repair of spalled concrete areas with high-quality mortar inevitably leads to the repair material falling off also! Corrosion is an electrochemical process whereby a metal undergoes a reaction with chemical species in the environment to form a compound. The chemical species are principally oxygen and water. Electrons are released from a point called the anode (think of anions) and are acquired at a point called the cathode (think of cations). Steel reinforcement has a natural tendency to set up anodic and cathodic points if access to oxygen is possible in a moist environment. This is because steel is formed of metals found naturally occurring as ores, to which they wish to revert. The durability of reinforced concrete requires conditions in which the dissolution of metal atoms is not supported and the reinforcement is inaccessible to oxygen and moisture. Two self-defence mechanisms are employed to achieve this. The first involves a naturally occurring protective film, or patina, on the reinforcement. This thin skin requires certain conditions for its survival. The second involves cover concrete of sufficient depth and impermeability. Regarding the first issue, the high pH level of fresh concrete leads to the formation of a passive skin on the surface of reinforcement. This skin prevents the initiation of corrosion by sealing the metal from contact with oxygen and moisture. The passive DOI: 10.1201/9781003261414-5

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film may be broken down in time through carbonation or chloride ingress reaching the steel. Regarding cover, the rate of corrosion depends on the rate at which oxygen and moisture may penetrate the cover. This has a two-fold influence. First, oxygen and moisture are required to feed the process. Second, the concrete must be sufficiently moist that its electrical resistance is low enough to allow the creation of an electrochemical cell. Damage to the structure is manifested in a number of ways: the cross-sectional area of reinforcement available for load-carrying is locally reduced; the bond between the steel reinforcement and the concrete is lost; the surface of the structure may exhibit cracking and spalling; rust staining on the surface may become apparent even in the absence of cracking. The cracked cover zone may become ineffective, leading to a reduced effective member cross-section for structural purposes. In the case of ferrous reinforcement, the compounds formed through corrosion are hydrated iron oxides, for example, ferrous hydroxide Fe (OH)2 or ferric hydroxide Fe(OH)3, and a secondary reaction forms rust. The initial reaction is generally the formation of ferrous hydroxide, and the secondary reactions produce a form of rust dependent on the environmental conditions. Examples include FeOOH, Fe2O3.H2O, αFe2O3, or γFe2O3. Schematic descriptions are as follows: 2Fe (OH )2 + O2

FeOOH

and 2FeOOH

Fe2 O3 + H2 O

Also 4Fe (OH )2 + O2 + 2H2 O

4Fe (OH)3

and 4Fe (OH )3

2Fe2 O3 . H2 O + 4H2 O

The resulting oxide has a lower density than the parent metal, and so a volume increase occurs when a mass of metal is replaced by the new compound. This leads to the additional hazard of spalling in reinforced concrete structures. Corroding reinforcement in concrete not only loses valuable load-carrying capacity but the cover concrete may also have to accommodate the increased volume of the iron oxides produced on the metal surface. The resulting rust may occupy a volume between two and five times that of the parent steel. If this cannot be accommodated,

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Figure 5.1 Stages in corrosion-induced damage where growth of corrosion products cause cracking and eventual spalling.

expansive pressures result. The low tensile strength of concrete may be exceeded by the pressure caused through the buildup of corrosion products. Cracking may ensue, and further development to the stage of delamination and spalling is not uncommon where the ratio of depth of cover to bar diameter is low. The first sign of distress might be pop-outs where individual cells propagate. However, as these become more widespread, the existence of long, thin cracks along the line of the reinforcement may be harbingers of the limit state of collapse. The various stages of this phenomenon are illustrated in Fig. 5.1 and an example of corroded reinforcement is illustrated in Fig. 5.2. The precise form of damage to the concrete depends on the bar diameter, bar spacing, and the depth of cover. The form of damage to the reinforcement depends on the proximity of anodic and cathodic reaction sites on the steel, leading to “microcell” or “macrocell” corrosion. The latter produces a form of attack described as “pitting corrosion”, as illustrated in Fig. 5.3. A pit develops on the reinforcement surface. This form is more insidious than spalling because it can lead to loss of cross-section without evidence of distress being apparent on the surface. The loss of effective cross-section is very significant compared with that occurring where there is general corrosion on the complete perimeter. Pitting can occur where small anodic sites are in combination

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Figure 5.2 Example of corroded reinforcement where the cross-sectional area has locally “necked” to the point of being incapable of carrying load.

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Figure 5.3 Pitting corrosion causes significant loss of cross-section at a point, making it a highly stressed weak link in the structural capacity chain.

with a large cathodic area. Such conditions can occur where moisture conditions vary across a structure, due to drainage conditions or leaking joints, leading to localised anodic sites. This mechanism of attack is usually related to the presence of chlorides and is dependent on the relative amounts of chloride and hydroxide ions. If the chloride ions predominate, the loss of Fe2+ ions is accelerated and a pit develops. If the OH– ions predominate, there will be precipitation of FeOH+ which repairs the passive oxide film on the reinforcement. The loss of cross-section in pitting corrosion can occur rapidly and critically reduce the load-bearing capacity of the reinforced concrete member. Unlike carbonation-induced corrosion, where cracking and spalling can serve as an early indicator of a problem, chloride-induced pitting corrosion may be quite advanced before evidence is apparent. Chen et al. (2020) examined 66 reinforcing bars extracted from elements suffering from corrosion, and among their findings was that pitting corrosion along a bar does not necessarily lead to longitudinal cracking. Deterioration in these cases may be so advanced at the time of discovery that the structure may be beyond economic repair.

ELECTROCHEMICAL PROCESS Measures to control corrosion of reinforcement require an understanding of the processes involved. Corrosion is an electrochemical process and, therefore, a basic understanding of electrochemistry and its application to the particular case of reinforced concrete is required.

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Oxidation and reduction Corrosion is related to the flow of electrons. It is necessary to recognise that elements may gain or lose electrons. An element or compound that loses electrons is said to be “oxidised”, and one that gains electrons is said to be “reduced”. The elements and compounds of interest in the case of reinforced concrete are iron, water, and oxygen, since the metal undergoes a reaction with chemical species in the environment ─ principally oxygen and water ─ to form a compound. When metal oxides are formed the metal atoms lose outer electrons. The metal atoms are oxidised and the oxygen is reduced. The following examples are presented in a form whereby the number of electrons is equal in each case, as this is a significant controlling factor in the consideration of the rate of corrosion. Oxidation example: Fe

Fe 2+ + 2e

cation

Reduction example: H2 O +

1 2

O2 + 2e

2OH anion

The electrochemical cell Oxidation and reduction processes will be detectable if conductors of electricity are immersed in a solution capable of conducting electric current. A classic example of a cell in which a flow of electrons is readily detectable is shown in Fig. 5.4. This illustrates two electrodes connected together and

Figure 5.4 Basic form of an electrochemical cell.

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immersed in an electrolytic solution. The cell may be easily constructed in a laboratory and is familiar as a form of battery. In this example, the positive electrode is the anode while the other is the cathode. An electrochemical cell may also occur with a so-called “mixed electrode”. This describes a condition in which the anode and cathode are on the same material. The occurrence of a mixed electrode may occur in reinforced concrete if a number of conditions apply: • the protective passive film on the reinforcement must be breached locally • the concrete must be moist enough to act as an electrolyte • the concrete cover zone must be permeable to oxygen. The consequent scenario is illustrated in Fig. 5.5. We assume that the permeability of the cover concrete allows moisture to ingress from the environment and a sufficiently saturated state develops at which the moist concrete may function as an electrolyte, providing a network of highways for ion flow dissolved in the pore liquid. In addition, the depth and permeability of the cover conspire to allow the penetration of carbon dioxide or chlorides to an extent that local depassivation of the reinforcement will occur. The next stage is the formation of anodic and cathodic sites on the reinforcing bar. Metal oxidation occurs at the anode, and the freed electrons flow through the pore water to the cathodic sites. Oxygen from the atmosphere penetrates the cover to feed the cathodic reaction, and the electrons are consumed in reduction processes with the formation of hydroxyl ions. In the example illustrated, a buildup of rust products signals an active corrosion cell. The distance between the anodic and cathodic sites can vary greatly. If they are in very close proximity, they form a microcell. On the other hand, a series of anodic sites might be widely spaced from a common cathodic site, which leads to the phenomenon of macrocell corrosion. For example, one layer of reinforcement may become the anode and another the cathode. Equally one part of the structure may be depassivated and become the

Figure 5.5 Mixed electrode in reinforced concrete.

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anode, while a much larger area may remain passive and become the cathode. Such high cathode to anode area ratios can lead to very significant increases in corrosion current and insidious deep pitting, especially in chloride-laden environments. An interesting aspect of the corrosion cell in reinforced concrete is the fact that damaged areas, for example, zones of spalling, may represent only the anodic sites. Patch repair of concrete around a spalled area may not be effective if the cathodic sites remain permeable – in time the patches will spall, since the cell is allowed to continue functioning. In such a scenario it may be prudent to enhance the concrete’s impermeability, for example, by coating the surface, to ensure an effective repair that stifles the corrosion cell’s activity at both the anodic and cathodic sites.

Corrosion rate The corrosion rate is a function of the maximum potential activity level at either the anode or the cathode – control one and you control the other. If they could act individually, the anodic and cathodic processes would lead to an accumulation of positive and negative charges, respectively, on the reinforcement. Such a scenario is not sustainable because the hydroxyl ions diffuse towards the anode, where they meet the counter-diffusion of ferrous ions. The resulting combination causes electrical neutralisation if the anodic and cathodic processes are coupled together in the form of a corrosion cell with no excess electrons. If there is no external source of electrons, the electrons produced by oxidation will be fully consumed by reduction. Thus the oxidation rate and reduction rate must be equal. The anodic reaction rate must be equal to the cathodic reaction rate. This equality controls the rate of corrosion. Thus, the rate of electron flow reflects the rate of corrosion. The rate of flow may be referred to as the “corrosion current density”. Essentially, it is the number of electrons flowing per unit area and is represented graphically in Fig. 5.6. A value at the upper end of the range would be 100 mA/m2. The rate of corrosion may be expressed as follows:

icorr =

Rc

where icorr = corrosion rate

Figure 5.6 Illustration of corrosion current density: the number of electrons flowing through a unit of area.

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β = Stern-Geary constant α = surface area of steel contributing to current flow Rc = charge transfer resistance. The charge transfer resistance may be determined by linear polarisation resistance measurement. Corrosion activity has been calibrated to the detected current through laboratory and on-site measurements. Andrade and Alonso (2001) reported that values in the range of 0.1–0.5 μA/cm2 indicate a low level of corrosion, while the 0.5–1.0 μA/cm2 range was classified as moderate. Passive reinforcement exhibited currents less than 0.1 μA/cm2. Robles et al. (2022) classified current in excess of 1.0 μA/cm2 as a very high-risk situation, while Andrade and Alonso noted that such high values are seldom measured in structures. Perhaps of greater concern in the context of the remaining service life of a deteriorating structure is the rate at which material is being lost from the reinforcement. A relationship may be established through Faraday’s Laws between the measured current flow and the annual material loss. Faraday’s First Law of Electrolysis states that the mass of substance liberated is proportional to the quantity of electric charge that liberates it. Faraday’s Second Law of Electrolysis concerns the masses of different substances liberated by the same quantity of electrical charge: the masses are proportional to the ratio of the atomic mass and the valence. Thus, the loss may be calculated from: m=

M C z F

where m = mass loss M = atomic mass C = electric charge z = valence F = Faraday constant (96487 coulomb). This ratio of atomic mass to valence is called the chemical equivalent. In the case of iron (Fe) and the ferrous ion (Fe2+), the atomic mass is 55.85 amu and the valence is two. Therefore, the chemical equivalent is 27.95 and oxidation of 27.95 g of Fe occurs per Faraday of electrical charge. Given that a coulomb is the electric charge conveyed in one second by a current of one ampere, it is interesting to review the effect of a corrosion current density of one milliamp per square metre. This equates to 0.001 coulombs/m2/s, therefore the rate of loss would be:

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(55.85)(0.001)(1) = 2.89 × 10 (2)(96478)

7

g/m2 /s

This would equate to a loss of about 10 g/m2/year. A corrosion current density at the upper end of the range, such as 100 mA/m2, could theoretically lead to an annual loss of approximately 1 kg/m2 of metal. Weight losses higher than those predicted by Faraday’s Laws have been detected in practice. Rodriguez et al. (1996) reported that the loss in bar cross-section can be estimated using the linear polarisation technique from the measurement of corrosion rate with the application of Faraday’s law, converting 1 μA/cm2 of the corrosion rate to 0.0115 mm/year of metal loss: = 0 x x = 0. 0115 Icorr t

where ϕ = residual bar diameter (mm) ϕ0 = nominal bar diameter (mm) α = coefficient depending on type of corrosion x = attack penetration (mm) Icorr = corrosion rate (μA/cm2) t = time from start of corrosion (years). The coefficient α has a value of two where the corrosion is homogenous. The value rises to a figure in the range of four to eight if pitting corrosion occurs.

Electrode potential The corrosion phenomenon depends on the ability of electrons to transfer across the interface between the metal and the electrolyte, and vice versa. This parameter is described as the “electrode potential” of a particular metal in a particular electrolyte. Thus, in reinforced concrete, the process is driven by the potential difference between the reinforcement and the cover concrete. The potential difference develops at the interface of the reinforcement and the concrete because of the tendency of the metal ions to dissolve and the difference in environment (metal and concrete) on either side of the surface. An excess positive charge builds up near the interface. The potential cannot be measured directly, so it is measured with respect to a reference electrode. Electrical potentials may be surveyed in existing structures by devices such as the half-cell, described later. Reference electrodes include, for example, saturated hydrogen electrode; saturated calomel electrode (SCE); and silver-silver chloride electrode (SSC). Where potentials are quantified, it is necessary to indicate the reference electrode

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type. The corrosion of steel in concrete typically occurs in the potential range from –450 mV SCE to –600 mV SCE.

Electrical resistance Flow between anodic and cathodic sites during corrosion depends on the electrical resistance. This is related to the resistivity of the material – its ability to withstand a flow of ions under the influence of an electrical field – and its geometry. Resistivity (ρ) is the product of resistance and distance, yielding a value in units of Ωm but more often reported in Ωcm. The current is conducted by ions in the pore water or, if necessary, through the gel water. Cement type influences the value. The presence of chloride ions in the pore water would also decrease the resistivity. The resistance is highly dependent on the moisture state of the concrete. Oven-dry concrete, for example, has such a high resistance that it may be regarded as an insulator, whereas resistivity at 50% relative humidity may be one-hundred-fold higher than saturated concrete (Lawrence 1990). Tests may be carried out at the surface or on a bulk sample such as a cylinder. The former is of greater interest in respect of an active corrosion cell, and in characterising the quality of the cover reinforcement as an electrolyte. Concrete that is too dry to support corrosion has been found to have a surface resistivity in excess of 100 Ωcm. Empirical tests reveal that resistivity values in the range of 50─100 Ωcm can support low levels of corrosion activity. The moderate to high risk of corrosion is found in the resistivity range of 10─50 Ωcm. Below 10 Ωcm, Andrade and Alonso (2001) state that resistivity is not the controlling parameter of corrosion rate.

POLARISATION CURVES, THE “EVANS DIAGRAM” An insight into the durability of reinforced concrete through the control of corrosion rate may be gained by consideration of “polarisation curves” and “Evans Diagrams” as demonstrated, for example, by Bentur et al. (1997). These curves graph the relationship between electrical potential and current. A polarisation curve may be generated by experimenting with two electrodes. A potential difference is applied between the two electrodes, and the current flow is monitored. If the potential difference is between the anode and the cathode, the cathode can be electrically positive with respect to the anode. The favouring of either anodic or cathodic reactions will be detectable at different potentials, but there exists a potential at which the reactions are balanced – this is described as the “reversible potential”. The more polarised the potential, that is the further the potential is from the reversible potential, the greater will be the current, as illustrated in Fig. 5.7. In the case of corroding reinforcement, it is of great interest to consider the potentials at the anode and cathode. The anode will initially have a certain potential difference with the electrolyte, and so will the cathode. A

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Figure 5.7 Polarisation curve showing the favouring of reactions at different potentials and the point at which the reactions are balanced ─ the reversible potential (Erev).

common electrode potential is generated between the steel and concrete which is intermediate between the individual anodic and cathodic systems. The corrosion process occurs at the potential where the rates of anodic and cathodic reaction are equal. It is very informative, therefore, to plot the relationship of potential and current in a manner that shows where equilibrium of current occurs. This is achieved by plotting curves on a common axis; that is, ignoring the sign of the current. Such a plot is known as an Evans Diagram. The potential that the metal adopts as it corrodes is determined by the point of intersection of the curves (Fig. 5.8). This electrode potential is the value detectable when the metal is corroding freely. The corresponding current may also be assessed. This potential (Ecorr) and

Figure 5.8 Evans Diagram showing the potential (Ecorr) adopted by a corroding metal.

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the corresponding corrosion current (icorr) may be determined by considering the kinetics of the anodic and cathodic reactions in a given situation. Design of durable structures involves a rudimentary appreciation of the way a corrosion cell behaves. Several consequences of the electrochemical nature of reinforcement corrosion are more readily understood by reference to an appropriate Evans Diagram. In this context, it is interesting to review the important phenomenon of passivity.

PASSIVITY The pore liquid phase in concrete is mainly a solution of sodium hydroxide (NaOH) and potassium hydroxide (KOH), due to the dominant alkalis in the cement. These are sodium oxide (Na2O) and potassium oxide (K2O). This yields a solution of high alkalinity. The high pH level of fresh concrete protects a passive skin formed on the surface of the reinforcement. This skin, of nanometre order of thickness, prevents corrosion occurrence by preventing further oxidation of the metal atoms. Reinforcement in fresh concrete initially begins to corrode, and the corrosion product forms the passive skin on the bar. The skin prevents the dispersal of ferrous ions and is a conductor of electricity. It is therefore unable to support an electrical potential difference, and the force driving the metal ions into solution is removed (Lawrence 1990). The film is formed through a reaction between iron and water. This reaction yields a twin-skinned film consisting of oxide (Fe2O3 or Fe3O4) with an outer hydrous oxide layer. The skin is stable at pH levels above 11.5. When anodic sites form on reinforcement, the reaction is that of oxidation of the metal ions to yield ferrous ions, and the anodic current rises. However, a point is reached where passivation occurs and the current reduces significantly. The Evans Diagram for a reinforcing bar in a passive environment is illustrated in Fig. 5.9. It may be noted that the intersection of the curves usually occurs at a point where the electrode potential is very high and the corrosion current very low. The corrosion current would be low enough to ensure an adequate service life, and the steel is considered to be in a passive state. Curiously, depassivation and therefore corrosion can occur where a very low rate of oxygen diffusion is encountered, such as in submerged structures. In such a case, the cathodic curve moves back to intersect the anodic curve at a low potential. However, the rate of corrosion is not significant in terms of the overall service life of structures.

CORROSION MECHANISM IN CARBONATED CONCRETE The phenomenon of carbonation is fully described in Chapter 6. In summary, a carbonation front delineates areas of differing pH. The front

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Figure 5.9 Evans Diagram for a reinforcing bar in a passive environment showing a significant drop in current as the anodic site becomes passive.

advances over time from the surface towards the passive reinforcement at a rate which is dependent on environmental factors and the material properties of the concrete. If peaks on the carbonation front reach the reinforcement, the passive film is locally destroyed, as it is unstable at low pH values. If the passive film breaks down, iron oxidises to form ferrous ions (Fe2+), with the following half-cell reaction: 2Fe

2Fe 2+ + 4e

The ferrous ions pass into solution. The electrons flow through the reinforcement to the cathode, where they are adsorbed by the electrolyte. The cathodic reaction produces hydroxyl ions (OH–) in the presence of oxygen and moisture as follows: 2H2 O + O2 + 4e

4OH

The hydroxyl ions move from the cathodic site through the moist concrete, the electrolyte, towards the anodic site where ferrous hydroxide Fe(OH)2 is formed. The anodic reactions may be represented as follows: 2Fe 2+ + 4OH

2Fe (OH )2

The ferrous hydroxide is unstable in the presence of oxygen, so the reaction may proceed to form rust. These reactions can be represented by the following:

Corrosion of reinforcement in concrete

2Fe (OH )2 + O2

127

FeOOH

2Fe (OH )2 + H2 O +

1 2

O2

2Fe (OH )3

Further advances in the carbonation front will increase the depassivated area. Widespread corrosion may then follow, with the development of cracking along the lines of the reinforcement.

CORROSION MECHANISM IN CHLORIDE-RICH CONCRETE The issue of chloride ingress is dealt with more fully in Chapter 7, but it is interesting at this point to review it in the context of Evan’s Diagram. The anodic reaction is of particular interest in the case of concrete subject to high chloride levels. As the chloride concentration increases, the anodic curve, vital to passivity, alters as illustrated in Fig. 5.10. It may be seen that the effect is to make the potential more negative with a consequent effect on the corrosion current. The process first involves the oxidation of iron to ferrous ions (Fe2+): 2Fe

2Fe 2+ + 4e

Figure 5.10 Evans Diagram showing the influence of chloride concentration in concrete.

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The cation combines with chloride ions to form chloride or oxychloride compounds FeCl2 and FeOCl, for example: 2Fe 2+ + 4Cl

2FeCl2

The process then becomes self-propagating, due to the acidic conditions created and the recycling of chloride ions. This occurs through the hydrolysing of the chloride compounds, for example: 2FeCl2 + 4H2 O

2Fe (OH )2 + 4HCl

Alternatively 2FeCl2 + 4H2 O

2Fe (OH )2 + 4H+ + 4Cl

and 2FeCl + 2H2 O

2Fe (OH )2 + 2Cl

There follows a consequent recycling of the liberated chloride ions. Although corrosion product is being produced, so too is hydrogen chloride and hydrogen (H+) or hydronium ions (H3O–). The increased acidity of the anodic area helps to prevent precipitation of corrosion product. The chloride and oxychloride compounds are therefore more stable and are free to migrate further. The increased acidity also encourages further oxidation of the iron. The rate of corrosion is influenced by the ability of the cement matrix to bind the liberated chloride, the resistivity of the electrolyte, and the availability of oxygen and moisture at the cathode. The binding ability is dependent on the characteristics of the cement. Sulfate-resisting cement, for example, has a lower binding capacity than, say, ordinary Portland cement. The issue of oxygen and moisture at the cathode is not straightforward because even low rates of oxygen supply may lead to severe pitting corrosion. This effect occurs because the anodic sites may be localised, but the corresponding cathodic sites may be spread out over a wide area. The cumulative effect of even low rates of oxygen supply to large cathodes may be significant. The fact that the corrosion product is discouraged from precipitation due to the acidic conditions, and the existence of highly active and localised anodic sites in combination with large area cathodic sites, accounts for the phenomenon of severe pitting without early warning through visible signs of deterioration at the concrete surface. When cracks do develop, the corrosion product will then be deposited along the crack, where the effect of the acidic zone will become less marked with distance from the reinforcement. By the time rust staining becomes apparent at the surface, the extent of reinforcement deterioration may be structurally significant.

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INFLUENCES ON CORROSION ACTIVITY Corrosion activity is influenced by local environmental factors including oxygen supply, relative humidity, and temperature.

Influence of oxygen supply The passive film may be broken down through carbonation or chloride ingress. The rate of corrosion will then depend on the rate at which oxygen may penetrate the cover. This is illustrated in Fig. 5.11. The oxygen supply to the reinforcement is a function of the permeability of the cover. Good-quality concrete with low permeability will restrict oxygen supply. Equally, a condition in which the concrete is saturated may not lead to corrosion, even in permeable concretes. Submerged parts of reinforced concrete structures in the ocean are not subject to the same risk of corrosion damage as those parts above water, despite the ingress of significant amounts of chloride. This is due to the restriction of oxygen ingress, because gaseous diffusion is very slow through the saturated pores and there may be little dissolved oxygen in the water. Nevertheless, the problems associated with large cathodic areas cannot be ignored. It may be noted from Fig. 5.11 that the cathodic reaction curve becomes steeper as the oxygen supply diminishes, leading to a reduced corrosion current as expected.

Figure 5.11 Evans Diagram showing the influence of oxygen supply in concrete.

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Relative humidity A corrosion cell cannot occur if the concrete is too dry to serve as an electrolyte or too wet to allow ingress of oxygen. Sufficiently dry conditions are typical inside buildings, and corrosion does not occur, despite the high probability of reinforcement depassivation through carbonation. Corrosion activity is most vigorous at relative humidity values above 80%, as illustrated in Fig. 5.12 using data from Andrade et al. (1986) and Parrott (1994). These levels are rarely found indoors. Air conditioned buildings are designed for an atmosphere of typically 50% relative humidity which, although ideal for promoting conditions conducive to carbonation, should not support the development of corrosion activity. Equally, corrosion activity reduces as the relative humidity approaches saturation due to the reduced permeability to oxygen, which cannot penetrate water-filled pores as readily as it can diffuse through dry pores.

Temperature The rate of corrosion increases with increasing temperature. The effect is not significant at typical European humidity levels, but it could be significant in hot humid climates. Browne (1988) reported that an increase in temperature from 20°C to 40°C could increase the rate of corrosion by a factor of five.

Figure 5.12 Influence of relative humidity on corrosion rate showing the need for enough moisture to feed the process but, approaching saturation, it stifles the required oxygen ingress.

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INFLUENCE OF CRACKING The topic of cracking in concrete is fully explored in Chapter 12, but it is interesting at this point to examine the influence of cracking on corrosion rate. It has been a subject of debate in the context of acceptable limits for crack widths. The jury is still out on the detailed impact of dormant cracks perpendicular to the reinforcement, but the Building Research Establishment (BRE 1993) and Concrete Society (2015) support the finding that live cracks parallel to the reinforcement represent a significant corrosion risk that requires urgent attention once detected. Various issues must be weighed up. On the one hand, there is the seemingly self-evident point that cracks which extend to the reinforcement provide easier access for carbon dioxide and chlorides, leading to depassivation, and to oxygen and moisture, thus encouraging the corrosion process. On the other hand, the observed rate of deterioration of cracked structures is not universally higher than non-cracked structures. Thus, the various influencing factors need to be considered in isolation. The first issue relates to whether or not the cracking occurs in the initiation or propagation phase of corrosion. If cracking results from the buildup of corrosion product, it is clearly limited to the propagation phase. Service life is the sum of time to initiation of corrosion, plus the time required for corrosion propagation to cause an unacceptable level of deterioration. The initiation phase will usually be relied upon to form the greater part of the service life. If the cracks are caused by factors other than corroding reinforcement, they could influence the initiation phase to some extent. On the other hand, cracks caused by the expansive forces generated by corrosion product obviously occur after depassivation and will therefore have minimal effect on the overall service life, if the initiation phase is of adequate duration. The second issue is the orientation of the cracks. If the cracks are parallel to the reinforcement and over it, they will pose a more significant problem than if they are perpendicular, due to the area of reinforcement being potentially exposed to harmful agencies. Fortunately, cracks parallel to, and over, reinforcement are more commonly associated with the effects of corrosion propagation rather than its initiation and should therefore be associated with the shorter phase of the service life. One notable exception is the case of plastic settlement cracks, which do contribute to a reduced initiation phase. Nevertheless, Chen et al. (2020) found that longitudinal cracks lead to extensive anodic areas, each capable of generating only low corrosion rates due to the distributed effort of the corrosion cells. A third issue is the dynamic state of the cracks. Cracks are capable of self-healing in certain circumstances – autogenous healing. This effect can be caused by the precipitation of calcium carbonate in the crack, leading to clogging of the crack and preventing further ingress of harmful species. Live cracks may also become clogged by debris, but are less likely to autogenously heal.

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Figure 5.13 Corrosion activity in cracked concrete: significance of uncracked zone over the cathode.

A fourth issue involves the density of cracking. Individual cracks that intersect the reinforcement locally may encourage anodic activity, but access to the cathodic areas will not be different from that pertaining to crack-free surfaces, as illustrated in Fig. 5.13. Therefore, the rate of corrosion should not be different, as it will continue to be influenced by the accessibility of the cathodic sites. This partly accounts for the missing link between cracking and a significantly reduced service life. It is self-evident that cracks allow greater ingress of carbon dioxide and chlorides to the reinforcement, which reduces the time to depassivation, thus reducing the initiation period. However, this does not necessarily replicate the situation in the propagation period. Chen et al. (2020) hypothesise that the nature of the crack pattern and crack density influence the location of anodic and cathodic sites. Available cathodic areas are limited by the distance between cracks when crack frequency increases. Thus the local corrosion rate at individual sites is lower in cracked concrete, because microcells predominate and macrocells are inhibited. A final issue is that of crack width. Other than noting that autogenous healing is more likely when the crack is narrower, it does not appear that crack width itself is a significant factor. Chen et al. (2020) found no relationship between crack width and rate of corrosion during the propagation phase. The maximum local corrosion level was not dependent on maximum crack width in a study of flexural cracks from 0.1─0.4 mm in width.

MODELLING THE RATE OF CORROSION Modelling the service life of a structure where end-of-life is due to corrosion-induced damage would ideally embrace both the initiation and propagation phases of deterioration. This combination is based on the service life concept of Tuutti (1982), where:

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tcr = tp + tcor

where tcr = service life (years) tp = initiation phase of corrosion (years) tcor = propagation phase of corrosion (years). The limit state of deterioration could be defined in several ways. Examples include the period required for cracking of cover concrete to reach a certain limit; the period required for the cross-sectional area of corroded reinforcement to increase (rust accumulation) or reduce (pit formation) by a defined percentage; and the period required for the element to reduce to a defined level of stiffness. However, accurately predicting the length of the propagation phase of a corroding structure is proving to be much more difficult than predicting the initiation phase. To date, therefore, proponents of the design life approach to durability management have conservatively assumed that only the initiation phase would define the design life. This conservative assumption has a sound basis in risk management but falls short of society’s expectations in respect of sustainable development, where we would drive our materials to their predictable limits. The propagation period can represent a substantial period, especially where corrosion rates are moderate or slow. At certain relative humidities, the propagation phase can be longer than the initiation phase, as illustrated in Fig. 5.14. This demands that we further pursue accurate modelling of the propagation phase. Even if the initiation period predominates, we can potentially extend the predicted service life by a decade or two with the propagation phase. That may allow producers to meet performance specification targets with binder contents reduced by a precious few percents, compared with the more conservative approach.

Figure 5.14 Relative distribution of initiation and propagation phases depending on relative humidity.

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The obstacles to accurate modelling have been identified by Otieno et al. (2011): • • • •

time-variant nature of corrosion rate influence of cover cracking on corrosion rate variability of results from different corrosion measurement techniques lack of validation of models derived from accelerated testing with inservice structures • insufficient data hinders the ability to characterise statistical distributions of input parameters (cover depth, binder type, concrete quality, etc.) for probabilistic modelling. Each of these five issues requires detailed research before there is international convergence on an acceptable model of corrosion rate. Many empirical models have been proposed based on linear regression analysis of laboratory and field data. An example based on 156 reported tests from the literature, and seven existing models, is that of Lu et al. (2019), who proposed a model for the corrosion rate in chloride environments taking account of the variation of current with time, resistivity, chloride level, and ambient conditions: icorr =

3

1 exp 1.23 + 0.618 ln Ct 1+t

3034 T (2.5 + RH )

5 × 10

3

where icorr = corrosion current density (μA/cm2) t = period of corrosion activity (years) Ct = chloride content (kg/m3) T = temperature (K). RH = relative humidity expressed as value from 0 to 1, where 1 is saturation ρ = resistivity (kΩcm). An early example of modelling the propagation period based on kinetics in the corrosion cell was developed by Bazant (1979). The model considered the expansion of metal due to rusting that ultimately resulted in cracking of the cover concrete. The parameters included the instantaneous corrosion rate of rust, density of the metal, and reinforcement diameter: tcor = jr

=

D D p jr W icorr F

cor

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where tcor = steady state corrosion or propagation period (years) ρcor = combined density factor for steel and rust (3600 kg/m3) D = diameter of rebar (mm). ΔD = increase in diameter of rebar due to rust formation (cm). p = perimeter of bar (mm). jr = instantaneous corrosion rate of rust (g/m2-s) icorr = corrosion current density (μA/cm2) W = equivalent weight of steel (27.925) F = Faraday constant (96487 coulomb). Lawrence (1990) and Parrott (1994) further examined this route, whereby the propagation period is a function of the changes in reinforcement dimensions and corrosion rate: tp = f {CD , CR,

}

where tp = propagation period CD = corrosion depth which causes the limit of acceptable damage CR = corrosion rate ϕ = reinforcement diameter. Parrott (1994) found that the corrosion depth that caused visible damage is about 100 μm. The corrosion rate varied from about 0.3 μm/year at 50% relative humidity to a maximum of about 50 μm/year at 98% relative humidity. Ferreira et al. (2016) constructed a limit state model based on a combination of the initiation and propagation periods. This allowed the satisfactory service life to be assessed in one analysis ─ from time of construction until cracking of the cover concrete ─ rather than as two separate limit state analyses. The framework for their model begins with the form of reliability analysis for a limit state that we examined in Chapter 3. In that case, we checked that the probability of failing to satisfy a limiting condition was less than a maximum allowable failure probability (Pf max) as follows: P (R

S < 0)

Pfmax

The “resistance” function R and “load” function S were independent and could be described by probability density functions fR(r) and fS(s), respectively. Ferreira et al. considered the probability of failure of the joint function fRS(r,s) to develop a convolution integral as follows:

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pf = P (R

S

0) =

pf = P (R

S

0) =

f (r , s) dr ds D RS s r fR (r). fS (s) dr

pf = P (R

S

0) =

FR (x). fS (x) dx

ds

where pf = probability of failure of a limit state condition R = durability resistance of element S = durability load on element, same units as R but independent of R fR(r) = probability density function describing R fS(s) = probability density function describing S fRS(r,s) = joint bivariate density function of fR(r) and fS(s) FR(x) = probability of R being less than its failure value x fS(x) = probability of S being between x and x + Δx as Δx approaches zero. Using the convolution integral, the authors set up a reliability analysis of a combination of two limit states. To illustrate the application of the model, they chose an initiation period where the limit state was the buildup of chloride to a critical level, and the propagation period limit state was the first appearance of a crack of defined minimum width on the surface of the concrete. Thus: fi, p (t) =

t 0

f p (t

) fi ( ) d

with t = 0 at the start of the initiation period, assuming chloride ingress begins at the start of service life and t = τ at the start of the propagation period where fi,p(t) = probability of failure of the combined initiation and propagation limit state fi(t) = probability density function describing the initiation period fp(t) = probability density function describing the propagation period t = time.

LIMIT STATE FUNCTION FOR PROPAGATION PHASE TIME TO CORROSION INDUCED CRACKING Modelling of the initiation periods for carbonation and chloride ingress are well researched, and many such examples are presented in Chapters 6 and 7. These models allow limit state functions to be readily constructed for determining the limit state of corrosion initiation. Modelling of the

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propagation phase is not as advanced, but Ferreira et al. (2016) have proposed the following limit state function for time to corrosion induced cracking of the cover concrete due to corrosion: rcrack (t)

rcorr (tcorr) > 0

where Δrcrack = critical increase of bar radius due to corrosion at time of cracking (mm) Δrcorr = increase of reinforcing bar radius due to corrosion (mm) t = concrete age and tcorr is the time of corrosion. The critical increase of bar radius due to corrosion product buildup at the time the crack appears depends primarily on the tensile strength and modulus of elasticity of the concrete, the diameter of the reinforcement bar, and cover depth to reinforcement. Other influences are Poisson’s ratio of concrete, a plasticity and cracking factor, local effects of pitting corrosion, and the percentage of reinforcement. This is encapsulated in the following relationship:

rcrack (t) =

fct (t) Ec, eff , D

.

(

) (1 + v) . k 1+( )

ds + 2c 1 +

c ds

3ds + 6c 2 3ds + 4c

nonlin .

klocal . k

where fct = tensile strength of concrete (MPa) Ec,eff,D = effective modulus of elasticity of concrete (MPa) ds = diameter of reinforcing bar (mm) c = cover (mm) v = Poisson’s ratio of concrete knonlin = factor related to plasticity and cracking klocal = factor related to localisation effects due to pitting corrosion kμ = factor related to percentage of reinforcement. The increase of reinforcement bar radius due to corrosion depends on the corrosion rate, the expansion factor of volume conversion from the parent metal to rust, the porosity and thickness of the transition zone accessible to the corrosion products, and the migration of rust into the concrete pores. The corrosion rate would be described by a model. The parameters are encapsulated in the following relationship: rcorr (tcorr) = tcorr . xcorr (tcorr). (

1)

d por (tcorr)

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where ẋ = a model of corrosion rate (mm/year) λ = expansion factor calculated as the volume of rust divided by that of steel dpor = function to take account of rust migration into the concrete pores (mm). The final term in the expression for the increase of reinforcement bar takes account of the time delay in stress buildup due to the relieving effect of corrosion products migrating into the voids and pores surrounding the reinforcement. The partial saturation of this transition zone occurs before stress begins to transfer into the cover concrete as corrosion continues. This is described by the following relationship: dpor (tcorr) = p . dtz. tanh

xcorr (tcorr). ( p. dtz

1)

. tcorr

where p = porosity of the transition zone dtz = thickness of the transition zone.

MONITORING CORROSION ACTIVITY Monitoring corrosion activity in concrete, especially in the field, can be achieved by detecting two key aspects of electrical activity: the potential and the resistance. The location of corrosion cells may be determined by electrical potential contour plots taken on concrete surfaces. A connection to the reinforcement must be made, and the technique relies on electrical continuity through the reinforcement. Measurement of electrical potential alone will indicate the location of corrosion but does not indicate the severity. Resistivity measurement in conjunction with potential mapping may better assist the investigator. High rates of corrosion are associated with large potential gradients and low concrete resistivities. In theory, knowledge of potential and resistivity would allow calculation of the current flow and, therefore, the rate of corrosion. In practice, it is difficult to get sufficiently accurate values, but estimates are informative.

Potential (half-cell) mapping The current flow between anodes and cathodes on the reinforcement leads to variation in potential that is detectable at the concrete surface. A map of equipotential lines on the surface can be used to detect active corrosion cells. The principle is illustrated in Fig. 5.15. The principle is based on bringing two

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Figure 5.15 Principle underlying the use of potential mapping on concrete surfaces to detect corrosion activity.

half-cells together. One half-cell is the reinforcement in the concrete, the other half comprises part of the test equipment. These are connected to form a full cell. The cell potential is measured. The part of the potential contributed by the test equipment is constant, and so variations detected across the circuit are attributable to the reinforcement/concrete half-cell. This permits the identification of areas where corrosion activity is highest. The test apparatus consists of a circuit formed by an electrical connection to the reinforcement, a high impedance voltmeter, a reference electrode, an ionic bridge, and a sufficiently moist concrete surface (Fig. 5.16). The ionic bridge is required between the reference electrode and the concrete surface. It could consist of a sponge soaked with an electrolytic solution such as water or a weak sodium hydroxide solution. Commercial devices are available in which the bridge is incorporated. If a circuit cannot be formed, due to a lack of continuity in the reinforcement, it may be possible to conduct a survey by using two electrodes on the surface – one fixed and one moving – connected through a voltmeter. The approach to investigation generally involves a cover survey to mark out the reinforcement location. A survey grid is then established at about one-metre intervals, the reference electrode is placed over the grid points, and the potential is noted. The grid is progressively reduced to a spacing of about 200─500 mm intervals where negative potentials of interest are detected. A commercially available device is available which can be wheeled across surfaces to produce a continuous plot of potentials. Results are

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Figure 5.16 Test set-up for half-cell potential mapping.

stored in a data logger, and a dot matrix printer can generate variable density contour plots. The potential can be measured only with respect to a reference electrode, such as copper-copper sulfate (CSE) and silver-silver chloride (SSC). It is necessary to indicate the reference electrode type where potentials are quantified. For example, a potential of +130 mV (SSC) corresponds to zero on the CSE scale. Contemporary practice favours silver-silver chloride electrodes because copper-copper sulfate electrodes require frequent recharging with solution to overcome contamination, and the CSE scale may still be used to express the results. Value ranges have been reported which have been calibrated with field observations to give an indication of the relationship between electrical potential and corrosion risk. The findings of Van Daveer (1975) are widely used, but other value ranges have been reported. A general summary is that electrical potentials more negative than –350 mV (CSE) indicate a very high risk of corrosion and that electrical potentials more positive than –200 mV (CSE) indicate a very low risk of corrosion. Judgement must be made on the extent of risk if intermediate values are encountered. Results obtained from dry soffits, saturated structures, or elements in contact with water, require careful interpretation. It is important to note that the equipment can identify only active corrosion sites. The absolute values of electrical potential can be affected by carbonation, oxygen starvation, chloride content, and temperature. Therefore, its primary strength is as a device for comparative studies of corrosion activity on similar members of a particular structure.

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Electrical resistivity measurement Surface resistivity measurements are useful for the relative assessment of corrosion risk in different parts of a structure, assisting decision-making in respect of repair strategy and prioritising areas for protection from corrosion or rehabilitation. The process involves the use of one, two, or four electrode devices. The measurement of resistivity basically involves passing electrical current via the electrodes through a defined length of the concrete at a measured voltage from which the resistance may be deduced, and thereby the resistivity. The electrodes may be a single disc on the concrete surface and a connection to the reinforcement cage, two electrodes on the surface, or four electrodes on the surface. The most commonly used device is the Wenner four-point apparatus, passing a current of 10 μA to 200 μA. The basis of the four-electrode technique is illustrated in Fig. 5.17, showing an arrangement of two outer and two inner probes at known spacings and in contact with the concrete surface. An alternating current is passed between two outer probes and the voltage between two inner probes is measured. The resistivity (ρ) can be determined from the following relationship: =2

a

V I

where a = probe spacing V = voltage between inner probes I = current between outer probes.

Figure 5.17 Test set-up for electrical resistivity measurement using a Wenner four-point apparatus in direct contact with a concrete surface, showing the significance of probe spacing.

