Geotechnical Aspects of Underground Construction in Soft Ground: Proceedings of the Tenth International Symposium on Geotechnical Aspects of ... Cambridge, United Kingdom, 27-29 June 2022 [1 ed.] 0367337339, 9780367337339

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M.Z.E.B. Elshafie, G.M.B. Viggiani & R.J. Mair, editors

GEOTECHNICAL ASPECTS OF UNDERGROUND CONSTRUCTION IN SOFT GROUND

PROCEEDINGS OF THE TENTH INTERNATIONAL SYMPOSIUM ON GEOTECHNICAL ASPECTS OF UNDERGROUND CONSTRUCTION IN SOFT GROUND, IS-CAMBRIDGE 2022, CAMBRIDGE, UNITED KINGDOM, 27-29 JUNE 2022

Geotechnical Aspects of Underground Construction in Soft Ground

Editors Mohammed Z.E.B. Elshafie Qatar University, Doha, Qatar

Giulia M.B. Viggiani University of Cambridge, UK

Robert J. Mair University of Cambridge, UK

Cover top photo credit: Tideway Bottom cover photo: © Copyright Crossrail

CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business © 2021 ISSMGE, London, UK Typeset by Integra Software Services Pvt. Ltd., Pondicherry, India All rights reserved. No part of this publication or the information contained herein may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, by photocopying, recording or otherwise, without written prior permission from the publisher. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. Library of Congress Cataloging-in-Publication Data Applied for Published by: CRC Press/Balkema Schipholweg 107C, 2316XC Leiden, The Netherlands e-mail: [email protected] www.routledge.com – www.taylorandfrancis.com ISBN: 978-0-367-33733-9 (Hbk) ISBN: 978-1-032-02769-2 (Pbk) ISBN: 978-0-429-32155-9 (eBook) DOI: 10.1201/9780429321559 https://doi.org/10.1201/9780429321559

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Table of contents

Preface to the current edition

xiii

Organisation

xv

Sponsors

xvii

Field case studies and sensing technologies Ovalisation of cast-iron tunnels in response to nearby tunneling work M. Alhaddad & K. Soga

3

Lubrication characteristics of pipejacking in soft alluvial deposits W.-C. Cheng, G. Li & D.E.L. Ong

12

A recent subway construction incident in soft alluvial deposits of Taiwan W.-C. Cheng, G. Li & M.M. Rahman

19

Settlements due to tunneling in the City of São Paulo A. Lopes dos Santos, W. Bilfinger & H.C. Rocha

26

Evaluation of geological/geotechnical geostatistical models for tunneling applications R. Gangrade, W. Trainor-Guitton, M. Mooney & J. Grasmick

35

Pillar uplift observed in braced excavation of a subway station Y.B. Gao, Z.Y. Yang, T.J. Yao & X.M. Tang

43

Solution implementation for a tunnel collapse in a weak embankment soil: The case of the Ali Boumendjel Metro station, Algiers, Algeria R. Hebib, Z. Derriche, B. Alloul & D. Belhai

51

Design of large twin-wall cofferdams for ship impact B.D. Jones, E. Murphy & P.J. Astle

60

Design of large temporary works ship impact protection structures B.D. Jones, E. Murphy & P.J. Astle

68

Large-scale improvement work of subway stations in soft ground S. Konishi, T. Moriya, R. Fukuda, T. Murakami & K. Okanoya

74

Comparison between predicted and measured performance of a deep excavation in soft clay in Gothenburg, Sweden J. Langford, K. Karlsrud, E. Bengtsson, C. Hof & R. Oscarsson Grouting for the rescue of stuck TBM in conglomerate boulder layer H.-J. Liao, S.-J. Weng, C.C. Ho & R. K.N. Wong

v

83 91

Tunnelling in Urban Areas and Pile Interception Challenges – A Case Study: Bank Station Upgrade Project (BSCU) A. Nasekhian, C. Anthony, B. Haig, M. Dewhirst, J. Ares & C. Barker

98

Impacts of new development on existing underground assets using greenfield model K. Neaupane & Y. He

108

Cementitious systems with carbon nanomaterials for underground infrastructure I. Papanikolaou, A. Al-Tabbaa, M. Goisis & T. Embley

115

Pipe jacking tunnel construction crossing the Bang Pakong River K. Piriyakul, S. Pochalard & E. Rungrueng

123

Nine Elms Station substructure – implementation of the Observational Method by progressive modification Nigel Pye, C. Lile, Anthony O’Brien & C. Penh

129

Evaluation of InSAR data for measuring the surface settlement during shield tunnel construction of the North-South Line in Amsterdam K.J. Reinders, F.J. van Leijen, R.F. Hanssen & M. Korff

137

Deep excavations for buildings in the Sabana Formation Bogota J.A. Rodriguez

153

EPB tunneling and shaft construction in soft deltaic deposits for a railway link to Barcelona airport S. Sánchez, E. Silva, J. Izquierdo, A. Gens & E. Alonso

145

Use of jet grouted columns for deepening of basements and underpinning existing structures at Valkyrien in Oslo T. Sandene, K. Karlsrud & A. Worren

161

Settlements induced by EPB TBMs tunneling, a case study of theoretical and monitoring values A. Siemińska-Lewandowska & R. Kuszyk

169

Monitoring of an existing concrete-lined tunnel at CERN excavated in the molasse rock K. Soga, V. Di Murro, L. Pelecanos, C. Kechavarzi, L. Scibile, J.A. Osborne & R.F. Morton

177

Achieving sustainability in tunnelling through innovation A.H. Thomas

184

Measured post-construction ground response to EPBM tunnelling in London Clay M.S.P. Wan, J.R. Standing, D.M. Potts & J.B. Burland

191

Gilgel Gibe II Hydropower Project in Ethiopia; TBM Tunnelling, when the rock turns into mud: Analysis of a major collapse, its causes and solutions E. Zoppis & A.M. Baldi

199

Modelling and testing of tunnels and deep excavations in soft ground Torque estimation for tunnel boring machine design in soft ground M. Ambrosi, K.B. Glab & W. Broere Variability of soil stress-strain non-linearity for use in MSD analyses evaluated using databases of triaxial tests on fine-grained soils M.E.W. Beesley & P.J. Vardanega

vi

209

217

Asymmetric pressure distribution in EPB shields: Evaluation of measurements and numerical simulations A. Bezuijen, T.S. Dang & G. Meschke

226

Pore pressures in front of a slurry TBM, the influence of plastering mechanisms A. Bezuijen

234

Face pressure and volume loss relationships for pressurized tunneling in Granular soils S.J. Boone & J.N. Shirlaw

242

Albert Embankment: Design of deep excavations in the River Thames foreshore O. Brown, P. Stewart, S. Thomson, B. Patel, A.M. Waller, S. Sismondi & F. Quesada

250

Centrifuge and numerical modelling of the influence of structural stiffness on basement heave in over-consolidated clay D.Y.K. Chan & S.P.G. Madabhushi

259

Influence of TBM geometry on lining loads of deep tunnels V. De Gori, A. de Lillis & S. Miliziano

266

Development of a new excavation technique for centrifuge testing in sand N.E. Faustin, R.J. Mair & M.Z.E.B. Elshafie

275

Interpretation of soil parameters used for numerical analysis with small strain model for deep excavation in loose to medium dense sand H.B. Bin-Chen, P. Khac Hai & C. Hung

283

Prediction of damage intensity of reinforced concrete tunnels and soil against blast loading K. Senthil, S. Rupali & L. Pelecanos

291

EPB-TBM tunnel under internal pressure: Assessment of serviceability N.A. Labanda, A.O. Sfriso, D. Tsingas, R. Aradas & M. Martini

300

A risk assessment of downdrag induced by reconsolidation of clays after upwards pipe jacking N.A. Labanda, A.O. Sfriso, D. Tsingas, R. Aradas & M. Martini

309

Small-scale modelling of pile drilling in sand – investigation of the influence on surrounding ground E.J. Lande, S. Ritter, E.J. Lande, H. Tyvold & S. Nordal

317

Excavation of an artificial tunnel using compressed air N. Losacco, M. Cafaro & R. Marazzita

325

The use of adaptive smoothed finite-element limit analysis to seismic stability of tunnels H.C. Nguyen

330

Upper bound analysis of seismic stability of tunnels using cell-based smoothed finite element H.C. Nguyen

337

Investigation of the seismic performance of the complicated tunnel sections with non-uniform heights H. Nitta, S. Ito, T. Otsuka, S. Konishi, K. Tsuno, S. Tsuchiya & K. Maekawa Influence of the annulus grout on the soil-lining interaction for EBP tunneling M. Ochmański, G. Modoni & G. Spagnoli

vii

343 350

Long term additional load on a shield tunnel in soft clay due to clay consolidation with water leakage S. Oka, J. Saito, Y. Ito, W. Li, S. Kaneko, A. Afshani & H. Akagi A study of a strut-free excavation system in deep excavations C.Y. Ou, A. Lim, P.G. Hsieh & S.C. Chien

357 365

Increasing the passive resistance of deep excavations in very soft soils to mitigate ground movements through centrifuge modelling J.P. Panchal & A.M. McNamara

371

Machine learning algorithms applied to the blowout susceptibility estimation around pressurized cavities in drained soil F. P.-Ramirez & C. Arson

379

Specifying and Testing fibre Reinforced sprayed concrete: Advantages and Challenges of some testing methods Benoit De Rivaz

385

Undrained seismic response of tunnels E.A. Sandoval & A. Bobet

395

Preliminary evidences on the influence of grains micro-structural features on the TBM tools wear D. Sebastiani, S. Miliziano, G. Guida & F. Casini

401

Front-face pressure drop during the standstill phase for EPB mechanized tunnelling in coarse-grained soils D. Sebastiani, S. Miliziano & A. Bezuijen

408

A miniature EPB TBM for use in a geotechnical centrifuge C.J. Shepheard, A.S.N. Alagha, G.M.B. Viggiani & S.K. Haigh

415

Urban tunnelling in glacial soils: Tunnel de Champel, Geneva W. Steiner, T. Witschi & A. Ferrari

421

Simplified stress-strain models applied to data from triaxial and pressuremeter tests on London Clay P.J. Vardanega, M.D. Bolton, S.K. Haigh, R.W. Whittle, A. Klar & M.G. Williamson

430

Propped cantilever wall stability design with the ‘What You Design Is What You Get’ method – background and development C.K.S. Yuen

438

Multi-objective optimisation design for composite tunnel linings using non-dominated sorting genetic algorithm W. Zhai, D. Chapman, A. Faramarzi, H. Huang & D. Zhang

444

Support pressure transfer at a slurry supported tunnel face due to time dependent decrease of soil permeability C. Zhao, Z. Zizka, B. Schoesser, M. Thewes & A.A. Lavasan

451

SEM deformation prediction and observation by 3D numerical analysis H. Zheng, M. Mooney, M. Gutierrez & C. Bragard

459

Face stability of slurry shield-driven tunnel in an aquifer T. Xu, W.H. Zhou & A. Bezuijen

467

viii

Numerical analysis of Double-O-Tube shield tunneling in Shanghai D. Zhou & L. Zdravković Physical modelling of transient processes at the slurry supported tunnel face during shield excavation Z. Zizka, B. Schoesser & M. Thewes

475

482

Ground movements, interaction with existing structures and mitigation measures Interaction between a newly excavated underground ramp and deep existing tunnels A. Afshani, G. Hassan, H. Akagi & K. Endou

491

Pile driving interaction with existing tunnel K.J. Bakker, R. Spruit & F.C.M. van Overstraten Kruijsse

501

Numerical modelling of framed structures with masonry infills affected by tunnelling­ induced deformation and damage D. Boldini, N. Losacco, A. Franza & S.M. Miraei

510

Analytical investigation for the circumferential behavior of the shield-driven tunnel adjacent to a braced excavation H.Z. Cheng, R.P. Chen, H.N. Wu & F.Y. Meng

517

Winkler model for axial deformation of compressible piles due to tunnelling J.J. Crispin

523

Bayesian inference for deep excavations W.J. de Wolf, W.J. de Wolf, M. Korff, A. van Seters, & J.H. van Dalen

529

An underground excavation in Barcelona and its interaction with existing structures A. Di Mariano, A. Varga & A. Gens

536

Equivalent frame model for the assessment of tunnel-induced damage to masonry buildings D.B. Gulen, S. Acikgoz & H.J. Burd