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The probe spacing must be such that the resistivity of the cover concrete is being determined. If the spacing is too wide, the electrical field will be interfered with by the reinforcement; if it is too narrow it will be overly influenced by surface effects. The resistivity readings are significantly influenced by the moisture state of the concrete. It is therefore important to survey the concrete in the state of greatest interest, taking account of the local climate. This can be difficult to achieve in practice. So much so that Cheytani and Chan (2021) have proposed that four individual metal probes be inserted to a depth of 15 mm from the surface of the element to reach a stable moisture level. The Wenner four-point apparatus would then be placed on the protruding inserts, rather than directly onto the concrete surface. Resistivity values in the range of 10–100 Ωcm can be used to interpret the likely risk of corrosion activity. However, resistivity measurements are critically dependent on the moisture level, as discussed earlier in this chapter, which can create difficulties in making accurate conclusions from site measurements subject to the prevailing weather conditions. Concrete absorbs water rapidly but can be slow to dry out. Thus, moist conditions can persist even when rainfall is sporadic, making resistivity a useful ─ but not perfect ─ tool. Equally, it cannot be used in sub-zero temperatures. The recommendation, therefore, is to use it as a relative measure or to calibrate absolute values against reference specimens of similar cement type in a controlled temperature and relative humidity. Other devices are available for measuring resistivity on-site, such as one commercially available as part of a linear polarisation resistance measurement system, and this tends to be the favoured apparatus among researchers for reliable data on corrosion cell activity. A recommendation on testing was made by RILEM Technical Committee TC 154-EMC (2000), also reported by Polder (2001). Surface resistivity measurements are distinct from “bulk” electrical resistivity, as measured, for example, by standard test methods ASTM C1876 (ASTM 2019f) and CSA A23.2–26 C (CSA 2019b). The bulk test provides an indication of resistance to the penetration of fluids and aggressive ions. These tests are described in more detail in Chapter 7 in respect of their correlation with rapid chloride penetration testing. In summary, these tests involve placing a bulk sample, such as a cylinder, between two plate electrodes and forcing an AC current across the sample. An average value of the bulk electrical resistivity of the concrete is determined. European standard EN 12390–19 (CEN 2023) includes both bulk and surface test methods, but it is expressly stated that the use of resistivity to assess the potential for corrosion of reinforcement in existing structures is not specified in this pre-standard.

Linear polarisation resistance measurement The half-cell mapping technique can be used to identify corrosion activity but does not give an indication of the rate of corrosion. Equally, resistivity

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measurements can only indicate the likely corrosion rate if it occurs. Taken together, the two techniques can yield very useful information, but if a measure of the corrosion rate is required, this can best be achieved through the polarisation resistance technique. The technique involves making an electrical connection to the reinforcement, followed by the application of a small current and measurement of the shift in potential. The corrosion rate may then be deduced. If a second electrode is introduced to a reinforced concrete system and an external current is applied, the anodic and cathodic potential is shifted, or polarised, from the corroding potential (Ecorr). An external corrosion current then exists between the two electrodes and can be measured. At very low external polarisation voltages, a linear relationship exists between external current and the polarisation. The polarisation resistance (Rp) is defined as follows: Rp =

E I

where δE = change in potential δI = applied current. The polarisation resistance is inversely proportional to the notional corrosion current at the mixed potential (Ecorr) and, therefore, the corrosion rate may be detected. Experience with a commercially available device that has been developed for on-site polarisation resistance measurements is described by Rodriguez et al. (1995). The device confines the applied current by means of a circular sensor with concentric counter electrodes (Feliu et al. 1990). The sensor houses a central reference electrode and a pair of electrodes between the concentric rings (Fig. 5.18).

Figure 5.18 Test set-up for linear polarisation resistance measurement.

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The polarisation resistance is determined and the corrosion rate (Icorr) is deduced from the following relationship: Icorr =

K Rp

where the constant K is taken as 26 mV for corroding reinforcement and double that figure for passive reinforcement. Rodriguez et al. (1995) report that rates less than 0.2 μA/cm2 indicate reinforcement in the passive condition; values up to 0.5 μA/cm2 indicate low to moderate rates; values in the range of 0.5 μA/cm2 to 1.0 μA/cm2 indicate moderate to high rates. Values greater than these are indicative of high corrosion rates.

Real-time corrosion monitoring in service The detectable electrical circuit characteristics of reinforcement corrosion in concrete have also been used as a non-destructive testing tool for monitoring the performance of structures in service. This early warning system allows preventative maintenance measures to be implemented, rather than later costly repairs. The initial investment is readily recouped in the maintenance of high-risk structures, such as multi-storey carpark decks subject to chloride exposure from the dripping underbodies of vehicles using de-iced highways. Corrosion sensors typically monitor the potential gradient between the surface and the reinforcement or the electrical resistance of the cover zone. The game-changer in recent years has been the ease with which the data can be accumulated and communicated through radio frequency identification (RFID) technology, mobile phone SIM cards, and the internet, allowing real-time remote monitoring and automated alert messaging. Details of the components of a system to automate data collection for monitoring the risk of corrosion, together with communication hardware and bespoke software to interpret the results, are exemplified by a system designed by Parthiban et al. (2006). A compact self-powered system has been described by Eriguchi (2022) that can be fully encased in new structures or retrofitted to existing structures. There are no external wires, which makes it ideally suited to exposed site conditions in maritime environments. Simplicity is key, with just a sensor and RFID tag. The data is retrieved by a read/write device held over the buried tag, which is powered by radio waves transmitted from the external reader. Many other examples of harnessing the detectable characteristics of corroding reinforcement married to data acquisition systems exist. These are based on the principles of the corrosion measurement tools outlined in the previous section.

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SUMMARY An understanding of the electrochemical process of corrosion is required to fully appreciate the characteristics of durable concrete. Prevention of corrosion is best approached by creating conditions where the reinforcement remains passive and where the ingress of moisture and oxygen is severely limited. The threat to passivity comes from carbonation and chloride ingress. Corrosion activity may take the form of homogenous corrosion or pitting corrosion. The latter is commonly associated with chloride ingress and represents a significant threat to durability. Ingress of oxygen and moisture may be limited by the specification and achievement of impermeable concrete through a low water/cement ratio, good compaction, and good curing, and in most cases, the influence of cracking will not be significant. Modelling of the propagation phase of deterioration due to corrosion is not yet as advanced as that for the initiation phase, but progress is being made on this in order to extend the predicted design life of structures in the context of achieving sustainable development through optimal use of resources. It is estimated that inclusion of the propagation period in durability design could extend the notional design life by a decade or more. Where corrosion activity occurs, it can be detected through several techniques that exploit the electrochemical characteristics displayed by an active corroding cell. REFERENCES Andrade, C. and C. Alonso. 2001. On-site measurements of corrosion rate of reinforcements. Construction and Building Materials 15, 2–3: 141–145. Andrade, C., Alonso, C. and J. González. 1986. Some laboratory experiments on the inhibitor effect of sodium nitrite on reinforcement corrosion. Cement, Concrete and Aggregates 8: 110–116. ASTM. 2019f. C1876, Standard test method for bulk electrical resistivity or bulk conductivity of concrete. West Conshohocken: ASTM International. Bazant, Z. 1979. Physical model for steel corrosion in concrete sea structures application. ASCE Journal of the Structural Division 105, 6: 1155–1166. Bentur, A., Diamond, S. and N. Berke. 1997. Steel corrosion in concrete: fundamentals and civil engineering practice. London: CRC Press. BRE. 1993. Concrete, cracking and corrosion of reinforcement. Digest 389. Watford: Building Research Establishment. Browne, R. 1988. Durability of reinforced concrete structures. In Proceedings, Pacific Concrete Conference, 847–886. Auckland. 8-11 November. Wellington: NZ Concrete Society. CEN. 2023. EN 12390-19. Determination of resistivity. Brussels: Comité Européen de Normalisation. Chen, E., Berrocal, C., Löfgren, I. and K. Lundgren. 2020. Correlation between concrete cracks and corrosion characteristics of steel reinforcement in precracked plain and fibre-reinforced concrete beams. Materials and Structures 53, 2: doi: 10.1617/s11527-020-01466-z

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Cheytani, M. and S. Chan. 2021. The applicability of the Wenner method for resistivity measurement of concrete in atmospheric conditions. Case Studies in Construction Materials 15: doi: 10.1016/j.cscm.2021.e00663 Concrete Society. 2015. The relevance of cracking in concrete to corrosion of reinforcement. Technical Report TR44. Camberley: Concrete Society. CSA. 2019b. CSA A23.2-26C, Test methods and standard practices for concrete: bulk electrical resistivity of concrete. Mississauga: Canadian Standards Association. Eriguchi, A. 2022. Wireless monitoring system for reinforcement corrosion detection. Concrete 56, 1: 17–18. Feliu, S., González, J., Feliu Jr., S. and M. Andrade. 1990. Confinement of the electrical signal for in situ measurement of polarization resistance in reinforced concrete. American Concrete Institute Materials Journal 87, 5: 457–460. Ferreira, M., Bohner, E. and O. Saarela. 2016. Designing concrete durability by coupling limit state of corrosion initiation and corrosion induced cracking of concrete cover. Nordic Concrete Research 54: 7–20. Lawrence, C. 1990. The mechanism of corrosion of reinforcement steel in concrete structures. Crowthorne. British Cement Association. Lu, Z-H., Lun, P-Y., Li, W., Luo, Z., Li, Y. and P. Liu. 2019. Empirical model of corrosion rate for steel reinforced concrete structures in chloride-laden environments. Advances in Structural Engineering 22, 1: 223–239. Otieno, M., Beushausen, H. and M. Alexander. 2011. Prediction of corrosion rate in RC structures - a critical review. In Proceedings, Joint fib-RILEM workshop: modelling of corroding concrete structures, ed.C. Andrade and G. Mancini. RILEM Bookseries. 5. Dordrecht: Springer. Parrott, L. 1994. Design for avoiding damage due to carbonation-induced corrosion. In Durability of Concrete. Proceedings of the Third International Conference, ed. V. Malhotra, 283–298. Special Publication SP-145. Farmington Hills: American Concrete Institute. Parthiban, T., Ravi, R. and G. Parthiban. 2006. Potential monitoring system for corrosion of steel in concrete. Advances in Engineering Software 37, 6: 375–381. Polder, R. 2001. Test methods for on site measurement of resistivity of concrete - a RILEM TC-154 technical recommendation. Construction and Building Materials 15, 2-3: 125–131. RILEM TC 154-EMC. 2000. Electrochemical techniques for measuring metallic corrosion: test methods for on site measurement of resistivity of concrete. Materials and Structures 33, 10: 603–611. Robles, K., Yee, J. and S. Kee. 2022. Electrical resistivity measurements for nondestructive evaluation of chloride-induced deterioration of reinforced concrete - a review. Materials 15, 8. doi: 10.3390/ma15082725 Rodriguez, J., Ortega, L. and A. Garcia. 1995. On-site corrosion measurements in concrete structures. Construction Repair 9: 27–30. Rodriguez, J., Ortega, L., Casal, J. and J. Diez. 1996. Corrosion of reinforcement and service life of concrete structures. Durability of Building Materials and Components 7. ed. C. Sjöström, 117–126. London: Routledge. Tuutti, K. 1982. Corrosion of Steel in Concrete. Stockholm: Swedish Cement and Concrete Research Institute. Van Daveer, J. 1975. Techniques for evaluating reinforced concrete bridge decks. Journal of the American Concrete Institute 72: 697–704.

Chapter 6

Carbonation

CARBONATION AND CORROSION Carbonation is a term used to describe the dissolution of carbon dioxide (CO2) gas in a liquid. Carbonation of concrete occurs when atmospheric CO2 diffuses into the permeable pore structure of the cover zone, dissolves in the pore solution and reacts with calcium hydroxide and calcium silicate hydrates with subsequent precipitation of calcium carbonate (CaCO3). Essentially, the motivation for the reaction is neutralisation of the bases in the concrete by acid. The term “neutralisation” is sometimes also used in the literature. This naturally results in a diminution of the alkaline environment in the cover zone, which exhibits a reduction in the pH of the pore solution in contact with the carbonated concrete. Unfortunately, this impacts on the potential durability of reinforced concrete structures by acting as a gateway to reinforcing steel corrosion, even though it does not weaken the concrete itself. Reinforcement in concrete is embedded in an oxygenated alkaline solution. The reinforcement will not corrode if the protection afforded by the passive film – a thin layer of oxide deposited on the steel – remains substantially intact. This insoluble oxide film prevents oxygen reaching the steel and inhibits corrosion. The reinforcement is said to be “passive” when it is in this state. Corrosion of reinforcement can commence, however, if the passive oxide film protecting the reinforcement is destroyed, the cover concrete is sufficiently permeable to oxygen and moisture, and the concrete is moist enough to serve as an electrolyte. The lowered pH in zones of carbonated concrete may threaten the continuity of the passive film. The simplest method of avoiding any possible problems with corrosion damage induced by carbonation is to specify cover concrete that is capable of resisting the penetration of the carbonation front as far as the reinforcement during the service life of the structure. However, corrosion may not automatically result from carbonation front penetration, and a lessconservative approach to durability management is recommended by Angst et al. (2020) so that low binder content innovative concretes may be used in carbonation exposure conditions. In summary, the propagation period is DOI: 10.1201/9781003261414-6

147

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their focus of attention, rather than the initiation period used up to now as the determinant of service life. The pH of the pore solution of fresh concrete is approximately 12.6. The alkaline nature of concrete is principally due to the presence of calcium hydroxide, Ca(OH)2, formed during the hydration of the cementitious binder. Dissolution of the Ca(OH)2 leads to the presence of hydroxyl ions in the pore water and this gives concrete its high pH. Calcium hydroxide is susceptible to reaction with CO2 from the atmosphere. The reaction proceeds in the presence of moisture as water provides a medium for the reaction. Concretes in service will almost always contain sufficient moisture for the reaction to proceed. The reaction involves the production of CaCO3. Conversion to CaCO3 influences the surrounding pore fluid pH, which falls to about 8.3. This fall in pH results in a carbonation front which divides low pH carbonated and high pH non-carbonated zones. The durability of reinforcement is compromised if the carbonation front advances to or beyond the depth of cover. This is because the passive ferrous oxide layer on the reinforcement breaks down when the surrounding pH falls below 9 – so called “depassivation”.

CHEMISTRY OF CARBONATION IN CEMENTITIOUS SYSTEMS The topic of carbonation of concrete as a durability concern has mainly been researched since the 1960s. However an awareness of the interaction between cementitious systems and CO2 dates back to the nineteenth century. The matter was being actively discussed at annual general assemblies of the German Portland Cement Manufacturers. The earliest such paper in the author’s collection is that of Tomei (1880) at the third meeting. A decade later, Passow (1896) addressed the 19th Assembly on the topic of mortar and CO2 while six experts spoke at the 20th Assembly on the topic of CO2 and cement (Schiffner et al. 1897). It has also been stated (Huhta 1960) that a patent for curing of concrete products through the use of CO2 was taken out in 1868. The curing of unreinforced concrete blocks by CO2 can aid early strength and reduce shrinkage in service. The carbonation reaction in concrete is only a threat to the durability of reinforced concrete through initiation of metal corrosion – unreinforced concrete structures are not weakened by the carbonation process. Indeed there is growing interest in the natural sequestration of carbon dioxide by concrete. Climate change policies impacting on the construction sector often incorrectly assume that the net contribution of cement and concrete production to greenhouse gas levels are based solely on CO2 emissions. It would be more accurate if life-cycle assessments factored in the amount of CO2 sequestered. Preliminary estimates range from 75 kg to 125 kg CO2 per tonne of cement, for a 100-year service life, thus representing about 20% of the total calcined CO2 per tonne of cement produced (Andersson et al. 2013; Fitzpatrick et al. 2015).

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149

Initially, CO2 diffuses through the surface of the concrete due to the concentration difference between the atmosphere and the concrete pore structure. A thin skin of carbonated concrete develops, which may be less than a millimetre in thickness. Further penetration is primarily a function of the concrete permeability and the amount of calcium hydroxide available for reaction. Carbon dioxide passes unhindered through the carbonated layer and is available for reaction with the next layer of calcium hydroxide. It may progressively penetrate further into the concrete over time and, ultimately, part of the carbonation front may reach the reinforcement and cause depassivation (Fig. 6.1). Carbon dioxide can react with the constituents of both unhydrated and hydrated cement. In the context of durability, the latter is of greater interest. The various reactions have been summarised by, for example, Aschan (1963), Berger and Klemm (1972), and Papadakis et al. (1989). The most significant reaction is that involving calcium hydroxide and therefore carbonation of concrete is generally summarised as follows: Ca (OH )2 + CO2

H2 O

CaCO2 + H2 O

The steps of the reaction are, first, the dissolution of calcium hydroxide, followed by reaction with dissolved CO2: Ca (OH )2

Ca2+ + 2OH

Ca2+ + 2OH + CO2

CaCO3 + H2 O

Additional reactions involving unhydrated constituents and calcium silicate hydrate are possible whereby the CaO in these compounds combines with CO2 to form calcium carbonate and hydrated silica: 3CaO. SiO2 + 3CO2 + yH2 O

SiO2 . yH2 O + 3CaCO3

2CaO . SiO2 + 2CO2 + yH2 O

SiO2 . yH2 O + 2CaCO3

Figure 6.1 Ingress of the carbonated zone to the reinforcement.

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The mineralogical forms of calcium carbonate are calcite, aragonite, and vaterite. Calcite, the most stable, is the ultimate form found in carbonated concrete (Gaze and Robertson 1956). Sauman (1971) found that the carbonation process initially produced some vaterite but that this was gradually converted to the more stable calcite. This may be influenced by the degree of hydration. It has been reported by Kondo et al. (1969) that the carbonation of hydrated tricalcium silicate (C3S) produced calcite, whereas unhydrated C3S produced vaterite. The vaterite level decreases if moisture is available to maintain the process of hydration. Cole and Kroone (1960) reported that vaterite is formed first, and that this is transformed into aragonite and finally into poorly crystallised calcite, following Ostwold’s law of successive reactions. The process of carbonation in concrete is not uncontrollable. It is possible to produce durable concrete in which the rate of carbonation may be so slow that, in engineering terms, the carbonation front may be regarded as having reached a limit. This limit may be of the order of a millimetre or less. This situation can occur through a further lowering of the diffusivity in highly impermeable concrete by deposition of carbonation products because CaCO3 occupies a greater volume than the reacting constituents.

PRIMARY FACTORS INFLUENCING CARBONATION RATE Carbonation rate is significantly influenced by several factors that interact and the combined effect of these may exacerbate or ameliorate the process. The primary factors, determined from both field observations and theoretical considerations, are: • • • •

diffusivity and permeability reserve alkalinity environmental CO2 concentration exposure condition.

Diffusivity and permeability The carbonation phenomenon is governed by the diffusion process whereby material migrates from an area of high concentration to one of lower concentration. Permeability, on the other hand, is generally characterised by flow under the influence of a pressure difference. Nevertheless, the permeability properties of a concrete are often used to indicate its quality with respect to carbonation resistance. The lower the permeability the greater the resistance to the inward diffusion of CO2 into the concrete, for a given moisture content. The requirements for achieving low permeability concrete are well-known and include low water/cement ratio, proper compaction, and adequate and

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151

timely curing. Strength is related to these three factors and so it may be postulated that the carbonation rate should be inversely proportional to strength. While strength is not a significant factor in itself, lower rates of carbonation have generally been observed in stronger concretes. The rate of carbonation in high quality concretes may be so low that the carbonation depth tends to a limit of a millimetre or less. Low depths of carbonation may also be detected on occasion in weak, highly permeable concretes exposed to rain. In such cases, the diffusion of CO2 will have been fortuitously hindered by water-filled pores. Carbon dioxide diffusion is 106 times slower in water than in air. Nevertheless, carbonation rates in tropical climates, such as in Singapore, have been found to be higher on average than in comparable concretes in temperate climates despite higher relative humidity levels. A possible explanation for this has been proposed by Roy et al. (1996), who note that the higher temperatures in tropical climates may influence the diffusion coefficient. It may also be noted that a 10°C rise in temperature is generally associated with a doubling of reaction rate in chemical reactions.

Reserve alkalinity The ability of a concrete to resist the progress of the carbonation front is related to the volume of calcium hydroxide present. This reserve of alkalinity may be considered as a buffer with the calcium hydroxide acting as a sacrificial defence system – the more calcium hydroxide present the more effective the filter against CO2. The calcium hydroxide volume is proportional to the calcium oxide content of the cementitious binders. The reserve alkalinity of concrete made from blended cements such as Portland fly ash cement is lower than that made with Portland cement. In relation to carbonation, however, this disadvantage is counterbalanced by the lower permeability of well-cured blended cement concretes. Indeed, the influence of the water/cement ratio on the carbonation rate is greater than the cement content. The benefits of materials such as pulverised fuel ash in the carbonation context are related to the reduction in water demand for a given workability. Further support for the proposition that strength plays a significant role in diminishing the carbonation rate comes from consideration of the role of cement content. Under favourable hydration conditions, strength is proportional to cement content. The higher the cement content the greater the amount of calcium hydroxide produced and, therefore, the higher the reserve alkalinity.

Environmental carbon dioxide concentration The rate of carbonation increases with increasing CO2 content. Highest rates, other than in particular industrial situations, are currently found inside

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Figure 6.2 Influence of environment on the rate of carbonation.

buildings, but there is obviously concern over the general rate of carbonation increasing in all environments due to the effect of greenhouse gas emissions. The level of CO2 in the atmosphere varies slightly from a minimum in coastal environments to a maximum in urban environments. Highest levels typically occur in the interior of buildings. This has been found to influence the rate of carbonation as illustrated in Fig. 6.2 using data from a study in Ireland (Richardson 1988). Atmospheric CO2 levels, currently quoted as 360 parts per million (ppm), are rising. Hayward (1997) reported that the concentration in 1800 averaged 280 ppm, but that this had risen to 340 ppm by 1990. Moss et al. (2010) developed four scenarios for atmospheric CO2 levels in the 21st century. The forecast is that the best case scenario will see a rise to a peak level of 490 ppm before declining. The worst-case scenario would see a rise to 1370 ppm CO2 equivalent level by the year 2100. Rates in the coastal environment are low, perhaps due to high humidity, but of course, chloride-induced corrosion would pose a greater threat to structures in coastal and marine environments. Rates in rural and suburban areas are typically lower than in urban areas, but in practical terms, the differentiation is not meaningful, as microclimates with high CO2 levels can occur in any area. The rates inside buildings are high due to low humidity levels and consequent semi-dry permeable pore structure as much as being due to the higher CO2 concentration. However, the low humidity levels discourage corrosion initiation, and so the internal environment is not always a cause for concern. It should be borne in mind that carbonation and corrosion are not inexorably linked. In certain concretes, the carbonation front may advance beyond the cover concrete, but the relative humidity of the atmosphere may not be high enough to sustain corrosion. However, each internal atmosphere in an industrial setting must be individually evaluated for its unique risk profile of a carbonation and corrosion combination. For example, Lv et al. (2022) found elevated temperatures and carbon dioxide levels in a steelworks averaging 32.7oC and 525 ppm respectively, which exacerbated carbonation rates, but the average relative humidity was 45.9%, which inhibited corrosion propagation.

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153

Figure 6.3 Influence of relative humidity on the rates of carbonation and corrosion.

Rates of carbonation are highest in the range 50 to 75% relative humidity. Below about 45% relative humidity, the water in the pores is not in a state that encourages dissolution of Ca(OH)2 or CO2 and this reduces the reaction rate. Above 75% relative humidity, the influence of water filling the pores becomes significant. The critical relative humidity level with respect to corrosion is about 80%, below which there is insufficient moisture to sustain the process. The trends are illustrated in Fig. 6.3. The impact of climate change will be to shift the annual average temperature and relative humidity over the coming century. Allied to increasing CO2 levels, this will impact on carbonation rates to varying degrees in different geographical locations. For example, Peng and Stewart (2016) predict that mean carbonation depths in Chinese cities may increase by 45% in 2100, based on case studies of the Jinan, Kunming, and Xiamen environments. Mizzi et al. (2018) predict a 40% rise in carbonation depths for structures in Malta by 2070 if worst case climate scenarios play out in the Mediterranean. Although the effect will vary from district to district due to microclimates, Bastidas-Artega et al. (2022) found that climate change will lead to significantly earlier initiation of corrosion by carbonation after 2050 if there is no change to current concrete practice. It should be noted that very low rates of carbonation have also been observed in outdoor concretes of all environmental categories. This further highlights the significance of exposure to moisture as a key parameter.

Exposure condition Low rates of carbonation have been observed in outdoor concretes of all strengths and in all environmental categories, emphasising the significance of exposure and moisture level in the permeable pore structure. It would

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Figure 6.4 Influence of exposure condition on the rate of carbonation.

appear that moisture plays a fundamental role in the phenomenon and greatly influences the carbonation rate. This is most probably due to the significant effect on the diffusion coefficient of the moisture content of the pore structure. The permeability of wet concrete is a fraction of that for dry or semi-dry concrete. This is illustrated by analysing data from a study of concrete in Ireland (Richardson 1988), categorised as follows: • external, exposed to rain • external, sheltered from rain • internal. The results are illustrated in Fig. 6.4. These trends agree with studies in other countries of temperate climates using the same categories. The external environment in Ireland includes moderate to high relative humidity. Rainfall in excess of one millimetre per day occurs, on average, half the time. It may be anticipated, therefore, that concrete in the “external, exposed to rain” category would be wet and rarely dry. The permeability of these concretes is therefore low, irrespective of the quality of the concrete. Concrete in the “external, sheltered from rain” category experiences moderate humidity. The durability of these concretes is critically dependent on the concrete quality. Concrete in the internal environment is usually dry, potentially leading to high permeability, and it is also exposed to high CO2 concentrations. Nevertheless, the likelihood of corrosion of reinforcement in such an environment is low. MATHEMATICAL MODELLING OF THE RATE OF CARBONATION The phenomenon of carbonation provides one of the earliest examples of the application of mathematical modelling to deterioration processes in concrete.

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155

A formula was published in 1928 by Uchida and Hamada, showing the depth of carbonation to be proportional to water/cement ratio and the square root of time. Since then many models have emerged which indicate that a square root relationship provides a reasonable basis on which to predict future behaviour, being somewhat conservative for most exposure conditions. Observations in the field have shown that carbonation rates lower than that predicted by the square root model are experienced where the concrete is subject to rain. Nevertheless, a significant aspect of the (approximately) square root relationship to specification and site practice is that a reduction of cover by a quarter could halve the time to depassivation.

Empirical formulae This square root relationship may be stated as follows:

x=k t where x = depth of carbonation k = ‘k-factor’ dependent on diffusivity, reserve alkalinity, CO2 concentration, and exposure condition t = time This approximate relationship has been verified by observations of concrete in the laboratory and in service. For example, Alexandre (1976) published the following model:

x=

2k1 t

where k1 = material coefficient Equally Alekseev and Rozental (1976) published the formula: x = An t

where A = coefficient n = 1.92 for water/cement ratio 0.6 n = 2.54 for water/cement ratio 0.7 Inclusion of the water/cement ratio is logical since it would have a significant influence on the permeability of the concrete and hence on the ease with which CO2 could diffuse into the concrete. The significance of water/cement

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ratio was long recognised. The formula of Uchida and Hamada (1928), referred to earlier, may be stated in the following form: x=

w0

0.3

t

0.3(1 + 3w0)

where w0 = water/cement ratio Similarly, formulae by Kishitani (1960) can be presented in the following forms for water/cement ratios (w) below 0.6 and equal to or above 0.6 respectively: x = x =

4.6w 1.76 t 7.2 w 0.25 0.3(1.15 + 3w)

t

It may be noted that the second of Kishitani’s formulae is basically that of Uchida and Hamada, but with w0 substituted with (w + 0.05). The difference may be accounted for by the moisture state of the sand used in their experiments. Further development of these formulae was published by Kishitani (1964) and Hamada (1969) with the introduction of a factor “R”, which took account of cement type, aggregate type, and agents such as air-entrainment admixtures. Tsukayama et al. (1980) published a formula along similar lines to that of Kishitani, but with a slightly lower value of the material factor. It may be stated in the following form: x = (1.187w

0.493) t

The foregoing set of formulae concentrated on the influence of water/ cement ratio, recognising its significant influence on permeability. The effect of cement content, which influences the permeability and reserve alkalinity was implied but not specifically included. This was addressed by Smolczyk (1969), who introduced the influence of cement content by inclusion of compressive strength in a formula for carbonation: x=a

w NT

b

t

where a and b are coefficients and NT represents the compressive strength at T days. Another variation is that of Weirig, published by the Commission Carbonatation (1972), as follows:

Carbonation

x=

582w + 1.12w s osf

1.47

157

t

where s is the loss on ignition and osf is the specific surface. Other empirically derived formulae have been published by, for example, Richardson (1988), which also demonstrate the enduring design of empirical models around the square root assumption.

Theoretical verification of the square root relationship The empirical models are limited in their usefulness because they cannot be extrapolated to predict performance in all circumstances. Therefore, a more fundamental approach to modelling the carbonation phenomenon has been undertaken by several researchers. A theoretical basis for the square root relationship has been proposed since the 1960s by researchers such as Meyer et al. (1967), Alekseev and Rozental (1976), and Engelfried (1977). The theory involves equating the two processes involved: diffusion of CO2 and the chemical reaction between CO2 and calcium hydroxide. The relevant parameters are listed below and illustrated in Fig. 6.5. n = quantity of CO2 diffusing through element (kg) D = diffusion coefficient (m/s2) A = area (m2) r = concentration of CO2 at surface (kg/m3) rx = concentration of CO2 at depth × (kg/m3) c = alkaline material available in a unit volume to buffer CO2 (kg/m3) t = time (s) The diffusion process is governed by Fickian diffusion. Thus, using standard sign convention, we may state:

Figure 6.5 Parameters in mathematical model of carbonation rate.

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Fundamentals of Durable Reinforced Concrete

dn =

D A

r

rs x

dt

or

dn = D A

r

rs

dt

x

The chemical processes may be stated as:

dn = c A dx Hence c A dx = D A

r

rs x

dt

and x dx = D

r

rs x

dt

By integration, we may solve for x. In addition, it may be assumed that the CO2 concentration at the depth of carbonation will be zero. Thus: x=

2 D r t c

The quantity of CO2 in the atmosphere (parameter r) is about 360 ppm, which equates to 7.2 × 10–4 kg/m3. The amount of alkaline material available in a unit volume of concrete to buffer the CO2 (parameter c) may be calculated as the product of the cement content, the fraction of calcium oxide in the cement (for example, 0.65 for Portland cement), and the ratio of molecular weights of CO2 (44.009) to calcium oxide (56.0774), a value of 0.785. A slight restatement of the above formula is that of Kondo et al. (1969):

x=

2 D c0

c

t

where the denominator is the product of the amount of reactant per unit weight and the density.

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Formula based on reaction engineering A fundamental model of carbonation of concrete was developed by Papadakis et al. (1989) based on reaction engineering principles. The model considers the mass conservation of CO2, calcium hydroxide, and calcium silicate hydrate. The formula was derived by considering the kinetics of the hydration and carbonation reactions, the decrease in porosity consequent on the fact that the products of carbonation have higher molar volumes than the calcium hydroxide and calcium silicate hydrate, and the influence of humidity. The model has been found to give good agreement with experimental results at typical relative humidity levels (Papadakis et al. 1990). For fully hydrated concrete, it is of the form: x=

2[CO2] De, CO2 [Ca (OH )2 ] + 3[CSH ]

t

where [CO2] = molar concentration of CO2 (mol/m3) [Ca(OH)2] = molar concentration of calcium hydroxide (mol/m3) [CSH] = molar concentration of calcium silicate hydrate (mol/m3) De, CO2 = effective diffusivity of CO2 (m2/s) A further 19 formulae are employed in the solution of the model involving parameters such as the cement composition, water/cement ratio, aggregate/ cement ratio, CO2 concentration, and the relative humidity. Certain simplifications can be applied to the model while retaining its applicability (Papadakis et al. 1992). These are as follows: [CO2 ] = 42 yCO2

[Ca (OH )2 ] + 3[CSH ] =

De, CO2 = (1.64 × 10 6)

33 000 1 + ( c / w)(w /c) + ( c / a)(a / c) 1.8 p

with

p

=

c w

w /c 1 + ( c/

0.3 )( w w / c)

1

RH 2.2 100

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Fundamentals of Durable Reinforced Concrete

where yCO2 = ambient CO2 content by volume ρc = mass density of cement (kg/m3) ρw = mass density of water (kg/m3) ρa = mass density of aggregate (kg/m3) w/c = water/cement ratio a/c = aggregate/cement ratio ∈p = porosity of hardened cement paste RH = relative humidity A simplified version of the expression for effective diffusivity, which was found to fit well with experimental data, is: 2 p

De, CO2 =

RH 2 100

1

where β = 1.2 ± 0.1 Thus for ordinary Portland cement concrete the depth of carbonation may be modelled by the following formula: xc = 350

w /c

c w

1+

c/ w

0.3

1

(w / c)

RH 100 0.5

c

1+

w

(w /c) +

c a

(a / c) yCO2 t

Deviation from the square root relationship Consideration of a range of material properties led Daimon et al. (1971) to publish a formula that took account of the deviation from the square root relationship through the concept of an induction period. The formula was of the form: x=

ke

where ke =

kt f

t

ti

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161

and kt =

2 D0 P T 0.75 2731.75 R c0

where ti = “induction time” T = exposure period ke = rate constant obtained experimentally kt = theoretical rate constant f = tortuosity D0 = diffusion coefficient at 0°C ∈ = porosity of specimen ΔP = partial pressure difference across the carbonated layer T = temperature R = gas constant c0 = amount of reactant in unit weight of concrete specimen ρ = density of specimen

Tendency to a limit In practice, particularly in permanently or cyclically wet environments, the carbonation depth tends to a limit. The limiting value of carbonation depth is thought to be due to two factors. The first involves the back diffusion of material from the pore water in the non-carbonated zone to the carbonated zone, due to the concentration difference. The second involves the reduction in diffusion coefficient caused by the deposition of calcium carbonate in the concrete pores. The reaction products of carbonation occupy a greater volume than the constituents that react. Thus, a limit may be reached when the amount of CO2 reaching the carbonation front is low enough to just about react with the counter diffusing hydration products. An alternative possibility is that the pores may become waterlogged for periods thus slowing the diffusion of CO2 to an extent that change in carbonation depth is not detectable. This has been addressed by several researchers, including Martin et al. (1975), Schiessl (1976), Frey (1993), Bakker (1994), and Sickert (1997). Schiessl (1976) demonstrated that the Fickian diffusion model could still be used taking these factors into account. First, the diffusion coefficient was not constrained to be constant and was allowed to decrease with time due to the reduction in permeability caused by deposition of calcium carbonate in the pore structure of concrete. The diffusion coefficient would also vary with time due to fluctuations in moisture level. Thus, the diffusion coefficient (D) was replaced by a mean value (Dm): Dm = D (1

f x)

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where f = environment factor Second, the counter diffusion of calcium hydroxide was taken into account by assuming that an additional amount of CO2 (b kg/m2) was required to enter the concrete to react with the counter diffusing calcium hydroxide. The equations for the physical and chemical processes therefore became: dn = D (1

f x) A

r

rx x

dt

and dn = c A dx + b A dt

Thus: c A dx + b A dt = D (1

f x) A

r

rx x

dt

and dx =

f x) r

D (1 c

rx

b dt c

x

The solution to the equation as time approaches infinity is the value: x =

Dr

rx d

where d = f Dr

rx + b

Thus the solution of these equations for the depth of carbonation approaches a limit as time approaches infinity, validating the link between theoretical and observed behaviour. The departure from the strict square root relationship is also accounted for. The considerations taken into account by Schiessl were also studied by Frey (1993) and further examined by Sickert (1997). Frey proposed the following formulation: dk = (e Aw

2

1)(1

e

)

w t 0.5

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Figure 6.6 Influence of wetting and drying cycles on the rate of carbonation.

where dk = depth of carbonation (mm) t = time (years) A = parameter (for example, 0.11 year–1) W = carbonation characteristic number (for example, 0.25 year–0.5) The carbonation characteristic number is dependent on the concrete composition. The formula reaches a limiting value as time approaches infinity of: dke = (e Aw

1)

2

Bakker (1994) further investigated the tendency towards a limit in circumstances where the concrete was subject to cyclical wetting and drying. The effect of wetting and drying cycles on carbonation is illustrated in Fig. 6.6. Bakker showed that if one considers the wetting time tw and related drying time td to be as follows: tw = tn

tn

1

xn = A teff

n

td =

xn 1 B

then:

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Fundamentals of Durable Reinforced Concrete

where teff

n

x1 B

= td1 + td 2 xn 1 B

... + tdn

2

...

2

and A =

2 Dc (C1 a

C2)

B =

2 Dv (C3 b

C4)

if xn = carbonation depth after n cycles (m) n = number of cycles a = amount of alkaline substance in concrete (kg/m3) Dc = effective diffusion coefficient for CO2 at a given moisture distribution in the pores (m2/s) C1 – C2 = CO2 concentration difference between air and carbonation front (kg/m3) tdn = length of nth drying period (s) Dv = effective diffusion coefficient for water vapour at a given moisture distribution in the pores (m2/s) C3 – C4 = moisture difference between air and the evaporation front (kg/m3) b = amount of water to evaporate from the concrete (kg/m3) td = length of drying period (s) tw = length of wetting period (s) The phenomenon has been described as “self termination” of carbonation by Woyciechowski and Sokolowska (2017), who proposed a hyperbolic rather than a parabolic model of carbonation of the form x = f (t

0.5)

with x=a

w + b (tec) + c (t c

0.5)

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165

where x = carbonation depth (mm) w/c = water/cement ratio tec = curing time (days) t = exposure time (years) a, b, c = coefficients mainly dependent on binder type and CO2 concentration APPLICATION OF MODELS TO SERVICE LIFE PREDICTION The use of mathematical models for service life prediction often involves the choice of a model that strikes a balance between accuracy with complexity and adequacy with simplicity. The empirical relationship models are limited in applicability but, on the other hand, durability design with probabilistic-based performance models are not yet readily applicable due to the difficulty in accurately determining the variability and statistical distribution of some parameters. The application of the models to practice has been explored by Parrott (1994), who considered in particular the diffusion aspect, and in the fib Model Codes, building on formulae proposed by the CEB, which introduced consideration of the chemical buffering effect of the binder. Parrott (1994) refined the basic square root formula to take account of the deviation from the square root relationship as relative humidity increases. The diffusion coefficient was modelled by the air permeability of the cover concrete. The amount of alkaline material was represented by a factor related to the calcium oxide. The relationship may be stated as follows, using the notation of Parrott: d=

a k0.4 tin c0.5

where d = depth of carbonation at the end of the initiation phase (mm) a = coefficient k = air permeability of the cover concrete (units of 10–16 m2) ti = duration of initiation period to start of corrosion c = calcium oxide content in hydrated cement matrix that can react with CO2 (kg/m3 of the cement matrix) n = power exponent. Value close to 0.5 for indoor exposure and decreases with increasing relative humidity The carbonation rate in the internal environment is therefore modelled by Parrott’s equation as a t relationship, which accords with observations under controlled conditions. The rate of carbonation is lower in exposed environments and departs from the t relationship. Hence, Parrott suggests

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the introduction of a power exponent n, which is close to 0.5 for indoor exposure but decreases with increasing relative humidity above 70%. Additionally, Parrott takes account of the variation in air permeability value (k) through a factor related to relative humidity. The use of a parameter related to calcium oxide content demonstrates that the relationship could be used in the design of durable concrete using parameters that are more readily determined than the more theoretical ones that form part of the classic carbonation formulae. The application of Parrott’s formula in practice requires evaluation of specific parameters, some of which may require calibration with extensive field experience. The depth of carbonation (d) at the end of the initiation phase would in practice be taken as the minimum depth of cover. The coefficient (a) is used to calibrate the equation with observations from structures in service. Parrott has assigned a value of 64 for the coefficient, based on an extensive literature review. The air permeability is dependent on the relative humidity but may be estimated from tests on a specimen dried at 60% relative humidity by the following relationship: k = m k60

where m = coefficient k60 = permeability of specimen dried at 60% relative humidity The value of m varies from 1.00 at 40% cent relative humidity to one hundredth of this value at saturation level as illustrated in Fig. 6.7. The calcium oxide content depends on cement composition, exposure condition, and proportion of cement reacted. It represents the alkaline buffer. For example, Parrott quotes values for a CEM I cement from 460 kg/m3 at

Figure 6.7 Decline in coefficient “m” in Parrott’s (1994) carbonation formula as relative humidity increases above test value.

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167

relative humidity levels in the range 40 to 70% rising to 610 kg/m3 at saturation level. The model initially explored by the CEB, leading into the later DuraCrete project (COWI 2000), was based on the following form of the carbonation relationship: xc =

2 k1 k2 Deff Cs a

t

t0 t

n

where xc = depth of carbonation (m) at time t (s) k1 = constant related to execution k2 = constant related to exposure condition Deff = effective diffusion coefficient (m2/s) Cs = concentration of CO2 a = chemical buffering capacity t0 = age at which Deff is determined n = constant related to the environment This was then woven into the fib Model Code for Service Life Design (Fédération Internationale du Béton 2006) limit state approach in an “inverse carbonation” form that was tied into values of carbonation resistance obtainable from natural or accelerated tests: xc =

1 2 ke kc (kt . RNAC + t). CS .

t . W (t)

where xc = depth of carbonation at time t ke = environmental function which accounts for prevailing relative humidity kc = execution function which accounts for influence of curing kt = parameter taking account of model uncertainty R−1NAC = inverse carbonation resistance from natural or accelerated tests ([mm2/year]/[kgCO2/m3]) εt = error term taking account of uncertainty ([mm2/year]/[kgCO2/m3]) Cs = concentration of CO2 (kgCO2/m3) t = target service life (years) W(t) = weather function factor related to probability of driving rain The weather function factor accounts for the reduction in carbonation rate during periods when concrete is wet, where a value equal to 1 results in no reduction. By way of example, Harrison (2019) determined W(t) values of 0.79

168

Fundamentals of Durable Reinforced Concrete

and 0.91 for a 50-year and a 100-year design life, respectively, following an analysis of 30 years of weather data across the regions of the United Kingdom. These examples show, on the one hand, the development of simple tools for informed decisions on specification that are more robust than the deemed-to-satisfy route and, on the other hand, the complex framework already in place for full probabilistic limit state durability design tools when the datasets are numerous enough to characterise variability in selected parameters. Once there is refinement of the input parameters and calibration with observations in practice, it will be a small step to create the software tools that will allow the safe and accessible introduction of probabilistic design for carbonation resistance in practice.