545

Use of underpinning, horizontal jet grouting and ground freezing for ground stabilization to control settlement of existing MRT tunnels during construction of a link-way and railway tunnels for a new mrt line in Singapore Kaoru Hashida, Tadashi Hashimoto, Yong Kwet Yew, Ramesh Nair & John Busbridge Tunnel lining behaviors of close proximity large diameter parallel shield tunnels T. Hashimoto, J. Nagaya, M. Isa & K. Fujiwara Bolu NATM Tunnel, changes to the support system due to soft ground conditions and deformation S. Işık & Z. Buket Use of spile umbrellas to reduce deformation A. Klinger & J. Fillibeck

554 563

569 576

Modified gap method for prediction of TBM tunnelling-induced soil settlement in sand - a case study B.T. Le, N.T. Nguyen, S. Divall, R.J. Goodey & R.N. Taylor

ix

584

Long-term behaviour of ground around tunnel due to groundwater level fluctuations Wei Li, Shigeaki Oka, Alireza Afshani & Hirokazu Akagi

590

Numerical investigation into time-dependent effects on short-term tunnelling-induced ground response in London Clay A.R. López, A. Tsiampousi, D.M.G. Taborda, J.R. Standing & D.M. Potts

597

Semi-coupled modelling of soil-structure interaction during tunnel construction: Two case studies from Bank Station Capacity Upgrade A. Luciano, M.N. Pascariello, E. Bilotta, S. Acikgoz & R. Mair

605

Centrifuge tests on tunnel-building interaction in liquefiable soil G. Miranda, V. Nappa, E. Bilotta, S.K. Haigh & S.P.G. Madabhushi

613

The effect of deep excavation on existing railway tunnel M. Mitew-Czajewska

620

Simplified modelling of the transient response of underground structures due to dynamic loads generated from underground tunnels L. Pelecanos, K. Senthil & S. Rupali

627

Mapping the risk of building damage due to excavation-induced displacements S. Ritter, L. Piciullo, A.O. Kydland Lysdahl, M. Kahlström, J. Langford & F. Nadim

632

A new approach for compensation grouting in highly permeable gravel M. Sailer, J. Fillibeck & S. Geuder

640

A case study on the effects of anchor drilling in soft, low sensitive clay and sandy, silty soils T. Sandene, S. Ritter & E.J. Lande

647

Prediction of long-term settlement in shield tunnel using GA-BP neural network Yi-Ming Shen, Dong-Mei Zhang, Jie Zhang, Dong-Mei Zhang & Jie Zhang

656

Tunnelling through a piled foundation: Interaction effects Davor Simic & Belén Martínez-Bacas

664

The use of protective structures to reduce tunnelling induced damage to buildings G. Song, A.M. Marshall & C.M. Heron

673

Evaluation method on ground movement using continuum ground model M. Sugimoto, J. Chen, P.T. Anh, K. Manabe, L.G. Lam & S. Chaiyaput

681

The response of the Europier Terminal Building to the excavation of the T2B Basement at Heathrow Airport G.R. Taylor Longitudinal structural deformation of shield tunnels induced by overlying excavation H.N. Wu, S. Chen, R.P. Chen, Y. Liu, H.Z. Chen, F.Y. Meng & S.L. Shen

689 696

Simulating the water-soil leakage induced deformation around the shield tunnel with material point method X.C. Xie, D.M. Zhang, M.L. Zhou, S.J. Feng, D.M. Zhang & S.J. Feng

702

Influencing factors and protection technologies of underground diaphragm wall and deep foundation pit construction on metro station H.F. Xing, L.L. Liu & H. Zhang

710

x

A centrifuge modelling study on the effect of foundation configuration on tunnel-frame interaction J. Xu, A.M. Marshall & A. Franza

719

Modelling ground response to TBM tunnelling with active face support T. Xu, W.H. Zhou & A. Bezuijen

727

Tunnel face stability considering drainage and surface settlements C. Yi, S. Senent & R. Jimenez

732

Impact of subsoil spatial variability on deformations of immersed tunnel X. Zhang, X. Wu & W. Broere

738

Coupled elastoplastic analysis of the soil-pile foundation interaction induced by deep excavations C. Zheng, A. Franza & R. Jimenez

746

Design and application of ground improvement for underground construction Microcapsule-based self-healing cement stabilised clay B. Cao, C. Litina, L. Souza & A. Al-Tabbaa Application of biopolymer hydrogel for ground hydraulic conductivity control under pressurized conditions I. Chang, G-C. Cho & A.T.P. Tran

753

761

Soil enhancement via microbially induced calcite precipitation C. Konstantinou & G. Biscontin

765

Base grouting against uplifting water for a deep excavation in Taipei Basin H.J. Liao, S.J. Weng, S.H. Cheng & R.K.N. Wong

773

The largest tunnels in freshly consolidated soft soil: Tuen Mun Chek Lap Kok Link Subsea tunnels, Hong-Kong T. Lockhart

780

Soil conditioning for EPB tunnelling in coarse grained soils based on laboratory model tests A.S. Merritt, S.A. Jefferis & R.B. Storry

788

Influence of mineral or polymeric modification on bentonite-based tunnel face support P. Mianji, W. Baille, T. Wichtmann, R. Verst & M. Pulsfort

796

Compensation grouting for conventional tunnelling with low overburden at the Oberau Bypass Tunnel E. Neun & S. Sabew

804

Parameter selection for ground improvement works: Lessons from an instrumented site and numerical analysis G.A. Pittaro, S.H. Goh, C.F. Leung & N. Mace

812

Jet grouting for tunnelling at London Victoria Station P. Rutty, C. Prangley, I. Heath & F. Mimnagh

821

Unconfined compressive strength of sand-fines mixtures treated with chemical grouts G. Spagnoli, W. Seidl, E. Romero, M. Arroyo, R. Gómez & J. López

829

xi

Permeability characteristics of coarse-grained soil conditioned with foam for EPB shield tunnelling S. Wang, S. Huang, Q. Hu & Z. Liu

836

Stabilisation of Singapore soft marine clay using a novel sustainable binder for underground construction H. Yu, Y. Yi, R. Liu & N. Jiang

836

Experimental study of pore water pressure development in soil when foam infiltrates into saturated sand D. Zheng, A. Bezuijen & M. Thewes

848

Author index

855

xii

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Preface to the current edition

Technical Committee 204: “Underground Construction in Soft Ground” of the International Society for Soil Mechanics and Geotechnical Engineering (ISSMGE) was first established as TC28 in 1989 to provide a forum for interchange of ideas and discussion amongst representatives from different countries with an active interest in tunnelling and deep excavations in the urban environment. In 1994, under the Chairmanship of Prof. Keiichi Fujita, TC28 organised its first symposium as a satellite event to the ISSMGE International Conference in New Delhi. Since then, the host society of TC28 was handed over from the Japanese Geotechnical Society to the British Geotechnical Society (under the Chairmanship of Prof. Robert Mair), then to the French Geotechnical Society (under the Chairmanship of Prof. Richard Kastner) and then to the Nether­ lands Geotechnical Society (Prof. Adam Bezuijen is the current Chairman of TC204). Over the years TC28 (and TC204) has always kept its commitment towards publishing cutting edge research and practice concerning geotechnical aspects of the design, construction and analysis of deep excavations, tunnels and large underground structures in the urban environment. Particular emphasis has traditionally been placed on the development, effects and control of ground move­ ments, their interaction with existing structures, mitigation measures and risk management. The success of the New Delhi symposium in 1994 led to the organisation of eight more International Symposia on “Geotechnical Aspects of Underground Construction in Soft Ground” since then. These were held every three years in London (1996), Tokyo (1999), Toulouse (2002), Amsterdam (2005), Shanghai (2008), Rome (2011), Seoul (2014) and Sao Paulo (2017). The 10th symposium (IS-Cambridge 2020) was planned to be held in June 2020 in Cambridge, United Kingdom. The call for papers drew an overwhelming response and 278 abstracts were received, resulting in 112 technical papers accepted for publication in the proceedings from over 25 countries. However, the COVID-19 outbreak in January 2020 and the very rapid spread of the virus to all corners of the globe, with all the associated health risks, eventually prompted the symposium organisers to postpone the conference to June 2022. The organisers also decided to publish two versions of the proceedings. A first version (Version 1), this volume, which contains only the accepted technical papers (112 in total), is published almost 1.5 years prior to the actual symposium to ensure that the hard and valuable work of the authors is duly recognised. In 2022, just before the actual symposium, a second version of the proceedings (Version 2) will be published. This new version will contain all the components traditionally found in TC204 symposium proceedings including: all papers in Version 1 plus the written versions of (i) the Fujita Lecture, (ii) the Special Lectures, (iii) the Session Reports, and (iv) the Bright Spark Lecture. Version 1 will go out of print as soon as the complete Ver­ sion 2 is published. The themes for the two versions of the proceedings are in line with the terms of reference of Technical Committee TC204 and include: • • • • • •

Field case studies Sensing technologies and monitoring for underground construction in soft ground Physical and numerical modelling of tunnels and deep excavations in soft ground Seismic response of underground infrastructure in soft ground Design and application of ground improvement for underground construction Ground movements, interaction with existing structures and mitigation measures xiii

The success of publishing this version of the proceedings must be attributed to the authors of the papers who shared their valuable work, the reviewers who diligently worked through the papers providing useful technical advice and to the teams at Cambridge, which conducted the editorial review of the papers and helped manage the complex logistics through very uncertain times. The work presented here would not have been possible without the efficiency and tre­ mendous support of Mrs. Tian Wu, Mrs. Anama Lowday and Dr Njemile Faustin who sup­ ported the conference organisers every step of the way. The strong support from the staff at Taylor and Francis, and from the Chairman and Secretary of TC204 (Prof. Adam Bezuijen and Prof. David Chapman) is also acknowledged. Mohammed Elshafie, Giulia Viggiani & Robert Mair (Editors)

xiv

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Organisation The Symposium of Technical Committee 204 on ‘Geotechnical Aspects of Underground Con­ struction in Soft Ground’ was organised by the Geotechnical and Environmental Research Group at University of Cambridge under the Auspices of the International Society for Soil Mechanics and Geotechnical Engineering. LOCAL ORGANISING COMMITTEE Professor Lord Robert Mair (Chair) Professor Giulia Viggiani Professor Gopal Madabhushi Professor Abir Al-Tabbaa Dr Stuart Haigh Dr Mohammed Elshafie (Secretary) Dr Giovanna Biscontin TECHNICAL COMMITTEE 204 Prof. Adam Bezuijen, Chair Prof. Chungsik Yoo, Vice Chair Prof. David Chapman, Secretary Weidong Wang Hirokazu Akagi Mitsutaka Sugimoto Kiwamu Tsuno Ilkka Vähäaho Chang-Yu Ou Yung-Show Fang Michael Mooney Richard Pang John Endicott Hervé LeBissonnais Rui Carrilho Gomes Markus Thewes Achim Hettler César Sagaseta Salvador Senent Anna SiemińskaLewandowska Monika Mitew-Czajewska D.E. Razvodovsky Tony Ho

Jochen Fillibeck Lou Areias Fabrice Emeriault Jacques Robert Irawan Firmansyah Mete Erdemgil Mustafa Nalcakan Mustafa Koc Gang Zheng David Masin Petros Fortsakis Stavroula Schina Atul Nanda Ashish Juneja Alejo Oscar Sfriso Hugo Acosta-Martinez Marlísio Oliveira Jr Jorge Gabriel Laiun Christopher Menkiti Graham Taylor Carlos Lam

xv

Nagyoung Kim Altaf Usmani Jamie Standing Mandy Korff Wout Broere Alessandra Sciotti Marco Barla Daniela Boldini Ilias Michalis Yuepeng Dong Giovanni Spagnoli Dongmei Zhang Sergio Sánchez Rodríguez Rafael Jiménez Jay Lee Stephan Jefferis Long Phung Huy Hung Tran Tuan Nghia Do Shuying Wang

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Sponsors Conference Sponsor

Session Sponsor

xvii

Field case studies and sensing technologies

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Ovalisation of cast-iron tunnels in response to nearby tunneling work M. Alhaddad Transport for London, London, UK

K. Soga Department of Civil Engineering, University of California, Berkeley, USA

ABSTRACT: The behaviour of existing cast-iron (CI) tunnels subjected to ground movements induced by new tunnelling works is a much discussed topic. In many cases new tunnels and shafts, particularly in highly populated urban areas, will need to be constructed adjacent to existing tunnels. This inevitably results in ground deformations that are transmitted to these structures. This paper investigates the response and the radial tolerance of CI lining to these imposed movements. This paper highlights the common ‘damage assess­ ment’ procedures in practice and emphasizes the need of paying equal attention to historical case studies when carrying out often complex analyses. A recent case study from the construction of Crossrail Line (Queen Elizabeth Line) in London has been presented to demonstrate this. In this example the CI lining is subjected to significant deformations caused by nearby tunnelling and excavation works.

calculate the magnitude of these imposed deformations so they can assess whether the lining can accommodate them ‘safely’. Safely here means a risk based approach to assess the likelihood of damage and the conse­ quences of deformations, or failures at extreme cases.