TEST METHODS TO DETECT THE CARBONATION FRONT There are many different methods for locating the carbonation front. Standard tests employ acid/base indicators. These are adequate for practical purposes in identifying the boundary between carbonated and uncarbonated layers. Some evidence of calcium carbonate may be found in advance of the front indicated by an indicator. Although this is not very significant for practical purposes, it has led researchers to employ a wide range of alternative techniques. These techniques tend to be dependent on specialist laboratory equipment, which is why they have not been included in standards documents for routine comparative testing or on-site testing. A useful summary of specialist techniques used in carbonation studies may be found in Revert et al. (2016). Techniques referenced include thermogravimetric analysis (TGA); X-ray diffraction (XRD); Fourier transform infrared spectroscopy; optical microscopy with crossed polarised light; scanning electron microscopy with energy dispersive spectrometry; neutron diffraction, gammadensimetry; and chemical analysis to detect calcium carbonate. Magic-angle nuclear magnetic resonance spectroscopy can detect changes in calcium silicate hydrates caused by carbonation through use of the 29Si isotope. Mercury intrusion porosimetry imagery has been used to investigate changes in porosity due to carbonation. One note of caution: Theophilus and Bailey (1984) report that measurements of ground material can give rise to false readings if unhydrated cement in carbonated zones is hydrated, releasing Ca(OH)2. The most commonly used acid/base indicator is phenolphthalein (C20H14O4), which changes colour gradually in the range of pH 8.3–9.2. This indicator solution method is the basis of almost all carbonation studies, especially due to its usefulness in field tests as it requires little skill in use and gives reproducible results. The indicator is colourless at low pH and changes colour quite strongly to purple at a pH of approximately 9. The test is carried out on freshly broken surfaces brushed free of dust and

Carbonation

169

Figure 6.8 Forms of carbonation profile encountered in practice.

sprayed with the indicator solution. Readings on structures in service are best carried out on cores. Alternatively, on-site testing may be carried out by drilling a 20 mm diameter hole and exposing the edges of the hole with a hammer and chisel. Phenolphthalein may then be sprayed onto the freshly broken surface. The smooth drilled surface is not amenable to testing. Concrete that is difficult to expose may be examined by drilling closely spaced holes and breaking out the concrete in between. The depth of carbonation is determined as the average distance from the surface of the concrete element to the zone where phenolphthalein indicator solution changes colour to purple, indicating that CO2 has not reduced the alkalinity of the hydrated cement in that zone. The depassivation threat from carbonation is adequately assessed by locating the front with phenolphthalein even if the indicator does not mark the border between calcium hydroxide-dominated and calcium carbonate-dominated zones with absolute precision. The carbonation front does not always advance at a constant rate, due to inhomogeneities in the concrete. Thus, it may be necessary to record both the average and the maximum depth of carbonation (Fig. 6.8). The average depth gives an indication of the quality of the concrete and the influence of the local environment. The maximum depth is important from the point of view of durability being threatened if a sufficient number of peaks on the front reach the reinforcement. Dense aggregates are not coloured by phenolphthalein and, for calculation of average carbonation depth, it is assumed that the front is a line connecting the carbonation depth on either side of the particle. Precise guidance on calculating an average value where air voids or porous aggregate leads to the front advancing locally by more than 4 mm is set out in the European standard for relative carbonation resistance CEN 12390–10 (CEN 2018b). In respect of safety precautions, if the carcinogenic warning in respect of phenolphthalein is flagged as a concern, the use of thymolphthalein (C20H14O4), with colour change in the range of pH 9.3–10.5, will also

170

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Figure 6.9 Influence on carbonation profile of biaxial penetration of carbon dioxide.

adequately identify the boundary between carbonated and uncarbonated layers for practical purposes. Caution is required when using the phenolphthalein test on-site if breaking corner pieces off columns and beams of concrete. The corners will experience biaxial penetration of CO2, as illustrated in Fig. 6.9, and the permeability of concrete placed in the corners of shutters may not be representative of the member due to compaction difficulties. The effect of biaxial penetration has been estimated by Hunkeler and Lammar (2012) to increase average depth by a factor dependent on the depth of carbonation along the side faces. Despite a lot of scatter (R2 of 0.36) they determined an interaction coefficient ranging from 1.1 at depths of over 40 mm up to 1.4 at depths of less than 10 mm. The value at 20 mm was 1.2.

STANDARD TEST METHODS FOR CARBONATION RESISTANCE Formulation of a representative standard test method for carbonation resistance is hampered by three practicalities. The first is that carbonation is a slow process. The market needs a test that specifiers, producers, and clients can jointly use in the context of performance-based specifications, but such a test must produce definitive results in a matter of weeks, rather than years. The second is that acceleration of the test through elevating CO2 concentration levels is limited by the need to keep the level low enough that physicochemical conditions remain similar to atmospheric conditions in service. If CO2 levels are too high, the rate of carbonation will reduce due to the pore blocking effect of reactions with the calcium silicate hydrates and increase in the moisture level of the permeable network by water released in the carbonation reaction. The third point relates to the

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171

applicability of results if laboratory environmental test conditions (relative humidity and temperature) differ significantly from prevailing regional weather conditions in place of use. This is especially significant for slowly reacting low carbon concretes that rely on prolonged moist curing. Because testing for carbonation is by necessity a lengthy process, it cannot be used as a routine control test in the same way as a 28-day compressive strength test. Nevertheless, great progress has been made in developing tests in natural and accelerated conditions that are leading practice into adoption of performance-based specifications. The seeds for this progress came through suggestions, for example, from the Concrete Society (1996), on how such tests could be used as a type approval test. This would involve tests on concrete samples to establish that the performance of a particular mix is satisfactory. Additionally, tests could also be used to establish the relative performance of a mix with one of known performance. In the 1990s, Harrison (1996) noted that such a type approval test could establish the relationship between concrete strength and depth of carbonation so that strength could be used as the control measure for a given production unit. Interestingly, his vision is now being realised in the 2020s through the “exposure resistance class” concept, his final project, which he worked on with exemplary dedication up to 2022, the year of his passing from this life. Extensive research on the repeatability and reproducibility of carbonation testing through RILEM Committee CPC-18 (1988) eventually led to the adoption of two standard tests in European practice. These are a test for carbonation resistance at atmospheric CO2 levels EN 12390–10 (CEN 2018b) and an accelerated test procedure EN 12390–12 (CEN 2020c). In addition, as discussed in Chapter 3, the South African approach to performance-based design employs three durability indices, and one of these, the oxygen permeability test is related to the assessment of carbonation resistance. The test is now a South African standard test, SANS 3001-CO3-2 (SABS 2022).

Carbonation resistance at atmospheric CO 2 levels The European reference test for carbonation resistance at atmospheric CO2 levels, EN 12390–10 (CEN 2018b), allows calculation of the carbonation rate (kc) from measurements taken over a year of storage. Storage in the natural atmosphere, sheltered from rain, is an option, but the reference method uses a climatically controlled chamber. The reference storage conditions are 400 ppm CO2 level (0.04% by volume), 65% relative humidity, and a temperature of 20oC. Measurements of carbonation depth are made after 3, 6, and 12 months in storage, following an initial period of 28 days for casting and curing. The test is extended to 24 months of storage if the depth of carbonation is less than 5 mm after 12 months. The method of carbonation front

172

Fundamentals of Durable Reinforced Concrete

detection is by acid/base indicator solution. The reference solution is 1% phenolphthalein, with one gramme of phenolphthalein powder per 70 ml of ethanol and 30 ml distilled water, but other solutions are permitted where safety concerns favour an alternative. The result (kc) is expressed in units of millimetres per square root year (mm/√a) through the slope of the regression line in a plot of average measured depth (y-axis) versus square root of time in storage (x-axis). The test may also be used to determine carbonation rate in a natural environment, including a curing regime that simulates strength development typical of site practice. Curing involves storage in sealed polythene bags to bring the specimens to at least half of the 28-day compressive strength that would be achieved under standard EN specimen preparation conditions. The specimens are stored in a natural atmospheric environment sheltered from precipitation through the use of a protective container that is best modelled on the Stevenson Screen used to house meteorological instruments. Thereafter, the procedure is the same as for the reference method.

Carbonation resistance at elevated CO 2 levels The design of tests that accelerate the carbonation process were under discussion for many years in respect of achieving the optimal CO2 level that would allow rapid results without diverging from the natural carbonation reaction that occurs at atmospheric CO2 levels. An early example was Portugal’s E-391 test (LNEC 1993b) using a carbonation chamber typically set at 5% CO2 level, 60% relative humidity, and 23oC, but a lower value of 3% CO2 is now preferred, and a consensus has been reached on a European standard for carbonation resistance at elevated CO2 levels: EN 12390-12 (CEN 2020c). The test allows calculation of the accelerated carbonation rate (KAC) from measurements taken over a 70-day period of storage. The storage conditions are 30,000 ppm CO2 level (3% by volume), 57% relative humidity, and a temperature of 20oC. Measurements of carbonation depth are made at the time of storage and at 7, 28 and 70 days thereafter. Preparation for storage involves a period of 28 days for casting and curing followed by 14 days preconditioning in a laboratory air environment. The method of carbonation front detection is by a slightly weaker solution of phenolphthalein than in the reference test – a 0.8% solution rather than 1%. Thereafter, the procedures, method of calculation of the result, and the expression of results are similar to the reference method EN 12390–10, except for an allowance to be made for any existing carbonation depth at time of initial storage. The atmospheric level test and the accelerated test serve different purposes. As reported by Shaik et al. (2016), there is not a direct correlation between the assessment of carbon depths under natural and accelerated carbonation conditions. The test under natural conditions provides a useful

Carbonation

173

carbonation rate value from which to predict future service. The accelerated test is very useful for assessing, in a reasonable timeframe of about four months, the relative accelerated carbonation rate of different binder types or mix proportions. Furthermore, Hunkeler (2016) has published a relationship between values from the accelerated test (KS) and a coefficient converted to an equivalent at natural atmospheric conditions (KN,ACC). This was achieved through a study of the relative carbonation coefficients (KRel) of concrete determined at elevated and at natural CO2 levels. For example, a multiplier of 2.6 was found from tests at 4% CO2 benchmarked against tests at a natural atmospheric level of 320 ppm using the following relationships: KN , ACC =

365 1

1 [CO2 ]S [CO2 ]N

KRelN KS KRelS

and KRel =

KSN KN

where KN,ACC = carbonation coefficient from the accelerated carbonation test converted to 400 ppm CO2 and corrected with the conversion factor obtained from relative carbonation coefficient values at elevated and natural CO2 levels (mm/year 0.5) [CO2]S = elevated CO2 content for accelerated carbonation (%) [CO2]N = natural CO2 content (%) KRel = relative carbonation coefficient KRelN = relative carbonation coefficient determined for natural CO2 content KRelS = relative carbonation coefficient determined for elevated CO2 content KS = carbonation coefficient under accelerated carbonation condition (mm/day0.5) KSN = carbonation coefficient measured under accelerated conditions and converted to the reference CO2 content of 400 ppm (mm/year0.5) KN = carbonation coefficient measured under natural atmospheric conditions (mm/year0.5) The equation includes a conversion factor from days to years because of the difference between the units for expressing KN,ACC and the units of measurement employed in the accelerated carbonation test. Irrespective of which test is used, Harrison (2019) argues that the somewhat low relative humidity used in standard climatic chambers disadvantages the slowly reacting cements of less than 55% Portland cement clinker.

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Fundamentals of Durable Reinforced Concrete

Concrete durability testing index – oxygen permeability test The oxygen permeability index (OPI) forms one of three durability indices of South Africa’s performance-based approach to durability (Ballim and Alexander 2018). It is determined in accordance with Standard SANS 3001-CO3-2 (SABS 2022). Although not specifically a test of carbonation resistance per se, the OPI combines with the other two indices to index the quality of the concrete on a performance scale within the context of a materials system in an interaction with a variable exposure environment. The OPI is calculated from the rate of pressure decline as oxygen under pressure flows through a disc specimen, from which a permeability coefficient is determined, leading to the OPI value. Salvoldi et al. (2015) demonstrated a strong correlation between the permeability coefficient from the test with the carbonation diffusion coefficient. Specimens are prepared in accordance with Standard SANS 3001-CO3-1 (SABS 2015). Concrete discs 30 mm thick are prepared by slicing 70 mm diameter cores recovered from cubes or a concrete element on-site. The surface 5 mm sections of the core are ignored in selecting segments for the test discs. Each disc is pre-conditioned by oven-drying at 50 ± 2oC for 7 to 8 days, followed by cooling in a desiccator to 23 ± 2oC for 2 to 4 hours. The disc is then enclosed in a rubber collar and rigid sleeve which permits only unidirectional flow of oxygen into and out of the parallel faces of the disc. One face is exposed to standard grade oxygen (99.8% purity) under a pressure of 100 ± 5 kPa and the test is initialised five minutes later. The rate of falloff in pressure to 50 ± 2.5 kPa is monitored. Highly impermeable concrete will exhibit such a slow decline in pressure that the minimum pressure may not be reached in a reasonable timeframe, in which case the test is terminated after six hours. The OPI is calculated as: OPI =

log10k

and k=

Vgdz RAT

where k = coefficient of permeability (m/s) ω = molecular mass of oxygen (kg/mol) V = volume of the permeability cell (m3) g = gravitational acceleration (m/s2) d = average specimen thickness (m)

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z = slope of linear regression line of ln(P0/Pt) against t through the 0,0 point (s−1) R = universal gas constant (Nm/kmol) A = cross-sectional area of specimen (m2) T = absolute temperature (K) with

() ln ( ) t

ln z=

P0 Pt

2

P0 Pt

and P0 = pressure at initial reading, time t = t0 (kPa) Pt = pressure at time t (kPa) t = time from initial reading at time t0 (s) MANAGEMENT OF DURABILITY IN THE CONTEXT OF CARBONATION The threat of durability failure due to carbonation is currently managed in one of three ways. The first is the prescriptive deemed-to-satisfy approach. The second is a performance-based approach. The third is the exposure resistance class approach, which embodies the simplicity of prescription underpinned by a performance-related methodology. These approaches are based on delaying the onset of corrosion through a prolonged initiation period that forms a substantial proportion of the design life. This conservative approach is questioned by Angst et al. (2020), who fear that innovative low carbon concretes might be disadvantaged by the emphasis on the initiation phase if the propagation of damaging corrosion products in carbonated concrete is not significant. Pending further research, the current approaches will find favour with specifiers requiring legal cover from codes and standards, or substantial case histories of successful innovation.

Prescriptive deemed-to-satisfy approach The prescriptive deemed-to-satisfy approach involves a material resistance typically based on minimum cement content and maximum water/cement ratio combined with a minimum depth of cover to reinforcement. A minimum compressive strength of concrete may also be prescribed. The requirements are typically benchmarked against advice in codes and standards for a 50-year design life. Trade-offs can then be made of increased

176

Fundamentals of Durable Reinforced Concrete

strength for decreased cover for a 50-year design life. In the case of a shorter or longer design life, the strength and/or cover may be reduced or increased against the advice for the 50-year benchmark. Four distinct exposure classes have been internationally identified in respect of corrosion induced by carbonation. In the wider context of concrete concerns it might be argued that differentiating exposure classes for separate carbonation environments is an unnecessary luxury if all may be trumped by other durability threats. Nevertheless, four classes have been identified as meriting slightly different management actions, and specifiers must consult the relevant advice on limiting values in national annexes or complementary standards valid in the place of use. The least onerous class covers reinforced and prestressed concrete that is either dry or permanently wet. The risk of corrosion, in either case, is low, but for differing reasons. In the case of dry concrete it is likely that carbonation rates will be high, but that corrosion will not occur due to the low humidity. In the case of permanently wet concrete, CO2 will not have ready access through the water-filled pores, and so carbonation rates will be low. Examples of dry concrete in this context include the surfaces of members inside buildings or structures where the relative humidity will remain low at all times during the service life. Processes or activities that lead to periods of high humidity within the building would require consideration of an alternative class. Examples of permanently wet concrete obviously cover the case of the surface of members that will be submerged at all times during the service life. Any likely change of use of the building or structure during its life would also need to be considered if this could alter the dry or permanently wet condition. Given the low risk to durability posed by carbonation in these conditions, the limiting values on concrete composition need not be onerous. Hobbs et al. (1998) demonstrated that, due to the low risk of corrosion, the requirement for a 100-year intended working life should not demand a change to the requirements for a 50-year intended working life. The next class covers reinforced and prestressed concrete that is wet, rarely dry. The risk of corrosion in such concrete is relatively low. Carbonation may occur during the drying periods but, since these will be rare, the overall rate of carbonation will be very low. Examples of wet, rarely dry concrete could include reservoirs or water towers if their service conditions kept them full most of the time, and members below ground level, including foundations. Typically, an increase in cement content of 20 kg/m3 and a decrease in the maximum water/cement ratio by 0.05 is recommended over the minimum values for carbonation classes. The third class, in ascending order of durability threat severity, covers reinforced and prestressed concrete that is exposed to moderate humidity. The risk of corrosion in such an environment is significant if the depth of carbonation can reach or exceed the depth of cover during the service life.

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Examples of such situations are concrete surfaces in the external environment that are sheltered from rain, and surfaces in the internal environment where humidity is higher than normal. The latter condition can arise, for example, in buildings housing processes related to the brewing industry, or in commercial laundries. Typically, this might require an increase in cement content of the order of 20 kg/m3 and a decrease in the maximum water/cement ratio by 0.10 in comparison with minimum standards for carbonation resistance. The study by Hobbs et al. (1998) demonstrated that, in the UK, a significant reduction in maximum water/cement ratio and increase in minimum strength class would be required to extend the intended working life from 50 to 100 years. The values were approximately 0.15 and 15 MPa, respectively. The most severe class for carbonation covers reinforced and prestressed concrete that is cyclically wet and dry. The severity of this category varies across regions, depending on the local climate. In certain regions, the drying cycles would allow carbonation to proceed at a reasonable rate, while the wetting cycles would promote rapid corrosion once depassivation has occurred. In other regions, the climate would generally be moist enough to reduce the drying effect and thereby limit the window of opportunity for carbonation. Examples of concrete covered by this class include surfaces in the external environment that are exposed to rain, and external or internal surfaces exposed to contact with water or high humidity on a cyclical basis through natural or industrial processes.

Performance-based approach The performance-based approach can harness the accelerated test to demonstrate the carbonation performance of a concrete mix against a required target resistance. For example, the condition to be satisfied for durability has been stated by Hunkeler (2016) as: dkmax

Kmax t 0.4

where dkmax = maximum allowable carbonation depth (mm) Kmax = maximum allowable carbonation coefficient (mm/year0.4) t = service life (years) Having specified the maximum allowable carbonation depth, for example, as 80% of the minimum cover as in Swiss standard SIA 262/1 (SNV, Swiss Association for Standardisation 2013), the target maximum value of the coefficient Kmax may be determined. Kmax = rk KSN

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where Kmax = carbonation coefficient from the accelerated carbonation test converted to 400 ppm CO2 and corrected with the conversion factor obtained from relative carbonation coefficient values at elevated and natural CO2 levels (mm/year0.5) rk = factor considering the influence of relative humidity, for example, 0.8 KSN = carbonation coefficient measured under accelerated conditions and converted to the reference CO2 content of 400 ppm (mm/year0.5) The inverse carbonation resistance has also been used to specify the required performance. In this approach, the limit state condition to be satisfied has been stated as follows, allowing this equation to be solved for the material resistance (R−1NAC) once the minimum cover to be used is decided upon: c min

r=

1 2 ke kc (kt . RNAC + t). CS .

t . W (t )

where cmin = minimum cover to reinforcement (mm) r = repeatability of the EN 12390–10 test (1.68 mm) 1 Denis (2022), for example reported that criteria were set at RACC −11 −11 2 (accelerated test) equal to 8 × 10 and 10 × 10 in units of [m /s]/ [kgCO2/m3] across a suite of concrete mixes employed in a section of London’s Thames Tideway Tunnel, the UK water industry’s biggest infrastructure project. Two different batches of each mix were tested in pre-construction trials, and hundreds of durability assessment trials were made during the construction phase.

Exposure resistance class (ERC) approach The exposure resistance class (ERC) concept, outlined in Chapter 3, is regarded as ideally suited for piloting in the case of the carbonation durability threat. There are suitable models from which to derive carbonation rates, and accepted test methods for determining carbonation resistance. This allows the required level of performance to be calculated in the form of the maximum acceptable rate of carbonation (mm/year0.5) for the environmental exposure and required service life. Based on the rate, the specifier may then select an ERC and the corresponding level of cover to reinforcement – the latter in respect of advice valid in the place of use. For example, a low level of carbonation, such as 0.4 mm/year0.5 for a structural element in a dry environment, would be satisfied by a class of XRC 0.5 combined with a minimal level of cover. At the other end of the

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scale, an anticipated high level of carbonation, such as 8.1 mm/year0.5 for an element in a cyclically wet and dry environment, would demand a class of XRC 7 and a very high level of cover. The producer would then supply a concrete mix that can be demonstrated to meet the requirements through standard test methods. The XRC class designation is based on the characteristic value of carbonation depth at the 90% fractile, assumed to be obtained after 50 years of exposure to a carbon dioxide concentration of 400 ppm in a constant environment of 20oC and 65% relative humidity, thereby aligning with the one-year reference test for carbonation resistance.

SUMMARY Reinforced concrete may deteriorate through corrosion of reinforcement induced by carbonation. Unreinforced concrete is not at risk because changes caused to the material by carbonation are not detrimental. Carbonation rate is significantly influenced by the moisture state of the permeable pore structure. In conditions of high humidity, the rate of carbonation is slow. Equally, in conditions of low humidity corrosion is unlikely even if the carbonation rate is high. Carbonation-induced damage, therefore, is not a significant problem in temperate climates. The quality of concrete required from the viewpoint of strength will generally assure resistance to carbonation. The carbonation depth often reaches a limit at very low values or may proceed at a rate that is slow enough to prevent the depassivation of reinforcement within the design life of the structure. Serviceability problems that do arise will generally be due to incomplete construction processes; for example, poor curing or inadequate cover due to errors in specification or construction. Consideration of the simplified square root relationship between depth of carbonation and time emphasises the importance of achieving the specified cover. A reduction of about one quarter in the cover could lead to a halving of the time to depassivation of the reinforcement. Rates of carbonation inside buildings will be significantly higher than outdoors and the carbonation front may proceed through the full depth of cover within a decade or two. However, corrosion may not ensue if the relative humidity indoors remains low. Caution is required, however, where the environment promotes significant wetting and drying cycles which promote cycles of high carbonation rate followed by conditions conducive to corrosion propagation. The maximum rate of carbonation may be up to double the average rate of carbonation due to local inhomogeneities. Therefore, isolated spalling may provide early warning of the potential for full depassivation of the reinforcement within the service life of the structure. Exceptional examples of a significant serviceability failure may be associated with unique industrial environments leading to CO2 levels above

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atmospheric norms and pessimum conditions of humidity that conspire to depassivate the reinforcement in conditions that are conducive to corrosion. Research on mathematical modelling of carbonation is well-advanced and there is international convergence on accepted models. Standard test methods have been adopted for determining the rate of carbonation in both reference atmospheric conditions and in accelerated tests. This is helpful in supporting specifiers in moving from a prescriptive mindset to durability design and performance-based specification, not least through the convenience of the ERC approach. The support of the research community is still required to build up significant databases on the variability of parameters in the models, which would then allow a full probabilistic approach to be adopted – the key to sustainable concrete. Serviceability failure due to carbonation-induced corrosion is rare in most climates, due to the mutual exclusivity of high carbonation rates and high corrosion rates, other than in cyclical wet and dry environments. Therefore, research is needed in order to shift the focus from initiation to propagation of corrosion in carbonation environments. This might allow greater innovation in low-carbon concretes that have the potential to provide durable concrete over a typical service life despite high carbonation rates, if the corrosion damage is internal and benign in the overall context.

REFERENCES Alekseev, S. and N. Rozental. 1976. The Rate of Concrete Carbonation. In RILEM International Symposium on the Carbonation of Concrete. Fulmer Grange. 5-6 April. Slough: Cement and Concrete Association. Alexandre, J. 1976. Vitesse de Carbonatation. In RILEM International Symposium on the Carbonation of Concrete. Fulmer Grange. 5-6 April. Slough: Cement and Concrete Association. Andersson, R., Fridh, K., Stripple, H. and M. Häglund 2013. Calculating CO2 uptake for existing concrete structures during and after service-life. Environmental Science and Technology 47, 20: 11625–11633. Angst, U., Moro, F., Geiker, M., Kessler, S., Beushausen, H., Andrade, C., Lahdensivu, J., Köliö, A., Imamoto, K., von Greve-Dierfeld, S. and M. Serdar. 2020. Corrosion of steel in carbonated concrete: mechanisms. practical experience. and research priorities – a critical review by RILEM TC 281-CCC. RILEM Technical Letters 5: 85–100. Aschan, N. 1963. Termogravimetrisk undersokning av karbonatiserings -fenomenet i betong. Nordisk Betong 7, 3: 275–284. Bakker, R. 1994. Prediction of service life of reinforcement in concrete under different climatic conditions at given cover. In Corrosion and Protection of Steel in Concrete, ed. R. Swamy. Sheffield: Sheffield Academic Press. Ballim, Y. and M. Alexander. 2018. Guiding principles in developing the South African approach to durability index testing of concrete. In Proceedings, Sixth

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international Conference on Durability of Concrete Structures, ICDCS2018 ed. P.A.M. Basheer, 36–45. Caithness: Whittles Publishing. Bastidas-Arteaga, E., Rianna, G., Gervasio, H. and M. Nogal. 2022. Multi-region lifetime assessment of reinforced concrete structures subjected to carbonation and climate change. Structures 45, 11: 886–899. Berger, R. and W. Klemm. 1972. Accelerated curing of cementitious systems by carbon dioxide (Part 2). Cement and Concrete Research 2, 6: 647–652. CEN. 2018b. EN 12390-10. Testing hardened concrete - Part 10: Determination of the carbonation resistance of concrete at atmospheric levels of carbon dioxide. Brussels: Comité Européen de Normalisation. CEN. 2020c. EN 12390-12. Testing hardened concrete - Part 12: Determination of the carbonation resistance of concrete - accelerated carbonation method. Brussels: Comité Européen de Normalisation. Cole. W. and B. Kroone. 1960. Carbon dioxide in hydrated Portland cement. American Concrete Institute Journal 31, 12: 1275–1295. Commission Carbonatation. 1972. Compte rendu de la premiere reunion. Paris, 29 Mars 1971. Materials and Structures 5, 25: 53–62. Concrete Society. 1996. Developments in durability design and performance-based specification of concrete. Special Publication CS109. London: Concrete Society. COWI. 2000. General guidelines for durability design and redesign: DuraCrete, probabilistic performance based durability design of concrete structures. Gouda: CUR Civieltechnisch Centrum Uitvoering Research en Regelgeving. Daimon, M., Akiba, T. and R. Kondo. 1971. Through pore size distribution and kinetics of the carbonation reaction of Portland cement mortars. Journal of the American Ceramic Society 54, 9: 423–428. Denis, P-E. 2022. Concrete mixes on the eastern section of the Tideway project. The Institute of Concrete Technology Yearbook 26: 14–26. Engelfried, R. 1977. Carbonatisation von beton, ihre bedeutung und ihre beeinflussung durch beschichtungen. Defazet 31, 9: 353–359. Fédération Internationale du Béton. 2006. fib Model code for service life design. fib Bulletin 34. Lausanne: Fédération Internationale du Béton. Fitzpatrick, D., Richardson, M. and É. Nolan. 2015. Sequestration of carbon dioxide by concrete infrastructure: a preliminary investigation in Ireland. Journal of Sustainable Architecture and Civil Engineering 1, 10: 66–77. Frey, R. 1993. Untersuchung der zwanzigjahrigen carbonatisierung von betonen. Beton 43, 3: 116–120. Gaze, R. and R. Robertson. 1956. Some observations on calcium silicate hydrate (1) tobermorite. Magazine of Concrete Research 8, 22: 7–12. Hamada, M. 1969. Carbonation of Concrete. In Proceedings, Fifth International Symposium on the Chemistry of Cement Volume 3, 343–384. Tokyo: The Cement Association of Japan. Harrison, T. 1996. The specification of durability by performance - when? In Concrete in the service of mankind: Radical concrete technology. ed. R. Dhir and P. Hewlett, 413–425. London: E & FN Spon. Harrison, T. 2019. Specifying resistance against carbonation-induced corrosion by performance. Magazine of Concrete Research 71, 7: 341–348. Hayword, D. 1997. Weather eye on revision. New Civil Engineer 13 February 1997. 26–27.

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Hobbs, D., Marsh. B., Matthews, J. and S. Petit. 1998. Minimum requirements for concrete to resist carbonation-induced corrosion of reinforcement. In Minimum Requirements for Durable Concrete, ed. D. Hobbs, 11–42. Crowthorne: British Cement Association. Huhta, R. 1960. Carbonation of concrete block: a look that two plants. Concrete Products. 63, 1: 34–35. Hunkeler, F. 2016. Swiss requirements for the carbonation resistance of concrete for the exposure classes XC3. XC4 and XD1. In Proceedings, 9th International Concrete Conference, Environment, Efficiency and Economic Challenges for Concrete, ed.M. Jones, M. Newlands, J. Halliday, L. Csetenyi, L. Zheng, M. McCarthy and T. Dyer, 890–903. Dundee: University of Dundee. Hunkeler, F. and L. Lammar. 2012. Anforderungen an den karbonatisierungswiderstand von betonen. Research report VSS 649. Zurich: Swiss Association of Road and Transportation Experts. Kishitani, K. 1960. Consideration on durability of reinforced concrete. Transactions of the Architectural Institute of Japan 65: 9–16. Kishitani, K. 1964. Uber die Bestandigkeit von Stahlbeton unter dem Einfluss von CO2. Zement-Kalk-Gips 4: 158–159. Kondo, R., Daimon, M. and T. Akiba. 1969. Mechanisms and kinetics on carbonation of hardened cement. In Proceedings, Fifth International Symposium on the Chemistry of Cement, Volume 3, 402-109. Tokyo: The Cement Association of Japan. LNEC. 1993b. LNEC-E391, Specification Concrete. Determination of carbonation resistance. Lisbon: Laboratório Nacional de Engenharia Civil. Lv, Y., Niu, D., Liu, X. and Y-C. Li. 2022. Corrosion damage and life prediction of concrete structure in a 41-year-old steelworks. Materials 15, 17. doi: 10.3390/ ma15175893 Martin, H., Rauen, A. and P. Schiessl. 1975 Karbonatisierung von beton aus verschiedenen zementen. Betonwerk + Fertigteil-Technik 41, 12: 588–590. Meyer, A., Wierig, H. and K. Husmann. 1967. Karbonatisierung von schwerbeton. Deutscher Ausschuss fur Stahlbeton 182: 1–33. Mizzi, B., Wing, Y. and R. Borg. 2018. Effects of climate change on structures: an analysis of carbonation-induced corrosion in reinforced concrete structures in Malta. fib Conference: Sustainable Concrete. IOP Conference Series 442. doi: 10.1088/1757-899X/442/1/012023 Moss, R., Edmonds, J., Hibbard, K., Manning, M., Rose, S., van Vuuren, D., Carter, T., Emori, S., Kainuma, M., Kram T., Meehl, G., Mitchell, J., Nakicenovic, N., Riahi, K., Smith, S., Stouffer, R., Thomson, A., Weyant, J. and T. Wilbanks. 2010. The next generation of scenarios for climate change research and assessment. Nature 463, 7282: 747–756. Papadakis, V., Vayenas, C. and M. Fardis. 1989. A reaction engineering approach to the problem of concrete carbonation. American Institute of Chemical Engineers Journal 35, 10: 1639–1650. Papadakis, V., Fardis, M. and C. Vayenas. 1990. Fundamental concrete carbonation model and application to durability of reinforced concrete. In Durability of Building Materials and Components. Proceedings of the Fifth International Conference, ed.J. Baker, H. Davies, A. Majumdar and P. Nixon, 27–38. London: Spon Press.

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Papadakis, V., Fardis, M. and C. Vayenas. 1992. Hydration and carbonation of pozzalanic cements. American Concrete Institute Journal 89, 2: 119–130. Parrott, L. 1994. Design for avoiding damage due to carbonation-induced corrosion. In Durability of Concrete. Proceedings of the Third International Conference, ed.V. Malhotra, 283–298. Special Publication SP-145. Farmington Hills: American Concrete Institute. Passow, H. 1896. Ueber die einwirkung von kohlensäure auf cementmörtel. Protocol der Verhandlungen des Vereins Deutscher Portland-CementFabrikanten 19: 128–139. Peng, L. and M. Stewart 2016. Climate change and corrosion damage risks for reinforced concrete infrastructure in China. Structural Infrastructure Engineering 12, 4: 499–516. Revert, A., De Weerdt, K., Hornbostel, K. and M. Geiker. 2016. Characterization of mortar with Portland cement and flyash, comparison of techniques. Nordic Concrete Research 54: 60–76. Richardson, M. 1988. Carbonation of Reinforced Concrete: its causes and management. Dublin: CITIS Ltd. 118–121. RILEM Committee CPC-18. 1988. Measurement of hardened concrete carbonation depth. Materials and Structures 21, 6: 453–455. Roy, S., Beng, P. and D. Northwood. 1996. The carbonation of concrete structures in the tropical environment of Singapore and a comparison with published data for temperate climates. Magazine of Concrete Research 48, 177: 293–300. SABS. 2015. SANS 3001-CO3-1, Civil Engineering test methods: Concrete durability testing index – Preparation of test specimens. Pretoria: South African Bureau of Standards. SABS. 2022. SANS 3001-CO3-2, Civil Engineering test methods: Concrete durability testing index – Oxygen permeability test. Pretoria: South African Bureau of Standards. Salvoldi, B., Beushausen, H. and M. Alexander. 2015. Oxygen permeability of concrete and its relation to carbonation. Construction and Building Materials 85: 30–37. Sauman, Z. 1971. Carbonization of porous concrete and its main binding components. Cement and Concrete Research 1, 6: 645–662. Schiessl, P. 1976. Zur frage der zulässigen rißbreite und der erforderlichen betondeckung im stahlbetonbau unter besonderer berucksichtigung der karbonatisierung des betons. Deutscher Ausschuss fur Stahlbeton 255: 1–175. Schiffner, F., Schott, F., Meyer, F., Hauenschild, H., Kosmann, B. and A. Hoch. 1897. Ueber die einwirkung von kohlensäure auf Portlandcement. Protokoll der Verhandlungen des Vereins Deutscher Portland-Cement-Fabrikanten 20: 136–149. Shaik, H., Bhuna, D. and S. Singh. 2016. Assessment of carbon depths under natural and accelerated carbonation conditions. Indian Concrete Journal 90, 11: 57–64. Sickert, G. 1997. Extended modelling for the advance of carbonation in concrete. Betonwerk-Fertigteil-Technik, 12: 74–88. Smolczyk, H. 1969. New aspects about carbonation of concrete. In Proceedings, RILEM International Symposium on Durability of Concrete. Prague. D59–D75.

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SNV, Swiss Association for Standardisation. 2013. SN 505 262/1 – SIA 262/1, Concrete Structures – Supplementary Specifications. Zurich: Schweizerischer Ingenieur- und Architektenverein. Theophilus, J. and M. Bailey 1984. The significance of carbonation tests and chloride level determination in assessing the durability of reinforced concrete. In Proceedings, Third International Conference on the Durability of Building Materials and Components, ed. T. Sneck. and A. Kaarresalo. Vol. 3. VTT Symposium No. 50: 209–238. Espoo: VTT Technical Research Centre of Finland. Tomei, A. 1880. Die absorption von kohlensäure durch Portlandcement bei belüftung. Protokoll der Verhandlungen des Vereins Deutscher PortlandCement-Fabrikanten 3: 14–18. Tsukayama, R., Abe, H. and S. Nagataki. 1980. Long-term experiments on the neutralisation of concrete mixed with fly ash and the corrosion of reinforcement. In Proceedings, Seventh International Symposium on the Chemistry of Cement. 3. IV: 30-35. Tokyo: Cement Association of Japan. Uchida, S. and M. Hamada. 1928. Durability tests of steel and concrete. Journal of Architecture and Building Science 516: 1–18. Woyciechowski, P. and J. Sokolowska. 2017. Self-terminated carbonation model as an useful support for durable concrete structure designing. Structural Engineering and Mechanics 63, 1: 55–64.

Chapter 7

Chloride ingress

CHLORIDE INGRESS AND CORROSION It is universally recognised that corrosion of reinforcement due to chloride penetration is the most significant threat to the existing reinforced concrete infrastructure of developed countries. For durability design purposes, the structures of concern are those located in the maritime environment and on highways subject to de-icing salt application. The marine environment, especially the tidal and splash zones, are recognised as high-corrosion-risk areas, as illustrated in Fig. 7.1. Inland structures near coastal regions can also deteriorate through corrosion induced by windblown salts. Structures subject to wetting and drying cycles created by weather or tidal patterns are particularly vulnerable to high chloride uptake. Concrete permanently submerged in seawater may allow significant chloride penetration, but significant corrosion may not occur due to the low level of oxygen supply. Structures such as tunnels for rail and road traffic may be exposed to saline groundwaters of similar characteristics to seawater. Reinforced concrete highway structures such as bridges, tunnel linings, retaining walls, and median barriers are at risk from either direct application of de-icing salts or chloride-laden spray generated by fast-moving traffic. Evidence of corrosion initiated by de-icing salts in post-tensioned grouted duct bridges caused major concern in the 1980s and 1990s, leading to a four-year moratorium on such techniques by the Highways Agency in the United Kingdom. In highway structures, drainage paths or leaking joints may channel runoff in a manner that exposes localised areas to high concentrations of chloride. Chloride-induced damage has also affected building structures. Car park ramps and decks have deteriorated due to water laden with de-icing salts dripping from cars. Regarding other sources of chloride, building façades have prematurely deteriorated due to corrosion induced by calcium chloride used as an accelerator in precast work. Kropp (1995) notes that chlorides may also be introduced in fire-damaged structures through exposure to thermally DOI: 10.1201/9781003261414-7

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Figure 7.1 Corrosion-induced damage in a tidal zone.

decomposed polyvinyl chloride (PVC) fittings and furnishings. Many buildings contain PVC, which liberates hydrochloric gas at temperatures in excess of 80°C. Specifications based on prescriptive advice in 20th century codes of practice may have underestimated the requirements for durable concrete in chloride environments, leaving a trail of structures at risk. Particular attention has been drawn (Bamforth 1994, Grantham 1999) to the potential inadequacy of Portland cement concrete to provide the required level of durability over a typical service life, and this has generated an interesting debate. The passivity of reinforcement, as discussed in Chapter 5, is dependent on the stability of the passive film formed on it when the steel is immersed in the alkaline environment of fresh concrete. The passive film is rendered ineffective in circumstances where the chloride level in the surrounding concrete exceeds a critical level. Only a low level of chloride can be tolerated in durable concrete. Chlorides can inadvertently be introduced as trace elements in, for example, brackish water, aggregates, and admixtures. Excessive levels occur over the service life when significant external sources of chloride are available in combination with concretes of inadequate chloride resistance. The internal sources of chloride are limited to tolerable levels by specification. Many international codes and standards limit the chloride content to a range of 0.10–0.40% in the case of concrete containing reinforcement. The strictest limitations apply to prestressed concrete. Aggregate standards also limit the chloride content of aggregate

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Figure 7.2 Process of pitting corrosion in a chloride-rich environment.

for use in concrete. The use of seawater, chloride-bearing aggregates, or admixtures such as calcium chloride accelerator, is thus strictly controlled. Chloride-induced corrosion is generally focused on a small area, which forms a pit surrounded by uncorroded reinforcement. The process is illustrated in Fig. 7.2. This can lead to rapid loss of cross-section and critically reduce the load-bearing capacity of the reinforced concrete member. Deterioration could be significantly advanced before damage is apparent at the surface. The mechanism of attack is described more extensively in Chapter 5. During the corrosion cycle, ferrous ions become available to combine with the chloride ions to form compounds such as ferrous chloride (FeCl2). Hydrolysis of these products over time releases chloride ions with a consequent reduction in the anode pH. The corrosion rate increases because the oxidation of the iron is encouraged in such acidic conditions. The recycled chloride ions exacerbate the problem. An influencing parameter is the chloride ion:hydroxyl ion ratio at the reinforcement. It has been found (Tritthart 1989) that the higher the hydroxyl ion concentration the greater the fraction of total chlorides represented by free chlorides. If the chloride ions predominate the loss of ferrous (Fe2+) ions is accelerated. If the hydroxyl ions predominate, precipitation of FeOH+ ions help to repair the reinforcement’s passive oxide film. The reduction in pH discourages precipitation of the corrosion product. The pit at the anode may develop rapidly due to the combination of a localised anode with a comparatively large cathode. An insight into the influence of chloride level on the electrochemistry of the situation may be gained through the departure from a passive condition illustrated in the Evans Diagram presented in Fig. 7.3. The possible mechanism of chloride ion interaction with the reinforcement and passive layer has not been fully resolved. The studies of Foley (1970, 1975) are often quoted, for example, by the American Concrete Institute Committee 222 (1985) and by Sagoe-Crentsil and

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Figure 7.3 Evans diagram showing departure from a passive condition due to the presence of chloride.

Glasser (1989). Three theories have been advanced – adsorption theory, oxide film theory, and transitory complex theory. The adsorption theory postulates that chloride ions are adsorbed onto the surface of the reinforcement in preference to dissolved oxygen and hydroxyl ions. The reaction rate of iron with the chlorides is higher, and soluble complexes are formed. The resultant dissolution promotes formation of a pit. The oxide film theory suggests that the passive oxide layer is more open to penetration by chloride ions than other anions. Defects and pores in the film are thought to allow chloride ions access to the reinforcement leading to pitting corrosion. The transitory complex theory postulates that a soluble complex of iron chlorides forms from the chloride ions and ferrous ions in competition with the ferrous hydroxide reaction. The diffusion of the iron chloride complex away from the anode is thought to destroy the passive film. The complex later breaks down with precipitation of iron hydroxide liberating the chloride ion. The chloride ion is then free to recommence the pernicious cycle. Chlorides may be present in three states: free chloride ions in the pore solution, chlorides strongly bound, and chlorides loosely bound. The free chlorides are the most significant contributors to the corrosion risk. They may be introduced from an external source or carbonation may release bound chlorides from internal sources. The aluminates can combine only with internally or externally introduced chlorides up to a certain

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concentration of chloride, helping to reduce those remaining as dangerous free chlorides. However, the later lowering of pH by carbonation releases some of these bound chlorides ions into the pore solution. The strongly bound chlorides are chemically combined, mainly as calcium chloroaluminates (3CaO.Al2O3.CaCl2.10H2O) and calcium chloroferrites (3CaO.Fe2O3.CaCl2.10H2O) through the calcium aluminate hydrates (for example 3CaO.Al2O3.6H2O). There is a two-stage reaction involving the production of calcium chloride from calcium hydroxide and sodium chloride. The loosely bound chlorides are adsorbed by the pore walls made up from calcium silicate hydrate such as 3CaO.2SiO2.3H2O.