1 INSTRUCTION 1.1 Cast-iron (CI) tunnels subjected to deformations For near two centuries CI tunnels have been built for a variety of reasons including for pedestrian usage, transports, mail delivery and to carry utilities. Most of today’s London Underground tunnels are made of cast-iron. A typical CI transport tunnel consists of a track­ bed that lies on the tunnel invert supporting the tracks. The lining is made of several segments, typic­ ally of the same size, in addition to a segment that is of a smaller size called key-segment. Each segment is made of a panel and four flanges, which are bolted to the adjacent segments, forming the rings trans­ versely and the body of the tunnel longitudinally. To note is that CI tunnels are not always uniform throughout their body and the diameter of the tunnel could change, when approaching the platforms for example. The area of separation is formed by a concrete block called the head-wall. There are also other features that disturb the uniformity, such as ventilation shafts and cross passages. When external forces are imposed on these tunnels, they get accommodated partly by the segments moving flexibly against each-other and partly by the deform­ ation of their structural elements: namely, the segments and the bolts. This paper focuses on the radial deform­ ation of the body of the tunnel. Designers carry out ‘damage assessments’ to predict the nature and to

1.2 Damage assessment The response of existing CI tunnels to tunnelling­ induced deformations is a complex issue and prac­ ticing engineers are faced with significant uncertainties about assessing their tolerances. The assessment is often numerically complex and therefore empirical methods are alternatively implemented to (i) predict these ground deformations and also to (ii) examine the tolerance of the affected existing tunnels. Depending on the outcomes, further complex assessments may deem to be necessary. (i) The ground movement caused by the tunnelling work is a fairly understood topic. It has been shown in numerous case studies as well as experi­ mental works (often using centrifuge tests) that soil behaves in a Gaussian looking shape transversely to the tunnelling direction, and in a bow-wave looking shape in the longitudinal direction. The shape and the extent of the movements are pre-determined based on given parameters that have been arrived at empirically over the years. These predictions are par­ ticularly more accurate for settlement measurements, especially for cities such as London, thanks to the abundance of case studies over the decades. How­ ever, there are more uncertainties with the prediction

DOI: 10.1201/9780429321559-1

3

Underground standards (London Underground 2007 and 2009). Li et al., (2015a) shows an example of this envelop (Figure 1) for a typical geometry of LU North­ ern Line CI tunnels (Diameter = 3.8 m and Depth = 19 mbgl at Euston Station area). When the lining is defected (visually inspected), reduction factors are applied.

of horizontal movements and even more so at the longitudinal direction ahead of the face of the tunnelling. Alhaddad et al 2017 and 2020 describe these in more details. The focus of this paper is not the accur­ acy of predicting these movements but rather the response of CI tunnels to accommodate them at radial directions. (ii) It is often conservatively assumed that CI tun­ nels deform freely to follow these calculated ground movements from stage (i). In many cases, this has been shown to be a reasonable assumption along the longitudinal axis, where the cast iron rings are more flexible to slide and bend against each other to accom­ modate those movements (Alhaddad 2016). However, transversely, this is subject to debate for mainly two reasons: 1- the calculated horizontal movements derived from the empirical methods are not as reliable as the settlement predictions and 2- the CI tunnel rings are less flexible to accommodate movements radially than they are longitudinally. Hence, the importance of case studies to provide evidence based ‘assessments’.

2.2 Determine existing hoop forces in the lining Hoop forces are determined using the ‘Elastic Con­ tinuum’ method after Duddeck & Erdmann (1985) where the maximum and minimum hoop thrusts are calculated using Equation 1 and Equation 2, respect­ ively. Existing hoop forces are calculated based on given input parameters. The imposed hoop forces are assumed to be zero unless it is thought that sig­ nificant loading changes occur.

1.3 Case studies There are a limited number of cases in which CI tun­ nels have been influenced by large enough deform­ ations to examine the extent of the longitudinal and radial deformations imposed on them. Crossrail (2007) lists the case studies recorded in the literature and Alhaddad (2016) presents several new case studies from recent years. Rarely damages are reported, and when they are, it is not clear if they have been caused by the longitudinal deformations or the ovalisation of the linings. This paper selects one of the new case studies where the authors have detailed information to derive reasonable conclusions. The case study is a disused Royal Mail Tunnel (RMT) that is built from cast iron lining. The tunnel was influenced by nearby Sprayed Concrete Lining (SCL) excavations (tunnel­ ling and shafts) at close proximities. It is the radial dis­ tortion of the CI lining that is the topic of this paper.

Where σv and σh = vertical and horizontal stres­ ses of the ground; R = Radius of the tunnel; ν = Pois­ son ratio of the tunnel; E = Young’s modulus of the tunnel lining; Ec = Stiffness of the ground; A = Sec­ tion area; J = Second moment of area; K = Earth Pressure Coefficient; and Ir = Reduced second moment of area of the tunnel lining in accordance with LU standards and is equal to J(4/n)2 where n is the number of segments.

2 DAMAGE ASSESSMENT OF RADIAL DISTORTIONS Radial movements are usually assessed in the form of imposed ovalisation on the lining. The ‘damage assessment’ process is described briefly in the fol­ lowing five steps. These steps have been adopted for the assessment of CI tunnels influenced by the con­ struction of the new Crossrail Line, also known as Elizabeth Line (Crossrail 2010 and 2011) and are generally used in similar recent assessments. 2.1 Determine permissible tunnel lining capacity envelope This is the moment-thrust permissible envelop of tunnel lining and is measured by the given parameters of CI lining, such as those provided by London

Figure 1. Tunnel lining permissible capacity envelop (Li et el., 2015a).

4

2.3 Determine existing bending moment in the lining

2.6 Circularity of the lining Wright (2009) states that typically CI tunnels have a squatting of 0.5% to 1%. When the existing circular­ ity of the tunnel is unknown, the LU standard requires the use of a value of 0.6%. If the assessment for 0.6% ovalisation fails, it is recommended to repeat the pro­ cedure and to adopt a more realistic approach taking into account that some of the existing ovalisation has been caused by the as-built imperfection of the lining during its initial construction. Assuming a correct dir­ ection of this imperfection (squat or elongation) is fundamental. The TLL Tube Line Report (2005) has carried out a comprehensive and an extensive laser scanning and FE modelling analysis of 33% of running tunnels. 93% of these experience around 1% squatting, which is believed to be caused by the imperfection during installation. When looking at a number of CI tunnels that are built adjacent and one after the other, the data show that one of these tunnels shows more squatting than the other (Wright, 2009; 2010). This suggests that at least this differential squat between the two tunnels (assuming the one that was squatted more was installed first) might have been caused by the changes in the stresses within the lining rather than by the as-built imperfections. This magnitude should not be disregarded. Almost all of the linings scanned by TLL had experienced squat (Wright, 2010) while FE modelling and soil parameters at typical LU tunnel depths should result in elongation of the tunnels (earth pres­ sure coefficient ‘K’ values are thought to be 1-1.25). This indicates that changes in horizontal stresses during the excavation and/or the drainage of soil in the long-term caused by the higher permeability of the tunnel has lowered the surrounding pore water pressure and resulted in the coefficient earth pressure ‘K’ values to be lower than 1. If we accept this hypothesis, then it is more likely that the recorded non-circularities are largely due to the external stres­ ses and less likely to be caused by the as-built imper­ fection. It will be shown in the presented case study that tunnel linings are more susceptible to imposed squatting than they are to elongation (cracks hap­ pened where rings experienced squatting). This indicates that current damage assessment procedures might need revising, especially where the CI lining is subjected to additional squatting. Aas­ suming that the non-circularity is a result of as-built imperfection is not completely justified. On the other hand, assuming that the ring is a rigid ring (con­ tinuum theorem) is not perfectly representative too. Li (2014) carried out a 3D FE modelling paramet­ ric study and suggests that bending moment is pri­ marily sustained by the contact of the circumferential flanges with little influence from the joints while hoop thrust is mainly transmitted through the joints. This should be taken into account and a ‘continuum theorem’ that assumes a rigid ring is therefore a less

The bending moment within the lining is calculated using Equation 3 after Morgan (1961). The existing bending moment is measured by assuming an exist­ ing ovalisation. For LU tunnels this is specified to be 0.6% of the tunnel diameter (the applicability of this assumption is debated in this paper). For example, a 3.56 m diameter tunnel will have a maximum radial deflection of 10.69 mm. This deflection is assumed to be either constant (Figure 2 right) or variable (Figure 2 left). For the variable case, the assessment considers both scenarios: squatting (horizontal egg shape) and elongation (vertical egg shape).

Where δ = Deflection (actual radius - nominal radius) 2.4 Determine increase in bending moment from lining distortion due to the new construction The total bending moment due to the construction work is measured by adding the existing deflection value to the maximum imposed value. In tunnelling related cases this value is calculated from assuming that the lining follows the imposed ground move­ ments freely and equates to the differential soil movements radially along the extrados of the lining. 2.5 Check the axial force and bending moment remain within the permissible capacity envelope The envelope is determined at stage 1 and if the imposed stresses calculated at following stages are outside the envelope further analyses (e.g. FE mod­ elling) and/or assessments will be needed. Mitigation measures are also recommended accordingly. An important input into specifying the tolerable ovalisation on the lining is the existing circularity of the lining and the assumptions that are made about its nature.

Figure 2. Deflection Shape of the Lining; squat (left) and uniform (right).

5

realistic approach. When modelling a flexible ring, Li (2014) observes that the ovalisation is first caused by the pressure of the ring’s self-weight and con­ tinues to change by soil loading to a value as high as 1% in the absence of imperfections. A value which is close to what has been witnessed in the majority of circularity surveys of tunnels. At complex scenarios such as cross-passage openings, this modelling becomes especially less relevant (Li et al., 2015b). It is noteworthy to mention that the scanned tun­ nels have not had concerning signs of damages and therefore have not necessarily exceeded their stress state envelopes. In other words, assuming an elastic continuous ring is not a realistic approach, although it is surely a conservative one. However, this becomes less conservative when the existing ovalisation is treated to be an as-built imperfection. More importantly, the direction of ova­ lisation (squat or elongation) becomes more relevant to making a more evidence based decision, as will be demonstrated in the case study below. Deploying mitigation measures, such as monitoring, become a pragmatic tool to manage the uncertainties and the risks associated with such assessments. 2.7

Figure 3. RMT’s Rings Condition and Geometry.

of on-going water ingress could be seen. This was confirmed by the many visits of the author to the tunnel. A typical schematic of the missing bolts and tunnel cross-section is shown in Figure 3. Note that the location of the key-segment varies and some­ times staggers between the adjacent rings. The invert is made of an in-situ cast concrete block. The quality and consistency of the concrete is not known and historic drawings and records show that there is an underlying layer of gravelly material beneath the concrete surface. 3.2

Description of Crossrail work

The Crossrail tunnels were constructed using SCL techniques and excavated in two stages: a pilot tunnel of approximately 6 m diameter followed by an enlargement of the diameter to approximately 11 m. The invert of the RMT is at an elevation of between +92 to +86 m ATD, falling from west to east over the extent of the interface with the Crossrail works. The RMT is located within London Clay, while the Crossrail works extend into the top of the Lambeth Group. The interface between the London Clay and Lambeth Group is at around +76 m ATD. The new tunnelling work crossed RMT at four locations (Figure 4). They are described below (based on a chainage reference at RMT):

Monitoring ovalisation

Monitoring helps with managing the risks during the construction work but also provides an insight that can be used for future assessments. Conventionally systems such as ‘Bassett Convergence System’, prisms read by automatic total stations and chain of tilt meter sensors have been used to monitor the ova­ lisation in real-time and more recently new technolo­ gies such as shape arrays and digital image correlation (DIC) techniques have been deployed successfully to measure these deformations. Alhad­ dad et al. (2020) describes deployment of a high pre­ cision DIC system within tunnel environments. 3 RMT AT LIVERPOOL STREET STATION (RMT) – CROSSRAIL WORK 3.1

Description of the tunnel

The Royal Mail Tunnel (RMT) extends beneath cen­ tral London. It is formed from bolted CI segments and was constructed between 1914 and 1917. Since 2004, the RMT has not been in use unless on occa­ sions for the purpose of running battery-operated locomotives to support inspection or maintenance visits and more recently as a tourist attraction. It has an internal diameter of 2.74 m (Figure 3). The lining is built of 0.5 m wide rings bolted to each other at their flanges. Before the Crossrail work started, an inspection carried out by the contractor confirms that the tunnel was in a good condition and no obvious cracks or deformations were reported, although many bolts were recorded missing and in many locations signs

Figure 4. RMT Layout and its Proximity to Crossrail Work.

6

1) 2) 3)

4)

Chainage 223 to 243: Construction of the SCL westbound platform tunnel (PTW) with 6m clearance. Chainage 263 to 288: Construction of the SCL central concourse passageway with 5m clearance. Chainage 298 to 313: Construction of SCL Finsbury Circus access shaft excavated around 2 m to the south of the RMT. The shaft was approximately 40 m deep; varying between 12.5 m and 15.7 m in diameter. Chainage 343 to 485: Construction of the SCL eastbound platform tunnel. To the east of chai­ nage 343 the clearance between RMT and the enlarged Crossrail platform tunnel is approxi­ mately 2 m decreasing to less than 0.5 m towards the end. From around chainage 460 to 480, the RMT gradually enlarges towards a step plate junction. The new tunnel was constructed parallel and directly beneath these chainages for more than 100 m. The photograph in Figure 4 shows the enlargement directly beneath the RMT at these chainages.

Figure 5. Propping Installed over 15 m of RMT.

After construction had finished, an inspection by the author and Crossrail’s Asset Protection team as well as the limited monitoring data in place con­ firmed that the props did not get engaged (they did not take any load meaning elongation of all of the rings were less than 10 mm or 0.4% ovalisation).

Other relevant sections to note are as follows: 1)

3.4 Monitoring layout

RMT Enlargement Chamber: This chamber, bordered by chainages 243 and 263, consists of 13 rings of 3.7 m diameter with an opening at the northern side to an unused shaft. The con­ nection between the larger rings and the smal­ ler diameter rings is a concrete head wall which is typical for such transitions. Blomfield Box: From chainage 440 to 475, a 36m deep box-retained cut excavation was formed around 5m to the south of the RMT. Propping: In accordance with the Damage Assessment for the POT, propping was installed to mitigate radial deformation; along a 15 m section between chainage 375 to 390.

3.3 Damage assessment and mitigation measures

Figure 6 shows the conventional monitoring that was carried out inside RMT. This included the use of auto­ mated total stations (ATS) monitoring 3- and 5-point radial prism arrays at generally 20 m centres. In add­ ition, levelling points (LP) were installed in the invert of the tunnel (track-bed) and monitored manually on a weekly basis. Looking west the arrangement of the prisms were; P1 at Crown, P2 and P4 at right-axis and right-invert and P3 and P5 at left-axis and left-invert, respectively. LP at invert level was monitored manually. Visual inspection of the RMT was also carried out generally on a weekly basis to check for any signs of distress or change in tunnel condition. Over a short section of the RMT above the con­ struction of eastbound platform tunnel, further moni­ toring was installed by Cambridge Centre for Smart

The damage assessment of radial and longitudinal movements was carried out as explained in Section 2, which concluded that some cracks might occur at the track-bed (invert) of the tunnel due to the exces­ sive imposed longitudinal curvature (not radial). Devriendt & Alhaddad (2015) explain the damage assessment and general observed movements. Given the level of uncertainty relating to the ground movement and structural impact assessment calcula­ tions, to mitigate radial impacts, a short 15 m section of the tunnel was propped at axis level (Figure 5). This took place where the tunnel was expected to first experience elongation and the propping was installed such that the loads would only be taken up if radial convergence at axis level exceeded 10 mm. Additional props were kept on standby in case they needed to be installed elsewhere (subjected to future observations and assessments).

Figure 6. Examples of Prism (P), Manual Monitoring (LP) and Automatic Total Station (ATS) Arrangement inside RMT.

2) 3)

7

Infrastructure and Construction (CSIC). This work is described in Alhaddad et al. (2014). The results of this monitoring are not relevant to this study except that no damage or cracks were seen in this section of the tunnel. The lining there was mainly subjected to elongation (0.4%) and longitudinal curvatures. 3.5

5) 6)

The settlements of the crown level and invert level, the horizontal convergence of the RMT measured from points 2 and 3 and the vertical convergence measured from points 1 and LP for each of these phases are shown in Figure 8. Vertical convergence is measured from differential values collected from two different sys­ tems (ATS and manual monitoring) with an estimated measurement error of up to 2 mm. Also, LPs only meas­ ure the settlement and are installed on the track bed and are assumed to follow the lining at the invert level. A maximum settlement of approximately 100 mm was measured for the construction of the westbound platform tunnel and approximately 90 mm for the construction of the eastbound platform tunnel. This

Construction progress and monitoring results

Figure 7 shows sketches of the construction progress at each of these stages. The significant movements to the RMT occurred at the following times: 1) 2) 3) 4)

Stage 5: Enlargement of the influencing crosspassages and the start of Blomfield Box exca­ vation – May 2014. Stage 6: End of the majority of the construction including Blomfield Box and the arrival of the Eastbound TBM – February 2015.

Stage 1: Excavation of the Finsbury Circus Shaft (FCS) – Completed before June 2012. Stage 2: Excavation of the Central concourse pilot tunnel and enlargement – Completed by early April 2013. Stage 3: Excavation of the Eastbound and Westbound pilot tunnels – Completed in October 2013. Stage 4: Excavation of the Eastbound and Westbound enlargement and influencing Crosspassages pilot tunnels – Completed by midMarch 2014.

Figure 8. Monitoring of Settlement and Convergence for RMT; (a) Settlement at Invert Level (Manual Monitoring), (b) Settlement at Crown Level (ATS Monitoring), (c) Con­ vergence of the Tunnel Horizontally Measured Between P2 and P3, (d) Convergence of the Tunnel Vertically Between P1 and LP.

Figure 7. Main Construction Stages Influencing the RMT.

8

exceeded the preconstruction assessment calculating a maximum of 70mm settlement and further assess­ ments had to be carried out during the works. This included carrying out a circularity survey to measure the tunnel’s existing circularity (Crossrail 2013). The horizontal and vertical convergences give an indication of the magnitude of the ovalisation imposed on the lining. Concluding from the direction of the movements, RMT experienced up to 10 mm of elongation (with no signs of damage) and up to 10mm of squat (where cracks were reported). These were above the 7 mm value calculated during the ini­ tial damage assessment procedure. 3.6

Summary of cracks

First reported cracks were between chainages 303 and 323 (Figure 9) and were most probably caused by the excavation of FCS and/or the excavation of the SCL cross-passage directly beneath the RMT. The 6 cracks were all on the circumferential flanges at bolt locations and were seen around January 2013 (eight months after the completion of FCS and days after the breakage of cross-passage enlargement). The cracks were all approximately on the same longitudinal line near the crown and emanate from the bolt position. The cracked flanges were all on the weaker segment (every other neighbouring segment on the next ring that is not cracked is a key segment). Figure 8 (d) shows that around 10 mm squat (0.4% ovalisation) was imposed on the tunnel by the exca­ vation of FCS. The deformed shape appears to be an egg-shape pointing towards the bottom of FCS and the approaching excavation of the cross-passage. The second series of the cracks were observed after stage 4 of the construction and were located to the west of the RMT enlargement chamber at around CH250. They can be seen on the CI lining and also on the invert and the connecting head-wall between the RMT enlargement chamber and the running

Figure 10. Cracks above PTW Excavation at CH 250.

tunnels (Figure 10). It is unclear when exactly they occurred, but the majority of the imposed movements were caused by the enlargement of PTW westwards from the end of January to early February 2014. The mechanism of deformation and the arrange­ ment of the cracks are very similar to those seen around FCS. The deformed shape is likewise an eggshape pointing towards the approaching excavation of PTW from the east. Also, here cracks occurred on the same position adjacent to the crown and on the weaker segments at approximately 0.5% ovalisation.

Figure 9. Cracks Adjacent to FCS at CH 303.

9

When looking at the convergence history of the closest monitored ring at CH243 (7 rings away from the nearest crack – where the furthest crack is one ring away from the headwall and is roughly at CH 250), it may be seen that there is a rapidly increasing trend of movements when PTW enlargement excava­ tion approaches the tunnel (Figure 11). The values are maximum on 28th of January 2014 when the tun­ nelling face was recorded to be directly beneath the RMT. The trends start to decrease as the tunnelling face crosses RMT and proceeds to the other side. Data are baselined on 15th of July 2013, a few days before the chainage was influenced by the PTW pilot excavation (the construction of the pilot tunnel then was paused until 1st of October 2013). It should also be noted that the divergence of axes 1-5 and the convergence of axes 3-4 start to return to their original shape after a peak corresponding to the PTW crossing underneath. The trend of returning to the original shape appears to be taking on a plastic

deformation shape. This trend of movements implies that the cracks or some of them may have happened around the peak (the invert cracks had been spotted on a visit on 7th of February 2014, hence the cracks must have happened before this date). 4 CONCLUSION CI tunnels deform longitudinally and/or transversely to accommodate external movements (such as those caused by nearby construction activities). The focus of this paper is the response of CI tunnels to those movements transversely and along their radial rings. This paper demonstrates the ‘damage assessment’ stages that are often followed to assess whether CI tunnels can accommodate those movements radially and briefly describes the mitigation measures where the deformation are assessed to be excessive; namely the use of props inside the tunnel, when possible. Mitigation measures are not often practical (for examples props are not possible within running tun­ nels) and it is important to understand the likely response of the tunnels to be able to evaluate the risks in advance of the start of any given construc­ tion work. Numerical assessments are complex, and results of such assessments are not guaranteed to be correct. Historical case-studies provide an important insight to enable evidence-based decision makings. This paper presents a case study where the CI tunnel was subjected to significant radial movements. It is argued that where constructed within London Clay, typical CI linings (uniform tunnels not influenced by other structures such as existing cross passages) are likely to be more susceptible to squatting than they are to elongating deformations. This is because the evi­ dence suggests that CI linings have experienced squat­ ting over their lifetime and as a result they would have less bending tolerance to accommodate squatting than elongation. The cracks in this case study have evidently been the result of radial deformation when the rings were squatting with about 0.4-0.5% ovalisation. The rings had elongated with similar ovalisation values with­ out any sign of damage. This reinforces the argu­ ment that CI rings experience squat over their lifetime. In this case the alignment of the segments and the staggering of the keys appear to also contrib­ ute to the radial tolerance of CI rings (by perhaps concentrating the loads over the weak axes). Designers must be more cautious and should not downgrade the existing circularities as ‘as built imperfections’ especially when the new constructions are likely to impose more squatting on the tunnel. This is, for example, when excavating shafts nearby or when carrying out tunnel excavations with a parallel and an offset axis from the existing linings. On the other hand, accepting that there is the risk of cracks happening on the radial flanges will open the conversation between the stakeholders early on and will provide a more pragmatic approach to

Figure 11. History of Monitoring of CH 243: Settlement (a), Horizontal Convergence (b) and Vertical Convergence (c).

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mitigating the risks. The cracks on their own are not likely to impose a critical risk to the operation of the rail-tunnels and at extreme scenarios they can be fixed/replaced after the construction work is com­ plete. Hence providing more extensive and better tar­ geted (locally targeted) monitoring solutions for such cases could become a more favourable approach.

Alhaddad, M. et al., 2020. Cast-Iron Tunnels’ Tolerance to Imposed Longitudinal Settlement Curvature Geotechni­ que Symposium in Print: Linear infrastructure. Crossrail, 2007. London Underground Interface, Case His­ tories, Prepared by Geotechnical Consulting Group for Crossrail Project, 1D0101-G0G00-01022A. Crossrail, 2010. Civil Engineering Design Standard – Part 7, Ground Movement Prediction, Version 6.0, CR-STD -303-7 (CEDS 7). Crossrail, 2011. Assessment of ground movement effects on the Post Office tunnel at Liverpool Street Station (PO/05), Rev 3.0, C122-OVE-C2-RAN-CR101-00016. Crossrail, 2013. RESULTS FROM LASER SCANNING THE POST OFFICE TUNNEL AT LIVERPOOL STREET STATION (PO/05), Rev 2.0, C122-OVE-C2-RG -50031. Devriendt, M. & Alhaddad, M., 2015. Construction impacts of Crossrail Liverpool Street Station on the Royal Mail Tunnel. XVI ECSMGE Geotechnical engineering for infrastructure and development 1: 831–836. Duddeck, H. & Erdmann, J., 1985. Structural design models for tunnels. International Journal of Rock Mechanics and Mining Sciences & Geomechanics 20(1): 83–91. Li, Z., 2014. Long-term Behaviour of Cast-iron Tunnel Cross Passage in London Clay. PhD thesis, University of Cambridge, Department of Engineering. Li, Z., Soga, K. & Wright, P., 2015a. Behaviour of cast-iron bolted tunnels and their modelling. TUNNELLING AND UNDERGROUND SPACE TECHNOLOGY 50: 250–269. Li, Z., Soga, K. & Wright, P., 2015b. Long-term performance of cast-iron tunnel cross passage in London clay. Tunnelling and Underground Space Technology 50: 152–170. London Underground, 2007. Manual of Good Practice, Civil Engineering – Deep Tube Tunnels and Shafts, G-055, Version A1, (LU-G055). London Underground, 2009. Standard 1-50 Civil Engineer­ ing Common Requirements, Version A1, (LU-1-50) Morgan, H.D., 1961. A contribution to the analyses of stress in a Circular Tunnel. Geotechnique 11(1): 37–46. Wright, P., 2009. Assessment of London Underground tube tunnels investigations, monitoring and analysis. Smart Structures and Systems 6(3): 239–262.