CRITICAL CHLORIDE LEVEL FOR CORROSION The depassivation of reinforcement in concrete by chloride ions is in competition with the repair of the passive oxide film by the hydroxyl ions. This has led to the concept of a critical chloride threshold level for the onset of corrosion. The threshold level of chloride for corrosion is not fully established, because the critical level at which passivity is lost and corrosion commences is difficult to establish. The reasons for this include the variable influences of environment, but measurement also causes difficulty. The nature of the chlorides is significant, but while free chlorides are most relevant, it is the total (free and bound) chloride level that is detected in tests. In addition, the chloride ion to hydroxyl ion ratio is an influencing factor governing corrosion activity, but it is not readily determinable in testing. The method of measurement (acid or water soluble) and expression (mass by weight of cement or concrete) can also make comparisons difficult. Thresholds quoted are therefore generally single point values of total chloride, not specifically the harmful free chlorides. The most common analytical techniques are generally based on acid-soluble chloride. Total acid-soluble chloride level, expressed as a percentage by mass of binder, is traditionally used in specifications. The literature includes comprehensive reviews of experimental studies of the critical chloride threshold level. Pettersson (1992) presented a review based on six published studies. The range was 0.06% to 2.5% by mass of cement. It was noted that different properties of the concrete and its service environment influenced the range. Variables included pH, cement type, curing regime, water/binder ratio and the use of admixtures. Bamforth (1996) published a review derived from over 20 sources including Funahashi’s review of seven studies in 1990. The values for critical chloride threshold levels varied from 0.06% to 2.2% by weight of cement. The wide range was partially accounted for by different methods of test. Glass and Buenfeld (1995) noted a range of published threshold values of 0.17% to 0.7% for field exposure tests and 0.4% to 2.5% for laboratory-based trials.

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They suggest that determination of a unique chloride threshold level applicable to a wide range of structures is not achievable. Extensive durability failures in highway structures subjected to de-icing salt application led American Concrete Institute Committee 222 (ACI 1985) to take a very conservative approach in framing recommendations. The Committee decided that the large amount of conflicting data on the corrosion threshold values, and the difficulty of defining the service environment throughout the life of a structure, necessitated such an approach. It suggested levels of 0.20% for reinforced concrete and 0.08% for prestressed concrete. Glass and Buenfeld (1995) suggest that it may be beneficial to consider the critical chloride threshold solely as an indicator of corrosion risk. It is long accepted (Browne 1980) that values greater than 2.0% chloride by mass of cement represent certain risk, while 1.0–2.0% chloride by mass of cement presents a highly probable risk of corrosion. The results of a study of bridges in the UK showed that a chloride content between 0.35% and 0.5% by weight of cement gives a corrosion risk of below 25%, whereas chloride contents in excess of 1.0% present a corrosion risk of over 70%. It was concluded that without further work, no improvement could be made to the suggested chloride threshold levels of 0.4% for buildings exposed to a temperate European climate and 0.2% for structures exposed to a more aggressive environment. Equally, the Concrete Society (1996) acknowledged that while the threshold level of chloride can lie between 0.17% and 2.5%, it is best to use criteria in the 0.2% to 0.4% range. It is informative to view the source of variation, even if the sensitivity of the influencing parameters is not clear. The factors of influence include: • • • • • • •

the chemistry of the binder, especially the C3A content the proportion of the total chlorides that are free chlorides chloride ion to hydroxyl ion ratio water/binder ratio hydroxyl ion concentration temperature and relative humidity electrical potential of the reinforcement.

The interaction of these factors is such that the critical chloride content cannot be generically determined due to its interdependence on the unique properties of a particular concrete and its service environment. For example, there is some conflict on the influence of pH. However, it has been noted that reducing the water/binder ratio increases both the corrosion threshold and the pH, whereas the addition of silica fume has the opposite effect in each case. Nevertheless, there is currently a large amount of published data on suggested design values for the chloride ion

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content, below which there is an acceptably low risk of corrosion to both normal and prestressed reinforcement. The latter is particularly at risk of sudden failure and is conservatively limited to half of the chloride threshold used for the former. The Concrete Society (2014) recommended existing structures be further investigated for reinforcement corrosion if the measured chloride content in the structure was greater than 0.6% by mass of cement. These figures were worked up into design values for limit state scenarios in durability design. A review of the literature, such as Everett and Treadaway (1980), BRE (1982), Concrete Society (1984), and Marques et al. (2012), reveals a trend of agreement that for Portland cement concrete a 0.4% chloride by mass of cement represents an acceptable benchmark for design purposes, while 0.6% chloride by mass of binder is appropriate for ggbs concretes. For water/cement ratios in excess of 0.4, a limit of 0.3% has been recommended for the harshest exposure class (LNEC 2007). Conservative values in the range of 0.1–0.2% chloride by mass of cement are recommended for prestressed concrete. There is additional concern in the case of concrete made with sulfate-resisting Portland cement. These concretes have a lower level of tricalcium aluminate available to bind chlorides. Also, if sulfate attack occurs, the free chloride ion levels may increase due to the breakdown of calcium chloroaluminates. A lower threshold, typically 0.2% chloride by mass of cement, is generally adopted. These are limits on the total chloride content of new works, not the free-chloride content, and therefore do not represent the corrosion threshold as such, which would be a higher value. Buenfeld (1986) stated that even 0.2% chloride by mass of cement can lead to corrosion if the chlorides are being introduced from an external source and a large proportion remains free in the pore solution. Although a unique threshold level of chloride for corrosion is thus not established, a possible relationship between critical chloride content, relative humidity, and concrete quality has been published by the Comite Euro-International du Beton (1989). This relationship was used as a guide in the design of the Western Scheldt Tunnel in the Netherlands (Breitenbucher et al. 1999) in an early use of mathematical models in a durability design performance-based specification. The trend of the relationship is indicated in Fig. 7.4. The precise elevation of the curve depends on the quality of the concrete and the extent of carbonation. The better the concrete quality the higher the critical chloride level. Carbonation of the concrete lowers the critical chloride level. The range of values covered by the graph in the Comite Euro-International du Beton (1989) guide supports the contention that a critical chloride level of 0.4% chloride by mass of cement for reinforced concrete is a suitable design level for good quality uncarbonated concrete.

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Figure 7.4 Typical critical chloride content for good quality uncarbonated concrete.

PRIMARY FACTORS INFLUENCING CHLORIDE INGRESS The rate of chloride ingress through the cover to reinforcement is dependent on multiple interactive mechanisms. The following material and environmental factors have been highlighted by researchers as being influential: • • • • • • • • • •

chloride diffusivity of the concrete chloride diffusivity of the aggregate sorptivity and wetting-drying cycles ability of concrete to bind chlorides carbonation water/cement ratio degree of exposure to chloride source temperature hydrostatic head (if applicable) stress from external loading.

Chloride diffusivity of the concrete The chloride diffusivity of concrete is clearly a key issue in its ability to remain durable in a chloride-rich environment. The uptake of chlorides at the surface is primarily a function of sorptivity but, once inside the concrete, the transport rate of the chloride ions is governed by diffusion. Diffusion occurs in pores that are totally or partially water filled. The process of chloride ion diffusion is often accompanied by some of the ions reacting with the cement phases and becoming bound. Potential durability is therefore often characterised by either a diffusion coefficient or an effective diffusion coefficient. In steady state conditions, a constant flux of chloride ions pass through a section. In non-steady state conditions, the

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diffusion coefficient varies over time due to chloride binding and age of the concrete. There is an order of magnitude in the difference between steady state and non-steady state coefficients. Bamforth (1994) drew particular attention to the shortcomings of Portland cement concretes in the chloride environment through estimation of typical chloride diffusivity using Fick’s second law. He applied the error function solution by Crank (1975) to chloride concentration profiles generated from data derived through surveys of structures in service. Blended cement concretes performed well, but the effective diffusion coefficients for Portland cement concretes were found to be at a level that would not afford adequate protection over a typical service life when used with practical levels of cover. This assertion was challenged by Spooner (1995) on a number of grounds. While the ability of normal Portland cement concretes to consistently provide durability in chloride environments is a source of debate, the potentially enhanced performance of secondary cementitious materials, reported earlier by, for example, Higgins (1986), is noteworthy. The decrease in chloride ion diffusion with age is more significant in the case of pulverised fuel ash (PFA) and slag concretes than in the case of Portland cement concretes. Fookes (1995) noted that in the severest microclimates, blended cements were necessary, otherwise a cover of over 200 mm would be required! In the case of very high chloride-level exposure conditions, specifiers may also consider the use of supplementary materials such as microsilica or metakaolin.

Chloride diffusivity of the aggregate The rate of chloride ion ingress into concrete is influenced by both the diffusivity of the cement paste fraction and the diffusivity of the aggregate. Hobbs (1999a) drew attention to the chloride diffusivity of aggregate and noted that consideration would need to be given to the aggregate volume in certain circumstances. Aggregates can have chloride diffusivities in excess of or lower than the cement paste fraction, and so the aggregate volume may need to be limited or increased to achieve the desired limit on chloride diffusivity. Controls may need to be considered to ensure that the aggregate selected for a structure in a chloride environment does not have a detrimental impact on the required chloride resistance of the concrete. Hobbs (1999a) states that highly permeable aggregate could increase the chloride ion diffusivity of concrete by a factor of ten.

Sorptivity and wetting-drying cycles The transport of chloride ions into concrete has a capillary suction component in addition to diffusion. Capillary suction of water containing chlorides occurs in air-filled pores near the surface zone. The effect of absorption is significant. Concrete subject to wetting and drying can

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experience a greater uptake of chlorides than concrete subject to diffusion alone. The salt-laden water can be absorbed by the initially dry pores. The pores near the surface may become saturated. A drying cycle will evaporate the water leaving behind chlorides, which can then diffuse inwards under the influence of the concentration difference. The surface pores are then available for further chloride absorption. A reduction in surface absorption can significantly enhance service life by lowering the uptake of chlorides. A technique that may achieve this is the use of controlled permeability formwork (CPF). Such formwork systems allow air and water to escape from the concrete under controlled conditions without the loss of cement or aggregate particles. This improves the quality of the surface and the performance of the outer layer of concrete by significantly reducing the water/cement ratio. Bamforth and Price (1993) reported that chloride ingress reduction through the use of CPF could at least double the service life. Nolan et al. (1995) found that a combination of CPF and silane produced concrete of very low sorptivity. The impetus of research on mixing geotextiles with concrete technology for use in chloride environments has receded since the 1990s despite its effectiveness, but could re-emerge under performance-based specifications if the interest is there.

Ability of concrete to bind chlorides Given that the threat to durability is posed by free chlorides, the greater the binding capacity of a concrete in respect of chlorides, the better it should perform. The ability of concretes to bind chlorides depends on the alkalinity of the binder, its fineness, and the tricalcium aluminate (C3A) and tetracalcium aluminoferrite (C4AF) content. The aluminates are very effective in binding chloride ions introduced at an early stage. They are not as readily available at later ages to bind externally introduced chlorides, because they will have reacted with other ions by then. Typically, the C3A content would be of the order of 7–12% in Portland cement. Sulfate-resisting cements contain less than 3.5% tricalcium aluminate and should therefore be less effective, although this has been questioned by Arya et al. (1990). The alumina content of PFA and slag (ggbs) is considerable, and this conveys benefits in respect of chloride binding. However, for all binders, the influence of C3A content becomes less significant as chloride concentration increases. The better performance of PFA concrete in comparison with Portland cement concrete may also be due to lower chloride diffusivity. It is thought that PFA concretes may have a more extensive capillary network, which allows greater absorption of chlorides, but lower diffusivity ultimately leads to less chloride at the reinforcement. Harrisson (1995) ranked the resistance to chloride penetration of concretes made with different cements as decreasing in the following sequence: Portland cement/ggbs; Portland cement/pfa; Portland cement; sulfate-resisting Portland cement. The beneficial effect of microsilica was also noted. Pettersson (1992) also demonstrated the beneficial effect of

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microsilica, and although the critical chloride concentration decreased in microsilica concretes, the corrosion rate was lower, leading to a net beneficial effect. Resistance to carbonation is also important in this context, because carbonation can lead to the release of bound chlorides.

Carbonation The impact of carbonation depends on the microclimate of greatest impact on a structure. On the one hand, carbonation products may densify the microstructure and reduce the rate of ingress of harmful ions. On the other hand, carbonation releases bound chlorides – Friedel’s salts ─ thus making free chloride ions available to ingress into the uncarbonated zones. This can increase the harmful free chlorides ion concentration in the cover zone and may bring the level at the reinforcement to a critical level (Wang et al. 2017). The chloride ions diffuse into concrete ahead of the slower ingress of the carbonation front, which separates carbonated and yet-to-be carbonated zones. Some of these chloride ions are physically bound by the calcium silicate hydrate gel. Others may react with the cement paste, producing Friedel’s salts through ion exchange with layer-structured hydrated calcium aluminate phases. These later react with the advancing carbon dioxide, releasing chloride into the pore solution as follows (Zhu et al. 2016): 3CaO · Al2 O3 · CaCl2 .10H2 O + 3CO2

3CaCO3 + 2Al (OH)3 + CaCl2 + 7H2 O

The amount of chloride redistribution caused by carbonation was reported to be considerable by Ye et al. (2016). They noted that the effect was pronounced in concretes incorporating supplementary cementitious materials due to the lower levels of Portlandite with which to restrain the rate of carbonation front ingress. In Portland cement concretes, the effect of carbonation was a pore blocking effect on small pores, which reduced chloride diffusivity. The result was a decrease in the total and free chloride contents in the carbonated areas and at the surface, but the creation of convective zones for chloride profiles which increased chloride levels at the carbonation front within the concrete.

Water/cement ratio Given the traditional link between water/cement ratio, permeability, and hence durability, it is unsurprising that chloride diffusion through high water/cement ratio concretes is more rapid than for dense low water/cement ratio concretes. Hobbs and Matthews (1998a) studied the relationship between effective diffusion coefficient (Dce) and water/cement ratio (w/c) using published data from ten studies of CEM I concretes. The coefficients

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were standardised through the Arrhenius equation to allow for temperature variation. The best-fit equation was determined as follows: Dce = 0.04(1166w / c ) × 10

12

Thus, the very significant influence of water/cement ratio is apparent.

Degree of exposure to chloride source Detrimental amounts of chloride may be expected to penetrate into high-quality concrete if the conditions of exposure are such that chloride-laden water is frequently in contact with the structure. Damage in bridges tends to be concentrated in locations where water containing de-icing salt drains or leaks. Pritchard (1986) suggests a range of details designed to minimise the problem. These include deck continuity to avoid joints, accessible deck joints, deck drainage, and drips. Wind speed and direction can lead to the buildup of chloride levels inland from coastal areas. For example, the surface chloride level on structures located even one kilometre from the coast in Portuguese practice (LNEC 2007) is taken as 60% of the value at the coast: CS = 2.5

w kver khor ktemp Cb c

where CS = surface chloride level w/c = water/cement ratio kver = coefficient from 0.7 to 1.4 depending on vertical height relative to sea level and exposure class khor = coefficient from 0.6 to 1.0 depending on horizontal distance from shore ktemp = coefficient from 0.6 (35oC) to 2.2 (0oC) depending on surface temperature Cb = surface chloride referenced to a salinity of 21 g/L at a temperature of 16±2oC, with a percentage of 2–3% depending on exposure class Member geometry may also influence durability. Columns are less durable than walls, and beams are less durable than slabs. Presumably, this is due to the differences between biaxial penetration and, effectively, uniaxial penetration.

Temperature The resistance of concrete to chloride ingress is temperature-related. The diffusion coefficient is influenced by temperature, and can double when average temperatures climb from about 10°C to 20°C. Dhir et al. (1993)

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report that the resistance of Portland cement concrete is dramatically less at 45°C than at 5°C. Higher temperatures result in deeper chloride depths at all relative levels within the concrete, thus maintaining the chloride profile but expediting its progress into the member (Yuan et al. 2008). Seasonal temperature variations may lead to bound chlorides taken up in the winter being released as free chlorides in the summer.

Hydrostatic head Parts of structures subject to a hydrostatic head in a chloride source will suffer a further force, driving chlorides into the cover concrete. Such conditions most commonly occur in sections of structures submerged in seawater.

Stress from external loading The effect of flexural loading on members in chloride environments is to increase chloride levels within the concrete. Ye et al. (2016) drew attention to this aspect in that the formation of micro-cracks allowed deeper penetration of the carbonation front, with its related impact on the chloride profile.

MATHEMATICAL MODELLING OF CHLORIDE INGRESS Chloride ingress from the external environment occurs by diffusion and by capillary suction. In the early stages of exposure, chlorides are transported into concrete by absorption. The absorption effect may reduce with time unless the concrete is subject to wetting and drying. Commonly used mathematical models of chloride ingress are primarily based on chloride diffusion, although attempts have been made to take absorption into account. The following review illustrates the variety of approaches to modelling chloride ingress that are being used as starting points in the development of service life prediction tools and performance-based specifications.

Fickian model with apparent diffusion coefficient Models based on consideration of diffusion alone are constructed around Fick’s second law of diffusion and the error function solution by Crank. Fick’s second law of diffusion concerns the rate of change of concentration with respect to time. Many models are based on the simplified assumption that the diffusion coefficient is constant. For most reinforced concrete surfaces, the diffusion may be assumed to be one dimensional, but Zhao et al. (2022) caution that modelling two dimensional situations has produced significant differences. Thus the dimensional ratio of each element needs to be carefully considered when selecting a model in durability design.

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The 1-D case may be stated as follows for diffusion in a semi-infinite, homogenous medium, where the diffusion coefficient (D) is independent of the dependent and independent variables: C (x , t) =D t

2 C (x ,

t)

x2

with initial and boundary conditions of: C (x , 0) = 0 and 0 < x < C (0, t) = Cs and 0 < t
tg t tL = i + tp

where tL = expected service life by calculation (years) tg = target (intended) service life period, typically 50 or 100 (years) ti = initiation period (years) tp = propagation period; minimum values of 40 and 80 in the case of exposure class XS2 for target service lives of 50 and 100 years respectively (years) γ = safety factor related to reliability class, 2.0, 2.3 or 2.8

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The initiation period is determined as follows, using values that the designer may choose from tables in LNEC E465:

ti =

2 erf dc

1

1

CR CS

Ci Ci

2

1 k D0 t0n

1 1 n

w kvert khor ktemp c kD, RH kD, T

CS = Cb 2.5 k = kD, c

where dc = cover to reinforcement as prescribed in Eurocode; use structural class S4 and S6 values for target service lives of 50 and 100 years respectively (m) erf−1 = inverse of error function CR = critical chloride content, depending on the water/cement ratio and exposure severity; 0.3% for w/c > 0.40 in XS3 to 0.6% for w/c < 0.30 in XS1 (% by mass of concrete) CS = surface chloride content (% by mass of concrete) Ci = initial chloride content in the concrete mix ((% by mass of concrete) Cb = surface chloride content calibrated to local salinity of seawater, 2% for XS1 exposure and 3% for XS2, XS3 exposure (% by mass of concrete) D0 = chloride migration coefficient measured at reference time t0 (m2/s) t0 = reference age of concrete; 28 (days) n = ageing factor depending on cement type and exposure class; 0.45 to 0.65 w/c = water/cement ratio kvert = coefficient related to height relative to sea level and exposure; 0.7 to 1.4 khor = coefficient related to distance from coast; 0.6 to 1.0 ktemp = coefficient related to concrete surface temperature, 0.6 to 2.2 kD,c = influence of curing on diffusion coefficient; 0.75 always wet to 2.4 for normal kD,RH = influence of relative humidity on diffusion coefficient; 0.4 XS1 to 1.0 XS3 kD,T = influence of temperature on diffusion coefficient; 0.4 at 0oC to 1.5 at 30oC The propagation period is determined as follows, again using calibrated values of the corrosion rate depending on the severity of the selected exposure class:

Chloride ingress

tp =

y ø0 1.15 Icorr

y =

74.5 + 7.3

dc ø0

17.4 ftd

225

0.2 ø0

where ø0 = initial diameter of reinforcement bar (mm) α = pitting corrosion factor; < 10 Icorr = corrosion rate, using values calibrated to exposure class (μA/cm2) y = empirical equation for relative reduction of reinforcement radius (%) ftd = concrete tensile strength from “Brazilian” splitting test, EN 12390-6 (MPa) Additionally, there are prescribed minimum values of the propagation period for exposure class XS2. This illustration of the management of the chloride threat, based on a careful blend of fundamental science and empirical research, represents a significant advancement from the prescriptive approach. Notwithstanding this, the process itself is not new because structural designers are very familiar with the responsible input of calibrated values from a design code or manual into design equations, while maintaining a strong sense of control over the design output. What is particularly helpful in the work of Marques et al. (2012) is their exploration of the ease with which the semi-probabilistic safety factor approach can be extended to a full probabilistic approach if values can be characterised as mean values with standard deviations and statistical distributions. The starting point is the same limit state function, but formulated as a performance function. An assessment is made of the probability of the expected service life (tL) being less than the intended service life period life (tg). The maximum acceptable probability may be taken as 6.7%, corresponding with a reliability index (β) of 1.5 as set out for Reliability Class RC2 members in Table C.2 of Eurocode 0, EN 1990 (CEN 2010). Pf = P [g (x) < 0] g (x) = tL tg

The Monte Carlo method is implemented using the equation:

g (x) =

2 erf dc

1

1

where λ = model uncertainty

CR CS

Ci Ci

2

1 k D0 t0n

1 1 n

+

y ø0 1.15 Icorr

tg

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Note that the propagation period is determined by the formula, without possible recourse to minimum values as in the partial safety factor approach. Deterministic values are used for the reference age (t0), curing factor (kD,c), relative humidity factor (kD,RH), diameter (ø0), and pitting corrosion factor (α). A log-normal distribution is used for the cover (dc) and normal distributions for other variables. Over time we may expect increased convergence of opinion on appropriate standard deviations and statistical distributions. In the interim, there are many examples in the literature from which the designer can consider a possible range of applicable values. To conclude with a case study, Denis (2022) reported the example of concrete mixes employed in a section of London’s Thames Tideway Tunnel, the UK water industry’s biggest infrastructure project. Chloride resistance was set at a maximum diffusion coefficient value of 6 × 10−12 m2/s, using the NT Build 492 test. More than 100 durability assessment trials were made during the construction phase, yielding average values for three different mixes in the 2.63–4.03 × 10−12 m2/s range.

Exposure resistance class (ERC) approach The exposure resistance class (ERC) concept, outlined in Chapter 3, is designed to manage the risk of depassivation of reinforcement during a defined service life. These performance-related concretes are matched to the European exposure resistance classes for chlorides from seawater (XS) or chlorides from de-icing salts and sources other than from seawater (XD). The ‘XRDS’ set of concretes, XRDS 0.5 to XRDS10, are based on the characteristic value of chloride penetration depth at the 90% fractile, corresponding to a reference chloride concentration of 0.6% by mass of binder after 50 years of exposure to one-sided penetration of reference seawater (30 g/L NaCl) at 20oC. The designer is guided through Eurocode 2 to specify the relevant XRDS concrete together with a minimum cover to reinforcement. Following specification, the next step is the production of a concrete that meets the test value criteria, using test EN12390-11. The producer may already have determined an acceptable concrete mix for each of the XRDS diffusion coefficient limits (10−13 m2/s).

SUMMARY The problem posed by chloride ion ingress represents the greatest threat to the durability of the world’s concrete infrastructure. The long service life requirement of coastal structures and highway bridges demand a level of chloride diffusion resistance that can only be achieved by concrete of the highest quality. This at least involves the use of low water/cement ratio concretes, but the benefits of secondary cementitious materials have also come to the fore in this context.

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Mathematical modelling of chloride ingress for use in service-life prediction and performance-based specifications has, to date, centred on the diffusion coefficient in Crank’s solution to Fick’s second law of diffusion. Recognising that this is an imperfect approach, others have attempted to produce enhanced models. The complex issues to be addressed include the significance of absorption, the estimation of typical surface chloride level for a given environment, the effect of chloride binding, the influence of seasonal temperatures, the change in diffusion coefficient with time, and the corrosion threshold level for a particular mix. Some of these aspects have yet to be definitively researched. Nevertheless, it is recognised that Crank’s solution to Fick’s second law provides a sound engineering basis from which to formulate chloride resistant concrete mixes benchmarked against diffusion coefficients derived from standard tests. The test methods used to evaluate the diffusion coefficient are a compromise between speed and faithful replication of natural conditions in service. Natural diffusion tests require almost a year to produce results. Therefore, accelerated tests have been designed to characterise the chloride resistance in a shorter time period. This is achieved by accelerating ionic flow through application of an electrical potential. Unfortunately, this can heat up specimens and thereby produce conditions that do not replicate natural service condition rates of ingress. Notwithstanding the drawback of accelerated tests, management of the chloride-induced corrosion risk is achieved through deemed-to-satisfy limits, specification of an ERC “XRDS” class or a performance-based specification based on a maximum diffusion coefficient tied to a particular test method. Sufficient confidence exists in modelling and testing to permit durability design by the partial safety factor method. Moving this to the next level, a full probabilistic based design, will require more research data on the mean values, standard deviations and distributions of variables. Nevertheless, the sustainability agenda demands that researchers and practitioners invest their energies together to embrace this possibility enthusiastically.

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Harrisson, A. 1995. Deleterious processes in concrete. Concrete 29, 6: 13–14. Helland, S. 1999. Assessment and prediction of service life of marine structures. A tool for performance-based requirements? CEN TC104/DuraNet Workshop. Design of Durability of Concrete. Berlin. 64–75. Higgins, D. 1986. Reducing the ingress of de-icing salts into concrete. Construction Repair and Maintenance 2, 1: 12–13. Hobbs, D. 1999a. Aggregate influence on chloride ion diffusion into concrete. Cement and Concrete Research 29: 1995–1998. Hobbs, D. and J. Matthews. 1998a. Minimum requirements for concrete to resist deterioration due to chloride-induced corrosion. Minimum Requirements for Durable Concrete, ed.D. Hobbs, 43–89. Crowthorne: British Cement Association. Kropp, J. 1995. Chlorides in concrete. In Performance criteria for concrete durability, ed.J. Kropp and H. Hilsdorf, 138–164. London: CRC Press. Liposcak. I. and A. Juraga. 2023. Application of MCI technology on Peljesca Bridge, Croatia. Concrete Engineering International 27, 1: 30–33. LNEC. 1993a. LNEC-E390 Specification Concrete. Determination of the resistance to penetration by chlorides. Immersion test. Lisbon: Laboratório Nacional de Engenharia Civil. LNEC. 2004. LNEC-E463, Specification Concrete. Determination of diffusion coefficient of chlorides by migration under non-steady state. Lisbon: Laboratório Nacional de Engenharia Civil. LNEC. 2007. LNEC-E465, Specification Concrete. Methodology for estimating the concrete performance properties allowing to comply with the design working life of the reinforced or prestressed concrete structures under the environmental exposures XC and XS. Lisbon: Laboratório Nacional de Engenharia Civil. Maage, M., Helland, S., Poulsen, E., Vennesland, Ø. and J. Carlsen. 1996. Service life prediction of existing concrete structures exposed to marine environment. American Concrete Institute Materials Journal 93, 6: 602–608. Marques, P., Costa, A. and F. Lanata. 2012. Service life of RC structures: chloride induced corrosion: prescriptive versus performance-based methodologies. Materials and Structures, 45, 1-2: 277–296. Nagesh, M. and B. Bhattacharjee. 1998. Modelling of chloride diffusion in concrete and determination of diffusion coefficients. American Concrete Institute Materials Journal 25, 2: 113–120. Nolan, É., Basheer, P. and A. Long. 1995. Effects of three durability enhancing products on some physical properties of near surface concrete. Construction and Building Materials 9, 5: 267–272. NordTest. 1995. NT BUILD 443, Concrete, hardened: accelerated chloride penetration. Espoo: Nordtest. NordTest. 1997. NT BUILD 355, Concrete, mortar and cement-based repair materials: chloride diffusion coefficient from migration cell experiments. 2nd Edition. Espoo: Nordtest. NordTest. 1999. NT BUILD 492, Concrete, mortar and cement-based repair materials: chloride migration coefficients from non-steady state migration cell experiments. Espoo: Nordtest. Onyejekwe, O. and N. Reddy. 2000. A numerical approach to the study of chloride ion penetration into concrete. Magazine of Concrete Research 52, 4: 243–250.

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Pettersson, K. 1992. Corrosion threshold value and corrosion rate in reinforced concrete. CBI Report 2:92. 19–20. Stockholm: Swedish Cement and Concrete Research Institute. Polder, R. and A. Hug. 2000. Penetration of chloride from de-icing salt into concrete from a 30 year old bridge. Heron 45, 2: 109–124. Pritchard, B. 1986. Road salt corrosion in U.K. concrete bridges - Part 2. Construction Repair and Maintenance 2, 6: 6–9. Roelfstra, P., Bijen, J. and T. Salet. 1996. Modelling chloride penetration into ageing concrete. In Concrete in the Service of Mankind: Concrete Repair. Rehabilitation and Protection, ed.R. Dhir and M. Jones, 245–255. London: CRC Press. SABS. 2015. SANS 3001-CO3-1, Civil Engineering test methods: Concrete durability testing index – Preparation of test specimens. Pretoria: South African Bureau of Standards. SABS. 2021. SANS 3001-CO3-3, Civil Engineering test methods: Concrete durability testing index – Chloride conductivity test. Pretoria: South African Bureau of Standards. Sagoe-Crentsil, K. and F. Glasser. 1989. Steel in concrete: Part 1. A review of the electrochemical and thermodynamic aspects. Magazine of Concrete Research 149: 205–212. Seatta, A., Scotta, R. and R. Vitaliani. 1993. Analysis of chloride diffusion into partially saturated concrete. American Concrete Institute Materials Journal 90, 5: 441–451. Siemes, T., Polder, R. and H. de Vries. 1998. Design of concrete structures for durability. Heron 43, 4: 227–244. Spooner, D. 1995. Chloride attack. Concrete 29, 2: 5. Streicher, P. and M. Alexander. 1995. A chloride conduction test for concrete. Cement and Concrete Research 25, 6: 1284–1294. Tang, L. and L. Nilsson. 1992. Rapid determination of the chloride diffusivity in concrete by applying an electrical field. ACI Materials Journal 89, 1: 49–53. Tang, L. and H. Sørensen. 2001. Precision of the Nordic test methods for measuring the chloride diffusion/migration coefficients of concrete. Materials and Structures 34, 8: 479–485. Tritthart, J. 1989. The influence of the hydroxide concentration in the pore solution of hardened cement paste on chloride binding. Cement and Concrete Research 19, 5: 683–691. Tuutti, K. 1982. Corrosion of Steel in Concrete. Stockholm: Swedish Cement and Concrete Research Institute. Wang, Y., Nanukuttan, S., Bai, Y. and P. Basheer. 2017. Influence of combined carbonation and chloride ingress regimes on rate of ingress and redistribution of chlorides in concretes. Construction and Building Materials 140: 173–183. Wu, J., Li, H., Wang, Z. and J. Liu. 2016. Transport model of chloride ions in concrete under loads and drying-wetting cycles. Construction and Building Materials 112: 733–738. Ye, H., Jin, X., Fu, C., Jin, N., Xu, Y. and T. Huang. 2016. Chloride penetration in concrete exposed to cyclical drying-wetting and carbonation. Construction and Building Materials 112: 457–463. Yuan, Q., Shi, C., De Schutter, G. and K. Audenaert. 2008. Effect of temperature on transport of chloride ions in concrete. In Concrete Repair. Rehabilitation and

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Retrofitting II, ed. M. Alexander, H-D. Beushausen, F. Dehn and P. Moyo, 345–351. London: CRC Press. Zhang, T. and E. Odd. 1996. Diffusion behavior of chloride ions in concrete. Cement and Concrete Research 26, 6: 907–916. Zhao, T., Zhang, Y., Kefei, L. and J. Wang. 2022. Probabilistic evaluation of chloride ingress process in concrete structures considering environmental characteristics. Structural Engineering and Mechanics 84, 6: 831–849. Zhu, X., Zi, G., Cao, Z. and X. Cheng. 2016. Combined effect of carbonation and chloride ingress in concrete. Construction and Building Materials 110: 369–380.

Chapter 8

Freeze-thaw effects

DISRUPTIVE FORCES IN FROZEN CONCRETE Concrete exposed to low temperatures on a cyclical basis is at risk of durability failure if appropriate measures are not taken to accommodate movement of water within the concrete. The problem arises due to concrete’s relatively low tensile strength. A common source of distress in concrete is its inability to withstand tensile stresses induced by the restraint of movement. Freezing of a proportion of the pore water in the cement paste or ice formation in susceptible aggregates is accompanied by expansion. Microcracking of the concrete will occur if the expansion is restrained, leading to an enlarged pore network. This may further increase the volume of water in the element and contribute to further ice-volume growth during the next freeze-thaw cycle. The cumulative effect of this vicious cycle over many freeze-thaw cycles may be serviceability failure at an age considerably less than the target service life. Fig. 8.1 illustrates the state of a cube after repeated freeze-thaw cycles in laboratory tests. The expansive forces are considerable. This can best be appreciated by filling a closed container with water beyond 92% of its volume and then freezing the container. Ice will wish to occupy a volume approximately 8% greater than the water from which it was formed. Pressure on the walls of the frozen container will become apparent if the critical saturation level of about 92% by volume is exceeded. The water-filled pore structure of concrete may be thought of in similar terms during a freeze-thaw cycle. The ice forms a moving front and pushes water ahead of it. The concrete will be durable if this movement can be accommodated until relief is provided from a thaw. However, expansive pressures beyond the tensile capacity of concrete will build up if movement is restrained by insufficient pore volume. The degradation process is more complex than just failure to accommodate the expanded volume (Pigeon and Pleau 1995), especially if de-icing salts exacerbate osmotic pressure. This aspect is discussed later in this chapter. The number of cycles is more significant than the absolute lowest temperature. Thus, concrete in continental Canadian Arctic regions, frozen DOI: 10.1201/9781003261414-8

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Figure 8.1 Freeze-thaw damage to a concrete cube following cyclical temperature variation in laboratory tests, showing scaling of surface layers.

for months at a time, is less at risk than concrete surfaces located in Icelandic Atlantic coastal conditions, which experience higher average temperatures but numerous cycles above and below freezing point each winter. The most vulnerable structures are those that have a high degree of saturation during cold weather. These principally include pavements and exposed elements of buildings, such as kerbs, parapets, copings, ledges, canopies, and corbels. Specialist infrastructure is also at risk, including freezer stores, tanks, and reservoirs.

MANIFESTATION OF DISTRESS Damage from the effect of freeze-thaw cycles is manifest in three ways: • internal cracking with subsequent D-line cracking or disintegration • surface scaling through delamination • pop-outs. Internal cracking and disintegration, often referred to as “D-line cracking”, results from repeated exposure of the cement paste to the damaging effects of expansive ice formation and related pressures, both hydraulic and osmotic. The terms “D-cracking” and “deterioration line” are also associated with this manifestation of distress. D-line cracking is typically

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Figure 8.2 Freeze-thaw damage pattern typically observable in an airfield apron subject to continental climate extremes but repeatedly cleared of snow for operational reasons.

first observed many years after construction, unless the process is accelerated through exposure to de-icing agents. Crescent-shaped lines appear on the surface of zones adjacent to free edges and joints in pavement slabs. Over time, further cracks are replicated with increasing distance from the joints, as illustrated in Fig. 8.2. This pattern, at the intersection of joints, has been characterised as “hourglass” by Thomas and Folliard (2007). The cracks initially appear near to and parallel to the edge or joint. Subsequent deterioration cycles cause additional cracking progressively further from the joint. Over time, the concrete surface within a distance of 0.5 m of the joint may exhibit spalling or disintegration which, on an airfield, would represent a serviceability failure due to the risk of foreign object damage to engines and airframes (Shahin 2009). The source of the distress is expansion of coarse aggregate particles at depth. Additionally, the water driven out of the aggregates may cause dissolution of components in the surrounding paste (Van Dam et al. 2002). Surface scaling is characterised by delamination of the surface due to the layered effect of frozen zones. Harrisson (1995) described the problems that occur in pavements where the ground under the slab may remain frozen for days at a time. The concrete at depth will remain frozen, but the surface layers may thaw during the day. Fall in temperature at night leads to the development of an ice front at the surface that moves towards the lower frozen layers. A lens of water is trapped between the two frozen zones. If the temperature continues to fall, the lens may expand and delaminate the surface. The process is illustrated in Fig. 8.3. Equally, problems may arise from unfrozen soils under a slab which feed water to the upper frozen layers in a wick effect. Scaling associated with de-icing salt application is another significant problem. First, the melting of thin ice sheets on the surface by salt requires energy, which is drawn from the concrete. This leads to a rapid temperature

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Figure 8.3 Delamination caused by the cyclical effect of a water lens trapped between a frozen subsurface and upper layers that alternately heat and cool in response to ambient air temperature.

drop near the surface. The consequent thermal shock may cause cracking and subsequent scaling. Second, the melt water, which contains de-icing salts, penetrates the surface. One consequence of this is the depression of the freezing point of the water near the surface. Rainwater in the uppermost surface layers will therefore freeze in advance of that below. This can lead to differential expansion of the surface layers with consequent scaling. Pop-outs are caused where freeze-thaw-susceptible coarse aggregate is used. Highly porous coarse aggregate particles may contain significant amounts of water when saturated and swell on freezing. Expansive pressures may be high enough when ice forms for the aggregate to fracture. Particles close to the surface may disrupt the cover concrete sufficiently for conical shaped pieces of concrete to detach from the surface. The differentiating factor from delamination is the presence of fractured aggregate at the base of the pop-out.

PROCESS OF FREEZING IN POROUS CONCRETE The disruptive process of a liquid freezing in porous media is related to a number of phenomena additional to the volume increase during the change of state. In the case of water in concrete, the other main aspects are the redistribution of solute concentration in the network and the influence of pore size on the temperature of freezing. Ice formation causes the flow of water within the network and the hydraulic pressure generated is dependent on the ability of the network to accommodate the volume of flow. The process involves both micro and macro aspects, including surface thermodynamics (Setzer 2013). Ice formation begins in the coarsest capillary pores while gel pores remain unfrozen. An ice body forms locally in a cavity from water already present and from water attracted to a growing ice crystal by osmotic pressure. This

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occurs because the pore solution contains alkalis and salts. The water expelled ahead of a developing ice crystal contains not only these existing solutes but also those driven from the now-frozen pure water. This increase in solute level leads to a concentration gradient and osmotic pressure (Powers 1975). This results in diffusion of water, including gel pore water, towards a developing ice body. The solute concentration aspect may be exacerbated in road slabs regularly subjected to de-icing salt application in cold weather. The excess unfrozen water and air are pushed out into the surrounding network. Resistance to this flow develops a hydraulic pressure, while the concentration gradient and osmotic pressure exacerbate the pressure buildup by feeding further ice growth. The solute impacts the freezing temperature of the remaining pore water. Also the growth of an ice body requires a seed ice crystal. These factors combine with others such that this pore water may become supercooled and remain liquid despite a continued fall in temperature below 0oC. The surface energy of the pore walls leads to adsorption of water molecules, reducing the pore water’s chemical potential, leading to a depression in the freezing point. The thin film of adsorbed water remains in a liquid state even after ice is formed in the pore. The greater the specific surface of the pore, the more the freezing point is depressed. Pores with a radius of 5 nm or less could remain unfrozen in temperatures above −20oC; less than half of the pore water in a typical concrete might be frozen by −30oC; and only two-thirds of the pore water may be frozen by −60oC (Comite Euro-International du Beton 1989; Pigeon and Pleau 1995). An obvious source of expansive pressures within frozen concrete is the volume increase of water as it changes state from liquid to solid. The hydraulic pressure buildup caused by the moving front formed by material forced out by local ice formation is also significant in creating microcracks. The flow may be impeded by insufficient pore volume, and resistance to flow is proportional to the length of the flow path in accordance with Darcy’s law for transport of fluids through a porous medium. The resulting expansive pressures may exceed the tensile capacity of the concrete, leading to microcracks which then become water-filled by capillary action. These cracks may experience growth if this water freezes in the next cycle.

PRIMARY FACTORS OF INFLUENCE The main factors of influence on freeze-thaw behaviour and damage are as follows: • degree of saturation • pore structure • aggregate characteristics

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• climatic conditions • the effect of de-icing salts.

Degree of saturation Saturation can occur in concretes that are permanently exposed to water and in concrete occasionally exposed to water under pressure. Concrete exposed to de-icing salts may also become saturated due to the effect of freezing point depression. The volume expansion of about 8% occurring on conversion of water to ice cannot be accommodated in fully saturated concrete. Clearly, the more air available in the pores, the less damaging the effect. The critical threshold of degree of saturation, based on a Belgian study cited by Neville (1995) and others, lies close to 90%. If the relative humidity of the atmosphere is less than 97%, some evaporation of water is possible during freezing, which may allow sufficient space to be created for expansion without deleterious effects (Comite Euro-International du Beton 1989). The greater the volume of water available, the greater the potential expansion, and hence the greater the risk of exceeding a concrete’s tensile strength. The total water volume is influenced by both the water present in the concrete after hydration and the subsequent availability of water from external sources. Low water/cement ratio concretes are beneficial in two ways in this context. First, the hydrated concrete contains less free water to begin with. Second, the impermeability of the pore network reduces the amount of external water capable of being introduced during the service life from external sources. The availability of water from external sources is influenced by the environment and by the detailed geometry of the structure. Features that promote a decrease in the contact time between water and concrete are thus beneficial. Details worthy of consideration are well-drained slabs, sloping wall tops, use of drips, and so forth.

Pore structure The diameter of a pore influences the temperature at which the water within it will freeze. The water in the largest pores is the first to form ice blocks at a given temperature. Further drops in temperature may cause the water in the capillary network to freeze, but the gel pore water can remain unfrozen. The volume and proximity of spaces into which the expelled water may escape greatly influence the degree of resistance to damage. Thus a balance is required between high porosity and low permeability. For example, Dhir et al. (2000) reported that the benefit of reducing the free water content to lower expansive forces on freezing may be partly offset by the restriction to movement caused by the improved microstructure. Air entrainment provides a controlled method of accommodating the water movement caused during ice formation. On thawing, the water in the entrained voids is drawn back out into the narrower pore network by capillary forces.

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Air entrainment represents one of those situations where we learn from nature that flexibility can be as important as strength. We observe after a storm that some of the strongest trees in a forest have snapped, while weaker trees survive unscathed. We observe after an earthquake that some old buildings remain standing, whereas some heavily reinforced structures exhibit significant signs of distress. Equally, in concrete technology, we might equate durability “strength” with low levels of porosity related to low water/cement ratio mixes. However, the concept of air entrainment is to increase the porosity to add “flexibility” to our concrete during a freeze-thaw episode. Water is provided with space to grow into during the freezing phase, which thereby prevents a buildup of pressure. Stress relief is provided by this flexibility. During the thawing phase, the water in the entrained voids is safely drawn back out into the narrower pore network by capillary forces. These entrained air bubbles are distinct from entrapped air, which causes uncontrolled permeable networks and is minimised by compaction of fresh concrete insitu. The effective use of air entrainment involves the achievement of a distribution of bubbles such that protected spheres overlap. Thus a large number of small air voids is effective, whereas an equivalent air volume in the form of a small number of large voids is not. Air entrainment is achieved by adding an admixture to the concrete that is capable of distributing discrete air bubbles. Three parameters are relevant – air content, average spacing, and specific surface. The admixture is a surfactant, a surface-active agent, which stabilises the air bubbles formed during mixing and distributes them uniformly through repellent forces. The air content required depends on the volume of frozen water to be accommodated. The volume of water is a function of the permeability and porosity and can be minimised through specification of a suitably low water/cement ratio. Additionally, the admixture increases the consistence and cohesiveness of fresh concrete, which enhances the durability of the near-surface layers. The mean air content required is approximately 9% of the cement paste volume. This equates to about 5% of the concrete volume in the case of concrete with a maximum aggregate size of 20 mm. Concretes with lower maximum aggregate size have higher cement paste volumes, and consequently higher air content requirements. The converse applies in the case of larger aggregate. Typical mean values are 7.5% and 4.5% for 10 mm and 40 mm maximum nominal upper aggregate size concretes, respectively, in Portland cement concretes. Although the total air content is significant, it is equally important that this volume is distributed in a manner whereby the path length to an air bubble is sufficiently short. Specific surface is the surface area to volume ratio. Thus the higher the specific surface, the smaller the air bubbles. It is recommended that the entrained and entrapped air should have a minimum specific surface of 25 mm−1. The entrained air bubbles have typical diameters of 50 μm.

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It has long been recognised that air entrainment leads to a slight loss of concrete strength (Wright 1953). A value of approximately 5% loss per 1% of entrained air is a good rule of thumb. This factor must be considered when framing specifications that include requirements in respect of air content and minimum strength requirement. A note of caution was recorded by Dhir et al. (1999) regarding air entrained PFA concretes. The chemical characteristics of the admixture need to be considered in respect of potential interaction with PFA. Vinsol resin admixtures performed satisfactorily, but the admixture demand increased by a factor of two or more compared with Portland cement concretes, depending on the loss on ignition of the ash. Hooton and Vassilev (2016) reported better performance with synthetic air-entraining admixtures than Vinsol resins in the case of slag concretes, although they also questioned the validity of current standard tests designed for Portland cement concretes when used with concretes containing secondary cementitious materials. A problem with such tests is the prescribed curing time of typically 28 days, which disadvantages concretes requiring longer hydration periods for maturity.

Aggregate characteristics Aggregates with high water absorption characteristics are at risk of contributing to freeze-thaw damage. Such aggregates may expand disruptively if they have a high water content in coincidence with freezing temperatures. Fracturing of shale, for example, is possible on freezing. Aggregate size also has an influence on freeze-thaw behaviour. The smaller the maximum aggregate size, the greater the fraction of cement paste per unit volume. This potentially increases the volume of pore water, the expansion of which would have to be accommodated. The precise relationship between aggregate porosity and freeze-thaw resistance is not easily established, because porous aggregate can also contribute valuable space to accommodate frozen product. A review by Thomas and Folliard (2007) identified pore size distribution as the single most important feature. They identify the size range of 0.1 to 5 μm as the greatest contributors to durability risk. Below 0.1 μm the freezing point is depressed. Above 5 μm the pores are not completely water-filled, and so there is room for expansion on freezing. Total porosity is also relevant, and aggregates with absorption capacities below 1.5% were reported as durable. Additionally, it is noted that the structure of the interfacial transition zone (ITZ) can be a significant factor in providing a possible space to accommodate water expelled from the aggregate without causing expansive pressures to build up. On the other hand, Liu et al. (2022) enhanced freeze-thaw durability by filling spaces, including the ITZ, by adding denitrifying bacteria to recycled aggregate concrete. The deleterious effect of using aggregates susceptible to durability failure in freeze-thaw conditions cannot be mitigated through air entrainment.

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Climatic conditions The absolute lowest temperature, period of freezing, rate of cooling, and frequency of freeze-thaw cycles are all influential contributors to durability risk. The cumulative effect of freeze-thaw makes the frequency of cycle particularly significant, but the other factors are noteworthy. The lower the temperature and the longer the freezing period the greater the volume of pores exposed to potentially damaging conditions. Concrete has a low rate of heat transfer; therefore, significant penetration of the ice front requires very low temperatures, long periods of freezing, or a combination of both. Climatic conditions that lead to a slow rate of cooling may mitigate damage by allowing dissipation of pressure build up from water movement. The number of freeze-thaw cycles per annum is more significant than the absolute temperature drop and period spent at low temperature.

De-icing salts De-icing salts can influence the degree of saturation and may also subject concrete surfaces to damaging thermal shock. The role of these agents in practice is to form a solution with a lower freezing point than water, thus expediting the thaw of snow and ice on concrete surfaces. This short-term gain may, however, have a long-term deleterious effect, leading to surface scaling in addition to the chloride ingress issues described in Chapter 7. The degree of saturation of pores in-service depends on both the exposure to water and the exposure to drying conditions. The smaller the pore the greater the surface tension holding in the water. Thus, for typical service drying conditions, the smaller pores remain more saturated. One effect of de-icing salts is to increase the surface tension forces, thus limiting the drying effect and increasing the likelihood of saturation. The critical relative humidity, below which drying is induced for a pore of given radius, increases as the solute concentration of the pore liquid increases (Harrison et al. 2001). The degree of saturation at a given temperature is therefore higher in salt-contaminated pore water than in concrete free from the influence of de-icing salts. Surface scaling results from the differential response of near-surface layers subject to repeated application of de-icing agents. Melt water contaminated by de-icing agents can penetrate the upper layers of a concrete slab. Rainwater wash-out of the near-surface layers reduces the concentration of de-icing agent in the upper layers compared with those below. Consequently, the reaction to the next cycle of low temperatures becomes stratified, with the saturated upper layers freezing while the contaminated layer below remains unaffected due to its depressed freezing point. In time, the upper layer becomes detached. Thermal shock is the rapid drop in temperature in the upper layers of concrete following the application of de-icing agents. The melting of surface

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snow and ice at low temperatures requires energy. Energy is drawn from upper surfaces of the concrete when de-icing agents are applied. A temporary drop in temperature occurs. Repeated application of de-icing agents will therefore encourage a series of temporary short-term stress episodes that cumulatively weaken the upper layers.

DEVELOPMENTS IN MODELLING FREEZE-THAW BEHAVIOUR Two immediate difficulties arise in attempts to model freeze-thaw behaviour in structures. First, there is the unpredictable pattern of damaging cycles. Second, it is difficult to correlate the sensitivity of mix parameter variations in laboratory tests with validated experience of structures in service. Regarding the first issue, freeze-thaw durability failure is event-driven, as illustrated in Fig. 8.4, and does not follow a predictable time-related pattern. The level of damage per cycle could vary from minor growth of internal cracks (for example, damage ‘x1’) to delamination of a piece of concrete of unpredictable size (for example, damage ‘x2’). Equally, the rate of damage will vary from year to year, depending on local climatic conditions. Although average temperatures are reasonably predictable, the number of freeze-thaw cycles in a period is much less certain. Additionally, in many locations, the dwell period between series of cycles (for example, period ‘td’) could vary from a few months to more than a year – a harsh winter versus a mild winter – such is the unpredictability of the weather. The second issue relates to translation of laboratory research findings into practice. Laboratory trials are generally conducted with daily cycles in a range from +20°C to –20°C. Although comparative performance of different concretes in laboratory trials is informative, the absolute performance in these trials cannot be readily translated to field experience due to climatic variability, which influences both temperature and degree of saturation.

Figure 8.4 Schematic representation of the event-driven nature of freeze-thaw damage.

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The possible direction of models for durability design and service life prediction has been outlined by Fagerlund (1997). The models are based on the concept that the concrete will fracture at a critical degree of saturation. This critical degree of saturation is a function of the material and is a consequence of the distance between a freezing site and the nearest air-filled space. The service life is also a function of the wetness of the environment. The capillary water uptake and long-term water absorption of the air pore system are the significant issues and are modelled by a time relationship.

Critical distance concept model The potentially damaging stress on the concrete pores due to freezing increases with the increasing distance between the freezing site and the nearest air-filled space. This gives rise to the concept of a critical distance (DCR) in respect of freeze-thaw damage. This can be envisioned as the thickest possible water saturated cement paste zone around an air-filled void, such as a sphere, which will remain undamaged if freezing occurs. This leads to the concept of a critical spacing factor (LCR), illustrated in Fig. 8.5. A relationship for the critical spacing factor was proposed by Powers (1949), related to Darcy’s law of water flow through a porous medium: 3 3 LCR 3LCR K T + = rb 2 U R

where LCR = thickness of critical zone around air void rb = radius of air void K = permeability coefficient of cementitious matrix T = tensile strength of cementitious paste U = quantity of water that freezes per unit fall of temperature R = rate of freezing.

Figure 8.5 Air entrainment: overlap of protected zones and concept of critical spacing factor (LCR).

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The “critical zone” thickness concept was studied by Fagerlund (1997), who related it to the “critical distance” and developed an equation for the air pore volume required to prevent the critical zone thickness being exceeded when the pores were water-filled. The starting point was the relationship between critical distance and the thickness of the critical zone around an air void: DCR = 2LCR =

s a

2 LCR +1 9

0.5

where DCR = critical distance α = specific surface of the enclosed air void s = total envelope surface of all air voids considered a = total volume of all air voids considered. Following on from Power’s equation, the spacing factor (L) of a system of spherical pores in a material matrix may be determined from: L=

3

1.4

VP +1 a

0.3

1

where L = spacing factor VP = volume of the cement paste. Thus, the volume of the air required to form the pore system may be stated as: a=

VP L 0.364 +1 3

3

1

Fagerlund (1997) demonstrated that the equation must be modified to take account of the fact that the air pores will not be completely dry. Therefore the required air content (a0) is as follows: a0 = aw + aCR + ab

where a0 = required air content aw = water-filled air pore volume

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aCR = air pore volume required to prevent LCR being exceeded when aw is reached ab = safety margin. Temporarily omitting the safety margin, the minimum air content required is therefore: (a0)min = aw + aCR

and the value may be determined from the following formulae: aw

= Sa (a0)min VP

aCR = 0.364

LCR 3

CR

+1

3

1

where Sa = degree of water-filling of the air pore system. The safety margin (ab) may then be added at a level that reflects the required service life. It has been determined experimentally that, for concretes with water/cement ratios in excess of 0.45, the mean critical thickness (DCR) is 1.2 mm for freeze-thaw in pure water and 1.8 mm in a 3% sodium chloride solution. These values yield critical spacing factors (LCR) of 0.40 mm and 0.54 mm, respectively, assuming a specific surface (α) of 15 mm–1.

Critical level of saturation model An alternative approach from Fagerlund (1993) is to consider the time required for suction to reach a point of capillary degree of saturation. Building on 25 years of research and his earlier publications on the topic, a detailed update of the theoretical calculation of service life and required air content is set out in Appendix 1 of his 2004 report (Fagerlund 2004). The potential service life (tp) is defined by the critical level of saturation (SCR) where: SCR = SCAP (tp)

and the capillary degree of saturation (SCAP) is modelled by the formula: SCAP (t) =

1

[

0

+C

D 0

a0 ( t)E ]

where SCR = critical level of saturation SCAP = the capillary degree of saturation

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tp = service life ∈ = total porosity ∈0 = porosity exclusive of air pores C, D, E = constants α0 = specific area of the air pore system a0 = total air content δ = diffusivity of dissolved air.

TEST METHODS FOR FREEZE-THAW RESISTANCE Tests methods related to freeze-thaw can be categorised into three groups: • tests on the proposed aggregates • tests on fresh concrete to verify air entrainment characteristics • tests on hardened concrete. Tests for frost-resistant aggregates form an important starting point for durable concrete. The fresh concrete tests are concerned with the testing of concrete prior to use on-site to verify that the specified level of air entrainment has been maintained from point of production to point of placement. Tests on hardened concrete primarily relate to the assessment of anticipated performance in service but have the potential to form the cornerstone of performance-based specifications. International development has converged on a small number of similar methodologies. An overview of these methodologies is presented in the summary review which follows drawing on standard tests, technical reports, and technical specifications published by ASTM International and CEN, the European Committee for Standardisation. Researchers may also be interested to study the work of Marks and Dubberke (1982) in respect of the Iowa Pore Index Test (IPIT) and development of the Washington Hydraulic Fracture Test (WHFT) reported by Janssen and Snyder (1994). The appropriateness of the current standard test methods for lower-carbon concretes rather than Portland cement concretes, and also high-performance concrete, is open to question. Further research on these topics is clearly of great interest, particularly in the context of performance-related specifications. A note of caution was sounded by the Concrete Society (1996), which noted that many freeze-thaw tests on hardened concrete are very severe and, in particular, non-air-entrained mixes that have good service records in the UK may fail such tests. This emphasises the need to take local conditions and experience into account when interpreting the results from tests, especially those outlined in this chapter.

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TESTING THE FREEZE-THAW RESISTANCE OF AGGREGATES Overarching requirements for aggregates in concrete may be found in standards such as ASTM C33 (ASTM 2018a) and EN 12620 (CEN 2008b). These include guidelines on frost-resistant aggregates. Screening of aggregates for durability in freeze-thaw exposure conditions begins with petrographic examination to identify pore structures that flag a potential durability risk. Petrographers working to standard guidelines such as ASTM C295 (ASTM 2019a) and EN 932-3 (CEN 2022a) can give early warning of frost susceptibility based on their expert knowledge and experience of aggregate performance in concrete in the place of use. Material that clears the initial screening might also be validated by absorption tests for coarse aggregate, such as ASTM C127 (ASTM 2015a) and EN 1097-6 (CEN 2022b). Absorption values below 1% are regarded as being resistant to freeze-thaw attack, but higher values may be satisfactory depending on petrography, and this is open to validation through cyclical exposure trials. Practitioners using novel materials will need to satisfy themselves as to the robustness of the aggregate combinations. A novel approach to the testing of fine recycled aggregate was proposed by Rodrigues et al. (2013) to overcome some drawbacks in tests designed originally for natural aggregates. Should circumstances merit further examination to validate the durability of an aggregate, it can be subjected to cyclical exposure under harsh conditions. A variety of standard tests, using different storage media, are available from which to ascertain the soundness or frost resistance. Examples of cyclical exposure include ASTM C88 (ASTM 2018b), CSA A23-2-24A (CSA 2004), EN 1367-1 (CEN 2007), EN 1367-2 (CEN 2009a), and EN 1367-6 (CEN 2008a). Abrasion tests may also have a role to play and include ASTM C131 (ASTM 2006) and EN 1097-2 (CEN 2020a), or the micro-Deval test ASTM D6928 (ASTM 2017d) and EN 1097-1 (CEN 2011c).

Cyclical freeze-thaw testing of aggregates The resistance to freezing and thawing test EN 1367-1 provides a good example of the basic methodology used in these exposure trials. The aggregates are washed, dried, and cooled to ambient temperature. The initial mass is recorded. The specimens are stored in metal cans and soaked in distilled or demineralised water for 24 hours. The aggregates in the cans are then subjected to ten freeze-thaw cycles in a temperature range of +20oC to −17.5oC in accordance with a prescribed time-temperature curve such that one cycle can be completed within 24 hours. An example of a single-cycle prescribed curve for a 24 hour period is illustrated in Fig. 8.6. The mass loss is determined at the end of the cycles by wet-sieving the contents of each can on a sieve that is half the lowest sieve size used in specimen preparation. The retained mass is determined after oven-drying and cooling. The mean value of

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Figure 8.6 Example of a time–temperature curve for one freeze-thaw cycle per day.

mass loss is used to determine a “freeze-thaw value” (F). The aggregates are then categorised. For example, “Category F2 aggregate” exhibits a mass loss greater than 1% but less than or equal to 2%.

Soundness tests as an indicator of freeze-thaw resistance Soundness tests expose aggregates to cycles of immersion in sulfate solutions. Sulfate crystallisation causes a buildup of expansive pressure, which is used to simulate the effect of ice crystallisation. Prior to the widespread availability of purpose-designed freeze-thaw chambers, the soundness test was used as an indicator of frost resistance. Performance of aggregates in a hostile soundness test was correlated with in-service experience of concretes in cold weather. The ASTM C88 test may be conducted with sodium sulfate or magnesium sulfate solutions. The EN 1367-2 test is solely based on the harsher magnesium sulfate solution. The aggregate is preconditioned by elimination of undersized and oversized pieces using 10 mm and 14 mm sieves, washed free of dust using distilled water, dried in an oven for 24 hours, cooled in a desiccator to ambient temperature, and transferred to mesh baskets. The baskets, each containing approximately 0.4 kg of aggregate, have a mesh size that is coarse enough to allow free circulation of the solution but fine enough to trap the aggregate particles at the start of the test. Five cycles are conducted, using immersion in sulfate solution for 17 hours followed by drainage, oven-drying, and cooling. Each cycle requires a period of 48 hours. A quantitative assessment of mass loss after five cycles allows determination of “magnesium sulfate values” (MS) calculated as the percentage mass loss.

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The aggregates are then characterised by one of four “MS” categories. For example, “Category MS35 aggregate” exhibit a mass loss greater than 25% but less than or equal to 35%.

Cyclical freeze-thaw testing of aggregates in salt solutions The quality of freeze-thaw resistant aggregates is an important consideration in continental Canadian climates. The CSA A23-2-24A test was developed to better simulate severe field conditions. This is achieved by soaking the aggregates in a 3% sodium chloride solution prior to five freeze-thaw cycles. The cycles involve 16 hours of freezing to −17.5oC followed by 8 hours of thaw at room temperature. Mass loss is determined after washing and oven drying. A maximum of 6% mass loss is recommended for aggregates to be used in severe exposure conditions and a limit of 10% otherwise. Control aggregates are available, if required. European practice has also adopted a severe freeze-thaw test, EN 1367-6, to assist in the specification of frost-resistant aggregates in concrete exposed to seawater and de-icing agents. The aggregates are stored in metal cans and soaked in a 1% sodium chloride solution. The specimens are subjected to ten freeze-thaw cycles in a temperature range of +20oC to −17.5oC. The mass loss is determined by sieving through a sieve that is half the lowest sieve size used in specimen preparation. The mean value of mass loss is used to determine a freeze-thaw value, distinguished from the EN 1367-1 “F value” by the subscript “NaCl” (FNaCl).

Abrasion testing of aggregates Potential durability of aggregates in cold climates may be derived from severe abrasion tests such as the Los Angeles test, ASTM C131 (ASTM 2006) and EN 1097-2 (CEN 2020a), or the micro-Deval test ASTM D6928 (ASTM 2017d) and EN 1097-1 (2011c). Mass loss is used as an indicator of durability, similar to interpretation of results from the soundness tests. The introduction of wet attrition to the abrasive process in the micro-Deval test has resulted in good reproducibility and correlation with other tests. An extensive test programme on aggregates of the North American continent by Lang et al. (2007) found that the micro-Deval test was an excellent method of identifying good-quality aggregates. The correlation with other tests was good, but they did not recommend its use as a definitive test for the rejection of an aggregate.

TESTING OF FRESH CONCRETE FOR AIR CONTENT Specifications commonly state the mean volume of air to be entrained as a percentage of the concrete. Although this is the specification requirement,

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what is intended in-service is both a controlled volume of air as a percentage of the cement paste and the distribution of this air in a manner that conforms to a narrow range of diameters and spacings. A common test method on fresh air-entrained concrete is to check the total air content using a pressure method which, through Boyle’s Law, can be directly related to air content. Such a test is useful for checking production consistency, but does not provide any indication of the manner of distribution of the air bubbles. A method has also been developed in Denmark that can measure air content, spacing factor, and specific surface in about 20 minutes. The physical characteristics of an air volume in concrete are the basis for three simple approaches that can determine a reliable estimate of total air content. Although these methods do not provide any information on the distribution of the air in the concrete, the tests serve as useful quality control techniques to affirm the presence and effectiveness of an admixture dosage in creating a controlled air volume. The tests are separately based on pressure, substitution of air by water, and unit weight. Tests designed for assessment of fresh concrete deliveries on-site are regarded as providing a stable result if three consecutive measurements are within ±1.5%.

Pressure method The pressure method is based on Boyle-Mariotte’s Law. It is assumed that solid constituents in fresh concrete and water are incompressible, so that volume change under pressure is due to the contraction of air voids. Volume change and difference in pressure are directly related to air content. There are various designs of the test, commonly described as the “water column”, “pressure gauge”, “Type A”, and “Type B” methods. The test is suitable only for relatively dense aggregates, and determination of the result involves the use of an aggregate correction factor. It does not take account of air in the aggregate pores and is therefore unsuitable for use with highly porous or lightweight aggregates. Examples of these recognised standard methodologies include ASTM C231 (ASTM 2009a) and EN 12350-7 (CEN 2022c).

Volumetric method The volumetric method measures air in the mortar fraction of the concrete. An apparatus (roller meter) allows water with isopropyl alcohol to replace air voids in agitated concrete while monitoring the displaced air volume. The quantity of alcohol to be added varies with air content, but experience with repetitive testing of similar mixes on-site would allow consistent amounts to be used over the course of a project to reliably detect deviations from specification. The kit is designed for use on-site, and annual calibration of the meter is recommended, lest wear and tear lead to inaccurate readings. The test is suitable for use with porous aggregates.

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The methodology is set out in ASTM C173 (ASTM 2016a), which is recognised internationally as the definitive method for lightweight aggregates.

Gravimetric method The gravimetric method is used to determine the density of freshly mixed concrete, from which it is possible to calculate the unit weight, yield or relative yield, cement content, and air content of the concrete. The difference between actual and theoretical unit weights is used to estimate the volume represented by air. An example of a standard test is ASTM C138 (ASTM 2017a).

Danish method The Danish method has been described by Price (1996). It is based on the principle that the rate of rise of an air bubble in water is related to its size. The technique involves sampling paste from fresh concrete and injecting it into a viscous liquid at the base of a column of water. The air bubbles are released and rise through the column, where they strike a plate. The change in buoyancy of the plate with time can be used to determine air content, specific surface, and spacing factor. By way of example, the test method allowed inclusion of air content limits (entrained and entrapped) and minimum specific surface value in the specification of concrete for the Storebaelt West Bridge project.

FREEZE-THAW RESISTANCE TESTS ON HARDENED CONCRETE Tests on hardened concrete in the context of freeze-thaw may be categorised into three distinct groups. The first is somewhat forensic, determining parameters of an air void system by microscopic examination. The second category relates to the internal structural damage, which reduces the dynamic modulus of elasticity. Three test methods are available to estimate the freeze-thaw resistance in respect of internal structural damage. The results can be used for comparative studies of the performance of proposed mixes against concretes of known resistance in local conditions. The third category relates to scaling. These comparative tests have evolved from a Swedish method using 50 mm thick slabs cut from cubes, a German method using cubes, and a RILEM method based on 70 mm thick specimens cast in cube moulds. The Swedish method is generally recognised as the reference test, and the others as alternatives. Current practice in Europe is substantially based on the co-ordinating work of RILEM Technical Committee TC 117-FDC (1995, 1996, 1997). The US standard related to scaling, ASTM C672, has not been

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updated within the normal five-year review period and was withdrawn in 2021. The European practice is based on a technical specification yet to be approved as a standard. The applicability of the tests to concretes containing secondary cementitious materials has been questioned and is a matter that will gain in importance among the research community as we move to greater use of low-carbon concretes.

Air void system test Spacing factor, air content, and specific surface can be determined through the ASTM C457 (ASTM 2009b) linear-traverse or modified point-count microscopical method. Enhanced repeatability and reproducibility have been reported by automation of the process (Jacobsen et al. 2006).

Tests for internal structural damage The test methods for the freeze-thaw resistance in respect of internal structural damage are set out in publications used internationally, such as technical report CEN/TR 15177 (CEN 2006b) and ASTM C666 (ASTM 2015b). Three test methods are set out in CEN/TR 15177, differentiated by specimen geometry and the temperature-time regimes for a freeze-thaw cycle. The geometries are as follows: 400x100x100 mm beams; or 50 mm thick slabs cut from 150 mm cubes; or 70 mm thick slabs formed by casting specimens in a 150 mm cube mould incorporating a centralised Polytetrafluorethylene plate. Each test has a prescribed freeze-thaw cycle, which falls somewhere between a maximum of 24oC and a minimum of −22oC. The tests are based on 56 cycles, but this may be regarded as a minimum value. The freezing medium is typically de-ionised water, but a mixture of tap water (97% by mass) and sodium chloride (3% by mass) may also be used to examine a harsher de-icing agent exposure scenario. Internal damage is monitored by a change in length (mm), fundamental transverse frequency (Hz), or ultrasonic pulse transit time (μs). This allows determination of the relative length change or calculation of a relative dynamic modulus of elasticity; for example, in the case of fundamental transverse frequency measurements: RDMFF , n = 100

fn

2

f0

where RDMFF,n = relative dynamic modulus of elasticity, fundamental frequency (%) fn = fundamental frequency after n freeze-thaw cycles (Hz) f0 = initial fundamental frequency (Hz).

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Microcrack development in the ASTM C666 test is detected by determining the change in dynamic elastic modulus by monitoring the fundamental transverse frequency (Hz). Cast prisms, cylinders, or samples cut from hardened concrete are used with cross-sectional dimensions in the range of 75–125 mm and 275–405 mm long. The specimens are exposed to 300 freeze-thaw cycles or until the relative dynamic modulus reduces to 60%. The cycles are short, with rapid freezing and cooling, either through storage in containers of water or frozen in air and thawed in water, over a temperature range of +4.4oC to −17.8oC. A “durability factor” is determined based on the relative dynamic modulus and the number of cycles in the test relative to the standard number of 300 thus:

DF =

P N M

where DF = Durability Factor P = relative dynamic modulus of elasticity, fundamental frequency (%) N = number of cycles at termination (typically either 300 or when P = 60%) M = maximum number of cycles specified at termination (typically 300).

Scaling resistance – Swedish slab test The reference test method for freeze-thaw resistance in respect of scaling in CEN/TS 12390-9 (CEN 2016) is based on the Swedish Standard SS 13 72 44 (Swedish Institute for Standards 2005). The test set-up is illustrated in Fig. 8.7. Four specimens 50 mm thick are sawn from individual 150 mm cubes. Rubber is glued to the faces other than the sawn test face, and protrudes

Figure 8.7 Test set-up for freeze-thaw resistance: Swedish slab method.

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above the test surface to form a reservoir. Preconditioning involves a period of 72 hours exposure to ponding of de-ionised water at 20°C. During the freeze-thaw cycles, the specimens are sealed into an insulated mould and covered by 3 mm of de-ionised water or 3% sodium chloride solution. A polyethylene sheet is used to prevent evaporation. One freeze-thaw cycle per day is achieved by exposing the specimen to a temperature range of +20(±4)°C to –20(±2)°C in accordance with a specified time-temperature curve. The freezing period covers 16 hours, allowing 8 hours to achieve a thaw before the next daily cycle. The mass of oven-dried scaled material is assessed after 7, 14, 28, 42, and 56 cycles. The cumulative value after 56 cycles is used for evaluating the scaling resistance (Sn) expressed in kg/m2.

Scaling resistance – German cube test One of two alternative test methods for freeze-thaw resistance in respect of scaling in CEN/TS 12390-9 (CEN 2016) is based on the German cube test published in the Deutschen Ausschuss fur Stahlbeton monograph (Bunke 1991) as a supplement to the German standard for testing of concrete DIN 1048. The German cube test involves immersing two pairs of 100 × 100 × 100 mm cubes in a freezing medium of de-mineralised water or 3% sodium chloride solution. Each pair of cubes is stored in a brass or stainless steel container, as illustrated in Fig. 8.8. Preconditioning involves immersion in the freezing medium for a period of 24 hours. Fifty-six daily freeze-thaw cycles are conducted to a specified time-temperature curve involving a rapid thaw. The temperature range is +20(±2)°C to –15(±2)°C. Scaling is monitored at 7, 14, 28, 42, and 56 days. Provision is made for a break in the daily cycle of tests for periods such as weekends, when a cycle can be extended by leaving the specimens in a frozen state at –15(±2)°C. Scaling resistance (P) is assessed by determining the oven-dried mass of the cumulative scaled material after 56 cycles and expressing this as a percentage by mass of the air-dry mass of the cubes prior to test.

Figure 8.8 Test set-up for freeze-thaw resistance: German cube test.

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Scaling resistance – CF/CDF test The second alternative test method for freeze-thaw resistance in respect of scaling in CEN/TS 12390-9 (CEN 2016) is based on the CF (Capillary suction and freeze thaw) and CDF (Capillary suction of de-icing chemicals and freeze thaw) tests recommended by RILEM Technical Committee TC 117-FDC (1996). The method involves casting two specimens per 150 mm cube mould by inserting a PTFE plate vertically at the midpoint of the sides. This yields 70 mm thick specimens with 140×150 mm faces. Five specimens are employed per test. The lateral surfaces of the specimens are sealed by foil-backed butyl rubber or by resin. The specimen is placed, test surface down, in stainless steel containers. The test surface, that cast against the PTFE, is raised off the base of the container by 5 mm spacers and immersed in the freezing medium, which fills the container to a depth of 10 mm, allowing capillary suction. The freezing medium is de-mineralised water or 3% sodium chloride solution for the CF and CDF test, respectively. The test set-up is illustrated in Fig. 8.9, showing each container suspended in the coolant of a temperature-controlled chest. The specimens are subjected to two freeze-thaw cycles per day by cycling from +20°C to –20°C in each 12-hour period in the time-temperature pattern illustrated in Fig. 8.10. Loosely adhering scaled material is recovered from the specimens by use of an ultrasonic bath after 4, 6, 14, and 28 freeze-thaw cycles in the CDF test or 6, 14, 28, 42, and 56 freeze-thaw cycles in the CF test. The scaled material is recovered by filtering and oven-drying. The cumulative value after 28 cycles (CDF Test) and 56 cycles (CF Test) is used for evaluating the scaling resistance (Sn) expressed in kg/m2.

Figure 8.9 Test set-up for freeze-thaw resistance: CF/CDF method.

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Figure 8.10 Example of time–temperature curve for two freeze-thaw cycles per day.

A somewhat similar test in China, GB/T 50082-2009 (MOHURD 2010), has criteria for passing the test in respect of both scaling and internal cracking. The mass loss should not exceed 1.5 kg/m3 and the relative dynamic modulus should be above 80%.

MANAGEMENT OF THE FREEZE-THAW DURABILITY THREAT The durability threat from cyclical freeze-thaw conditions is typically addressed by a strategy of avoidance rather than a controlled rate of deterioration over a target service life. Resistance is achieved through aggregate selection if possible, and by controlling the pore distribution in concrete by one of two diverse approaches. The first approach involves reducing the permeability and porosity of the concrete to such an extent that the free water content will be low enough that the amount of expansion will not be deleterious. The second approach involves increasing the porosity of the concrete in a controlled manner through air-entrainment so that the pores can act as safety valves, providing space for the freezing water to expand without stressing the concrete. The approaches carefully manage either the amount or the accommodation of free water in the concrete. One way or another, the specification of moderately low water/ cement ratio concretes is key. The prescriptive deemed-to-satisfy route therefore endures as the primary approach to managing the freeze-thaw threat to durability. Performance-based specifications based on mathematical models are not strongly favoured as a way forward, but a “halfway house” solution may emerge in time through the equivalent durability procedure (EDP) allied to internationally recognised test methods. There is a possibility that the exposure resistance class (ERC) concept in Europe might be extended to the XF exposure classes, but there are many hurdles to overcome before that could become a reality.

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Although the emphasis is on management of durability through careful decision making at design and specification stage, it is of course vital to protect immature concrete from freezing during the construction phase. The heat of hydration should be exploited to keep the temperature of the concrete above 5oC for the first 48 hours. If possible, concrete should not be placed if the ambient temperature drops below 2oC on a falling thermometer. Postponement of a pour may be impractical in certain geographical regions, in which case the necessary protective steps are most probably an integral part of local good practice. Precautions to be exercised during cold weather are well-publicised in guides and manuals, for example, Pink (1978). Climate change is leading to extremes of weather, whereby longer spells of unbroken freezing weather are occurring in otherwise temperate winter conditions. Basic principles of cold weather concrete practice from countries such as Canada may be helpful in this regard, for example, as set out by Turenne and Sereda (1975).

Aggregate selection The annual weather conditions in a region are reasonably predictable in respect of the mean temperatures and occurrence of freeze-thaw cycles. This has allowed experience to be built up on the satisfactory history of use of selected aggregates from which to deem an aggregate acceptable for local use. Guidelines are available if more specific criteria are required in specification, for example, in an informative annex to European standard EN 12620 (CEN 2008b). Advice is available on the specification of suitable “F” or “MS” category aggregates for several combinations of climate and end use. For example, airfield pavements in a Continental climate require at least category F1 or MS18, whereas category F2 or MS25 would suffice in a Mediterranean climate. These requirements can form part of the specification, and the producer can demonstrate compliance through certified test results. In high-risk exposure conditions, it is possible that crushing to a finer size will mitigate risk by providing a shorter exit path for water on freezing, thereby reducing expansive pressure buildup. This strategy is not universally accepted; however, especially in the case of carbonate aggregates. In the case of airfield pavements, the risk of D-cracking can be reduced by avoiding carbonate rocks of sedimentary origin. The technique of crushing to a smaller-sized aggregate is not recommended with carbonate aggregates (Dryden and Chapin 2009).

Prescriptive deemed-to-satisfy approach The prescriptive approach tends to take one of two approaches. The first is based on air entrainment. The second involves specifying a low water/ cement ratio concrete without entrained air, for which there is prior evidence of freeze-thaw durability in the place of use. A potential difficulty

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with the prescriptive approach is that our experience to date has been based on Portland cement concretes. What little is known of the interaction between air-entraining agents and secondary cementitious materials is not entirely encouraging. Some specifications are based on a target mean air content, but it is more common to prescribe a minimum value. A maximum value is also advisable, whereby durability is assured without excessive loss of strength. This may be formulated as a minimum air content together with an upper limit on air content as the specified minimum value plus 4% absolute. It is important to avoid ambiguous specifications, which include too many criteria for the producer to comply with. Yet it is common to see specifications with both a minimum strength class and a minimum air content. Exposure classes and attendant advice on limiting compositions in codes and standards provide the framework for specifiers seeking to select durable concretes in freeze-thaw environments. The guidance attempts to take account of the degree of saturation and the presence, if any, of external sources of salt from de-icing agents and seawater. However, international practice varies greatly on the choice of exposure class, especially if there is lack of clear guidance on judging the degree of saturation. A sloped surface and a vertical surface may have very different saturation levels, even if located in the same “environment”. In European practice, the degrees of saturation are “moderate” and “high”. At first sight, the use of such terms for differentiating exposure classes may cause some confusion. Moderately saturated concrete should not be at risk during a freeze-thaw cycle, because the water would have room to expand without deleterious effect. Harrison (2000) explained that CEN intended the term “moderate saturation” to imply a moderate risk of damage. This better differentiates the moderate risk classes (XF1 and XF2) from the higher risk classes (XF3 and XF4). Hobbs et al. (1998) noted that the differentiation could be on the basis of a lower number of freeze-thaw cycles per annum or a lower risk of freeze-thaw when saturated. There is lack of international unanimity on a requirement for freeze-thaw resisting aggregates in the least-severe freeze-thaw exposure classes, and this underlines the differences in interpretation of the exposure classes benchmarked against each other from region to region. Guidance in the most onerous conditions may direct the specifier to use a high-strength concrete with a prescribed high value of minimum cement content, but some authors have questioned the need for this. Dhir et al. (2000), for example, failed to correlate cement content with freeze-thaw resistance in non-air-entrained Portland cement mixes. Regarding working life, Hobbs et al. (1998) argue that the limiting values for a 100-year life should be the same as for 50 years. This is due to the event-dependence of the process and the fact that concrete capable of resisting freeze-thaw events on an ongoing basis should be durable, irrespective of age. The foregoing remarks highlight the difficulties for specifiers in achieving a balance between low-risk sustainable durable concrete and no-risk

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conservative overspecification when prescribing freeze-thaw resistance. Unfortunately, the task is further complicated by the difficulty of predicting the changing pattern of extreme weather events, and a lack of satisfactory history of use of low-carbon concretes. It is highly probable that a conservative prescriptive approach will continue to dominate practice.

Performance-based approach A performance-based approach based on theoretical models of deterioration, from which to determine the required material performance parameters, is not yet a feature of practice. The pathway to this has been clearly set out by Fagerlund (2004), but advice exclusively based on the prescriptive approach has tended to endure in national codes and standards. A step in the direction of nudging specifiers towards performance-related methods is the equivalent durability procedure (EDP). An example of EDP from Finland, on freeze-thaw resistance, is set out in an informative annex to CEN Technical Report CEN/ TR 16563 (CEN 2013b). The principle of the EDP method is to compare the performance in tests of a candidate concrete with a reference concrete of known satisfactory history of use in similar exposure conditions. This approach endorses that recommended by Siebel (1999) in respect of the freeze-thaw environment. He argued that the measured values in scaling tests are not sufficiently calibrated with behaviour in service, and therefore only comparative testing is possible. Harrison (2000), who had extensive firsthand insight into the development of international standards, did not expect it to be a widely adopted route. This has been the case so far, due to the absence of infrastructure in most countries to support a nationally curated set of reference materials. The criteria for acceptance of a candidate concrete are performance equal to or exceeding that of a reference concrete, and no adverse indicators. The first step is definition of the performance value. A reference concrete would then be produced to a recipe with a local tradition of providing freeze-thaw resisting concrete. Reference values from testing at a particular age could then be established. The comparative performance of candidate concretes could be established through reference values or the testing of reference concretes of similar age. The example from Finland provides an insight into how the influence of specimen age on a test result can be considered. The performance requirement for the maximum allowable scaling of the candidate concrete is determined as follows: mcand =

kbinder, kbinder,

kbinder = 1

cand ref

mref

(0.020 sf + 0.008 bfs + 0.004 fa)

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where mcand = maximum allowable scaling of the candidate concrete in test (g/m2) mref = scaling of the reference concrete in test (g/m2) kbinder = ageing effect factor sf = proportion of silica fume in binder (%) bfs = proportion of blastfurnace slag in binder (%) fa = proportion of fly ash in binder (%). The informative annexes to CEN/TR 16563 also include examples from Germany, Italy, and Norway in the use of EDP for managing freeze-thaw resistance.

SUMMARY Freeze-thaw damage represents a significant proportion of concrete durability failure in countries where the annual number of freeze-thaw cycles is significant. The parts of structures most at risk are those which can be saturated at time of freezing. Freeze-thaw resistant aggregates and air entrainment are used to control the risk of durability failure. Testing of aggregates to ensure suitability in cold climates is well-supported by a variety of standard test methods. Standard tests on fresh concrete to verify the presence of the correct volume of air entrainment do not include a check that the distribution of the air is appropriate. A Danish method has been devised to overcome this shortcoming. A suite of test methods is available on the performance of hardened concrete in freeze-thaw conditions, from which to assess the predicted performance of proposed materials and mixes. Mathematical modelling of freeze-thaw behaviour in concrete is based on the concept of a critical degree of saturation allied to the critical thickness of a protected zone around an air void. The maximum distance that water can travel to reach an air void without building up unacceptable expansive pressures in the concrete is defined as a critical distance. The critical distance therefore defines the required spacing of entrained air bubbles. However, the use of these models in performance-based specifications has yet to be exploited. On the other hand, comparative tests for freeze-thaw performance are leading the way for the introduction of the performance-related equivalent durability procedure (EDP). Nevertheless, the prescriptive approach to durability, using limiting values for composition and minimum levels of air entrainment, is expected to dominate practice for some time. Regional guidance based on local experience and satisfactory history of use of materials and mixes is invaluable in most situations and is especially important in the case of managing the risk of serviceability failure in freeze-thaw environments.

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Research is required on the freeze-thaw performance of concretes with medium to high levels of secondary cementitious materials, recycled aggregate concrete, and the interaction of air-entraining admixtures with additions. A note of caution has been sounded on the validity of several harsh regime test methods ─ designed when Portland cement concretes were the norm ─ for use with low-carbon concretes. Further research is required to advance the sustainability agenda of concrete technology in these areas.

REFERENCES ASTM. 2006. C131, Standard test method for resistance to degradation of smallsize coarse aggregate by abrasion and impact in the Los Angeles machine. West Conshohocken: ASTM International. ASTM. 2009a. C231, Standard test method for air content of freshly mixed concrete by the pressure method. West Conshohocken: ASTM International. ASTM. 2009b. C457, Standard test method for microscopical determination of parameters of the air-void system in hardened concrete. West Conshohocken: ASTM International. ASTM. 2015a. C127, Standard test method for relative density (specific gravity) and absorption of coarse aggregate. West Conshohocken: ASTM International. ASTM. 2015b. C666, Standard test method for resistance of concrete to rapid freezing and thawing. West Conshohocken: ASTM International. ASTM. 2016a. C173, Standard test method for air content of freshly mixed concrete by the volumetric method. West Conshohocken: ASTM International. ASTM. 2017a. C 138, Standard test method for density, unit weight, yield, and air content (gravimetric) of concrete. West Conshohocken: ASTM International. ASTM. 2017d. D6928, Standard test method for resistance of coarse aggregate to degradation by abrasion in the Micro-Deval apparatus. West Conshohocken: ASTM International. ASTM. 2018a. C33, Standard specification for concrete aggregates. West Conshohocken: ASTM International. ASTM. 2018b. C88, Standard test method for soundness of aggregates by use of sodium sulfate or magnesium sulfate. West Conshohocken: ASTM International. ASTM. 2019a. C295, Standard guide for petrographic examination of aggregates for concrete. West Conshohocken: ASTM International. Bunke, N. – editor. 1991. Prüfung von beton empfehlungen und hinweise als ergänzung zu DIN 1048 [Testing of concrete: recommendations and information as a supplement to DIN 1048]. Deutschen Ausschuss fur Stahlbeton. 422. Berlin: Beuth Verlag GmbH. CEN. 2006b. CEN/TR 15177. Testing the freeze-thaw resistance of concrete internal structural damage. Brussels: Comité Européen de Normalisation. CEN. 2007. EN 1367-1. Tests for thermal and weathering properties of aggregates – part 1: determination of resistance to freezing and thawing. Brussels: Comité Européen de Normalisation. CEN. 2008a. EN 1367-6. Tests for thermal and weathering properties of aggregates – part 6: determination of resistance to freezing and thawing in the presence of salt (NaCl). Brussels: Comité Européen de Normalisation.