ACKNOWLEDGEMENT This work was carried out in CSIC and would not have been possible without the support of Arup, CH2M Hill (now Jacobs), London Underground and Royal Mail Group and thanks to Engineering and Physical Sciences Research Council (EPSRC) for their financial support. The authors would like to namely acknowledge Robert Mair, Mohammd Elsha­ fie, Michael Devriendt, Frances McDonnell, Mathew Wilcock and Chang Ye Gue for their contribution throughout this work.

REFERENCES Alhaddad, M. 2016. Photogrammetric monitoring of castiron tunnels and applicability of empirical methods for damage assessment. PhD Thesis, Department of Engin­ eering, University of Cambridge, Cambridge, UK. Alhaddad, M. et al., 2014. Multi-Suite Monitoring of an Existing Cast Iron Tunnel Subjected to Tunnelling­ induced Ground Movements. Tunneling and Under­ ground Construction - Geo-Shanghai. American Society of Civil Engineers: 293–307. Alhaddad, M. et al., 2017. Imposed Longitudinal Settle­ ment on a Cast-iron Tunnel from the Excavation of a New Tunnel Beneath. 9th International Symposium on Geotechnical Aspects of Underground Construction in Soft Ground, IS – Sao Paulo: 343–353.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Lubrication characteristics of pipejacking in soft alluvial deposits Wen-Chieh Cheng & Ge Li Xi’an University of Architecture and Technology, Xi’an, China

Dominic E.L. Ong Griffith University, Queensland, Australia

ABSTRACT: The impacts of the lubrication upon the pipejacking works are significant and necessary to be considered in tunnelling design. This study described a method that can be used to evaluate the lubrication performance by the reduction in the frictional coefficient μ as a function of injection type, soil and lubrication natures, and misalignment. The results of an application of the proposed method to a pipejacking project in soft alluvial deposits were presented. The effect of misalignment increased the frictional stress τld to 12.5 kPa, reducing the reduction in the μ value to 71%. The combined effects of misalignment and varying face resistance contributed to the τld value of 4.0 kPa and were deemed as the main cause leading the low reduction of 84%. The misleading reduction of 60% was attributed to the inability of the occasional gravel to develop lower face resistance. The effects had greater influence on the lubrication performance than the injection type.

frictional stress was reduced from 45 to 90%. How­ ever, the distribution characteristics of injected lubri­ cant and their effects on the reduction of the frictional resistance has rarely studied by literature. The objectives of this study are (1) to present the distribution characteristics of injected lubricant with reference to the presented case history of pipejack­ ing, (2) to evaluate the lubrication performance of pipejacking in soft alluvial deposits using the base­ line technique and (3) to investigate the impacts of the distribution characteristics of injected lubricant, soil and lubricant natures and pipe deviation on the lubrication performance.

1 INTRODUCTION The pipejacking technologies are specially favour­ able for urban pipeline systems construction because of their low cost and short construction period. Inappropriate thrust force and/or inadequate jacking capacity is often seen while pipejacking, causing damages to jacked pipe string and adjacent proper­ ties. This would hold true especially for longdistance pipejacking in coarse soils. Effective lubri­ cation enables long-distance pipejacking to be con­ ducted. The friction resistance usually constitutes the major component of jacking loads, which can be approximated from the minimum bound to the total jacking load and exhibits a cumulative nature while pipejacking in coarse soils. There have been many factors affecting the friction resistance and amongst the factors, lubricant plays a vital role in achieving effective lubrication. Generally, lubricant should be maintained within overcut annulus, and the percent­ age reduction of frictional stress appears to be closely linked to the volume of injected lubricant. Cui et al. (2015) reported that due to loss of lubrica­ tion fluid into surrounding fissures, the effectiveness of slippery film was decreased, leading that the thrust force of Line 2 of a pipe ramming project con­ veying water from southern to northern Jiangsu was much higher than the other three lines (Lines 1 and 3-4). Pellet-Beaucour & Kastner (2002) indicated that for volumes varying from 25 to 170 litre/m,

2 METHODOLOGY 2.1

Lubricant distribution characteristics

The injection of lubricant can be categorised into three types which are single-point injection, two-point injec­ tion and multi-point injection. The injection into the overcut annulus through a single injector is termed single-point injection. The injection into the overcut annulus by two injectors at two different distances from the face is termed two-point injection, as depicted in Figure 1. Injector 2 at a distance more close to the face would work with Injector 1 as the friction during pipe ramming gets higher than a single injector can manage. While jacking RCP 14, the overcut annulus at

DOI: 10.1201/9780429321559-2

12

recommended by Stein et al. (1989) for soil-pipe (unlubricated) interface. The described method that can be used to evaluate the lubrication performance can be briefed in short as follows: (1) preparing the baseline of jacking loads, (2) calculating the fric­ tional stress τld for lubricated drives, (3) calculating the normal contact pressure σz, (4) evaluating the frictional coefficient μld for lubricated drives, and (5) assessing the percentage reduction in the μ value. 3 PROJECT BACKGROUND 3.1

Background

The slurry shield was adopted to perform excavation of the two pipejacking drives in soft alluvial deposits. The alignment for all the drives is straight. During pipe ramming, the bentonite slurry with unit weight of 10.6 kN/m3 was used to stabilise the excavation face and transport tunneling soil spoils to decantation chambers. Since the tunnels were excavated via the cutter wheel of 1500-mm in diameter, an overcut annulus of 30 mm was formed by using the smaller concrete pipe of 1440 mm in diameter. The two drives characteristics are detailed in Table 1. Figure 1. Distribution characteristics of injected lubricant for two-point injection.

3.2

4 m and 11 m distances from the face is saturated through Injectors 2 and 1 by injecting 0.13 m3 (20.9 %) and 0.50 m3 (79.1 %) lubricant, respectively. Figure 1 also provides the details about the ramming of RCPs 15 and 16. Since effective lubrication cannot be easily sustained for long-distance pipe ramming, multipoint injection may be used to prevent inadequate lubrication from occurring. 2.2

Engineering geology

The stratigraphic profile is established with reference to five 15-m deep geological boreholes penetrating through the 6-m thick silty sand layer into the 7.5 m thick poorly-graded to well graded sand and gravel layer. Based upon the standard penetration test and triaxial test results, the soil physical and mechanical properties are listed in Table 2. 4 ANALYSIS AND DISCUSSIONS 4.1

Lubrication performance

Development of jacking loads baseline

At Drive C, the shield spanned at a depth of 10.8 m and the associated activities are presented in Figure 2. The baseline of jacking forces consisted of the first 2-8 m section (gravel), 8-21 m (gravel) and 21-40 m (gravel) sections, and final 40-75 m (clayey gravel) section was established through the minima of

Lubrication can largely affect pipejacking works, which can be assessed through the reduction of the frictional coefficient μ. Pellet-Beaucour & Kastner (2002) indicated that local variations of total jacking load are generally linked to the varying face resist­ ance. A line constituted by the minima of total jacking load is thus referred to as baseline of jacking loads. The jacking forces from both ends of a baseline sec­ tion divided by the section length leads to the average jacking force favg. The average jacking force favg div­ ided by the circumference of pipe yields the frictional stress τld. There are many national standards available for calculating the pressure acting upon the pipe’s crown (termed normal contact pressure σz hereafter) but based upon different safety concepts. Dividing the τld value by the calculated σz value gives the frictional coefficient μld for lubricated drives. The percentage reduction can be calculated by (μnd-μld)/μnd ×100 where the value of μnd is

Table 1.

Characteristics of two pipejacking drives. Cutter Length Depth dia.

Soil cover

Face resist.

V. inj

Parameters m

m

m

m

kPa

l/m

Drive C Drive D

10.8 10.8

1.5 1.5

10.1 10.1

555 555

552 534

75 102

Note: Cutter dia = cutter wheel diameter, Face resist. = face resistance, V.inj = volume of injected lubricant.

13

F 1962 (ASTM 2011), and GB 50332 (MOC 2002) are deemed to be the well-developed models that can be used to calculate the normal contact pressure σz. The models were modified with reference to the Terzaghi arching model founded upon active trap-door experi­ ment (Terzaghi 1936) where the shear bands arise from the outside of tunnel cross sections along oblique lines, with a horizon included angle equal to 45°+ϕ/2, and then they turn to vertical lines after passing the tunnel crown’s level and finally arrive at the ground surface, as illustrated in Figure 4. Equation 4 in the modified models was derived from the limit equilibrium of a horizontal slide (Ter­ zaghi 1943):

Table 2. Summary of soil physical and mechanical properties. Thickness SPT-N γ Layer

m

Backfill Silty sand Gravel/Sand Silty sand

2.0 2.8 >4.5 >3.0

c’

ϕ’ qu

kN/m3 kPa °

2-9 >100 9-18

17.6 18.1 20.1 19.1

0 0 0

kPa

28 35 6.7e3 30

Note: γ = unit weight, c’ = cohesion, ϕ’ = friction angle, qu = uniaxial compression strength.

Figure 2. Pipejacking activities at Drive C.

where B1=silo width, γ=unit weight of soil, K=soil pressure ratio, c=soil cohesion, τf=shear strength of soil, and σz=normal contact pressure. Equation 4 is a single order ordinary differential equation. Inte­ grating Equation 4 and considering boundary condi­ tion σz=0=q at the surface deduces the σz value at any level:

jacking forces, with the average jacking forces favg at 1.6, 56.5, 0.5, and 12.3 kN/m, respectively. While their frictional stresses τld were calculated as being 0.4, 12.5, 0.1, and 2.7 kPa, respectively. The pipe ramming of Drive D traversed at 10.8 m depth and its activities are shown in Figure 3. The pipejacking results deter­ mined the baseline of jacking forces corresponding to the first 5 m (gravel) section (from 6 to 11 m), the sub­ sequent 11-24 m (clayey gravel) and 24-31 m (clayey gravel) sections, and the final 31-102 m (clayey gravel) section, with the average jacking forces favg at 2.0, 18.1, 49.0, and 3.5 kN/m, respectively. The favg values divided by the circumference of pipe yielded the frictional stresses τld equal to 0.4, 4.0, 10.8, and 0.8 kPa, respectively. 4.2

In fact, the σz value should be calculated one stra­ tum by one stratum from the surface to the bottom. Equation 5 can thus be rewritten as Equation 6 for

Evaluation of lubrication performance

JMTA (JMTA 2013), ATV A 161 (German ATV Rules and Standards 1990), BS EN 1594 (BS 2009), ASTM

Figure 4. Terzaghi arching model (after Terzaghi 1943).

Figure 3. Pipejacking activities at Drive D.

14

Table 3.

Parameters used in the modified models. Silo width

δ

Parameters

m

°

Terzaghi (1943) JMTA (2013) ATV A 161 (1990) BS EN 1594 (2009) ASTM F 1962 (2011) GB 50332 (2002)

De×[1+2tanα] (De+0.08)×(secα+tanα) 1.732De (ϕ=30°) De×(1+2tanα) 1.5De De×(1+tanα)

ϕ ϕ ϕ/2 ϕ ϕ/2 Katanϕ=0.19

K

c’ kPa

1 1 K0=0.5 K0 tan2(45°-ϕ/2)

C C None c (with verification) None None

Note: δ = friction angle in shear plane, α = 45°-ϕ/2, De = outer pipe diameter, Db = tunnel bore diameter, ϕ = soil friction angle, c = soil cohesion, K0 = soil pressure ratio at rest.

calculation of environment.

the

σz

value

in

multistratum

where i=stratum numbering from 1, 2,.to n, hi=thickness of ith stratum. Parameters used in the modified models are summarised in Table 3. Equa­ tions to calculate the normal contact pressure acting upon pipeline should refer to the national standards. The modified models assume the fully developed shearing bands. Zhang et al. (2016) measured the normal contact pressure acting upon the 17th, 24th and 26th pipe respectively during the construction of Gongbei Tunnel to verify their proposed new calcula­ tion model. The stratum properties utilised in the cal­ culation of the σz value against the national standards (Terzaghi, JMTA, ATV A 161, BS EN 1594, ASTM F 1962, GB 50332) are shown in Table 4. Compari­ sons of the normal contact pressure σz on the 17th pipe between the calculated normal contact pressures and the measured ones were conducted, as shown in Figure 5. The soil prism weight would give most safe, robust but uneconomical pipeline designs, as indi­ cated in Figure 5. It was evident that ATV A 161 and ASTM F 1962 give estimations in good agreement with field

Table 4. 2016).

Figure 5. Comparison of normal contact pressure between Terzaghi, JMTA, ATV A 161, BS EN 1594, ASTM F 1962, GB 50332 and measured result.

measurements for jacked pipes. Thus, ATV A 161 and ASTM F 1962 were selected as the preferred models for calculating the normal contact pressure σz. Dividing the frictional stress τld by the calculated normal contact pressure σz led to the frictional coef­ ficient μld. The percentage reduction was obtained , as listed in Table 5. using 4.3

Soil properties for 17th pipe (after Zhang et al. Thickness γ kN/m3

c’

ϕ’

kPa

°

Parameters

m

Medium to coarse sand with clay Silty clay

9.3

17.6

0

30.2

17.6

18.3

10.6

6.9

Effect of lubricant injection type

Two drives (Drives C & D) were analysed because of their good data completeness. The cumulative vol­ umes of injected lubricant for Drives C & D were retrieved from Figures 2 and 3 and estimated relying only upon lubricant injections into overcut annulus of the analysed baseline section (Tables 6-7). For instance, the analysed baseline section for Drive C was 2-8 m (Section 1), 8-21 m (Section 2), 21­ 40 m (Section 3) and 40-75 m (Section 4). While the

15

Table 5. Summary of reduction in μ value against each baseline section.

Drive C

Drive D

favg

τld

σz

kN/m

kPa

kPa

1.6 56.5 0.5 12.3 2.0 18.1 49.0 3.5

0.4 12.5 0.1 2.7 0.4 4.0 10.8 0.8

111.1 111.1 111.1 102.0 111.1 102.0 102.0 102.0

μld

Red. μ %

0.004 0.1 0.0009 0.03 0.004 0.04 0.1 0.008

98 71 99 88 98 84 60 97

analysed baseline section for Drive D was 6-­ 11 m (Section 1), 11-24 m (Section 2), 24­ 31 m (Section 3) and 31-102 m (Section 4). A 30 mm overexcavation outside the 1.44-m diameter pipe justified the overcut ratio of 0.02 cor­ responding to the theoretical overcut annulus of 0.138 m3/m. The lubricant injection type for the four sections of gravel (first three sections at Drive C and first section of at Drive D) refers to Tables 6-7. The averaged injection volume can be calculated as being the cumulative injection volume divided by the length of analysed baseline section. Their injec­ tion volumes averaged 0.139 m3/m (0.84 m3/6 m), 0.466 m3/m (6.06 (=4.8+1.26) m3/13 m), 0.619 m3/ m (11.77 (=7.06+2.35+1.18+1.18) m3/19 m), and 0.811 m3/m (4.05 m3/5 m), respectively, and were greater than 0.138 m3/m. The excessive injection volumes not only indicated a strong intention to

Summary of cumulative volume at Drive C.

RCP 02

RCP 09

RCP 19

RCP 29

RCP 38

RCP 48

RCP 58

0.84 (100) 4.8 (79.2) 7.06 (60) 9.08 (40)

1.26 (20.8) 2.35 (20) 4.54 (20)

1.18 (10) 2.73 (12)

1.18 (10) 2.73 (12)

1.82 (8)

0.91 (4)

0.91 (4)

Summary of cumulative volume at Drive D.

RCP 02

RCP 09

RCP 19

RCP 29

RCP 38

RCP 48

RCP 58

4.05 (100) 2.90 (63) 3.34 (63) 14.67 (36.2)

1.45 (31.5) 1.67 (31.5) 8.79 (21.7)

0.25 (5.5) 0.29 (5.5) 3.48 (8.6)

3.48 (8.6)

3.48 (8.6)

1.66 (4.1)

1.66 (4.1)

Table 7. (cont’d). Summary of cumulative volume at Drive D.

Note: Each drive has four baseline sections and values shown here are associated to the sections. favg = average jacking force, τld = frictional stress, σz = calculated normal contact pressure, μld = backanalysed frictional coefficient, Red. μ = percentage reduction in μ value.

Table 6.

Table 7.

RCP 68

RCP 78

RCP 88

1.66 (4.1)

0.81 (2)

0.81 (2)

Note: Number in bracket indicates the percentage volume of injected lubricant. RCP = reinforced concrete pipe. RCP02 is farther from the face than RCP88 and thus has larger percentage volume of injected lubricant.

saturate the overcut annulus to sustain effective lubrication conditions, but also showed a significant loss of lubricant while pipe ramming. Their percent­ age reductions in the μ value from ATV A 161 were calculated as being 98% (0.004 vs. 0.35), 71% (0.1 vs. 0.35), 99% (0.0009 vs. 0.35), and 98% (0.004 vs. 0.35), respectively. The buoyancy of 17.3 kN greater than the pipe self-weight of 12.6 kN and the satur­ ated overcut made a 71% reduction (the second lowest in this study) from 0.35 (the average μ value recommended by Stein et al. (1989) for gravel-pipe interface) to 0.1. The main cause to lead to the second lowest reduction was attributed to the excessive pipe deviation (Figure 2). The μld value of 0.1 matched the lower limit recommended by Stein et al. (1989) for lubricated drives, suggesting that the 8-21 m section of gravel was well-lubricated. Also, the overcut full of lubricant and the enough buoy­ ancy made the reductions high enough for the other three sections, with the μld values smaller than 0.1, indicating that lubrication during pipe ramming of the other three sections was very effective. Tables 6-7 details the lubricant injection type for the other four sections of clayey gravel (final section at Drive C and final three sections at Drive D). Their injection volumes averaging 0.649 m3/m (22.72 (=9.08+4.54+2.73+2.73+1.82+0.91+0.91) m3/35 m), 0.354 m3/m (4.60 (=2.90+1.45+0.25) m3/13 m), 0.757 m3/m (5.30 (=3.34+1.67+0.29) m3/7 m), and 0.570 m3/m (40.5 (=14.67+8.79+3.48×3+1.66×3 +0.81×2) m3/71 m), respectively, were also in excess

Note: Number in bracket indicates the percentage volume of injected lubricant. RCP = reinforced concrete pipe. RCP02 is farther from the face than RCP58 and thus has larger percentage volume of injected lubricant.

16

of the theoretical overcut annulus of 0.138 m3/m. The excessive injection volumes made the percent­ age reductions to reach to 88% (0.03 vs. 0.25), 84% (0.04 vs. 0.25), 60% (0.1 vs. 0.25), and 97% (0.008 vs. 0.25), respectively. The enough buoyancy and the lubricant-saturated overcut caused an 84% reduc­ tion from 0.25 suggested by Stein et al. (1989) for clay-pipe interface to 0.04. The varying face resist­ ance induced by jacking into the clayey gravel and the excessive pipe deviation were deemed to be the main cause to lead to the reduction of 84%. The μld value equal to 0.04 was far less than the lower limit, which also indicated adequate lubrication. It is worth to note that jacking into gravel at 31 m distance caused the reduction of 60% (the lowest in this study). This phenomenon was most likely because of the gravel not being long enough to develop lower face resistance. The use of the excessive injection volumes accompanied with the justified buoyancy made the percentage reduction far less than 0.1 for the other two sections, which also indicated adequate lubrication. To short, the effect of misalignment and the varying face resistance contributed to the low percentage reductions despite the excessive injection volumes. The occasional gravel led to the misleading reduction. The effects on the lubrication perform­ ance outweighed the effect of injection type. 4.4

times larger than 0.4 kPa of the 2-8 m section in the same gravel. While traversing through the 11­ 24 m section of gravel at Drive D, the jacking force increased by 637 kN to 1617 kN (Figure 3). The increase of 637 kN reduced the percentage reduction to 84% from 98%, most likely because of the exag­ gerated frictional stress τld of 4.0 kPa, which is 5 times larger than 0.8 kPa of the 31-102 m section in the same clayey gravel. The main cause to lead to the reduction of 84% was attributed to the combined effects of misalignment and varying face resistance, resulting from tunnelling into the clayey gravel. 5 CONCLUSIONS The method that can be utilised for assessing the lubri­ cation performance using the percentage reduction in the frictional coefficient was described. The effect of lubricant injection mode, soil and lubricant natures and misalignment on the lubrication performance was investigated. Some main conclusions can be drawn as follows: (1) The excessive volumes of injected lubricant made the overcut full of lubricant and led to enough buoyancy. Despite the excessive vol­ umes, the significant pipe deviation and/or the varying face resistance had implications on the percentage reduction in the frictional coefficient μ. The occasional gravel could lead to the mis­ leading reduction. The said effects on the lubri­ cation performance outweighed the effect of injection type. (2) The excessive volumes of injected lubricant were either attributed to permeable ground or to the inability of the injected lubricant to develop a filter cake of low permeability. The highly vis­ cous lubrication with Marsh cone viscosity of 38 mins reduced the friction resistance to viscous resistance, leading to the reductions greater than 88%. (3) The types of injection and the distribution charac­ teristics of lubricant may vary between pipejack­ ing projects. Notwithstanding that, this study provides an access of evaluating the lubrication performance for pipejacking works and the pre­ sented results would be useful in managing the lubrication performance for upcoming pipejack­ ing project.

Effect of soil and lubricant natures

The permeation of lubricant into the surrounding geology would mitigate the effort of establishing a lubricating layer at soil-pipe interface. This is most likely because of the inability of the lubricant to develop a filter cake of low permeability. Such a permeable overcut could also result in injection volume in excess of the theoretical overcut. The phe­ nomena discovered in this study resulted in the rela­ tively large injection volume of 0.552 m3/m at Drive C (including mostly the gravel) than 0.534 m3/m at Drive D (including mostly the clayey gravel). On the other hand, the continuous injection of the lubricant with Marsh cone viscosity of 38 mins into the overcut annulus effectively reduced the friction resist­ ance to viscous resistance. This led the reductions in excess of 88%, which is consistent with Staheli et al. (2006). 4.5

Effect of misalignment

The effect of misalignment significantly increased the friction resistance and had implications on the lubrication performance. There was an increase of the jacking force of 771 kN while spanning between 8 and 21 m distance at Drive C (Figure 2). This increase in the jacking force caused the percentage reduction in the μ value to reduce by 27% from 98%. The main cause was not because of the overcut not full of lubricant, but because of the exaggerated frictional stress τld of 12.5 kPa induced by the pipe deviation in excess of 60 mm, which is almost 30

REFERENCES ASTM. 2011. F1962-11 Standard Guide for Use of MaxiHorizontal Directional Drilling for Placement of Poly­ ethylene Pipe or Conduit under Obstacles Including River Crossings. West Conshohocken: PA. British Standards. 2009. BS EN 1594-09 Gas Supply System-Pipelines for Maximum Operating Pressure over 16 Bar-Functional Requirements. Brussels: UK.

17

Cui, Q.L., Xu, Y.S., Shen, S.L., Yin, Z.Y. & Horpibulsuk, S. 2015. Field performance of concrete pipes during jacking in cemented sandy silt. Tunneling and Underground Space Technology 49: 336–344. German ATV Rules and Standards. 1990. ATV-A 161 E-90. Structural Calculation of Driven Pipes. Hennef: Germany. Japan Microtunnelling Association (JMTA). 2013. Micro­ tunnelling Methods Serious II, Design, Construction Management and Rudiments. Tokyo: Japan. Pellet-Beacour, A.L. & Kastner, R. 2002. Experimental and analytical study of friction forces during microtunneling operations. Tunneling and Underground Space Technol­ ogy 17(1): 83–97. Stein, D., Möllers, K. & Bielecki, R. 1989. Microtunneling: Installation and Renewal of Nonman-Size Supply and Sewage Lines by the Trenchless Construction Method. Berlin: Germany.

Staheli, K. 2006. Jacking Force Prediction: An Interface Friction Approach Based on Pipe Surface Roughness. Ph.D. Thesis. Georgia Institute of Technology. Terzaghi, K. 1936. The shearing resistance of saturated soils and the angle between the planes of shear. In Pro­ ceeding of the 1st International Conference on Soil Mechanics and Foundation Engineering, June 1936. Cambridge: MA. Terzaghi, K. 1943. Theoretical Soil Mechanics. New York: USA. The Ministry of Construction of the People’s Republic of China (MOC). 2002. GB 50332-02 Structural Design Code for Pipeline of Water Supply and Waste Water Engineering. Beijing: China. Zhang, H., Zhang, P., Zhou, W., Dong, S. & Ma, B. 2016. A new model to predict soil pressure acting on deep buried jacked pipes. Tunneling and Underground Space Technology 60: 183–196.