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CEN. 2008b. EN 12620:2002 + A1:2008. Aggregates for concrete. Brussels: Comité Européen de Normalisation. CEN. 2009a. EN 1367-2. Tests for thermal and weathering properties of aggregates – part 2: magnesium sulfate test. Brussels: Comité Européen de Normalisation. CEN. 2011c. EN 1097-1. Tests for mechanical and physical properties of aggregates - part 1: determination of the resistance to wear (micro-Deval). Brussels: Comité Européen de Normalisation. CEN. 2013b. CEN/TR 16563. Principles of the equivalent durability procedure. Brussels: Comité Européen de Normalisation. CEN. 2016. TS 12390-9. Testing hardened concrete - part 9: freeze-thaw resistance with de-icing salts - scaling. Brussels: Comité Européen de Normalisation. CEN. 2020a. EN 1097-2. Tests for mechanical and physical properties of aggregates - part 2: methods for the determination of resistance to fragmentation. Brussels: Comité Européen de Normalisation. CEN. 2022a. EN 932-3. Tests for general properties of aggregates - part 3: procedure and terminology for simplified petrographic description. Brussels: Comité Européen de Normalisation. CEN. 2022b. EN 1097-6. Tests for mechanical and physical properties of aggregates - part 6: determination of particle density and water absorption. Brussels: Comité Européen de Normalisation. CEN. 2022c. EN 12350-7: 2019 + AC:2022. Testing fresh concrete - part 7: air content - pressure methods. Brussels: Comité Européen de Normalisation. Comite Euro-International du Beton. 1989. Durable concrete structures design guide. London: Thomas Telford. 112pp. Concrete Society. 1996. Developments in durability design and performance-based specification of concrete. Special Publication CS109. London: Concrete Society. CSA. 2004. CSA A23.2-24A, Test method for the resistance of unconfined coarse aggregate to freezing and thawing. Mississauga: Canadian Standards Association. Dhir, R., McCarthy, M., Limbachiya, M., Sayad, H. and D. Zhang. 1999. Pulverised fuel ash concrete: air entrainment and freeze/thaw durability. Magazine of Concrete Research 51, 1: 53–64. Dhir, R., Tittle, P. and M. McCarthy. 2000. Role of cement content in specifications for durability of concrete - a review. Concrete 34, 10: 68–76. Dryden, J. and L. Chapin. 2009. Evaluation of D-cracking preventive measures in Ohio test pavement. Transportation Research Record 2113, 1: 92–98. Fagerlund, G. 1993. The long time water absorption in the air-pore structure of concrete. Report TVBM. Vol.3051. Lund: Division of Building Materials. LTH. Lund University. Fagerlund, G. 1997. On the service life of concrete exposed to frost action. In Freeze/thaw Durability of Concrete, ed. J. Marchand, M. Pigeon and M. Setzer, 23–41. London: E & FN Spon. Fagerlund, G. 2004. A service life model for internal frost damage in concrete. Report TVBM. Vol.3119. Lund: Division of Building Materials. LTH. Lund University. Harrison, T. 2000. Resisting freeze/thaw attack. Concrete 34, 6: 46–47. Harrison, T., Dewar, J. and B. Brown. 2001. Freeze-thaw resisting concrete – its achievement in the UK. CIRIA Report C559. London: CIRIA.

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Harrisson, A. 1995. Deleterious processes in concrete. Concrete 29, 6: 13–14. Hobbs, D., Marsh, B. and J. Matthews. 1998. Minimum requirements for concrete to resist freeze/thaw attack. Minimum Requirements for Durable Concrete, ed.D. Hobbs, 91–129. Crowthorne: British Cement Association. Hooton, D. and D. Vassilev. 2016. Evaluation of modifications to the ASTM C672 deicer salt scaling test for concrete containing slag cement. Advances in Civil Engineering Materials 5, 2: doi: 10.1520/ACEM20160034 Jacobsen, U., Pade, C., Thaulow, N., Brown, D., Sahu, S., Magnusson, O., De Buck, S. and G. De Schutter. 2006. Automated air void analysis of hardened concrete - a Round Robin study. Cement and Concrete Research 36, 8: 1444–1452. Janssen, D. and M. Snyder. 1994. Resistance of concrete to freezing and thawing. Strategic Highway Research Programme SHRP-C-391. Washington: National Research Council. Lang, A., Range, P., Fowler, D. and J. Allen. 2007. Prediction of coarse aggregate performance by micro-Deval and other soundness, strength, and intrinsic particle property tests. Transport Research Record 2026, 1: 3–8. Liu, Z., Chin, C. and J. Xia. 2022. Novel method for enhancing freeze-thaw resistance of recycled coarse aggregate concrete via two-stage introduction of denitrifying bacteria. Journal of Cleaner Production 346. doi: 10.1016/ j.jclepro.2022.131159. Marks, V. and W. Dubberke. 1982. Durability of concrete and the Iowa Pore Index Test. Transportation Research Record 853: 25–30. MOHURD - Ministry of Housing and Urban-Rural Development. 2010. GB/T 50082-2009, Standard for test methods of long-term performance and durability of ordinary concrete. Beijing: General Administration of Quality Supervision, Inspection and Quarantine of the People’s Republic of China (AQSIQ). Neville, A. 1995. Properties of Concrete. 4th Edition. Harlow: Longman. Pigeon, M. and R. Pleau. 1995. Durability of concrete in cold climates. London: E&FN Spon. Pink, A. 1978. Winter concreting. Crowthorne. British Cement Association. Powers, T. 1949. The air requirement of frost-resistant concrete. In Proceedings, Highway Research Board 29: 184–211. Powers, T. 1975. Freezing effects in concrete. ACI Special Publication SP-47. Detroit: American Concrete Institute. Price, W. 1996. Measuring air voids in fresh concrete. Concrete 30, 4: 29–31. RILEM TC 117-FDC. 1995. Draft recommendations for test methods for the freezethaw resistance of concrete - slab test and cube test. Materials and Structures 28, 180: 366–371. RILEM TC 117-FDC. 1996. TC 117-FDC Recommendation - CDF test - test method for the freeze thaw and deicing resistance of concrete - tests with sodium chloride (CDF). Materials and Structures 29, 193: 523–528. RILEM TC 117-FDC. 1997. TC 117-FDC: Freeze-thaw and deicing resistance of concrete. Materials and Structures 30, 196: S3–S6. Rodrigues, F., Evangelista, L. and J. de Brito. 2013. A new method to determine the density and water absorption of fine recycled aggregates. Materials Research 16, 5: 1045–1051.

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Setzer, M. 2013. Fundamental aspects of frost damage in hardened cement paste. Restoration of Buildings and Monuments 19, 6: 433–442. Shahin, M. 2009. Concrete surfaced airfields. PaverTM distress identification manual. Champaign: U.S. Army Corps of Engineers. Engineering Research and Development Centre. Siebel, E. 1999. Performance testing freeze-thaw Resistance. CEN TC104/DuraNet Workshop. Design of Durability of Concrete. Berlin. 46–48. Swedish Institute for Standards. 2005. SS 13 72 44, Concrete testing - hardened concrete- scaling at freezing. Stockholm: Swedish Institute for Standards. Thomas, M. and K. Folliard. 2007. Concrete aggregates and the durability of concrete. In Durability of concrete and cement composites, ed. C. Page and M. Page, 247–281. Cambridge: Woodhead Publishing Limited. Turenne, R. and P. Sereda. 1975. Winter concreting – Canadian practice. Division of Building Research Paper 848. Ottawa: National Research Council Canada. Van Dam, T., Sutter, L., Smith, K., Wade, M. and K. Peterson. 2002. FHWA-RD01-163, Guidelines for detection. analysis. and treatment of materials-related distress in concrete pavements. Volume 1. Final Report. Washington: Federal Highway Administration. Wright, P. 1953. Entrained air in concrete. Proceedings, Institution of Civil Engineers 2, 3: 337–358.

Chapter 9

Chemical attack Sulfates

INTRODUCTION Deterioration of concrete may arise through contact with sulfate salts. These salts are chemical compounds formed by reactions between negatively charged anions (SO42−), with metal positively charged cations, such as Ca2+, Mg2+, and Na+ which dissociate in water into their constituent ions. The deleterious effects of these sulfate ions are manifest in a number of ways, most commonly through disruptive expansion of the concrete. In certain circumstances, calcium silicate hydrates may also be attacked, catastrophically weakening the material. Three phenomena can be differentiated through chemical or chronological characteristics: • conventional ettringite sulfate attack • thaumasite sulfate attack • delayed ettringite formation. Another sulfate action is seawater attack of concrete, which occurs with acids. It is discussed in Chapter 10. Identification of the threat is primarily based on a chemical analysis of the environment, especially the concentration of sulfates in substances that the concrete will be exposed to. Resistance is typically derived from the chemistry of the binder, allied to a high degree of impermeability through low water/cement ratio concretes.

CONVENTIONAL ETTRINGITE FORM OF SULFATE ATTACK The conventional, classically researched, form of sulfate attack is associated with buried concrete, such as foundations exposed to soils or to groundwater containing soluble sulfates. Deleterious conditions may also arise in sewers and tunnel linings. Significant durability problems occur if the solution is above a critical concentration level. Durability failure occurs through a DOI: 10.1201/9781003261414-9

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combination of expansive disruption and deterioration of the cement paste. Tricalcium aluminate hydrate and calcium hydroxide react with sulfate ions in solution to form solid products that occupy a larger volume than the source constituents, leading to disruptive expansion. The disruptive expansion can be accompanied by strength loss consequent on the chemical deterioration of cement paste and damage to the aggregate interface bond. Concrete may crack parallel to the surface or become friable. The principal compounds formed are ettringite and gypsum. The source of the sulfate ion is typically sodium sulfate, calcium sulfate, magnesium sulfate, or potassium sulfate. Ammonium sulfate may also be encountered. The reaction occurs between the salts in solution and the products of hydrated cement. Salts in solution are ionised, that is, dissociated into their constituent ions. For example, potassium sulfate dissociates as follows, leading to magnesium sulfate attack: K2 SO4 = 2K+ + SO42

These sulfates may be naturally occurring in soils, particularly clays, and groundwaters. Industrial processes, including agricultural, may also introduce an interface between concrete and sulfates. Sulfate attack may occur if the aggregate contains accessible gypsum. The Building Research Establishment (BRE 1996) noted that disturbance of clays bearing pyrite (iron sulfide, FeS2) may also raise the sulfate and acid concentrations in the groundwater through oxidation: 14FeS2 (s) + 14Fe3+ + 8H2 O = 2SO42 + 15Fe 2+ + 16H+

The sulfate ion concentration depends on the solubility of the salts. Sodium and magnesium sulfates are highly soluble, while calcium sulfate is not. Aggregates with sulfur contents and potential sulfate contents by mass of cement of 1.3% and 4.0%, respectively, were reported as being unlikely to give rise to abnormal expansion when subject to moisture exposure (Hobbs 2000). Calcium hydroxide and calcium aluminate hydrates are the most vulnerable products of hydration. Calcium silicate hydrates may also be affected if the calcium hydroxide becomes depleted. Sulfate ions react with calcium hydroxide to form gypsum (calcium sulfate, CaSO4.2H2O), while the reaction with the calcium aluminate hydrates forms calcium sulfoaluminate hydrates. Calcium sulfoaluminate hydrates are formed at the hydration stage and are found as ettringite (3CaO.Al2O3.3CaSO4.32H2O) and monosulfates (3CaO.Al2O3.CaSO4.12H2O). The ettringite formed may not be pure and its phase may be what is referred to as “AFt”, or “alumino-ferrite-tri”. The formation of the monosulfate occurs through reaction with the ettringite if

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more tricalcium aluminate is available than sulfate. The monosulfate is referred to as “AFm”, or “alumino-ferrite-mono”. If external sulfates are introduced, any unreacted tricalcium aluminates may react to form ettringite, while AFm phases may form AFt. The formation of ettringite may be described by the reaction: 3CaO. Al2 O3 + 3CaSO4 (aq) = 3CaO. Al2 O3 .3CaSO4 .32H2 O

The formation of monosulfate may be described by the reaction: 3CaO . Al2 O3 .3CaSO4 .32H2 O + 2(3CaO. Al2 O3) = 3(3CaO. Al2 O3 . CaSO4 .12H2 O)

The deterioration mechanism is generally considered to involve the generation of disruptive forces due, first, to the fact that the reaction products occupy a greater volume than the components causing the reaction, and second, due to the possible adsorption of water (Neville 1995). The reaction-product volume increase can be by a factor of two to five, with the higher values being associated with the effect of ettringite formation. However, Skalny and Pierce (1999) point out that the mechanism is a complex sequence of physical and chemical processes. The proportion of ettringite formed may not relate to the amount of expansion. Ettringite and gypsum crystal growth have been observed in cracks of sulfate-damaged concrete. It is not clear whether pressure from the crystal growth causes the cracks or whether the cracks are sites for the deposition of ettringite and gypsum from solution. The ettringite first formed may be in a colloidal form and subsequently expands when it imbibes water. Other hypotheses have also been proposed. The reaction with sulfate ions may be one of equilibrium or may go to completion. For example, the reaction with sodium sulfate could reach equilibrium when only a portion of the calcium hydroxide has changed to calcium sulfate. In simplified terms: Ca (OH )2 + Na2 SO4

CaSO4 .2H2 O + 2NaOH

The intense attack that occurs when a reaction can go to completion is illustrated by the effect of magnesium sulfate. In simplified terms: Ca (OH )2 + MgSO4

CaSO4 .2H2 O + Mg (OH )2

Lea (1998) provides an example of ettringite formation: 2(3CaO. Al2 O3 .12H2 O) + 3(Na2 SO4 .10H2 O) 3CaO. Al2 O3 .3CaSO4 .31H2 O + 2Al (OH )3 + 6NaOH + 17H2 O

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Sodium sulfate (Glauber’s salt, Na2SO4.10H2O) reacts with calcium hydroxide and with calcium aluminate hydrate to form gypsum and ettringite, respectively. The calcium silicate hydrates may also be attacked. Ammonium sulfate, (NH4)2SO4, can react to form gypsum. Calcium sulfate (gypsum, CaSO4.2H2O) reacts with calcium aluminate hydrate to form ettringite. The solubility of calcium sulfate is considerably lower than the solubility of other sulfates and so it can be relatively harmless in soils. The sulfate reaction may, however, be initiated by calcium sulfate formed from reaction with calcium hydroxide. Magnesium sulfate (Epsom salt, MgSO4) reacts with calcium hydroxide, calcium aluminate hydrates, and calcium silicate hydrates to form gypsum, ettringite, brucite (magnesium hydroxide, Mg(OH)2), and magnesium silicates. It is an intense attack because the brucite is insoluble in water and, therefore, the reaction may go to completion. Hobbs (2001) notes that the attack on the calcium silicate hydrates may be intensified when brucite is formed due to lowering of the pore solution pH.

THAUMASITE SULFATE ATTACK Thaumasite is a complex mineral that can render concrete soft, weak, or mushy. It is likely that modern diagnostic tools have allowed differentiation of thaumasite from other forms of sulfate in mixed crystals, providing a new insight into a pre-existing phenomenon, first identified by Erlin and Stark (1965). The loss of strength is due to attack of the calcium silicate hydrates. This differentiates it from the conventional sulfate attack, which targets the calcium aluminate hydrates. Thaumasite formation results from a reaction between sulfates, calcium silicates in the cement, and calcium carbonate (Stark 2003). The carbonate may originate in limestone aggregates, limestone filler in cement, or groundwater percolation. The sulfates typically derive from external sources. Alternatively, it may form from a reaction involving ettringite, calcium silicate hydrates, and calcium carbonate. If this is the case, ettringite formation is a necessary precursor, implying that normal sulfate attack may be a prerequisite. Initial reactive alumina is another necessary condition for the reaction. Involvement of the calcium silicate hydrates in the reaction leads to loss of strength and cohesion in the concrete, leading to disintegration. Thaumasite has been described by Hartshorn and Sims (1998) as a complex sulfate-bearing mineral with the composition CaSiO3.CaCO3.CaSO4.15H2O. The material has a similar crystal structure to ettringite, but thaumasite is a silica-bearing phase as opposed to an aluminate. The composition has been described in an expert group report (DETR 1999) as a calcium silicate sulfate carbonate hydrate, Ca6(Si(OH)6)2(SO4)2(CO3)2.24H2O. Thaumasite formation appears to occur more readily in the presence of magnesium sulfate than sodium sulfate (Hartshorn et al. 1999), but is less

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likely to form in calcium sulfate solutions. Samples of deteriorated concrete may clearly show white haloes of thaumasite around the affected aggregate particles (Wallace 1999). Damage to ten motorway bridges in an area of west England, UK heightened awareness of this threat to durability. Hobbs and Taylor (2000) postulated that in certain instances, oxidation of pyrites in excavated clay backfill led to a pH reduction in the groundwater through formation of sulfuric acid. The sulfuric acid could become depleted by neutralisation within the backfill, by contact with concrete, and by washout. However, the acid’s sulfate ions may remain in solution in the clay’s interstices unless sufficient carbonate is present to both neutralise the acid and precipitate its sulfate: MgCO3 (s) + H2 SO4 (aq) + H2 O

MgSO4 (s) + 2H2 O + CO2 (g)

The groundwater sulfate level could rise, depending on the sulfates formed, leading to gypsum formation. It is possible that further reactions could include reaction of gypsum with calcium silicate hydrates, and calcium carbonate to form thaumasite. Most of the thaumasite cases in the UK’s motorway bridges involved the presence of pyrite.

DELAYED ETTRINGITE FORMATION Delayed ettringite formation is a form of internal sulfate attack that may occur at an advanced age in particular concretes. Although durability failure may occur through expansive pressures, as in conventional sulfate attack, the differentiating factor is that the sulfate source is exclusively internal. Portland cement contains both internal sulfates and gypsum, which is added to influence setting and early strength characteristics. Without the gypsum, the reaction between tricalcium aluminate and water would lead to a flash set. Ettringite, an expansive compound involving tricalcium aluminate and gypsum, can be harmlessly formed at the hydration stage while the material is plastic and can accommodate the resultant strains. However, in particular conditions, this internal sulfate may cause a phenomenon at the microscale later in-service, when the hardened concrete cannot withstand the internal expansion. Problems arise if ettringite is not allowed to develop at the plastic stage due to high temperature curing or conditions during hydration where the temperature is above 70°C for a prolonged period. Subsequent wet conditions in service may encourage ettringite formation in mature concrete with consequent expansion of the cement paste and cracking. Three conditions are required (Taylor et al. 2001). These are chemistry, which dictates the volume of ettringite that can be formed; paste microstructure,

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which dictates the degree of accommodation and expansive stresses; and concrete microstructure, which dictates the ability of the material to contain the stresses without durability failure. Ettringite is formed during hydration in plastic concrete except in high temperature situations. It is unstable at temperatures in excess of 70°C. Such high temperatures can be experienced during the curing period of large pours from the heat of hydration or through steam curing of pre-cast units. In these cases, ettringite is not formed. The sulfate is instead absorbed by the calcium silicate hydrate, forms small amounts of monosulfate, and some may go into solution in the pore fluid. The alumina content of the calcium silicate hydrates also becomes elevated. These phases are not stable at ambient temperatures. Thus, when the concrete cools in-service, ettringite crystals can begin to form in the paste structure. This occurs only in wet concretes. Expansion of the cement paste may result, followed by associated cracking. Hobbs (1999b) reports that the cracks are often of uniform width and proportional in width to the aggregate particle that they surround. The cracks may fill with ettringite.

FACTORS INFLUENCING SULFATE ATTACK Long-term tests by the United States Department of Agriculture, United States Bureau of Reclamation, the United Kingdom Building Research Station/Building Research Establishment, and the British Cement Association have contributed significantly to advances in knowledge and experience on resistance to sulfate attack. The US work pioneered the interest in low tricalcium aluminate sulfate resisting cements, while the UK research helped to broaden the range of factors taken into account in design and specification. A 25-year exposure trial of concrete specimens buried below the water table of a sulfate-bearing soil in west London was completed in the 1990s. The average SO42- content of the groundwater was about 3100 mg/L. Results were influential in the drafting of Digest No. 363 (BRE 1996) and Special Digest No. 1 (BRE 2005). Detailed guidance on the influence of hardcore containing sulfates on concrete floors was subsequently published (BRE 2008).

Influences on conventional form of sulfate attack Research has identified the following factors that have an influence on the intensity of sulfate attack: • • • •

sulfate concentration solubility of sulfates groundwater mobility chemical composition of the concrete

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Figure 9.1 Influence of groundwater sulfate concentration on relative resistance requirements.

• • • •

concrete permeability wetting and drying cycles evaporation degree of carbonation prior to exposure.

A threshold value of sulfate concentration is required to initiate attack. Attack intensity increases with increasing concentration up to a certain level above which the rate of increase diminishes. This may be illustrated by the relative minimum cement content requirement that was applied in the UK during the 1990s for resistance to sulfate in groundwater (Fig. 9.1). The solubility of sulfate salts in groundwater contributes to the critical sulfate concentration. Sulfate concentration in-service, however, may on occasion differ from that determined in the tests to classify the site. This is because the method of determination, by acid extraction, yields a value for the total sulfates. Thus the method does not distinguish between relatively innocuous calcium sulfate, for example, and the highly aggressive magnesium sulfate. Water-soluble sulfates can provide a better guide to the risk of attack on a particular site. This is reflected in UK practice that includes a water/soil extract test. Groundwater mobility influences the extent of sulfate supply in a reaction. Static groundwater conditions are therefore less problematic than those in which groundwater movement leads to refreshment of sulfate supply. Equally, a site experiencing groundwater movement could be classified as low in sulfates at one time, but experience elevated sulfate levels at a later stage. European Standard EN 206 differentiates static and mobile groundwater through a threshold soil permeability value of 10–5 m/s. The composition of the concrete influences the reaction. Control is exercised on the sulfates introduced at time of mixing. Sea-dredged aggregates are a concern. Testing, for example, to the methodology of

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ISO 24684 (ISO 2023) allows determination of the soluble sulfates. Concrete with a high content of calcium aluminate hydrate is highly vulnerable. The ingress of sulfate ions depends on the quality of the concrete. Resistance to sulfate attack is critically dependent on reducing the rate of sulfate ion ingress. Poorly compacted concretes are not sufficiently impermeable to provide protection. Highly impermeable concrete is especially required where the surfaces are in contact with sulfates in solution under the influence of a hydrostatic head. It is unsurprising that sulfate resisting concretes are generally characterised by low water/cement ratios. Concrete in sulfate conditions subject to water table level fluctuation may experience higher rates of deterioration if wetting and drying cycles are experienced. Wetting and drying promote greater sulfate ingress than conditions in which the concrete is constantly saturated. Sulfate attack is unlikely to be significant if the water table remains predominantly below the underside level of the concrete during the service life. The rate of attack is higher in members where moisture is lost through evaporation. Thick members are therefore more robust than thin ones. Elements with exposure to sulfates on all sides will deteriorate less than those which allow moisture loss from one or more faces. Ground floor slabs can allow evaporation from the top surface encouraging further ingress of sulfates from the fill below. Carbonated concrete is not subject to the same intensity of sulfate attack as non-carbonated concrete. The reason for this is that carbonation of the surface layers reduces the calcium hydroxide content through conversion to calcium carbonate. The insoluble calcium carbonate does not react with the sulfates. Thus, if concrete is carbonated prior to sulfate exposure, the rate of attack can be lowered. This may be of significance in practice where pre-cast concrete units or concrete blocks are used in contact with sulfate-bearing soils or groundwater. Similarly, exposure of insitu concrete to the air for a period after moist curing but before contact with sulfates should be beneficial.

Influences on thaumasite sulfate attack The primary factors that must simultaneously be present for thaumasite sulfate attack are as follows (Hartshorn and Sims 1998; DETR 1999; BRE 2008): • • • • •

sulfates and/or sulfides in the ground mobile groundwater source of carbonate pH 10 or above low temperature.

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The sulfates may pre-exist in the soil, but thaumasite may also occur through the oxidation of pyrite. An abundant supply of water is required. Cold conditions are required, with temperatures at least below 15°C. An alumina content is required, even if only at a low level. Indeed, alumina contents encourage ettringite formation, whereas lower amounts may facilitate reaction with the carbonate and calcium silicates.

Influences on delayed ettringite formation The potential occurrence of delayed ettringite formation is predicated on the temperature during hydration and the availability of moisture in service. The principal influences on delayed ettringite formation are as follows: • • • •

air temperature during hydration size and geometry of pour cement content cement chemistry and fineness.

Delayed ettringite formation will not occur if the concrete can be kept below a temperature of 70°C during hydration. Air temperature combined with heat of hydration will obviously contribute to the temperature value, and hot-weather concreting raises the risk of reaching the threshold temperature limit. Steam-curing introduces particular risks in this regard. Large concrete pours are at risk if the geometry of the formed element is such that the heat loss during hydration is less than the rate of rise. The effect of heat of hydration in these conditions is such that the temperature within the element may be considerable, and thereby increase the risk. This may conflict with one of the approaches to the control of early thermal shrinkage cracking that involves temporary insulation. This is sometimes referred to as the “if you cannot keep it cool, keep it hot” approach. Concern about crack control must be balanced against the risk of inducing delayed ettringite formation. Cement content influences heat of hydration at a contributory rate of about 13°C for each 100 kilograms of cement per cubic metre of concrete. Thus, high cement content mixes in large pours and/or during hot-weather concreting operations may push the absolute concrete temperature above the critical threshold for delayed ettringite formation. The cement chemistry and fineness are influential, both in terms of heat rise during hydration and their effect on the microstructure of the paste. Blastfurnace slag and PFA have obvious benefits over Portland cements in this regard. Sulfate resisting cement also appears to have a lower risk of delayed ettringite formation in comparison with Portland cement, due to the lower tricalcium aluminate level. The effects of individual parameters and fineness were studied by Lawrence (1993) and Kelham (1996). Hobbs (1999b) reports that the following equation gave a reasonable fit:

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Exp = k1 (f . MgO) + k2 SO3 + k3 Na2 O + k4

where Exp = ultimate expansion (%) f = cement fineness (m2/kg) MgO = magnesium oxide (% by mass of cement) SO3 = total sulfate (% by mass of cement) Na2O = sodium oxide (% by mass of cement) k1, k2, k3, k4 = coefficients (0.00085, 0.30, 0.56, –1.4, respectively). However, not all cements low in the parameters identified are immune to delayed ettringite formation, nor are cements that are high in two of the compositional parameters necessarily susceptible. TESTS FOR SULFATE RESISTANCE The possible approaches available to assess the sulfate resistance of concrete include expansion, strength loss, mass loss, change in dynamic modulus, and change in appearance. Testing of the sulfate resistance of concretes has traditionally been based on monitoring the performance of the cement paste fraction through mortar prisms stored in sulfate solutions. Concrete cylinders were specified in a test method developed by the United States Bureau of Reclamation, but even the time required to complete accelerated tests on mortar bars is long. Expansion measurement is thought to offer the most reliable basis of assessment. Testing the change in appearance can be rather subjective, and the other tests are either destructive and necessitate a high number of samples or can become problematic as the specimens deteriorate. Standard tests for expansion include the American standards ASTM C452 (ASTM 2021a) and ASTM C1012 immersion test (ASTM 2018c); Canadian standard A3004-C8 (CSA 2018); French standard test NF-P-18-837 (AFNOR 1993); and Chinese standard GB/T 749-2008 (MOHURD 2008). Work is progressing to explore a unified European standard test method, but this would first require convergence on the agreed characteristics of European sulfate resisting cements. The ASTM C452 test, which is applicable only to Portland cements, involves constructing 25×25×285 mm mortar bars to a prescribed recipe with an elevated sulfate level. The mix is brought to a sulfur trioxide (SO3) level of 7% by mass through use of gypsum. Expansion is measured after 14 days immersion in water at 23oC. The test can yield early results but is not representative of materials in service. The ASTM C1012 test is applicable to a wide range of cements, such as blends of Portland cement and pozzolans or slags. Length change of mortar bars is measured after immersion in a 5% sulfate solution such as Na2SO4. The test period is

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lengthy despite the fact that this is an accelerated test. Readings are taken at intervals and continue for up to 12 or 18 months. Many other international tests are based on this methodology. The Chinese test, GB/T 749-2008, expresses the deterioration as a “corrosion coefficient”. The coefficient is the ratio of the flexural strength of a specimen following a period in sulfate solution compared to a control specimen stored in water. Many approaches have been researched for accelerating the effect, building on the work of Koch and Steinegger (1960) with small specimens of 10×10×60 mm and that of Wittekindt (1960), whose methodology using a flat prism specimen of 10×40×160 mm has endured. Ferraris et al. (2005) experimented with small 10×10×40 mm specimens, using the same solution as ASTM C1012, monitored over a minimum of 14 days or until deterioration commenced. The search for a definitive test continues. For example, use of tensile strength loss as an indicator has been found to be promising by Haufe and Vollpracht (2019). Aguayo et al. (2020) vacuum soaked the bars in the solution to increase the rate of sulfate penetration. These tests, and many other variations, provide early results and are very useful research tools. However, there is a desire to develop a suitable test for use in performance-based specifications. Comprehensive reviews of available test methods on pastes, mortars, and concretes were reviewed by CEN Technical Committee TC51: WG12, which published its findings in CEN/TR 15697 (CEN 2008c), and by Jabbour et al. (2022). Both reports contain recommendations on the desired parameters that would guide design of a future appropriate test method. They seek a methodology that could discriminate between sulfate resisting and non-sulfate-resisting cements while being more representative of in-service conditions than many existing methods. The option to accelerate the test is also fully explored. Following a detailed study, the CEN Committee recommended that flexural and compressive tests be conducted on 40×40×160 mm mortar prisms solution after 1 and 2 years of immersion in sodium sulfate. Curing should be 28 days in limewater. A maximum sulfate solution of 10 g/L was recommended, together with the option to test at the 3 g/L level. Temperature of test was considered to be a significant factor. A value of 10oC, or even 5oC if possible, was proposed. Irrespective of the method used, Taylor et al. (2001) caution that expansion tests where specimens with different mix parameters are contained in the same tank are unsound. Each specimen should be in a separate container.

MANAGEMENT OF THE DURABILITY THREAT FROM SULFATES Prescription for avoidance of deterioration is used to manage the threat. Time-dependent models with international consensus are not available. The

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approaches to specification and design practice for sulfate resistant concrete elements involve consideration of the severity of the environment, the permeability of the concrete, the control of sulfates in the mix, and the choice of cement type. These measures are prescriptive and primarily relate to control of conventional sulfate attack. However, comparator performance-based tests are under development.

Severity of the environment Specification requirements and design practice are based on an assessment of the risk to durability as reflected in the severity of the sulfate environment. The severity of the environment may be determined by reference to the concentration of sulfate (SO42−) or magnesium (Mg2+) in water or soil. Analysis reports may record the SO3 concentration rather than SO4 but values can be converted based on the SO4/SO3 ratio of 1.2. The classification for a particular situation is based on the chemical characteristics of the groundwater, soil, or seawater. Prescriptive limiting values are presented in guidance documents based on an assumption of water/soil temperature between 5°C and 25°C and a water velocity slow enough to approximate to static conditions. A special study of the environment would be required for more severe conditions. This may lead to the specification of physical protection to the concrete if other options based on the composition and chemistry of materials are ruled out. The analysis of groundwater may include testing for sulfate ion (SO42−) free ammonium ion (to determine NH4+), magnesium ion (Mg2+), and aggressive carbon dioxide concentration (mg/L), in addition to pH value. Groundwater values expressed in milligrammes SO4 per litre are used to classify the severity of environment in European practice. The range of threat from least to most aggressive is 200 to 6000 mg/L. A special study may be needed to establish the relevant exposure condition where the limits are outside the range. Values in grammes per litre are traditionally quoted in some countries while values in parts per million may also be found, for example, 1.5 g/L and 1500 ppm, respectively. The severity of concentrations in soils may be expressed in milligrammes per kilogramme. The European exposure classes range from 2000 to 24000 mg/kg. Less onerous conditions in respect of concrete composition may be permissible in clay soils with permeability below 10–5 m/s. Values in milligrammes per litre or in percentage terms may also be encountered, with the latter in a range typically from 0.1 to 0.2%. Analysis of soil may require determination of sulfate ion (SO42−) concentration (mg/kg) and soil acidity (ml/kg). The reference test method for sulfate concentration is the acid extraction method in EN 196-2 (CEN 2013a), but a water extraction method is permitted if experience is available in the place of use. This is particularly significant for UK practice.

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For example, the Concrete Society (1996) notes that the acid extraction test would classify London clay above the upper limit of the most severe European exposure class. This overestimates the aggressiveness of the environment, because the sulfates present are mainly calcium sulfate. A 2:1 water:soil extract test, British Standard BS 1377-3 (BSI 2021b), is in common use in the UK which yields values requiring different interpretation limits. It is understood (Mure 2000) that the 2:1 test was specifically designed for soils frequently encountered in UK clays, where calcium sulfate is the main constituent. Detailed guidance on the appropriate approach to classification of chemical classes and concrete composition requirements for use in UK practice is presented in a Building Research Establishment (BRE 2005) guide that takes account of recent research.

Concrete permeability An essential aspect of producing durable concrete in sulfate-rich environments is the achievement of low permeability concrete to hinder the ingress of sulfates and thereby limit the attack to surface layers. Concretes with water/ cement ratios below 0.55 are required, and maximum values of 0.45 are prudent in the severest conditions. The role of cement in minimising the water/ cement ratio is more significant in this context than the absolute value of the cement content. This is because the constituents of the hydrated cement are an integral part of the fuel for the reaction, and increasing cement content eventually leads to diminishing returns.

Cement type Considerable attention is paid to the use of a type of cement which yields optimum performance in the aggressive conditions created by sulfates. These include sulfate-resisting Portland cement and blended cement mixes containing blastfurnace slag or pozzalana, such as pulverised fuel ash (PFA). Sulfate-resisting cements are characterised by limits on the oxide and compound composition in respect of the SO3 content and C3A content. The SO3 limits in Europe range from 2.3–4.0% and the C3A limits range from 3–5%. In the US, cements with tricalcium aluminate contents of less than 8% are considered to be moderately resistant, while those with contents less than 5% are deemed highly resistant. The difference in practice is apparent rather than real – the methods of calculating the tricalcium aluminate content being dissimilar. Sulfate-resisting cements eliminate the risk of damage from formation of ettringite. The concrete may, however, still be vulnerable to attack on the calcium hydroxide content and potentially on the calcium silicate hydrates. The use of blastfurnace slag cements has been found to be beneficial at high slag contents. Blastfurnace slag content concretes are not in themselves

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chemically resistant to sulfates, but may be sufficiently impermeable to provide an effective defence. A minimum slag content of about 65% by mass of cement is required. Low slag levels decrease the sulfate resistance. O’Connell et al. (2012) reported that the addition of 50–70% of ground granulated blastfurnace slag resulted in considerable reductions in sulfate induced expansion and even a slight improvement in performance relative to sulfate-resisting Portland cement binders. Hobbs and Matthews (1998b) drew attention to the possible difference in the sulfate resistance of factory-produced slag cements from blends produced at the mixer. The former may produce more resistant concrete. This is due to the enhanced SO3 level typical of factory-produced cement compared with a blend. Research indicates that increased SO3 levels enhance the sulfate resistance of slag cements. In the case of blends produced at the mixer, the tricalcium aluminate content of the clinker and the alumina content (Al2O3) of the slag are particularly relevant. Slags of high alumina content are less resistant, particularly if combined with high C3A content Portland cements. Therefore, in the case of Portland cement and slag combinations, it is necessary to limit the tricalcium aluminate content of the Portland cement if the alumina content of the slag exceeds a certain value. For example, a limit of 10% C3A is imposed in some countries if the alumina content of the slag exceeds 14%. Another approach is to limit the potential production of calcium hydroxide through appropriate pozzolanic mixes. This can be achieved using a natural pozzolan, silicious fly ash, or a combination of both at clinker substitution levels of 20–55%. In addition to the chemical restrictions on the clinker, the CEM IV/A-SR has an allowable maximum clinker content 10% less than that of CEM IV/A. The cement must also pass the pozzolanicity test EN 196-5 (CEN 2011a) at eight days. The concrete itself needs to be mature before exposure to the aggressive medium. Table 9.1 provides examples from global practice of specifically manufactured sulfate resisting cements. The main constituents are drawn from combinations of clinker (K), natural pozzolana (P), slag (S), and siliceous fly ash (V). Cements to EN197-1 and ASTM C150 standards include allowance for minor additional constituents to a maximum of 5%. Satisfactory performance has also been reported with other material combinations, including burnt shale. Ryle (1999) quotes results of a laboratory study that showed good sulfate resistance in mortar prisms containing 15% metakaolin. Hobbs and Matthews (1998b) cite Sellevold and Nilsen in reporting that similar levels of silica fume have demonstrated good performance in Norway. Table 9.2 illustrates this divergence of practice in Europe, where many cements that do not fully meet the EN 197-1 conditions for “SR” are nevertheless considered by national standards to be valid in the place of use in sulfate exposure conditions.

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Table 9.1 Examples of specifically manufactured sulfate resisting cements Classification Sulfate resisting Portland cement

Sulfate resisting Portland-Slag cement

Sulfate resisting PortlandPozzolana (Fly Ash) cement

Sulfate resisting PortlandComposite cement

Sulfate resisting blast furnace cement

Sulfate resisting pozzolanic cement

Indicative requirements for sulfate resisting cement production in European and American practice EN 197-1: CEM I – SR0, 95–100% K with C3A = 0 and SO3 ≤ 3% EN 197-1: CEM I – SR3, 95–100% K with C3A ≤ 3% and SO3 ≤ 3% EN 197-1: CEM I – SR5, 95–100% K with C3A ≤ 5% and SO3 ≤ 3.5% ASTM C150: Type II, 95–100% K with C3A < 8% and SO3 ≤ 3% or maximum 0.020% expansion at 14 days in ASTM C1038 test when moderate sulfate resistance required ASTM C150: Type V, 95–100% K with C3A < 5% and SO3 ≤ 2.3% or maximum 0.040% expansion at 14 days in ASTM C452 test when high sulfate resistance required ASTM C595 IS (MS), SO3 ≤ 4.0% if S ≥ 70%, SO3 ≤ 3.0% if S < 70%, maximum 0.10% expansion at 180 days in ASTM C1012 test ASTM C595 IS (HS), SO3 ≤ 4.0% if S ≥ 70%, SO3 ≤ 3.0% if S < 70%, maximum 0.050% expansion at 180 days and 0.10% at 1 year in ASTM C1012 test ASTM C595 IP (MS), Class F fly ash (siliceous, low calcium content: 5–10%), SO3 ≤ 4.0%, maximum 0.10% expansion at 180 days in ASTM C1012 test ASTM C595 IP (HS), Class F fly ash (siliceous, low calcium content: 5–10%), SO3 ≤ 4.0%, maximum 0.050% expansion at 180 days and 0.10% at 1 year in ASTM C1012 test ASTM C595 IT (MS), L < 5%, SO3 ≤ 4.0% if S ≥ 70%, SO3 ≤ 3.0% if S < 70%, maximum 0.10% expansion at 180 days in ASTM C1012 test ASTM C595 IT (HS), L < 5%, SO3 ≤ 4.0% if S ≥ 70%, SO3 ≤ 3.0% if S < 70%, maximum 0.050% expansion at 180 days and 0.10% at 1 year in ASTM C1012 test EN 197-1: CEM III/B -SR, 20–34% K, 66–80% S EN 197-1: CEM III/C -SR, 5–19% K, 81–95% S A comparable cement is ASTM C595, IS(x) with high levels of ASTM C989 slag cement and limits on expansion at 180 days in ASTM C1012 test as indicated for ‘Portland-Slag cement’ above EN 197-1: CEM IV/A (P) - SR, 65–79% K with C3A ≤ 9% and 21–35% P, total SO3 ≤ 3.5% EN 197-1: CEM IV/A (V) - SR, 65–79% K with C3A ≤ 9% and 21–35% V, total SO3 ≤ 3.5% (Continued)

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Table 9.1 (Continued) Examples of specifically manufactured sulfate resisting cements Classification

Performance specification

Indicative requirements for sulfate resisting cement production in European and American practice EN 197-1: CEM IV/A (P-V) - SR, 65–79% K with C3A ≤ 9% together with 21–35% combination of P and V, where P > V, total SO3 ≤ 3.5% EN 197-1: CEM IV/B (P) - SR, 45–64% K with C3A ≤ 9% and 36–55% P, total SO3 ≤ 3.5% EN 197-1: CEM IV/B (V) - SR, 45–64% K with C3A ≤ 9% and 36–55% V, total SO3 ≤ 3.5% EN 197-1: CEM IV/B (P-V) - SR, 45–64% K with C3A ≤ 9% together with 36–55% combination of P and V, where P > V, total SO3 ≤ 3.5% ASTM C1157 Type MS, moderate sulfate resistance ASTM C1157 Type HS, high sulfate resistance ASTM C1157 cements are produced to performance attributes and have standard physical requirements but no standard chemical requirements, (ASTM, 2008).

Specific approach in respect of thaumasite attack The thaumasite form of attack can be avoided by good concrete practice in respect of the production of low permeability concrete through control of water/cement ratio and good workmanship (Crammond 2002). The DETR (1999) Expert Group report on thaumasite sulfate attack provides recommendations in respect of practice to minimise the risk. Conservative recommendations cover modification of the existing advice in the case of conventional sulfate conditions in excess of 0.24% total sulfur. This involves implementation of additional measures including designed drainage, surface protection, and use of aggregates of limited carbonate content. Guidance on the practical application of the Group’s recommendations has also been published by the Quarry Products Association (Harrison 1999). In relation to Portland limestone cements, Bensted (2000) urged continuation of the precautionary principle that such cements would be restricted to non-sulfate conditions, pending further research.

Specific approach in respect of delayed ettringite formation Control of delayed ettringite formation requires management of the maximum temperature reached during the hydration and curing process. The risk can be minimised by keeping the concrete temperature below 70°C during the hydration period. This may be achieved through several measures including consideration of the combined influences of ambient

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Table 9.2 Examples of common cements that are valid in the place of use as sulfate resisting Classification Portland cement Portland-Slag Portland-Fly Ash

Type CEM CEM CEM CEM

I II/A-S II/B-S II/A-V

CEM II/B-V CEM CEM CEM CEM CEM CEM CEM CEM CEM

II/A-W II/B-W II/A-P II/B-P II/A-L II/A-LL II/B-L II/B-LL II/A-D

Portland-Composite cement

CEM CEM CEM CEM

II/A-T II/B-T II/A-M II/B-M

Blast furnace cement

CEM III/A

Pozzolanic cement

CEM IV/A CEM IV/B CEM V/A

Portland-Pozzolana Portland-limestone cement

Portland-Silica Fume Portland-Burnt Shale

Composite cement

CEM V/B

National jurisdictions Denmark Austria, France, Italy, Portugal, Spain Austria, France, Italy, Portugal, Spain Austria, Denmark, France, Hungary, Italy, Portugal, Spain Austria, Denmark, Italy, Poland, Portugal, Spain, United Kingdom Italy Italy France, Italy, Portugal, Spain Italy, Portugal, Spain Italy, Portugal Italy, Portugal Italy Italy Austria, Italy, Portugal, Spain, Switzerland Italy Italy Austria, France (S-V), Italy, Portugal Austria, Italy, Switzerland (D, LL, S, T, V) Austria, France, Italy, Poland, Portugal, Spain, United Kingdom Italy, Portugal, United Kingdom Italy, Portugal, United Kingdom Belgium (S-V), France, Italy, Poland, Portugal, Spain France, Italy, Poland, Portugal,

temperature at time of curing, effect of geometry of pour on heat loss during hydration, cement content, and cement type. The cement content of large pours should be selected with a view to limiting the absolute concrete temperature to a value below 70°C during hydration and cure. In this regard Hobbs (1999b) recommended a maximum fresh concrete temperature of 10oC for a 500 kg/m3 cement content mix and progressively less restrictive for lower cement contents, with a 30oC maximum at 360 kg/m3. Further information and guidance are available in Quillan (2001).