18

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

A recent subway construction incident in soft alluvial deposits of Taiwan Wen-Chieh Cheng & Ge Li Xi’an University of Architecture and Technology, Xi’an, China

Md Mizanur Rahman University of South Australia, Mawson Lakes, Australia

ABSTRACT: A recent water leak incident taken place throughout the parallel tunnels LUO09 construction in the soil alluvial deposits in Kaohsiung, Taiwan was analysed and discussed in this study. Dumping quick-set cement was intended to ease the water leak incident but in vain. The water leak was initiated by the piping and the associated ground loss caused two large surface cave-ins. The existing underpass caused the jet-grout col­ umns installed not exactly in plumb. Their overlapping was estimated to be less than the design value of 60 cm, developing seepage-prone weak zones. The hydraulic gradient being equal to 12.1 and existence of seepageprone weak zones were deemed as the main cause initiating the water leak incident. The pinhole test results highlighted not only the nonplastic nature of the Kaohsiung silt but also its vulnerability to piping under large hydraulic gradients. Some bullet points that indicate engineers should do or avoid were summarised.

1 INTRODUCTION

incident often occurs, accompanied with significant casualties and economic losses, while developing infra­ structure in urban areas and relates mostly the phenom­ enon of water leak despite several protective measures available. In spite that serious attention has been drawn, consensus on prevention and mitigation meas­ ures has not reached yet. The objectives of this study are: (1) to analyse and discuss the water leak incident occurred while deepening the middle sump pit of the parallel tunnels LUO09, (2) to reveal the triggering mechanism, with reference to the testimony from the men at work and the field measurements, and (3) to suggest preventive measures against similar leaking incident.

Braced excavation and/or shield tunnelling in soft ground with high piezometric levels leads to a high potential to trigger water leak incident. Such water leak into excavation pit or sewage pipeline can result in ground collapse and damages to adjoining facilities. Prevention of such leaking incident has thus been deemed to be the key in developing sustainable infra­ structure in urban areas. Jo et al. (2016) indicated that the migration of the fine particles mainly contributed to the ground sinking incident occurred in Seoul, Korea. Hou et al. (2015) declared that the combined effects of the sewage pipeline corrosion, soil strength deterior­ ation, and construction disturbance triggered a 20-m long, 14-m wide and 12-m deep cave-in throughout soft soil tunnelling of Beijing metro Line 10. Chen et al. (2015) indicated that the ground collapse occurred in a 15.7-m deep excavation in very sensitive clay in Hangzhou, China caused damages to the retaining structure and the water main and led to ground sinking of a road next to the excavation. Feng and Lu (2016) reported that the bi-slurry grouting failed to stop the seepage causing a failure of retaining structure installed in very thick sand layers with a high phreatic surface during a metro station excavation in Nanchang, China. Tan & Lu (2017) indicated that a flawed slab connector embedded in the silt and sand layers has been deemed as being the main cause to lead to a sudden outburst of groundwater while deepening a subway excavation in Shanghai, China. The above studies indicate that

2 BACKGROUND Two 837-m long, 6.1-m wide parallel tunnels LUO09 were excavated using the shield tunnelling method. A 6.24-m diameter earth pressure balance (EPB) shield was responsible for tunnel boring operations. Their ends are linked to the subway stations O7 and O8 of Kaohsiung metro system, and a cross passage with a middle sump pit at 32.6 m depth was built for safety purpose. Several boreholes that penetrate through a 42­ m thick alternating layer of soft clay and silty sand and into a more than 30-m thick silty gravel were installed providing a detailed description of the field geological conditions. Figure 1 shows the soil properties profile. The groundwater level was at 5-6 m depths below the

DOI: 10.1201/9780429321559-3

19

Figure 2. Scenario for progressing the water leak incident. Figure 1. Soil properties profile.

ground surface. A series of 3.2-3.5 m diameter jetgrout columns at the depth of 35 m, with a mutual overlapping of 60 cm, were installed in the vicinity of the cross passage using the super jet-midi (SJM) method because of the surrounding silty sand of high permeability. The grouting tubes, while boring oper­ ations, were installed with horizontal included angles because of the existing overlying vehicle underpass. Upon completion of the cross passage, the 3.3-m diameter sump pit was excavated. 3 INCIDENT, DAMAGE AND REMEDIAL MEASURE

Figure 3. Two surface cave-ins.

It was noticed that an outburst of mud water presented while excavating to near the bottom of the sump pit. Authorities urgently responded to the water ingress by dumping sand bags and quick-set cement but in vain. Men at work soon heard sounds of breaks at liner joints of the tunnel on the south side, with water leaked into the tunnel from the ripped liner joints. The water also carried away fine particles and the tunnel due to this reason lost contact with the ground leading to stepwise surface settlement in the longitudinal dir­ ection. More soil, while the settled cone expanded upwards, fell by slipping along the vertical wall of the vehicle underpass and passing through the ripped liner joints into the tunnel. Not only serious ground loss, but also two undermined water mains transferred the soil into more flowable debris, causing a large surface cave-in on the south side. The ground loss also caused the tunnel on the north side to sink, accompan­ ied by the offsets of the liners. The soil fell into the tunnel mainly from the liner offset at the junction between the tunnel and the cross passage, leading to another surface cave-in. Figures 2-3 show the scenario for progressing the water leak incident and the two surface cave-ins respectively. The amount of earth and quick-set cement dumped to fill the surface cave-ins was estimated to be as much as 12,000 m3, whereas the water inflow, resulting from the two undermined water mains, was approximately 2,000 m3.

Subsequently a 2-year rehabilitation depicted in Figure 4 was performed to resume the traffic and its activities are detailed as follows: (1) Two watertight plugs in the tunnels at some dis­ tance from the point of water ingress were to secure the watertight effectiveness with add­ itional jet-grout blocks; (2) Porewater within the surrounding soil was frozen using the ground freezing method;

Figure 4. Schematic illustration for progressing the rehabilitation.

20

(3) Diaphragm walls of 1.5 m in thickness to retain the excavation-induced lateral loads were con­ structed at the depth of 60 m; (4) Dewatering wells screened at the bottom of the tunnels were constructed to secure the excavation face dry; (5) The removal of the undermined vehicle under­ pass was conducted upon completion of Region “II” excavation; (6) New tunnel liners were erected following com­ pletion of Region “I” excavation; (7) Controlled low strength materials were back­ filled to the bottom of new vehicle underpass, allowing for new vehicle underpass installation; (8) Controlled low strength materials were back­ filled to surface to resume traffic.

Figure 5. Layout of monitoring instruments and locations of surface cave-ins.

4 FIELD INSTRUMENTATION DATA Such significant ground loss caused two surface caveins to be developed on the south and north sides, respectively. The tunnel on the south side and the vehicle underpass due to loss of the contact with sur­ rounding ground significantly settled 2.7 m and 1 m, respectively. The tunnel on the north side showed a relatively small settlement of 0.16 m. Additionally, the adjacent railway also measured some settlements and track relative displacements. As discussed, the diaphragm walls with the frozen ground and soilcrete blocks aimed to secure the watertight effectiveness while progressing the rehabilitation. The rehabilitation including the removal of undermined vehicle under­ pass and the erection of new tunnel liners as well as the installation of new vehicle underpass was per­ formed when the water inside the walls was drained using the pumping wells and when the earth inside was removed. Thus, the proposed layout of monitor­ ing instruments depicted in Figure 5 intended not only to observe the groundwater level and surface settle­ ment variations while progressing the rehabilitation, but to assess the associated environmental impacts. To verify the watertight effectiveness of the dia­ phragm walls, frozen ground and soilcrete blocks, a group-well pumping test was performed using eight pumping wells installed inside the diaphragm walls prior to progressing the rehabilitation. Figure 6 pre­ sents the variations of the groundwater levels and the number of pumping well opened during the groupwell pumping test. The group-well pumping test con­ sisted of the pumping phase and recovery phase and each phase continued for 7 consecutive days. Table 1 lists the monitoring instruments used in the test. It can be seen from Figure 6 that the groundwater levels in the observation wells OW-S1, OW-S2, OW-04, and OW-10 declined sharply to the 33-35 m depths in the very beginning of the pumping phase and main­ tained almost constant until the end of the pumping phase. A similar tendency, but with smaller variation, was observed from the observation wells OW-02 and

Figure 6. Layout of monitoring instruments and locations of surface cave-ins.

Table 1. Details for the monitoring instruments used in the group-well pumping test. Location Inside d-wall Inside d-wall Inside d-wall Inside d-wall Inside d-wall Inside d-wall Outside d-wall

Water well type

Instrument number

Depth m

OW OW OW OW OW OW

OW-S1 OW-S2 OW-02 OW-04 OW-08 OW-10

40 45 45 45 45 45

OW

OW-14

40

Note: d-wall = diaphragm wall, OW = observation well.

Depth indicates where monitoring instrument is installed.

OW-08. There was a sudden increase in the ground­ water levels right before the completion of the pump­ ing phase, most likely because of the recharge effect resulting from a rainfall. The recovery rates of the

21

for SM-01 to 07. While the pit excavation appeared to have minimal influence on the ground surface settlements. It is worth to mention that when the ground was unfrozen prior to the installation of new vehicle underpass, the ground surface settlements increased very quickly, especially for SM-01 to SM­ 07. Despite the increases in the surface settlement, the surface settlements then soon reached a steady state condition.

observation wells OW-S1, OW-S2, OW-04, and OW­ 10 gradually declined from 1.2 m/day in the begin­ ning of the recovery phase to 0.1-0.2 m/day in the end of the recovery phase. Except the duration of rainfall, the groundwater level in the observation well OW-14 remained almost constant, indicating that the group-well pumping appeared to have negligible influence on the groundwater level outside the walls, which also indicated good watertight effectiveness. Figure 7 shows the variations of the ground sur­ face settlements while progressing the rehabilitation. The surface settlement point SM-T1 measured the smallest settlement, compared with the surface settlement points SM-T2, SM-T3, SM-T4, SM-A, SM-B, and SM-C. Similarly, the surface settlement points SM-01, SM-02, SM-04, SM-06, and SM-07 measured the relatively small settlements, compared with the surface settlement points SM-03 and SM­ 05. This phenomenon is due to the fact that the indi­ cated surface settlement points were all close to the short side of excavation pit that possesses larger bending stiffness than the long side, exhibiting smal­ ler lateral deflection and surface settlements. Despite the nonnegligible contribution to the ground surface settlements from the effects of the diaphragm wall construction and the strut removal, the associated increases were typically 4.8-9.4 mm and 5.5-7.5 mm, respectively, for SM-A to C and SM-T1 to T4 and 1-13 mm and 8.2-13.1 mm, respectively,

5 ANALYSIS AND DISCUSSIONS 5.1

Triggering mechanism

The piping phenomenon was responsible for initiating the water leak incident, with reference to the testi­ mony of workmen. Notwithstanding that, the initiation of piping was conceived variously. The existing vehicle underpass, in fact, led to some difficulties in installing the jet-grout columns exactly in plumb. Thus, the overlapping of jet-grout columns was esti­ mated to be less than the design value of 60 cm and seepage-prone weak zones were thus developed. On the other hand, as the groundwater level was at 5-­ 6 m depths below the ground surface, the head differ­ ence between the groundwater level and the 34 m depth where the piping initiated was 29 m. The length of shortest path for groundwater seepage was from the bottom of the soilcrete body to the point of water ingress and measured at about 2.4 m. This came out with the hydraulic gradient i being equal to about 12.1 (i=29 m/2.4 m=12.1). The critical hydraulic gradient icr for triggering piping within gapgraded sand-gravel mixture could be as low as 0.2-0.3 as compared to icr=0.9-1.0 for clean sands (Skempton & Brogan 1994). In spite that this might not be true for the Kaohsiung soil with the finer particle size range, the soil particles could easily be detached or washed away as subjected to such high hydraulic gra­ dient i of 12.1. It is evident that the seepage-prone weak zones allowed the water to flow through the jetgrout columns and into the sump pit and that this high hydraulic gradient made the water inflow even greater. The combined effects of high hydraulic gradient and existence of seepage-prone weak zones were deemed as the main cause to initiate the water leak incident. 5.2

Soil erodibility

Sherard et al. (1976) described the pinhole test where water flows through a 1-mm diameter hole punched in a specimen of compacted clay and the water emer­ ging from dispersive clay carries a suspension of col­ loidal particles (Bell & Walker 2000). Sherard et al. (1977) then utilised the pinhole test to investigate ero­ sion in clay-silt mixtures in response to the problem emerging from occasional failure of low-height earth dams which occurred upon water filling in Australia and U.S. The pinhole test provides a direct qualitative measurement of the dispersibility or deflocculation

Figure 7. Variations of ground surface displacements for progressing the rehabilitation: (a) SM-A to C and SM-T1 to T4 and (b) SM-01 to 07.