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SUMMARY Sulfates may lead to concrete durability failure through attack from external sources (conventional sulfate attack or thaumasite sulfate attack) or from internal sources (delayed ettringite formation). Test methods to ascertain the sulfate resistance of concrete are primarily designed to test the effectiveness of the cement in producing a durable material. Research has been conducted on pastes, mortars, and concretes. It has been found that mortars are the preferred route, due to the lengthy duration of tests on concrete. Researchers are striving to design tests of reasonable duration that could be used as comparator performance-based tests. A prescriptive approach to the problem of external sulfate attack is the norm, with particular attention to limiting maximum water/cement ratio so that impermeable concrete will result. Cement type is another significant consideration in severe sulfate environments. Many cements are regarded as sulfate resisting, but there is a lack of uniformity across the globe on criteria for classifying the sulfate resistance of a cement. The practitioner must refer to recognised cements and test methods in the place of use. Delayed ettringite formation is a rare phenomenon that can be avoided by keeping the concrete temperature below 70°C during hydration and cure.

REFERENCES Aguayo, F., Funez, O., Drimalas, T., Folliard, K. and R. Lute. 2020. An alternative method to evaluate the sulphate resistance of cementitious binders. In External Sulphate Attack - Field Aspects and Lab Tests ed. Menéndez and Baroghel-Bouny. RILEM Bookseries. 21: 93–105. AFNOR. 1993. NF P18-837, Special products for hydraulic concrete construction. Hydraulic binder based needling and/or sealing products. Tests of resistance against seawater and/or water with high sulfate contents. Paris: Association Française de Normalisation. ASTM. 2008. C1157, Standard performance specification for hydraulic cement. West Conshohocken: ASTM International. ASTM. 2018c. C1012, Test for length change of hydraulic-cement mortars exposed to a sulfate solution. West Conshohocken: ASTM International. ASTM. 2021a. C452, Standard test method for potential expansion of Portland-cement mortars exposed to sulfate. West Conshohocken: ASTM International. Bensted, J. 2000. Thaumasite review. Concrete 34, 8: 11. BRE. 1996. Sulphate and acid resistance of concrete in the ground. Digest 363. London: Construction Research Communications. BRE. 2005. Concrete in aggressive ground. Special Digest 1. 3rd Edition. Watford: BRE Press. BRE. 2008. Sulfate damage to concrete floors on sulphate-bearing hardcore. London: RIBA Publishing. BSI. 2021b. BS 1377-3:2018+A1:2021. Method of test for soils for civil engineering

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purposes: chemical and electrochemical tests. London: British Standards Institution. CEN. 2008c. CEN/TR 15697. Cement - performance testing for sulfate resistance State of the art report. Brussels: Comité Européen de Normalisation. CEN. 2011a. EN 196-5. Method of testing cement, Part 5: Pozzolanicity test for pozzolanic cement. Brussels: Comité Européen de Normalisation. CEN. 2013a. EN 196-2. Method of testing cement. Chemical analysis. Brussels: Comité Européen de Normalisation. Concrete Society. 1996. Developments in durability design and performance-based specification of concrete. Special Publication CS109. London: Concrete Society. Crammond, N. 2002. The occurrence of thaumasite in modern construction – a review. Cement and Concrete Composites 24, 3-4: 393–402. CSA. 2018. CSA A3004-C8, Test method for determination of expansion of blended hydraulic cement mortar bars due to external sulphate attack. Mississauga: Canadian Standards Association. DETR. 1999. The thaumasite form of sulfate attack: risks, diagnosis, remedial works and guidance on new construction. Report of the Expert Group. Rotherham: Department of the Environment, Transport, and the Regions. Erlin, B. and D. Stark. 1965. Identification and occurrence of thaumasite in concrete. Highway Research Record 113: 108–113. Ferraris, C., Stutzman, P., Peltz, M. and J. Winpigler. 2005. Developing a more rapid test to assess sulfate resistance of hydraulic cements. Journal of Research of the National Institute of Standards and Technology 110, 5: 529–540. Harrison, T. 1999. Practical application of the DETR Expert Group report on thaumasite sulfate attack on buried concrete. London: Quarry Products Association. Hartshorn, S. and I. Sims. 1998. Thaumasite - a brief guide for engineers. Concrete 32, 8: 24–27. Hartshorn, S., Sharp. J. and R. Swamy. 1999. Thaumasite formation in Portlandlimestone cement pastes. Cement and Concrete Research 29: 1331–1340. Haufe, J. and A. Vollpracht. 2019. Tensile strength of concrete exposed to sulfate attack. Cement and Concrete Research 116: 81–88. Hobbs, D. 1999b. Expansion and cracking in concrete associated with delayed ettringite formation. In Ettringite: The Sometimes Host of Destruction, ed. B. Erlin. ACI Special Publication SP-177, 159–181. Farmington Hills: American Concrete Institute. Hobbs, D. 2000. Chemical attack on concrete. In Proceedings of the tenth BCA annual conference on higher education and the concrete industry, 93–102. Birmingham: The University of Birmingham. 93–102. Hobbs, D. 2001. Concrete deterioration: causes, diagnosis, and minimizing risk. International Materials Review 46, 3: 117–144. Hobbs. D. and J. Matthews. 1998b. Minimum requirements for concrete to resist chemical attack. In Minimum Requirements for Durable Concrete, ed. D. Hobbs, 131–162. Crowthorne: British Cement Association. Hobbs, D. and M. Taylor. 2000. Nature of the thaumasite sulfate attack mechanism in field concrete. Cement and Concrete Research 30: 529–533. ISO. 2023. ISO 24684, Aggregates for concrete – test methods for chemical properties – part 2: determination of soluble sulfates. Geneva: International Standards Organisation.

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Jabbour, M., Omikrine Metalssi, O., Quiertant, M. and V. Baroghel-Bouny. 2022. A critical review of existing test-methods for external sulfate attack. Materials 15. 21: doi: 10.3390/ma15217554. Kelham, S. 1996. The effect of cement composition and fineness on expansion associated with delayed ettringite formation. Cement and Concrete Composites 18: 171–180. Koch, A. and H. Steinegger. 1960. A rapid method for testing the resistance of cements to sulphate attack. Zement-Kalk-Gips 7: 317–324. Lawrence, C. 1993. Laboratory studies of concrete expansion arising from delayed ettringite formation. Crowthorne. British Cement Association. Lea, F. 1998. The chemistry of cement and concrete, ed. P. Hewlett. London: Edward Arnold. MOHURD - Ministry of Housing and Urban-Rural Developmen. 2008. GB/T 7492008, Test method for determining capability of resisting sulfate corrode of cement. Beijing: General Administration of Quality Supervision, Inspection and Quarantine of the People’s Republic of China (AQSIQ). Mure, N. 2000. Concrete in aggressive chemical environments. Concrete 34, 6: 44–45. Neville, A. 1995. Properties of concrete. 4th Edition. Harlow: Longman. O’Connell, M., McNally, C. and M. Richardson. 2012. Performance of concrete incorporating GGBS in aggressive wastewater environments. Construction and Building Materials 27, 1: 368–374. Quillan, K. 2001. Delayed ettringite formation: in-situ concrete, Information Paper IP 11/01. Bracknell: IHS BRE Press. Ryle, R. 1999. Metakaolin - a highly reactive pozzolana for concrete. Quarry Management December: 27–31. Skalny, J. and J. Pierce. 1999. Sulfate attack: an overview. In Materials science of concrete: sulfate attack mechanisms. ed. J. Marchand and J. Skalny, 49–63. Westerville: American Ceramic Society. Stark, D. 2003. Occurrence of thaumasite in deteriorated concrete. Cement and Concrete Composites 25, 8: 1119–1121. Taylor, H., Famy, C. and K. Scrivener. 2001. Delayed ettringite formation. Cement and Concrete Research 31, 5: 683–693. Wallace, J. 1999. Strengthening thaumasite-affected concrete bridges. Concrete 33, 8: 28–29. Wittekindt, W. 1960. Sulphate-resistant cements and their testing. Zement-Kalk-Gips 13: 565–572.

Chapter 10

Chemical attack Leaching, acid, and seawater attack

ESSENTIAL FACTS Concrete in contact with liquids may deteriorate. This is due to leaching, acid, and seawater attack. Concrete is typically the most economic material for civic infrastructure, even in situations where the material is called upon to act outside its comfort zone. The material is placed into a battle, where it must remain serviceable as long as possible, even as its ability to fight on is constantly ebbing away. The annual cost of maintenance and repair of concrete in water and wastewater facilities, for example, exceeds the investment in new infrastructure (O’Connell et al. 2010). Every 1% improvement we can make in the durability and sustainability of investment in new infrastructure will therefore have an ongoing positive global impact. Leaching of calcium and hydroxide ions out of the concrete matrix occurs when concrete is immersed in water. Such conditions can occur in water-retaining tanks, pipes, and structures in lake water, such as dams and bridge piers. The dissolved ions in the pore structure diffuse out due to the high concentration difference, especially calcium ions encountering soft water. The effect is to increase the porosity of the concrete and also to reduce its strength. This makes the material more vulnerable to the ingress of deleterious agents. Well-cured and compacted concretes of low to moderate water/cement ratio are generally capable of providing a reasonable service life in conditions of chemical attack caused by acidic solutions, if the rate of attack is acceptably slow. However, it is a reality that all the main compounds in concrete paste may be dissolved by acids. Durability failure through contact with acids occurs by dissolution of the cement paste and certain aggregate types. The compounds affected include the calcium silicate hydrates, calcium hydroxide, tricalcium aluminate hydrate, and ettringite. Concrete in contact with chemicals in a dry state is not usually affected, but acid attack occurs if aggressive chemicals are in solution and contain acidic ions above a critical concentration level. The effect of dissolution may be manifested through surface erosion or complete DOI: 10.1201/9781003261414-10

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disintegration of the concrete. Concrete in contact with flowing acids is more vulnerable than that in contact with static fluids. A specific form of chemical attack is that caused by seawater. It is distinguished from other sources of attack in that it involves processes other than dissolution. These include chemical reactions with magnesium and sulfate ions, physical erosion, and even freeze-thaw attack. Examples abound of chemical environments that may promote attack. The challenge is ever-increasing in urban areas, as the need arises to redevelop former industrial sites that operated under lower environmental standards than today and which are now deemed as polluted sites. The implications of chemical attack must particularly be considered in the redevelopment of contaminated sites such as old gas works. Sewers may contain hydrogen sulfide, which oxides to sulfuric acid. Robery (1988) drew attention to bacterially active soils containing acid-forming bacteria such as thiobacillus concretivorous. Sulfuric acid may also occur in soils and groundwater through the oxidation of iron sulfide minerals in the form of pyrites or marcasite (Harrison 1987). Various industrial processes in the manufacturing sector can give rise to aggressive environments through a requirement to use specific chemicals in the production process. Animal and vegetable oils contain free acids that can increase with exposure to air, leading to problems in the event of frequent spillage on concrete floors. A range of potentially deleterious chemical environments can be found in rural areas (Braam and Frénay 1997). Agricultural examples include acidic effluent produced during silage making; naturally occurring waste from housed animals; and produce in storage sprayed with fungicide-containing acids. Carbonated water may also cause attack. Water streams can become acidic through dissolved carbon dioxide, and such waters are described as “soft”. The carbon dioxide is available from the atmosphere and may dissolve to form carbonic acid with a pH of about 5.6. It is not possible to produce a definitive list of the chemical effect on concrete of every potentially harmful compound, due to the varied chemistry of concrete itself, but useful summaries have been published. Table 10.1 gives a brief overview of the primary harmful agents. More Table 10.1 Effect of selected chemicals on concrete Category

Effect on good-quality concrete Disintegration

Acids

Salts and Alkalis

Hydrochloric; Hydrofluoric; Muriatic; Nitric; Sulfuric; Sulfurous Ammonium nitrate

Low level of attack Acetic; Carbonic; Carbolic; Humic; Lactic; Phosphoric; Tannic Chlorides of ammonia, copper, iron, magnesium, mercury, zinc

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comprehensive listings may be found in references such as Lea (1998) and the American Concrete Institute manual (ACI 1985). The potential susceptibility of a concrete member to chemical attack may be determined by reference to the threat level classified by environments generally described as “slightly”, “moderately” and “highly” aggressive. These classifications are based on a quantitative analysis of the chemical characteristics of natural ground conditions and seawater. In general terms, specification of concrete that is to be in contact with acids of pH 6.5 and lower requires consideration. Very harsh environments may involve chemical conditions beyond the scope of normal concrete composition and require specification of special protective measures. In broad terms, the specifier needs to give consideration to potential chemical attack in the following cases: • • • • •

concrete in acidic soils and groundwaters concrete in industrial projects specifically involving chemical processes concrete for use on farms concrete sewers structures in seawater.

The specific case of chemical attack from sulfates in soils and groundwater is considered separately in Chapter 9.

PHYSICO-CHEMICAL ASPECTS

Acidic attack By definition, acids are substances that yield hydrogen ions in solution by dissociation. The rate of attack on concrete depends on the amount of hydrogen ions formed. This in turn depends on the nature of the acid (strong or weak) and the concentration of the acid in solution (concentrated or dilute). Strong acids such as hydrochloric (HCl) and sulfuric acid (H2SO4), have a high degree of dissociation, whereas weak acids such as acetic acid (CH3COOH), do not. The externally introduced acids react with the bases in concrete to form a calcium salt and water. The salt may be soluble, and solid product is therefore lost. The reaction of calcium hydroxide, for example, with hydrochloric and sulfuric acid, respectively, is as follows: Ca (OH )2 + 2HCl Ca (OH )2 + H2 SO4

CaCl2 + 2H2 O CaSO4 .2H2 O

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In the second example, sulfuric acid reacts to form gypsum, which can be easily removed from the member surface exposing lower layers which are progressively attacked. Ammonium cations exchange with calcium ions, initially from calcium hydroxide but later from the calcium silicate hydrate, for example: 2NH4 NO3 + Ca (OH)2

Ca (NO3)2 + 2NH3 + 2H2 O

xNH4+ + xCaO . ySiO2 . nH2 O xNH3+ + xCa2+ + xOH + ySi (OH)4 + (n 2y) H2 O

Similarly for magnesium: MgCl2 + Ca (OH)2

Mg (OH )2 + CaCl2

xMg 2+ + xCaO. ySiO2 . nH2 O xMg (OH )2 + xCa2+ + ySi (OH )4 + (n x 2y) H2 O

Attack by nitrates can form nitroaluminates that disrupt the concrete. Concentrated chloride solutions can form chloroaluminates that may also cause disruption. Concrete in use in agricultural applications may be attacked by silage effluent consisting mainly of lactic acid and acetic acid. These form calcium lactate and calcium acetate, which are largely soluble. For example, with acetic acid: 2CH3 COOH + Ca (OH )2 2CH3 COOH + CSH

Ca (CH3 COO)2 + 2H2 O

SiO2 + Ca (CH3 COO)2 + H2 O

Soft water containing free carbon dioxide may strip off the surface layers of concrete exposing the aggregate. The reaction then tends to slow down. The carbon dioxide dissolves to form carbonic acid: CO2 + H2 O = H2 CO3 H2 CO3 + H2 O = H3 O+ + HCO3

The amount of dissolved carbon dioxide that forms carbonic acid is small, but when the HCO3 is removed by reaction, more carbonic acid is formed and so on. The attack initially involves production of insoluble calcium carbonate (CaCO3) from hydration products, but this further reacts to form soluble calcium hydrogen carbonate:

Chemical attack

CaCO3 + H2 CO3

289

Ca (HCO3)2

The attack associated with oxidation of iron sulfide minerals in soils has been described by Hobbs (2000). Two possible mechanisms are presented: 2FeS2 + 2H2 O + 7O2

2FeSO4 + 2H2 SO4

or 4FeS2 + 15O2 + 8H2 O

2Fe2 O3 + 8H2 SO4

The mechanism of attack in sewers has been described by van Mechelen and Polder (1990). There is anaerobic production of hydrogen sulfide (H2S) from soluble sulfides. This can become dissolved in moisture above the sewage level in pipes and manholes. It is oxidised by aerobic bacteria to produce sulfuric acid. Harrisson (1995) reports that this type of attack can lead to complete loss of concrete strength.

Seawater attack Seawater is a mildly aggressive solution, principally due to high levels of sulfate from highly soluble magnesium sulfate. Chemical reactions in seawater are different from those occurring in sulfate-bearing groundwater due to the presence in seawater of chlorides. The soluble salt composition is approximately 3.5% by mass, including chloride ions (20 g/L), sodium ions (11 g/L), and magnesium sulfate (4 g/L). Physico-chemical effects caused by contact with seawater include magnesium ion and carbonic acid attack on the products of hydration, leaching, and pressure from salt crystallisation within the concrete. Potential confusion between physical salt-weathering attack and chemical sulfate attack has also been reported (Mehta 2000). An example of seawater attack is illustrated in Fig. 10.1. Magnesium sulfate reacts with the cement matrix to form calcium sulfate and magnesium hydroxide: MgSO4 + Ca (OH )2

CaSO4 + Mg (OH )2

The insoluble magnesium hydroxide can provide protection below lowwater level but higher up it is lost through wave action. Below low-water level, ettringite may also be formed by reaction of sulfate with tricalcium aluminate and calcium silicate hydrate. It does not develop to the point of damaging expansion because the ettringite is soluble in the chloride-rich environment of seawater. Concrete just above the high-water mark can absorb salts by capillary attraction. The effect of evaporation is to

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Figure 10.1 Example of structure damaged by seawater attack.

concentrate the salts with eventual crystallisation on rewetting. This leads to disruptive pressures in the concrete. The phenomenon is known as salt weathering.

FACTORS INFLUENCING ATTACK Factors that influence the rate of chemical attack include the following: • concrete permeability • pH

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• • • • •

291

solubility of reaction products flow rate of acid characteristics of cementitious components aggregate type temperature.

Concrete permeability The rate of attack depends on the rate of penetration of the acid. Therefore, a prerequisite for durable structures in acidic environments is that the concrete is dense and impermeable. Chemical attack in concrete involves acids dissolved in solution. Impermeable concrete resists the ingress of acidic solutions and restricts damage to the surface layers. Obviously, low water/cement ratios are required to achieve low-permeability concrete. In the context of acid attack, the use of superplasticisers represents a better approach than increasing the cement content since extra cement merely adds fuel to the reaction. The rate of attack by some acids has also been found to decay with time due to a beneficial effect of insoluble reaction products. These products may act as pore blockers and reduce permeability. For example: CSH + HCl

CaCl2 + H2 SiO3

The silicic acid (H2SiO3) may polymerise to form silica gel (SiO2) and restrict further acid ingress.

pH The pH value may be used as an initial screening test of potential acid attack. If the pH of the acid exceeds 6.5, the possibility of attack is remote. Although each unit decrease in pH represents a ten-fold increase in acidity, the degree of dissociation is more significant. For example, concrete exposed to contact with oxalic acid of pH 3 reacts to form calcium salts that are insoluble, and durability is not impaired. The concentration of acid and the quantity in contact with concrete also influence the degree of attack.

Solubility of reaction products The effect of the attack on concrete durability depends on the solubility of the reaction products. Calcium sulfate, for example, is less soluble than calcium chloride. Reaction of concrete with humic acid – a complex of organic acids generated by the decay of organic matter in soil – can lead to the deposition of insoluble calcium humate with beneficial effects in the case of acid attack in static conditions.

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Flow rate of acid Flowing acids are more harmful than static fluids. Where concrete is in contact with flowing acid the option of neutralising the acid by the cementitious components is not available, and both soluble and insoluble reaction products are continuously carried away. The loss of insoluble reaction products prevents the formation of a protective skin. Aggressive water in impermeable clay soils may not attack concrete in contact with it because the static acidic solution is neutralised at the concrete boundary.

Characteristics of cementitious components An early study by Halstead (1954) showed some slight benefit in the use of high alumina cement in long-term tests of concrete in soft water of pH 4.4, but in general, the cement type does not seem to influence the resistance of the concrete over a typical economic service life requirement. Nevertheless, some enhancement of performance has been reported in respect of concretes incorporating fly ash, slag, silica fume, or metakaolin. Inclusion of pozzolanic materials is generally beneficial in two ways. Firstly, they can be used to reduce the permeability of the concrete. Secondly, they lead to a reduction in the content of calcium hydroxide Ca(OH)2 and thus reduce the quantity of vulnerable material. However, it appears that in the case of slag or PFA concretes, the extent of any improvement over Portland cement concretes of comparable water/binder ratio is not significant (Hobbs and Matthews 1998b). Slight improvements are always worthy of consideration and, for example, the rate of leaching of calcium ions was slowed down through 30% fly ash inclusion or through 40–60% of slag (Rozière and Loukili 2011; Tang et al. 2016). Silica fume has been found to be advantageous (Berke 1989; Durning and Hicks 1991). In addition to reducing permeability and the calcium hydroxide content, it produces a highly polymerised calcium silicate hydrate paste that is more stable and resistant to acid attack. The extent of the advantages gained through use of silica fume depends on the microsilica content and the type of acid. Improvement has been demonstrated with contents from 7.5% to 30%. Resistance to attack by acetic acid and formic acid has been more pronounced than against phosphoric acid and sulfuric acid. Mehta (1985) reported good results in tests with lactic acid. Knutsen and Obuchowicz (1997) reported optimum results at a level of 10% by weight of cement, in agricultural applications. Metakaolin, the purpose-made pozzalana made from kaolinitic clay, reacts with calcium hydroxide during hydration to form stronger and less-soluble hydrates with enhanced acid resistance. Martin (1997) studied the influence of metakaolin in tests with silage effluent. The effluent mainly consisted of lactic acid, with acetic acid also present. He reported that inclusion of 15% metakaolin resulted in one-third less damage compared with Portland cement

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concrete. Further improvement resulted from the use of ternary blends of Portland cement, slag, and metakaolin. The inclusion of metakaolin and latex co-polymer was also very effective. Silverlock (1999) reported that the use of metakaolin was specified in an industrial water-diversion scheme handling large volumes of acidic effluent.

Aggregate type Aggregate type influences the rate of attack and the effect on serviceability. Limestone aggregate is particularly vulnerable to acid attack, but this has potential advantages. First, it can contribute to the neutralisation reservoir in the case of static acidic waters in contact with concrete. Second, if used in pipes, it decays in tandem with the cementitious component, thus maintaining a hydraulically efficient smooth surface. The use of other types of aggregate in pipes not only leads to rough surfaces, but the aggregate becomes detached over time, leaving debris in the pipe that could impede serviceability. Protruding acid-resistant aggregate in the floor slabs of horizontal silage silos is undesirable from the viewpoint of animals using the facility in a self-feed manner. Limestone aggregate is preferable to maintain a serviceable, if deteriorated, surface. A case of “if you can’t beat them, join them”!

Temperature High temperature may compound problems of acidic attack because of the well-established relationship between rate of reaction and temperature. For example, acid attack in sewers is more pronounced in countries with hot climates. MATHEMATICAL MODELLING OF ACID ATTACK The rate of acid attack often decays with time, but in other cases the deterioration-time relationship is linear. The reduction of attack rate may often be related to the square root of time. The trends are illustrated in Fig. 10.2. The reduction in decay of attack rate with time is due to the formation of a protective layer of silica gel in place of the dissolved concrete. This impedes access by the acid to the soluble hydrated components. This protective layer may remain intact in the case of static acidic waters, but otherwise may be lost. A mathematical model proposed by Grube and Rechenberg (1989) tracks the development of the affected layer with time through Fick’s first law. Concrete in contact with a solution containing dissolved carbon dioxide forms, in the first instance, a thin layer of calcium carbonate, CaCO3, but further reaction produces soluble calcium hydrogen carbonate, Ca (HCO3)2. This soluble material diffuses out into the water. In the mathematical model, the thickness of the affected layer is estimated as a function of the following four parameters: the diffusivity of the protective

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Figure 10.2 Typical patterns of acid attack rate.

layer with respect to the calcium hydrogen carbonate; the composition of the concrete (mass of soluble matter per unit volume and relative proportion of insoluble matter); the concentration difference of calcium hydrogen carbonate in the concrete and surrounding solution (which is an indicator of the aggressiveness of the environment); and time. These parameters are represented in the model as follows: x = thickness of deteriorated or removed surface layer (cm) D = diffusion coefficient, for Ca(HCO3)2 or Ca2+ ions, of the gel layer (cm2/s) as/at = ratio of soluble matter area to total area ms = mass of soluble matter (g) of CaO per cm3 of concrete cs = Ca(HCO3)2 concentration, or highest concentration of Ca2+ ions, in the solution at the face of intact concrete (g/cm3) cm = Ca(HCO3)2 concentration, or highest concentration of Ca2+ ions, in the solution unaffected by the concrete (g/cm3) t = time (s). The quantity of material diffusing may be stated as: dl = D as

cs

cm x

dt

From a chemical reaction viewpoint, the quantity may be stated as: dl = ms at dx

Hence ms at dx = D as

cs

cm x

dt

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and x dx =

D as (cs ms at

c m) dt

By integration, the thickness of the deteriorated or removed layer may be determined at time ‘t’: x=

2 D (as / at )(cs ms

c m)

t

Grube and Rechenberg estimate the value of the diffusion coefficient (D) at 1.8×10–6 cm2/s for SiO2 gel of hardened cement paste. The ratio as/at can be determined from the ratio of the volume of soluble constituents to the total volume. If insoluble aggregates are used, the volume of soluble constituents is equal to the volume of hardened cement paste. The relationship may be restated as: x=a t

where “a” is a constant for a particular case, thus leading to a parabolic curve. However, if the protective layer is removed by mechanical action, for example by turbulent flowing acid, the removal rate becomes linear:

x=

t

The value of α can be determined in short-term tests.

MANAGEMENT OF CHEMICAL ATTACK The approaches to specification and design practice for serviceability failure due to chemical attack recognises the reality that concrete as a material is highly vulnerable in acidic environments. Therefore, the objective is to reduce the rate of attack to an acceptable level rather than to strive for immunity. In most cases, this involves the specification of concrete of adequate composition but, in the severest environments, other protective measures may be required. The first step in the prescriptive approach is to quantify the chemical characteristics of the environment and classify it into one of typically three levels. These exposure classes are defined by upper and lower chemical bounds within which limiting compositions in respect of minimum cement content and maximum water/cement ratios will produce serviceable concretes. The classification for a particular situation is based on the

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chemical characteristics of the groundwater, soil, or seawater. The class is determined by the most onerous value for any single chemical characteristic. The characteristics used in the case of groundwater are pH, and the concentration (mg/L) of sulfate (SO42−), ammonium (NH42+), magnesium (Mg2+), and aggressive CO2. Soil is classified according to its sulfate content (SO42- mg/kg) and degree of acidity (mL/kg) according to Baumann Gully. The reference test method is DIN 4030-2 (Deutsches Institut für Normung 1991). The threshold is a value in excess of 200 mL/kg Baumann Gully. In respect of pH, the minimum value for normal concrete compositions is pH 4. A special study is required for the specification of concrete in lower pH conditions. However, it is worth emphasising that pH value alone is not an adequate measure of aggressivity, although Harrison (1987) suggests that concrete may not be a suitable material in conditions where the pH value is below 2.5. Guidance on limiting values of concrete composition for an exposure class is normally based on an assumption of water/soil temperature between 5°C and 25°C and a water velocity slow enough to approximate to static conditions. Special precautions may be needed outside these conditions. Such environments, and others outside the upper bounds, require special measures. Examples of special measures include surface protection by external tanking or acid-resistant surface coating; improvement of site drainage to reduce the durability load; the use of a neutralising calcium carbonate backfill (limestone/chalk) as a sacrificial buffer between the environment and the concrete surface; or the use of section sizes that are initially above the minimum required for structural resistance and diminish over time. In the case of seawater attack, a report by American Concrete Institute Committee 201 (1962) recommended use of a Portland cement with not more than 8% tricalcium aluminate. More recently Osborne (1994) once again drew attention to the somewhat inferior performance of high tricalcium aluminate concretes. Hobbs (2001) noted that freeze-thaw resistant concretes should give adequate resistance to seawater attack if the water/binder ratio was 0.45 or lower and the tricalcium aluminate content was below 12%. He extended the range of resistant concretes to include the following combinations with Portland cement: 20–40% fly ash, 70–80% slag, and less than 10% silica fume. National annexes and national complementary standards provide definitive guidance for their jurisdictions in respect of limiting concrete compositions to ensure durability in conditions of potential chemical attack. It should be noted, however, that not all national documents classify exposure conditions by the same criteria. For example, the UK uses a system of “design chemical classes” derived from a consideration of an “aggressive chemical environment for concrete” (ACEC) class, the groundwater hydraulic gradient, the type and thickness of the concrete element, and its intended working life. Guidance on appropriate

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concrete composition requirements is presented in the Building Research Establishment Special Digest 1 (BRE 2005).

SPECIFICATION BY PERFORMANCE

Leaching Rozière and Loukili (2011) demonstrated the possible use of the equivalent performance method in the context of leaching. They set up a comparative test that assessed the leaching of hydroxide ions and calcium. Specimens were immersed for 60 days in a solution maintained at pH 5.0 through use of 0.5 mol/L nitric acid. The quantity of hydroxide ions leached was detected by the volumes of nitric acid that needed to be added. Leached quantities of calcium were detected by atomic absorption spectroscopy, from which equivalent performance may be assessed from a benchmark level.

Acid attack Performance-based design requires both a predictive model and a test method. An approach for developing a mathematical model for acid attack, as noted earlier, has been published by Grube and Rechenberg (1989). However, no standard test method exists for testing the acid resistance of concrete. This is due in part to the difficulty of accelerating in laboratory tests the deterioration that may take place in the field. Attempted acceleration by the use of unrepresentatively high concentration levels in laboratory tests would fail concretes that may give adequate service under field conditions due to the difference in chemical reactivity being replicated.

Seawater attack Testing the sulfate resistance of concrete was discussed in Chapter 9. The favoured method in developing a European standard test involves monitoring the expansion of mortar prisms stored in sulfate solutions, but future use of the method for testing resistance to seawater attack could possibly be achieved by substituting the sulfate solution by artificial seawater.

SUMMARY There is universal agreement that the achievement of dense impermeable concrete is a prerequisite requirement for durability in acidic environments. However, concrete is not chemically resistant to acid attack. Therefore, concrete may not be an ideal material in conditions where the pH value is

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below 2.5, but structures in milder acidic environments may achieve their required length of service through concrete compositions that have sufficiently slow rates of deterioration. This allows the use of concrete in urban and rural infrastructure vital to public health and food security, despite the harsh environment. In the most extreme environments, concrete may need additional surface protection. Exposure to seawater also leads to chemical attack, but the process is complicated by the physical effects of wave action and abrasion. Prescriptive guidance is widely available on limiting compositions for concrete exposed to chemical attack. Specifiers should be mindful, however, that not all national approaches to threshold limits that define exposure classes are the same. Consultation of local guidance is required. Tentative mathematical formulations have been advanced for modelling acid attack. Development of performance-based specifications is somewhat hampered by the difficulty of developing standard test methods for acid resistance. Performance tests for durability often require laboratory tests using harsh conditions to accelerate deterioration. This approach may not provide a reliable basis on which to predict field performance based on laboratory findings in the case of acid attack. Nevertheless, further research on the ability of future low-carbon concretes to provide serviceable infrastructure over an economic design life is warranted.

REFERENCES ACI Committee 201. 1962. Durability of concrete in service. American Concrete Institute Journal 59, 12: 1771–1820. ACI. 1985. American Concrete Institute Manual of Concrete Practice. Part 5. Detroit: American Concrete Institute. Berke, N. 1989. Resistance of microsilica concrete to steel corrosion, erosion and chemical attack. In Proceedings of the third international conference on fly ash, silica fume, slag and natural pozzolans, ed. V. Malhotra, 861–886. ACI SP114. Detroit: American Concrete Institute. Braam, C. and J. Frénay 1997. Durability of concrete in agriculture. In Proceedings of concrete for a sustainable agriculture, agro-. aqua- and community applications, ed. E. Berge, H. Magerøy and K. Berg, 51–68. Oslo: Norse Betongforening. BRE. 2005. Concrete in aggressive ground. Special Digest 1. 3rd Edition. Watford: BRE Press. Deutsches Institut fur Normung. 1991. DIN 4030-2, Assessment of water, soils and gases for their aggressiveness to concrete: collection and examination of water and soil samples. Berlin: Deutsches Institut fur Normung. Durning, T. and M. Hicks. 1991. Using microsilica to increase concrete’s resistance to aggressive chemicals. Concrete International 13, 3: 42–48. Grube, H. and W. Rechenberg. 1989. Durability of concrete structures in acidic water. Cement and Concrete Research 19, 5: 783–792.

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Halstead, P. 1954. An investigation of the erosive effect on concrete of soft water of low pH value. Magazine of Concrete Research 6, 17: 93–98. Harrison, W. 1987. Durability of concrete in acidic soils and waters. Concrete 21, 2: 18–24. Harrisson, A. 1995. Deleterious processes in concrete. Concrete 29, 6: 13–14. Hobbs, D. 2000. Chemical attack on concrete. In Proceedings of the tenth BCA annual conference on higher education and the concrete industry, 93–102. Birmingham: The University of Birmingham. 93–102. Hobbs, D. 2001. Concrete deterioration: causes, diagnosis, and minimizing risk. International Materials Review 46, 3: 117–144. Hobbs. D. and J. Matthews. 1998b. Minimum requirements for concrete to resist chemical attack. In Minimum requirements for durable concrete, ed. D. Hobbs, 131–162. Crowthorne: British Cement Association. Knutsen, K. and M. Obuchowicz. 1997. Microsilica for concrete durability in agriculture applications. In Proceedings of concrete for a sustainable agriculture, agro-. aqua- and community applications, ed. E. Berge, H. Magerøy and K. Berg, 73–82. Oslo: Norse Betongforening. Lea, F. 1998. The chemistry of cement and concrete, ed. P. Hewlett. London: Edward Arnold. Martin, S. 1997. Metakaolin and its contribution to the acid resistance of concrete. In Proceedings of concrete for a sustainable agriculture. agro-. aqua- and community applications, ed. E. Berge, H. Magerøy and K. Berg, 94–106. Oslo: Norse Betongforening. Mehta, P. 1985. Studies of chemical resistance of low water/cement ratio concretes. Cement and Concrete Research 15: 969–978. Mehta, P. 2000. Sulfate attack on concrete: separating myths from reality. Concrete International 22, 8: 57–61. O’Connell, M., McNally, C. and M. Richardson. 2010. Biochemical attack on concrete in wastewater applications: a state of the art review. Cement and Concrete Composites 32, 7: 479–485. Osborne, G. 1994. The durability of SRPC/GGBS concretes in aggressive sulphate, acidic and marine environments. In Euro-cements. Impact of ENV197 on concrete construction, ed. R. Dhir and M. Jones, 169–182. London: CRC Press. Robery, P. 1988. Protection of structural concrete from aggressive soils. Chemistry and Industry 13: 421–426. Rozière, E. and A. Loukili. 2011. Performance-based assessment of concrete resistance to leaching. Cement and Concrete Composites 33, 4: 451–456. Silverlock, R. 1999. Strengthening applications. Concrete Engineering International March. 61–63. Tang, Y-J., Zuo, X-B., He, S-L., Ayinde, O. and G-J. Yin. 2016. Influence of slag content and water-binder ratio on leaching behaviour of cement pastes. Construction and Building Materials 129: 61–69. van Mechelen, A. and R. Polder. 1990. Degradation of concrete in sewer environment by biogenic sulfuric acid attack. In Microbiology in Civil Engineering: Proceedings of the Federation of European Microbiological Societies Symposium, ed. P. Howsam, 146–157. London: CRC Press.

Chapter 11

Alkali-aggregate reaction

FORMS OF REACTION Aggregates in concrete are generally considered to be inert. Under certain conditions, however, some components in aggregates, especially siliceous minerals, may react with the sodium and potassium hydroxides in the alkaline pore solution of concrete. A porous silica gel of high surface area is formed, which can imbibe water. The reaction products have a molar volume in excess of the reactants, which leads to swelling. The resultant pressures cause expansion of the hardened concrete and, if the tensile strength of the concrete is exceeded, extensive cracking may become apparent on the surface. The gel produced by the reaction exudes from the cracks and changes colour to white when exposed to the atmosphere due to carbonation, as illustrated in Fig. 11.1. This phenomenon is known as alkali-aggregate reaction (AAR). The term “alkali-aggregate reaction” is used to encompass three forms of reaction. The predominant of these throughout the world is alkali-silica reaction (ASR). Other forms are alkali-carbonate reaction and alkali-silicate reaction. The alkali-carbonate reaction (ACR) occurs when alkalis react with fine-grained argillaceous dolomitic limestone aggregate containing calcite and clay. The alkali-carbonate reaction has been a greater cause for concern in Canada and China than elsewhere. It has been referred to as “so-called ACR”, in that it is a form of ASR (Sims and Poole 2017). However, the distinction is helpful in directing the investigator to distinct screening tests for carbonate aggregates. The alkali-silicate reaction is significantly less prevalent and less well researched. Attention in most continents, including Europe, is focused on alkali-silica reaction and, therefore ASR is more universally referred to than AAR, almost to the extent of excluding consideration of the other forms. HISTORICAL AND GEOGRAPHICAL ASPECTS The incidence of ASR is widespread geographically, but the number of deleterious occurrences in any one country is very minor compared with 300

DOI: 10.1201/9781003261414-11

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Figure 11.1 Example of a structure damaged by alkali-silica reaction, showing carbonated gel exuded from cracks as a white substance.

other causes of durability failure. The reaction occurs to a limited extent in many concretes, but catastrophic damage due to ASR is exceedingly rare. The reaction may become apparent at some time in the life of a structure, but may slowly exhaust itself, typically leaving the structure in a deteriorated but ongoing functional condition. The first reference to the problem was probably that by Poulsen (1914) to the Institution of Danish Civil Engineers in January 1914. Wider interest in the phenomenon is regarded as having been initiated by researchers in the US in the 1930s, exemplified by the publication of a paper by Stanton (1940). Expansion and cracking had been observed at the time in several Californian structures. Stanton demonstrated through tests on mortars that the high-alkali cement used locally could react expansively with opaline silica in the fine-aggregate sand. The problem has since progressively appeared in many countries, although it was several decades before the extent of the phenomenon became fully apparent. Notwithstanding Stanton’s work and extensive research in Australia, the UK, and Denmark, the aggregate sources of many countries still appeared to be innocuous despite siliceous components. Later, circumstances conspired to cause reaction. Diagnosis of a first case in a country soon led to the identification of several others. For example, the prevalence of opaline flint or chalcedonic flint in Danish aggregates gave rise to problems that were diagnosed in the 1950s. Across the border in northern Germany, severe damage due to ASR was identified in a bridge in the late 1960s, and further ASR-induced damage was identified in structures in the mid-1970s.

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Interestingly, the problems were greatest in northern Germany, due to the use of a local opaline sandstone. This regional aspect of the problem was later encountered elsewhere. For example, occurrences in Sweden have been primarily in the Scania region. Interest in the problem grew in the 1970s, fuelling media coverage about “concrete cancer”, which overstated the scale of the problem. In the UK, ASR damage was identified in a dam in Jersey in 1971, and in three power stations in southwest England in 1976. Further cases were identified in about 200 structures, primarily in the southwest and the Midlands. In Iceland, a combination of highly reactive volcanic aggregates and the introduction of a high-alkali-level (1.5%) home-produced cement caused ASR problems in 1960s structures. These problems were first diagnosed in 1976, and a survey in 1979 revealed that the problem was extensive. The first positive identification of ASR damage in South Africa was also in 1976, and many cases have since been identified in the Western Cape region. In France, ASR damage was identified in dams in the late 1970s, and a later survey of 140 bridges in northern France revealed that 29% exhibited AAR, but only 5% of the total exhibited signs of significant degradation. In the 1980s, in the south of Norway, a survey of more than 400 dams, hydropower plants, and road bridges identified ASR in 31 structures. Equally, a survey of over 400 New Zealand bridges commenced in 1989, initially raising more than 100 suspect cases, but AAR was positively identified in only about one-third of these. Nevertheless, five of these structures required major repair work, including the replacement of one structure. Meanwhile, ASR damage was identified in some Italian industrial structures and pavements, mainly on the Adriatic coast, and ASR problems were also identified in Japan during the decade. Further instances of ASR damage were first reported in other countries during the 1990s. Damage due to ASR was identified in a 30-year-old viaduct in the Netherlands in 1990, and other cases have since come to light where an indigenous flint aggregate was used. ASR damage has been identified in six Portuguese dams and in a viaduct, while pavement cracking due in part to alkali-silicate reaction was identified in 1993 at the airport of Santa Maria, Azores. Thus the problem continues to appear in countries previously thought to be unaffected. To date, no cases of deleterious reaction have been reported in certain countries, such as Austria (Krispel 2008) and the Republic of Ireland (Richardson 2003). Experience in Finland led Lahdensivu et al. (2018) to postulate that countries hitherto regarded as free of reactivity most probably have very slowly reacting sites that will collectively become apparent in structures of similar age once the first deleterious case occurs. However, it must be remembered that it is not potentially reactive aggregates alone that determine the risk of deleterious reaction in a region. A combination of factors must be present. Sims and Poole (2017) note that, in the UK, the deleterious cases in the south and south-east can be traced back to an unfortunate coincidental mixture of

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chert (or flint) bearing fine aggregate together with non-flint coarse aggregate and a high-alkali Portland cement. In a change to previous practice, it was the comparative dilution of the chert or flint content in the aggregate combination that triggered the reaction, probably related to the “pessimum” effect, discussed later. The geographical and chronological pattern of recognising ASR damage in structures indicates that the reaction probably occurred in specific countries during slight changes in concrete practice. These changes unwittingly brought together hitherto uncombined material and environmental conditions conducive to ASR at a time before the reaction was fully understood. Examples of change include higher alkali levels consequent to change in cement chemistry, or the rise in cement contents of concrete. The use of new aggregate sources such as sea-dredged aggregates or the more extensive use of de-icing salts, may also have been a trigger. For example, the first reported case of ASR in Hong Kong (Wong and Koirala 1992) led to the identification of five cases of deleterious expansion, all of which were in structures constructed in the 1980s. Guidelines for minimising the risk of deleterious ASR have been progressively introduced and updated in most countries from the 1980s onwards. A comprehensive review of the situation across 11 major regions of the world has been collated by Sims and Poole (2017).

MANIFESTATION OF THE PROBLEM The reaction may occur in many concretes, but does not always cause damage. The aggregate particle decomposes at the reaction site and a distinctive gel is generated. The gel may then replace the decomposed aggregate, permeate into voids, or migrate into microcracks caused by expansion. The reaction sites can be at the aggregate-cement interface or within the aggregate particle. Significant damage due to ASR is an even more rare occurrence than ASR itself. Indeed some cases of damage previously attributed to ASR have subsequently been diagnosed as delayed ettringite formation or freeze-thaw damage. Although ASR has been detected in a few deteriorated concrete structures that have required demolition, other durability factors in these cases were often more significant, such as corrosion due to chloride ingress. The reactive nature of Danish aggregates is such, for example, that ASR gel will be found in cracks of most deteriorated structures, but ASR will not necessarily have been the cause of the cracking. Manifestation of the problem as a serviceability failure generally takes the form of cracking or differential movement. Some commentators also feel that there are wider durability effects of ASR cracking, aside from the aesthetic problem. ASR-induced cracking will slightly reduce the strength of the material in compression, tension, and flexure.

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Visible damage The crack pattern depends on the level of stress in the member and the disposition of reinforcement. Crack depths are normally in the range of 25–50 mm but can be higher. Cracks initially develop in a three-pronged manner to produce characteristic “Manx” cracks (Hobbs 1988). In an unreinforced zone, the cracks further develop to join up with one another and display the characteristics of map cracking illustrated in Fig. 11.2. This pattern results from the differential movement of the interior and exterior of the concrete member. The crack pattern is more rectangular where reinforcement lies close to the surface, as illustrated in Fig. 11.3, due to the influence of the main direction of stress. The greater the amount of reinforcement present in a member, the smaller the induced strains in the steel due to ASR expansion. This may restrain the amount of expansion and

Figure 11.2 Map cracking pattern associated with ASR in unreinforced zones.

Figure 11.3 ASR crack pattern associated with influence of reinforcement.

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will also influence the crack pattern. An Institution of Structural Engineers (1992) report emphasised the need for three-dimensional, well-anchored reinforcement detailing if severe ASR expansions are to be limited.

Impact on serviceability The expansion and cracking associated with ASR do not usually degrade its structural adequacy. Hobbs (1988) reported that the loss of compressive, tensile, and flexural strength due to ASR is of the order of 10–30%. The reduction in elastic modulus consequent on expansion ranges from 20 to 50%. The performance of ASR-damaged structures in full-scale load tests has been satisfactory. The Institution of Structural Engineers (1992) issued guidance on appraisal of existing structures. The report contains a flow chart for the appraisal and guidance on strength estimation, dependent on the level of expansion to date or predicted in the time frame of interest. Stresses may be induced by ASR expansion if restraints exist. Differential movement may lead to misalignment of members, and differential expansion may take place in separate pours of the same member. The expansion of individual members may not be uniform even if cast in one pour. Asymmetrically reinforced members may hog as a result of the bending moment induced by differential restraint. The durability question is a difficult one to address. It has been suggested that the introduction of cracks due to ASR in the cover zone of reinforced concrete members should render the reinforcement more vulnerable to corrosion. This is because the permeability of the concrete will rise, and depassivating influences such as carbon dioxide and chlorides will have easier access. However, the debate on the relationship between corrosion rate and cracking is still unresolved. This is because the corrosion rate is influenced by the location of anodic and cathodic sites on the bar. The latter will not necessarily occur at the worst location (i.e. at a crack). This is discussed further in Chapter 12. A related problem is that of freeze-thaw behaviour where ASR-induced cracks may allow greater water ingress, which then has the capacity to cause disruption if it freezes.

MECHANISM OF EXPANSION AND REACTION The mechanism of expansion is accepted as being the adsorption of water by the gel produced in the reaction. The water in the gel has a lower free energy than the water in the pore solution surrounding it and so a flow into the gel results. The reaction itself is less well understood, but a number of key aspects have been identified by Diamond (1976, 1989), Dent Glasser and Kataoka (1981), Chatterji et al. (1987), Hobbs (1988), Chatterji (1989a, 1989b), and Chatterji and Thaulow (2000).

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The reaction may be idealised as follows: 4SiO2 + 2NaOH

Na2 Si4 O9 + H2 O

The reaction is more complex but the formula above gives an indication of the primary compounds involved. The chemistry involves stages including the formation of negatively charged SiO– ions from the aggregate that attract positively charged ions including Na+ and K+ from the pore fluid. The alkali ions that diffuse into the aggregate grains exceed the counter diffusion of silica (SiO2), and so expansive pressures build up. The driving force is the concentration of alkali hydroxides. Diamond showed that, for a given water/cement ratio, there is a linear relationship between hydroxyl ion concentration and the alkali content of cement, expressed as sodium oxide equivalent. This is significant in two ways. First, the concentration of alkali hydroxides in the pore solution is related to the hydroxyl ion concentration, and this correlates with the finding that ASR occurs in conditions of high pH. Second, sodium and hydroxyl ions penetrate reactive silica more easily than calcium ions. High alkali cements result in pore solutions that contain significant amounts of dissolved sodium hydroxide (NaOH) and potassium hydroxide (KOH) compared to calcium hydroxide (Ca(OH)2). The solution is capable of attacking reactive aggregates to form dissolved silica. The structure of quartz involves a framework of silicon–oxygen tetrahedra. Each oxygen is shared between two silicons, while each silicon is bonded to four oxygen atoms. If the framework is well-crystallised, the attack takes place on the surface and the reaction is slow. However, if the material is poorly crystalline, the metal alkalis penetrate and rupture the silicon-oxygen bonds through hydroxyl attack. This may be represented as follows, where the symbol ≡Si-O represents a tetrahedra with three other oxygen atoms connected: Si

O

Si

+ OH =

Si

OH + Si

O

Chatterji and Thaulow argue that the silica grain attracts cations more strongly because of the surface charge density on the grain. They postulate that the cement-aggregate-pore solution system results in negatively charged cement particles and silica grains surrounded by positively charged cations. The silica grains are larger and attract the Na+, K+, and Ca2+ cations more strongly. The negative Si-O– attracts more Na+ and OH– ions into the silica, further loosening the structure of the silica lattice. The penetrating hydroxyl and sodium ions cause a breakdown of the Si-O-Si bonds. This liberates silica but also opens the grains to penetration by further ions including larger ones such as calcium. The diffusion of silica out of the reacting grains is controlled by the calcium ion content and

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expansion results as the ions penetrating the aggregate exceed the outward diffusion. This generates expansive forces. In the presence of calcium, an insoluble gel is formed and over time the residual concentration of alkali hydroxide reduces. The gel forms as the Na+ ions join the negative oxygen on the silica surface: Si

O

Si

+2Na+ + 2OH = 2

Si

O

Na + 2H2 O

The essential role of calcium in the reaction was highlighted by Chatterji et al. (1987) and in research by Struble, cited by Diamond (1989). First, the presence of calcium hydroxide is essential for sodium and other ions to penetrate into the reacting grain. Second, the reaction products would not form a gel unless there was a source of calcium – the silica would otherwise simply remain in solution. The gel contains substantial amounts of calcium. The resultant complex then imbibes water and causes swelling. Dent Glasser and Kataoka (1981) observed that the maximum volume expansion of silica gel in sodium hydroxide solution occurs at an intermediate level of total SiO2/Na2O mole ratio. This may account for the pessimum effect observed in some concretes, whereby damaging expansion occurs at an intermediate silica level. PRIMARY FACTORS INFLUENCING THE REACTION Four components are required for a deleterious alkali-silica reaction to occur. These are as follows: • • • •

an adequate supply of moisture reactive silica a sufficiently alkaline pore solution proximity of soluble calcium.

All four must be present simultaneously. The moisture is usually from the environment. The reactive silica is usually from the aggregate. The alkalinity of the pore solution primarily derives from the cement, but can be increased by sources such as admixtures or external salts. Development to the point of damaging expansion requires calcium in the cement paste to sustain the ongoing formation of gel.

Moisture Moisture plays a dual role in the reaction. First, it acts as a transport route for the reactive ions. Second, the gel produced by the reaction will imbibe

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water if sufficient moisture is available. This leads to expansion that may cause damage. Thus the level of moisture controls the difference between the occurrence of ASR as such, and the occurrence of deleterious ASR. Deleterious ASR has generally been observed in cases where an external supply of moisture was available. Opinions vary on the minimum relative humidity value for damaging expansion, but a figure of 80% is commonly quoted.

Reactive silica Most aggregates used in the production of concrete are not pure, in the sense that they may contain a proportion of other mineral components. These other components may include a small percentage of reactive silica. This is especially true of fine aggregate. Indeed it has been estimated by Glasser (1991) that the average chemical composition of the accessible portion of the earth’s crust is about 65% silica by weight. Silica, besides occurring as a solid crystalline oxide such as quartz, also occurs in the form of a silicate whereby silicon and oxygen are combined with other elements such as calcium, potassium, and sodium. The silica in the reaction usually comes from the aggregates, but it must be noted that not all forms of silica are reactive. The degree of reactivity of silica minerals depends on the crystal structure, with disordered structures being the most vulnerable. This is because the more disordered the structure, the greater the surface area available for reaction. The crystalline structure will depend on the method of formation and geological history of the aggregates. Vulnerable silica structures include amorphous, glassy, crypto-crystalline, and microcrystalline. Unstrained quartz, the most abundant form of silica, is well-ordered, stable, and unlikely to react. Particular caution is required where petrographic reports indicate the presence of potentially reactive constituents: opaline silica, crystobalite, tridymite, microcrystalline quartz, crypto-crystalline quartz, chalcedony, and volcanic glass. Petrographic reports may indicate the presence of flint, chert, or greywacke. Damaging ASR in some countries has been linked to these. Crypto-crystalline quartz and chalcedony may be present in flint and chert. Microcrystalline quartz may be present in greywackes. A useful review of the nature of these reactive forms of silica was published by Diamond (1976). The issue of differentiating the reactivity of cherts and flints helps to emphasise the point that it is the crystalline structure of minerals in aggregates that is more significant than the aggregate type. For example, Strogen (1993) found significant differences in the crystallinity of English flints of cretaceous origin and Irish carboniferous cherts three times older. The highly reactive cretaceous flints had been subject to temperatures up to

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200oC, whereas the carboniferous cherts did not react in-service, having been effectively annealed through exposure to temperatures up to 350oC. The cherts were more crystalline, with quartz crystallinity indices averaging 7.0 compared with an average of 2.4 for the flints. The difference was reinforced in further studies of domain size by Strogen (McNally et al. 2004), which indicated 487 to 801 Å for the cherts and 271 to 319 Å for the flints. While being a useful tool in the detection of large-scale differences between well crystalline and poorly crystalline silica, the authors observed that domain size alone is not enough to reliably predict expansion due to ASR. Marinoni and Broekmans (2013) further studied the crystallinity topic and derived microstrain values from the diffraction profile. Based on a study of 12 Italian sedimentary rocks, they found that large microstrain values and small crystallite size indicated a high potential for alkali reactivity and, conversely, that large crystallite size and small microstrain values are an indicator of a low potential to develop ASR. Taken together, the Irish and Italian studies reinforce the importance of involving expert petrographers with experience in aggregates for concrete in the analysis of candidate sources for construction materials in the context of both durability and sustainability. It is worth noting that it is the aggregate combination – fine and coarse – that must ultimately be assessed. This is because the proportion of reactive silica in the combined aggregate determines the degree of reactivity. In certain cases, reaction will occur only if the proportions lie within a critical band. Proportions less than and more than the critical amount will not lead to reaction (the ‘pessimum effect’). One must be conscious of the potential abuse of the term “reactive aggregate”, commonly used to describe one that contains a potentially reactive component. The term gives the impression that the aggregates are in some way substandard, but the reaction will occur only if sufficient moisture and a high level of alkalis prevail. Even “reactive aggregate” will not react unless the other components are in place, yet one never finds reference to “reactive cement” or “reactive moisture”!

Alkali level The reaction will proceed only if a certain level of alkalinity in the pore fluid is reached. Even at very low water/cement ratios, water may be found in micropores in the cement paste and at interfaces of aggregate and hydrated cement particles. This pore fluid is distinct from the large diameter pore structure characteristic of higher water/cement ratio concretes. The alkalinity (hydroxyl ion concentration) of the pore fluid is primarily influenced by the sodium and potassium alkali metals in the cement and by the cement content. The alkali metals primarily derive from the characteristics of the raw feed materials used in the manufacture of cement.

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Secondary influences, such as external sources of salt and admixtures, are also contributory. The alkali content of concrete is determined in kilogrammes per cubic metre through the summation of alkalis contributed from the cement, aggregates, admixtures, and additions: [A]concrete = [a]cement + [a]aggregates + [a]admixtures + [a]additions

where [A]concrete = alkali content of concrete (kg/m3) [a]x = alkali contribution of component ‘x’ (kg/m3) In the context of ASR, the reactive alkali content of a Portland cement is represented by the “acid soluble equivalent sodium oxide content”. It is a convention to express the alkali metal concentrations in terms of their oxides. This value, commonly referred to as the alkali content of cement, is calculated through a summation of the weight percentage of sodium oxide (Na2O) and weight percentage of potassium oxide (K2O), while taking account of their relative atomic masses. Thus the alkali content of cement, expressed as a percentage, is: (Na2 O)equiv = (Na2 O) + 0.658(K2 O)

where (Na2O)equiv = acid soluble equivalent sodium oxide content of cement (%) (Na2O) = sodium oxide content of cement (% by weight) (K2O) = potassium oxide content of cement (% by weight). The value of 0.658 derives from the relative atomic masses of Na2O (61.979 amu) and K2O (94.230 amu). The reason for this approach is related to the significant role of the hydroxyl ions in the reaction. The concentration of hydroxyl ions produced by an alkali metal oxide is proportional to the molecular weight. The alkali content contributed by the cement is calculated as follows: [a]cement = (Na2 O)equiv (c /100)

where [a]cement = alkali content of cement (kg/m3) (Na2O)equiv = acid soluble equivalent sodium oxide content of cement (%) c = cement content (kg/m3).

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The acid soluble equivalent sodium oxide content can be reported by the cement manufacturer in a number of ways, and may or may not include an allowance for variability. A common method would be to report the “certified average alkali content”, defined as the average of the last 25 determinations of alkali content carried out on consecutive daily samples. Should an allowance for variability be required, this may be included by adding a factored number of standard deviations, typically 1.64. The value may safely be regarded as a maximum, as not all the alkalis are necessarily available for reaction. Alternative approaches also exist. For example, a cement manufacturer may use a “declared mean” approach, whereby the alkali level would not be continuously reported, but the user would be notified if alkali levels crept above a predetermined value. The alkali content of Portland cements produced throughout the world lies in the range of 0.3–1.5%, and awareness of the ASR phenomenon has generally led to a decrease in the average value. In the Republic of Ireland, for example, the average values for domestic production of Portland cement have progressively declined from values in excess of 1% to about 0.6%. Cements with alkali values of 0.6% and less are described as “low alkali” cements. It has been the practice in many countries to assume that concretes made with “reactive aggregates” would not suffer ASR-induced damage if used with a low-alkali cement because the hydroxyl ion concentration may be one-tenth of that found in concrete made with a high-alkali cement. The alkali contribution from aggregates needs to be taken into account if they contribute to the alkalis through, for example, sodium chloride content or feldspar. Sea-dredged aggregates would be an obvious source of concern, and some countries now insist on washed aggregates, but careful monitoring would also be required in the case of coarse recycled aggregate and coarse recycled concrete aggregate. Some national guidelines state that the contribution from this source may be ignored if the chloride ion content is less than 0.01%. The contribution, if any, may be taken into account by determining the percentage by weight of chloride ion and factoring this by 0.76. The value of 0.76 derives from a consideration of the composition of seawater. This figure may then be multiplied by the mass of the aggregate to yield the alkali contribution of the aggregate. Thus: [a]aggregate = [a]fine + [a]coarse [a]fine

= 0.76 (Cl )fine (Fagg /100)

[a]coarse

= 0.76(Cl )coarse (Cagg /100)

where (Cl-)fine = percentage by mass of chloride ion, dry fine aggregate (%)

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(Cl-)coarse = percentage by weight of chloride ion, dry coarse aggregate (%) Fagg = fine aggregate content (kg/m3) Cagg = coarse aggregate content (kg/m3). Potential problems with the release of alkalis from feldspar have been highlighted by Poulsen et al. (2000) and awareness of alkalis released from aggregates is specifically addressed by a RILEM-recommended test method (Menéndez et al. 2021). The contribution of alkalis from admixtures is included by reference to its sodium oxide equivalent and the dosage rate. Thus, for example, if the admixture was used at the recommended dosage rate, the expression might be: [a]admixture = {(Na2 O)equiv }admixture (c /100)

where {(Na2O)equiv}admixture = sodium oxide equivalent of admixture in kilograms per 100 kilogrammes of cement when used at the recommended dosage rate (kg/m3). The issue of alkalis contributed by additions has been a source of difference in guidelines for minimising the risk of damaging ASR. In the past, some guidelines recommended that the alkali contribution of PFA and slag be ignored; others recommended that it be taken as the water-soluble alkali content. Most current guidelines require that it be determined from the total acid soluble alkali content of the addition. The precise fraction of the total acid soluble alkali content of the addition depends on various circumstances, such as the proportion of the addition relative to Portland cement. Examples of the fraction to be used are discussed further in a later section on management of the risk.

Soluble calcium The role of soluble calcium is in a process referred to as “alkali recycling”. The ongoing formation of gel is a slow deleterious process as the concrete ages, as opposed to the reaction between the alkalis and the reactive aggregates being a one-off early effect in the service life of a concrete element. Rajabipour et al. (2015) explained that as a portion of the gel migrates away from the aggregate particle, it encounters a calcium-rich environment in the cement paste. A cation exchange occurs, with calcium replacing the sodium and potassium in the gel. These alkalis are thereby recycled and available to regenerate the reaction, which continues slowly until the available components are exhausted. The rate of reaction is slow enough to be sustained for decades, all the time adding to the deleterious expansive gel.

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ADDITIONAL CONSIDERATIONS INFLUENCING ASR Besides the four fundamental issues of alkalis from the cement, reactive silica from the aggregate, soluble calcium, and moisture level, the following factors are of note: • • • • •

the pessimum effect the effect of fly ash, slag, microsilica, metakaolin, and lithium salts external salts temperature effects aggregate porosity.

Pessimum effect The pessimum effect is exhibited by expansions caused by ASR. At low- and high-reactive silica concentrations, the expansions may be low. The maximum expansion occurs at an intermediate concentration. This phenomenon is illustrated in Fig. 11.4. This phenomenon complicates the design and interpretation of expansion tests used to assess the suitability of an aggregate or aggregate combination. Nevertheless, variations of standard tests have been formulated to allow the investigation of a possible pessimum effect for an aggregate combination. There are several possible explanations for the phenomenon. One is that at low concentrations the volume of gel produced is too low to exhibit significant expansion, whereas at high concentrations the reaction is vigorous but exhausts itself before the concrete has fully hardened. The pessimum proportion arises from the combination of coarse and fine aggregate. Thus an inert aggregate may contribute to the problem if it is combined with a reactive aggregate such that the combination results in a pessimum level. The answer to avoiding damaging ASR in cases where

Figure 11.4 Pessimum effect, showing a lower expansion at both low- and high-reactive silica contents than at an intermediate level.

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aggregate contains significant quantities of chert or flint may be to use a combination of coarse and fine aggregate that possesses a chert content high enough to exceed the pessimum.

Effect of fly ash, slag, microsilica, metakaolin, and lithium salts Secondary cementitious materials, such as PFA, ground granulated blastfurnace slag (ggbs), silica fume, metakaolin, and lithium salts can be beneficial in controlling expansion due to ASR if used in specific proportions. The effect of mineral and chemical admixtures was reviewed by Hobbs (1989) and by the Building Research Establishment (BRE 2002). Consideration of guidelines on the role of PFA and slag in minimising damaging ASR has been reviewed by Blackwell (1997). The silica in the PFA and ggbs can react with alkalis in cement with consequent lowering of the alkalinity, which in turn reduces the risk of damaging ASR. These materials can contribute some alkali to the pore fluid, but much is too tightly bound and will not be released. Chatterji (1989b) stated that the effect of the pozzolan was to combine with the calcium hydroxide, thus lowering the OH– and Ca2+ ion concentrations and consequently decreasing ASR activity and expansion. Furthermore, such additions would lower the calcium hydroxide content and allow more silica to diffuse out of the reactive grains. The total alkali content of fly ashes covers a wide range from about 0.7–7.8%, expressed as equivalent sodium oxide. At low replacement levels, the effective alkali contribution from a fly ash can be high and could have a detrimental effect on expansion. However, at intermediate replacement levels of approximately 20–25%, the effective alkali contribution is about one-sixth. At higher levels, there may be no effective alkali contribution, but research at present is limited to PFA with a maximum equivalent sodium oxide level of about 5%. The coarser the PFA the less effective it is, requiring minimum replacement levels of 30% for coarse material. The range of the alkali content of slags is about 0.3–2.6%, expressed as equivalent sodium oxide. At low replacement levels, the effective alkali contribution from a slag can be high enough to have a detrimental effect on expansion. At intermediate replacement levels of approximately 25–40%, the effective alkali contribution is about one-half. At higher levels, there may be no effective alkali contribution, but research knowledge is biased toward slags with equivalent sodium oxide levels below 1%. Microsilica is very effective in reducing the concentration of alkali ions. The range of equivalent sodium oxide content of microsilica is 0.3–5.5%, and the material contains between 60% and 98% by mass of amorphous silica. This has the beneficial effect of promoting a rapid reaction with the formation of a calcium-alkali-silicate hydrate, if used at a content of 5–20% microsilica by weight of cement. The difficulties caused by highly reactive

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aggregates in Iceland were overcome by the inclusion of silica fume in Icelandic cement. An initial level of 5% was temporarily increased to 7.5% before a more holistic approach was taken to blend aggregates of differing reactivity for use in concrete, allowing the level of silica fume in the cement to stabilise around 6%. A similar strategy was proposed for the highly reactive aggregates of Turkey, where research with 8% silica fume demonstrated satisfactory performance. Lewis and Bayrak (2016) reported a 50% reduction in expansion during 14-day tests for every 2% addition of silica fume. This opens the door to the potential use of 9–11% silica fume in combination with Turkey’s most reactive aggregates, well above the 6–8% value generally regarded as optimal in international practice. An alternative would be to research the effectiveness of a blend with slag and PFA at a lower level of microsilica. Significant reductions in expansion were reported in tests with an Icelandic reactive aggregate, when 10% silica fume and 25% natural pozzolan were combined (Gudmundsson and Olafsson 1999). Metakaolin, a purpose-made pozzolana, when used with Portland cement in concrete reacts with the calcium hydroxide to produce extra calcium silicate hydrates and calcium alumino-silicate hydrates. Jones et al. (1992) demonstrated that the reduction in calcium hydroxide content would suppress damaging expansion due to ASR. The replacement level would typically be at least 10%, but up to 15% can be used to beneficial effect. There is evidence of the beneficial influence of lithium nitrate (LiNO3), lithium carbonate (LiCO3), and lithium hydroxide monohydrate (LiOH.H2O) additions. Research spanning 60 years was reviewed by Rajabipour et al. (2015), who postulated that lithium reduces the dissolution rate of silica, impedes formation of silica gel, and reduces swelling. The lithium is incorporated into the alkali-silica gel. The dosage of lithium needs to be matched to that required in controlling the reactivity of the mineralogy of the aggregate. This equates to a molar ratio of lithium to sodium plus potassium in the range 0.7–1.2. Tremblay et al. (2010) postulated that the presence of LiNO3 led to a higher chemical stability of the silica.

External salts Alkalis from external sources may be introduced into concrete. Nixon et al. (1987) confirmed that addition of sodium chloride to cement paste rapidly converts to alkali hydroxide. This adds to the alkali load, and so exacerbates the threat of ASR. Hobbs (1988) notes that sodium chloride in the mix reacts in one of two ways that increase the hydroxyl ion concentration. Sources of external alkalis include road de-icing salts, seawater, and industrial alkalis. Foundations and other buried concrete may be subjected to alkalis in groundwater. Chatterji (1989a) describes the reaction as being of the following type:

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Reactive silica + 2Na+ + 2Cl + Ca2+ + 2OH + aq . = solid alkali – silica hydroxyl complex + Ca2+ + 2Cl + aq .

where aq. = aqueous solution. The smaller Na+ ions and OH– ions penetrate the reactive aggregate and break down the ≡Si-O-Si≡ bonds, thereby opening the grains to penetration by more Na+, OH–, and the larger Ca2+ ions, allowing part of the reactive silica to go into solution. It may be seen that the presence of sodium chloride (NaCl) contributes to the process.

Temperature effects Reaction rates and expansion rates are high at elevated temperatures, but decline with time. At low temperatures, the rates are slower, but total expansion may eventually reach or exceed that at higher temperatures (Fig. 11.5). The accelerating effect of elevated temperature is a feature of tests used in aggregate assessment. Hobbs (1992) found that the reaction occurred seven times faster for specimens stored at 38°C than for those stored externally at an average temperature of 9°C. The rate was four times faster than for samples stored at 20°C. The reaction generally tends to mature and cease in about 20 years, but longer periods may be expected in colder climates and shorter periods in hot countries.

Aggregate porosity The porosity of the aggregate can be significant. It has been found that if a potentially reactive situation occurs through an aggregate combination at

Figure 11.5 Influence of temperature on ASR-induced expansion, showing that overall expansion at high temperature may be lower than that at low temperature, despite high initial rates.

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pessimum level (BRE 2004a), that low-porosity coarse aggregate will produce more damaging expansion than a high-porosity material.

MODELLING AND SERVICE LIFE PREDICTION Modelling the rate of alkali-silica reaction and the form of the expansion curve has been studied by, for example, Hobbs and Gutteridge (1979) and Hobbs (1981). However, the application of mathematical models of ASR to specifications and service life prediction is not seen as being appropriate at this stage. The approach envisaged in the medium to long term remains prescriptive. Indeed, even the prospect of harmonising the prescriptive approach in Europe, let alone internationally, is thought to be unrealistic due to the influence of local factors on the reaction. A performance-based approach is eventually envisaged, through a combination of aggregate assessment tests combined with locally-derived acceptance criteria on expansion, or on alkali levels that trigger a reaction.

TEST METHODS FOR DETECTION OF ASR AND AGGREGATE REACTIVITY The rather surprising flood of deleterious ASR cases some decades after its first diagnosis led to a catch-up scenario in respect of diagnostic tests for existing cases and designing tests to minimise the risk of its occurrence in future works. Many countries responded with test methods based on international research. A significant contribution to harnessing these international developments and converging best practice in testing was conducted by successive RILEM technical committees steered by Dr. Philip Nixon (chair) and Dr. Ian Sims (secretary) over 25 years. The first of these was RILEM Technical Committee TC106-AAR, which produced an integrated assessment scheme (Sims and Nixon 2001). A summary of the recommendations from successor committees was published in 2016 as a state-of-the-art report, edited by Nixon and Sims (2016). This work continues through such RILEM technical committees (Wigum et al. 2016; Rønning et al. 2021b). Commonly used tests are based on the expansion of mortar-bars or concrete prisms. Broadly, testing may be considered under the following headings: • diagnosis of ASR as the source of damage • assessment of aggregate for potential reactivity by petrographic examination • assessment of aggregate for potential reactivity by mortar-bar tests • assessment of aggregate for potential reactivity by concrete prism tests

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• assessment of potential alkali-carbonate reactivity • performance testing of aggregate-binder combinations for resistance to ASR • determination of alkalis releasable by aggregates in concrete.

Diagnosis of ASR as the source of damage Confirmation that ASR is the cause of cracking in a structure requires expert investigation, as it involves detection of three key features: • presence of gel in cracks • presence of aggregate particles petrologically identifiable as belonging to a “potentially reactive” classification • presence of internal cracks in one of two patterns recognisable to experts in the field of ASR diagnosis. Diagnosis of ASR is difficult because many of the features of ASR cracking are common to other causes, such as drying shrinkage, freeze-thaw damage, leaching of calcium hydroxide, and efflorescence. The potential for confusion with efflorescence arises even though the gel produced by ASR exudes a transparent or brownish colour, because on exposure to the atmosphere the gel carbonates and turns white. A commercially available kit for the detection of ASR gel is available, but on-site testing alone is not generally adequate to conclusively prove ASR. A number of points are worthy of consideration in relation to ASR laboratory testing. Laboratory staff of geologists, scientists, technicians, and engineers should be experienced in the diagnosis of ASR. Tests involving elevated temperatures and humidity to accelerate expansion may not reliably model in-service performance of every aggregate combination. Finally, the costs of a thorough examination may not be economically justifiable. Excellent references on the topic of diagnosis have been published by the British Cement Association (1992) and RILEM (Godart et al. 2013). The procedure may be considered under four distinct headings: preliminary investigation, site investigation, sampling, and laboratory testing. Preliminary investigation involves the determination of pertinent matters from records covering the age of the structure, sources of aggregate, mix design, and exposure conditions including rainfall and exposure to salts. The site investigation comprises an examination of the crack patterns and checks for discoloration in the form of a light-coloured zone either side of the crack, with a dark colour along the edges of the crack itself. Gel exudation is checked for. Examination of hardened concrete to ASTM C856 (ASTM 2020b) includes a method for detecting ASR gel through application of uranyl-acetate and viewing under ultraviolet light. The presence of any pop-outs should be examined to see whether the base of the spalled area coincided with a reactive aggregate particle. A survey may reveal differential movement. Sampling

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extends to recovery of cores representing sound, typical, and damaged concrete. The samples should be wrapped in clingfilm for despatch to the laboratory. The BCA recommends that laboratory testing first involves wetting the surface of the core under running water and wrapping it in clingfilm for overnight storage. The next morning, it would be unwrapped and allowed to dry. The presence of any sweaty patches may indicate local ASR reaction sites. Microscopical examination would then be used to examine the material and internal crack patterns. Definitive confirmation of ASR is a highly specialised process. Stereomicroscopy and examination of thin sections must be entrusted to a petrographer with experience of aggregates for concrete and the AAR phenomenon.

Assessment of aggregate for potential reactivity by petrographic examination Petrographic examination is an essential tool in identifying the presence and concentration of any potentially reactive constituents. The importance of using a petrographer experienced in concrete technology ─ and alkali-aggregate reaction in particular ─ cannot be overemphasised. Methods commonly used for petrographic examination are in accordance with standards such as ASTM C295 (ASTM 2019a), BS 812: Part 104 (BSI 2021a) together with BS 7943 (BSI 1999b). A RILEM test procedure AAR-1 (Nixon and Sims 2016) is also available, supported by a comprehensive petrographer’s atlas (Fernandes et al. 2016). The procedures are based on visual examination of the sample to identify and quantify rock and mineral constituents that might be alkali-reactive. Two principal techniques are employed. The choice of technique is usually dependent on the particle size and the complexity of the sample, but both may have been used in many cases. The first technique involves visual inspection of the sample, and the second involves point counting within a thin section. In the case of aggregate where not all of the mineral constituents can be identified by these two techniques, further testing may be carried out, for example, by chemical analysis or x-ray diffraction analysis. The RILEM AAR-1 procedure proposes that the aggregate combination be classified into one of three categories: • Class I: unlikely to be alkali-reactive • Class II: potentially alkali-reactive • Class III: very likely to be alkali-reactive. The Class I aggregates are automatically cleared for use in concrete. The Class II and III aggregates could additionally be subjected to rapid screening tests and long-term confirmatory tests to assess the level of risk in their use. To follow this route, the aggregates would need to be further categorised under subclasses, based on the extent to which the material was siliceous

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(II-S, III-S), carbonate (II-C, III-C), or containing particulate constituents that pose a threat of silica and carbonate reactivity (II-SC, III-SC). The results would dictate the direction of further investigation in respect of selection of the relevant mortar-bar or concrete prism tests. The categorisation of aggregates in some countries is based on reactivity, defined in terms of the threshold alkali level. Three categories have been defined based on experience in-service: • low reactivity • normal reactivity • high reactivity. For example, in the UK, a listing of rocks and minerals by reactivity class was published by the Building Research Establishment (BRE 2004b).

Assessment of aggregate for potential reactivity by mortar-bar tests Research in South Africa led to the development of an accelerated test that serves as a rapid screening test (Oberholster 1983; Oberholster and Davies 1986). This has served as a basis for international practice in standards development. Examples include ASTM C1260 (ASTM 2022e), ASTM C1567 (ASTM 2022g), ASTM C441 (ASTM 2017b), AS 1141.60.1:2014 (Standards Australia 2014a), and Canadian standard A23.2–25 A (CSA 2000a). The RILEM assessment recommendation is AAR-2 (Nixon and Sims 2016). The accelerated mortar bar test (AMBT) involves constructing mortar bars to a prescribed mix recipe to test a single aggregate. The bar size is either 25 × 25 × 285 mm or 40 × 40 × 160 mm. Fine aggregate is prepared for test by washing. Coarse aggregate is crushed to yield defined percentages of different sizes. A grading of 0–4 mm is used. The test is conservatively severe. The specimens are immersed in sodium hydroxide solution at a temperature of 80°C for 14 days. Expansion is monitored, and aggregates are classified according to local limits in the place of use. The nature of local cements and aggregates requires the setting of local interpretation limits on a country-by-country or region-by-region basis. A review of international practice, illustrated in Fig. 11.6, correlates reasonably well with guidance in documents such as ASTM C1778 (ASTM 2022h) and Canadian standard A23.2–27 A (CSA 2014b). The following limits at 14 days, for 25 mm square bars, are indicative of widely accepted interpretation of results: < 0.10% expansion at 14 days: non-reactive 0.10–0.30% expansion at 14 days: moderately reactive 0.30–0.45% expansion at 14 days: highly reactive > 0.45% expansion at 14 days: very highly reactive.

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Figure 11.6 Indicative limits, accelerated mortar bar test.

The limit for non-expansive classification using 40 mm square bars was put at 0.08% maximum expansion. An example of a variation to take account of local conditions is the case of Iceland. The upper limit for innocuous classification is 0.20% if Portland cement is used in the test, rather than the Portland cement-silica fume combination often used in practice to suppress reactivity. The test has been found to be an effective screening tool for slowly reacting aggregates or those that produce expansion late in the reaction. One note of caution has been highlighted – the test fails to detect the reactivity of aggregates containing more than 2% porous flint. The lack of expansion may be due to the accommodation of the reaction product within the mortar bar. Although designed to test a single aggregate, an interesting variation of the method is to set up a series of tests with the aggregate to explore the possible existence of a pessimum level. This can be achieved by combining the aggregate under test with an unreactive one in proportions ranging from 5% to 100%. The ASTM C1260 test uses Portland cement in the screening of an aggregate’s potential reactivity. The ASTM C1567 test permits evaluation of the aggregate with pozzolan or slag binders. A limit of 0.10% maximum expansion is suggested for prediction of innocuous behaviour of the aggregate in-service if combined with a secondary cementitious material. This could then be evaluated in a more definitive concrete prism test. RILEM has designed a version of the test to detect possible alkali-carbonate reactivity as RILEM method AAR-5 (Nixon and Sims 2016). This is discussed later in this chapter.

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The mortar bar test has also been adapted to screen the effectiveness of a pozzolan or slag binder in controlling ASR. The ASTM C441 test explores this by combining the binder with a reactive silica in the form of borosilicate glass as the “aggregate”. The storage conditions and duration differ from the accelerated tests. Building on experience with the ASTM C227 test, now withdrawn, the specimens are stored over water in a sealed container and monitored at intervals from 14 days to 12 months, or longer if required.

Assessment of aggregate for potential reactivity by concrete prism tests Assessment of individual aggregates for their potential reactivity is very helpful in screening out potentially deleterious components of concrete. However, such a conservative approach on its own is unduly restrictive in circumstances where alternative aggregates are not readily available. Similar geological formations may cover vast areas and be relied upon as a source of building materials. Unduly restricting the use of “potentially reactive” aggregates may not be appropriate for reasons of economy and sustainable development. Thus, a suite of concrete prism tests (CPT) has been designed as reference tests on individual aggregates, and to further investigate the reactivity of aggregate-binder combinations from which safe mix designs may be determined. The long-term reference tests involve monitoring the expansion of concrete prisms stored for a year or more in an atmosphere conducive to alkali-aggregate reaction. The tests are of necessity long term to serve as the reference method. Similar versions are available in standards such as ASTM C1293 (ASTM 2020d), AS 1141.60.2 (Standards Australia 2014b), BS812: Part 123 (BSI 1999a), and Canadian standard A23.2–14 A (CSA 2000b). A similar methodology is included in the RILEM assessment recommendations as AAR-3 (Nixon and Sims 2016). An accelerated version of the test has been proposed by RILEM as AAR-4 (Nixon and Sims 2016). The concrete prism test also forms the basis of a series of RILEM tests, each designed to explore the effect of a particular parameter, such as additions (AAR-10), alkali levels (AAR-10), candidate aggregate-binder combinations (AAR-11), and external alkalis (AAR-12). Provision may be made to compensate for the possible effect of leaching on the specimens behaviour through a wrapping technique (AAR-13). In the reference tests, prisms of 75×75×250 mm are made up to a standard recipe, including an allowance for addition of a controlled amount of NaOH to bring the alkali content of the mix up to a defined level, expressed as (Na2O)equiv. multiplied by mass of Portland cement. The aggregate under test is combined with a non-reactive one. A coarse aggregate under test would be combined with an innocuous fine aggregate which demonstrated satisfactory performance in a screening test ( 0.240% expansion: very highly reactive.

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Figure 11.7 Indicative limits, concrete prism test.

As always, the specifier and producer must be mindful of requirements in the place of use. The RILEM accelerated concrete prism test AAR-4 (Nixon and Sims 2016) is based on work by French researchers. Although similar in many ways to the methodology of AAR-3, the storage temperature is much higher, at 60°C. The duration of the test is greatly reduced, at 15 to 20 weeks. Indicative interpretation of the limits is: < 0.03% expansion: non-expansive > 0.03% expansion: precautions required to prevent deleterious expansion in-service.

Assessment of potential alkali-carbonate reactivity The potential reactivity of carbonate rocks is a matter of concern in a number of geographical regions. The Canadian Standards Association has developed test method A23.2–26 A (CSA 2014a) to highlight chemical compositions that might indicate the risk of reactivity. A test on rock cylinders has been developed as ASTM C589 (ASTM 2019b). A concrete prism test to investigate aggregate binder combinations in the context of ACR is also available as ASTM C1105 (ASTM 2016b). RILEM has developed an accelerated screening test based on mortar bars as RILEM AAR-5 (Nixon and Sims 2016). The Canadian Standard test A23.2–26 A is used to investigate the composition of dolomitic rocks (CaMg(CO3)2). The first step involves determination of the content of calcium oxide (CaO), magnesium oxide (MgO), and aluminium oxide (Al2O3). The next step is to plot the CaO/MgO ratio against the percentage content of aluminium oxide (Al2O3).

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Interpretation of the result allows screening of the potential reactivity by plotting the result on a graph that has been calibrated with in-service experience. A reference concrete prism test may then be used to confirm the potential non-reactivity of an aggregate produced from the rock. The ASTM C589 test is conducted on small rock cylinders, 35 mm long and 9 mm diameter. The specimens are immersed in 1 N NaOH solution at room temperature. Ideally, the test duration is more than a year, but expansion after 28 days serves as a useful rapid screening test. Expansions above 0.10% at 28 days indicate potential for deleterious expansion in-service. The ASTM C1105 concrete prism test is a useful tool in the further investigation of material that has shown the potential for deleterious expansion due to ACR in the ASTM C589 test. A proposed aggregate-binder combination may be tested, and the results indicate whether preventive measures should be considered in-service. The specimens are stored in standard moist cabinet conditions of 23°C and relative humidity in excess of 95%. Expansion is monitored for a year, preferably, but some conclusions can be drawn after three and six months. The following guidelines are indicative of potentially deleterious expansion: > 0.015% after three months > 0.025% after six months > 0.030% after one year. The RILEM AAR-5 test results for ACR are interpreted in conjunction with the RILEM AAR-2 test for ASR when 40x40x160 mm specimens are used in both tests. The AAR-5 test uses only the 40x40x160 mm specimens, whereas two sizes are available in AAR-2. The aggregate is graded into a 4–8 mm size, as opposed to the 0-4 mm used in the AAR-2 test. Otherwise, the test methodology is similar to AAR-2. The results are interpreted as follows: < 0.08% in AAR-2 and expansion in AAR-5 < AAR-2: innocuous < 0.08% in AAR-2 and expansion in AAR-5 ≥ AAR-2: ACR reactivity > 0.08% in AAR-2 and expansion in AAR-5 < AAR-2: potential ASR > 0.08% in AAR-2 and expansion in AAR-5 ≥ AAR-2: potential ASR and ACR.

Performance testing of aggregate-binder combinations for resistance to ASR The basic RILEM test to determine the suitability of a candidate aggregate-binder combination is AAR-10 (Rønning et al. 2021a). The test is lengthy but there is a possibility of accelerating the process through the AAR-11 test procedure (Borchers et al. 2021). A further parameter, the effect of external alkalis, may be explored through the AAR-12 test (Borchers 2021). The AAR-12 test is a harsher version of the AAR-11 test. It has been designed to investigate the

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performance of aggregate-binder combinations which will be subject to de-icing salts in-service, where the criteria for acceptance in the AAR-11 test has not been calibrated with field performance in salt-laden environments. The RILEM AAR-10 test uses 100x100x400 mm concrete prisms and the humid storage conditions are kept at a temperature of 38°C. The duration of test is one or two years, depending on the variable being studied. Investigating a candidate binder, the key parameters under scrutiny are either the required minimum content of secondary cementitious material or the maximum alkali load permissible to achieve an acceptably low level of expansion with the aggregate proposed for the project. The RILEM AAR-11 and AAR-12 concrete prism tests are accelerated through humid storage conditions of 60°C and smaller specimen size of 75 × 75 × 250 mm. This allows a reduction of the test duration to 20 weeks. The AAR-12 test includes 10 cycles of exposure to sodium chloride solutions to mimic the additional effect of external alkalis from de-icing salts.

Determination of alkalis releasable by aggregates in concrete An additional threat to concrete durability in the context of AAR is the alkali releasable from aggregates into the pore solution over the long term, in addition to that calculated from the binder and other sources. The RILEM AAR-8 test (Menéndez et al. 2021) gives an indication of this threat by classifying aggregates into low-, medium- and high-threat groups. The test measures the alkali released from sodium and potassium ions from crushed fine and coarse aggregate (