22

that there is no necessity to perform the test through to 1020 mm of head. The purpose to further increase the head is to classify dispersive soils, which is not within the scope of this study. The effluent turbidity and size of pinhole at the end of each test were recorded. If the effluent remains clear and the pinhole size unenlarged, then the soil is non-dispersive. Con­ trarily, the effluent is turbid and the pinhole size is enlarged for dispersive soil. Table 3 summarises the results of the pinhole test. For the specimens which were categorised as D, they were dispersive soils. There was also a common char­ acteristic. Most of the specimens primarily included silt with a fines content varying from 52 to 96% (Lu et al. 2019) and their Unified Soil Classification System (USCS) symbols were ML and SM-ML. The pinhole size enlarged in a range of 2-4 mm after the test. The specimen S-19 categorised as ND was nondispersive soil with a fines content of 94% and mainly consisted of clay. The USCS classified this specimen as CL, which was an indication of low plas­ ticity clay. The pinhole size for the specimen S-19 remained unenlarged after the test. The test results revealed a fact indicative of the tendency of the silt from Kaohsiung, Taiwan to possess high dispersion, or very low strength, or both. The particle-size distri­ bution analysis indicated that the Kaohsiung soil belonged neither to the gap-graded gravelly sands sus­ ceptible for segregation piping nor to the dispersive clay prone to internal erosion. In fact, the particle size for the Kaohsiung soil lay midway in the range of silt and sand. The pinhole tests were regarded as an effective means to identify not only the nonplastic nature of the Kaohsiung silt contained in the sand but also its high vulnerability to piping or internal ero­ sion. It may be conclusively mentioned that the seepage-prone weak zones and high hydraulic gradient initiated the water leak incident and that water from the two undermined water mains transferred the Kao­ hsiung silt into more flowable debris and its nonplas­ tic nature aggravated the collapse even further. Despite an apparent inadequacy in dealing with simi­ lar incident in other sites, the local sandy silt was found to be prone to the piping at deep depths under large hydraulic gradients and the two undermined water mains enhanced its flowable nature, thereby enlarging the scale of the water leak incident.

and consequent erodibility of clay soils as described in ASTM standard D4647-93 (ASTM 2006). There were seven 33-mm diameter, 25-mm height speci­ mens tested in the pinhole test for which their proper­ ties are summarised in Table 2. The average grain size D50 for the specimens was about 0.075 mm. A steel nipple was pushed into the specimen and the hole was punched through the nipple as a guide hole using a 1-mm diameter steel needle. Wire screen and pea gravel were placed in the pinhole cell on either side of the specimen, as shown in Figure 8. As the pinhole cell assembled, distilled water was allowed to percolate through the specimen under constant heads of 50, 180, 380, and 1020 mm, in accordance with ASTM standard D4647-06. The principal differenti­ ation between dispersive and non-dispersive soils, however, is given by the test results under 50 mm of head, as suggested by Sherard et al. (1976), indicating

Table 2.

Summary of soil properties in the pinhole test. Sand

Silt

Clay

Water content

LL

Specimen

%

%

%

%

%

S-14 S-15 S-16 S-17 S-18 S-19 S-20

19 48 15 4 20 6 15

76 52 82 86 74 56 84

5 0 3 10 6 38 1

22.4 25.6 24.4 19.5 21.7 42.9 25.1

27 21.4 27.5 35.2 26.5 43.9 26.1

PI

4.1

0.8

1.2

6.4

3.3

20

1.2

Note: LL = liquid limit, PI = plasticity index.

5.3

Lessons learned

Quite often subway construction failures are not explained nor published, which makes the same mis­ takes to be made another time with great casualties and economic losses. Some bullet points that indi­ cate what engineers should do or avoid are learned and summed up as follows: (1) Quality control for grouting: Since in-ground obstructions such as wastewater and electricity mains can result in some difficulty in installing jet-grout columns exactly in plumb, a trial

Figure 8. Layout of monitoring instruments and locations of surface cave-ins.

23

Table 3.

Summary of the pinhole test results. Final head

Specimen

mm

S-14 S-15 S-16 S-17 S-18 S-19 S-20

50 50 50 50 50 180 50

Colour cloudiness

Moderately dark Barely visible Moderately dark Dark Dark Clear Moderately dark

Pinhole size

Specific surface area

mm

m2/g

3 2 3 4 4 1 3

3.20 3.27 3.11 2.95 2.84 6.18 3.07

Dispersion classification

D D D D D ND D

Note: Pinhole size = Pinhole size measured after the test, D = dispersive soil, ND = non-dispersive soil.

grouting should be performed prior to the formal one to verify the design grouting parameters. (2) Seepage-prone weak zone: Except for trial grouting, additional soilcrete columns, result­ ing from chemical grouting, should be con­ sidered during design phase and constructed next to jet-grout columns, which not only increases the length of path for the ground­ water seepage, but also prevents formation of the seepage-prone weak zone. (3) Local soil nature: The Kaohsiung silt contained in the sand has proved to possess high disper­ sion, or low strength, or both. There is a necessity to investigate the nature of local soils while progressing upcoming subway con­ struction project if they are found to exhibit peculiar behaviour differentiated from common sense in soil mechanics.

seepage was from the bottom of the soilcrete body to the point of water ingress; that is, 2.4 m, corresponding to the hydraulic gradi­ ent i of 12.1. (3) The existing vehicle underpass led to some dif­ ficulties in installing the jet-grout columns exactly in plumb and some seepage-prone weak zones were thus developed. The seepage-prone weak zones allowed the water to flow through the jet-grout columns into the sump pit and such high hydraulic gradient made the water inflow even greater. The combined effects were deemed as the main cause initiating the water leak incident. (4) The results of the pinhole test identified not only the nonplastic nature of the Kaohsiung silt but also its vulnerability to piping or internal erosion under large hydraulic gradients. Water from the two undermined water mains and the nonplastic nature of the Kaohsiung silt aggra­ vated the collapse even further. More detailed investigation is considered to be necessary to address the explored issue with different aspects of view.

6 CONCLUSIONS The main cause to lead to the water leak incident was investigated. The pinhole test results highlighted not only the nonplastic nature of the Kaohsiung silt but also its vulnerability to piping under large hydraulic gradients. Some main conclusions can be drawn as follows:

REFERENCES ASTM. 2006. D4647-93 Standard Test Method for Identifi­ cation and Classification of Dispersive Clay Soils by the Pinhole Test. West Conshohocken: PA. Bell, F.G. & Walker, D.J.H. 2000. A further examination of the nature of dispersive soils in Natal, South Africa. Quarterly Journal of Engineering Geology and Hydrogeology 33: 187–199. Chen, R.P., Li, Z.C., Chen, Y.M., Ou, C.Y., Hu, Q. & Rao, M. 2015. Failure investigation at a collapsed deep excavation in very sensitive organic soft clay. Journal of Performance of Constructed Facilities 29(3): 04014078. Hou, Y.J., Fang, Q., Zhang, D.L. & Wong, L. 2015. Exca­ vation failure due to pipeline damage during shallow tunnelling in soft ground. Tunnelling and Underground Space Technology 46: 76–84. Jo, Y.S., Cho, S.H. & Jang, Y.S. 2016. Field investigation and analysis of ground sinking development in

(1) The piping was responsible for initiating the water leak incident. Dumping sand bags and quick-set cement failed to ease the water leak. Mud water carried away fine particles in the ground and together flowed into the tunnels through the ripped liner joints leading to serious ground loss. Such ground loss caused two sur­ face cave-ins. The surface cave-ins not only impeded traffic, but also caused damages to adjacent properties. (2) The hydraulic pressure at the point of water ingress was estimated to be about 300 kPa. The length of shortest path for groundwater

24

a metropolitan city, Seoul, Korea. Environmental Earth Sciences 75: 1353. Lu, J., Wang, T.H., Cheng, W.C., Yang, T. & Luo, Y. 2019. Permeability anisotropy of loess under influence of dry density and freeze-thaw cycles. International Journal of Geomechanics 19(9): 04019103. Sherard, J.L., Steele, E.F., Decker, R.S. & Dunnigan, L.P. 1976. Pinhole test for identifying dispersive soils. ASCE Journal of Geotechnical Engineering Division 102: 69–85.

Sherard, J.L., Dunnigan, L.P. & Decker, R.S. 1977. Some engineering problems with dispersive clays. In: Sherad, J.L. & Decker, R.S. (eds.), ASTM STP 623; Pro­ ceedings of Symposium on Dispersive Clays, Related Piping, and Erosion in Geotechnical Projects, Chicago, 27 June–2 July 1976. Tan, Y. & Lu, Y. 2017. Forensic diagnosis of a leaking acci­ dent during excavation. Journal of Performance of Con­ structed Facilities 31(5): 04017061.

25

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Settlements due to tunneling in the City of São Paulo A. Lopes dos Santos Navier Géotechnique, École des Ponts ParisTech, Champs-sur-Marne, France

W. Bilfinger Vecttor Projetos Ltda., São Paulo, Brazil

H.C. Rocha Companhia do Metropolitano de São Paulo - Metrô-SP, São Paulo, Brazil

ABSTRACT: Major tunneling works in the city of São Paulo have started in the 1970´s, with the construc­ tion of the 1st Metro Line. Since then, tunnels with different cross sections have been constructed regularly, using either mechanized or conventional tunneling methods. From a geological point of view, the city is mainly located in a sedimentary basin, with its subsoil partly composed by residual soils overlaying bedrock. Experience has shown that the induced settlements in the residual soils are higher and more variable. Historic registers of tunneling induced settlements measured for different constructive techniques and soil types are presented on this paper. Focus is given to the measurements during the construction of Metro Line 4, where approximately 5 km of line tunnels were excavated using NATM in residual soils, saprolite and rocks, and around 8 km were excavated in tertiary sediments, using an EBM TBM. Typical values for volume losses on this context are proposed.

were excavated using NATM in residual soils, sapro­ lite and rocks, and around 8 km were excavated in tertiary sediments, using an EBM TBM. Finally, typical values for volume loss evaluated in different geological environments and different constructive methods will be proposed.

1 INTRODUCTION São Paulo is one of the so-called Megacities of the world and, for this reason, the use of tunnel is the only lasting solution to solve several problems, including transportation and utilities. Historical registers show that in the early 20th century some tunnels were built. But major tunneling started in the 1970´s, with the construction of the 1st Metro Line. Since then, tunnels with different cross sections have been built regularly, using either mechanized or con­ ventional tunneling methods. São Paulo is located, from a geological point of view, mainly in a sedimentary basin, with tertiary clays and sands, locally overlain by quaternary soils and manmade ground. Partially, the subsoil consists also of residual soils overlaying bedrock. Tunnel construction in these completely different environ­ ments generates the need for different constructive techniques and experience has shown that the induced settlements in the residual soils are propor­ tionally higher and more variable. The paper will present historic registers of tunnel­ ing induced settlements and analyses of volume losses measured for different constructive tech­ niques. Focus will be the settlements measured during the construction of Line 4 of Metro São Paulo, where approximately 5 km of line tunnels

2 SETTLEMENTS DUE TO TUNNELLING 2.1

Fundamental aspects

The first fundamental aspect about settlements related to tunneling is that they will certainly occur. Any relevant ground excavation will result in ground movements, of which the causes, magnitude and consequences depend on many factors, constituting a very complex phenomenon briefly discussed in the following paragraphs. The main causes of settlements associated to tun­ neling are related to soil stress state disturbance due to the excavation process. Every part of this process may have an influence on the induced ground move­ ments, from groundwater lowering, to the crosssection opening itself. Ground pre-conditioning, lining stiffness and its installation distance from the face also have an influence. In conventional tunneling, the settlements induced ahead of the tunnel face are directly related to the

DOI: 10.1201/9780429321559-4

26

2.2

control of the face stability (Leca (2006)). Temporary support also plays an important role by means of its relative stiffness and the time required for its installa­ tion. The stiffer it is, and the closer it is installed from the tunnel face, the less settlements should occur, since the soil mass is less released. The construction staging influences not just the magnitude, but the amount of settlement to take place ahead and behind the face. The bigger the cross-sectional area, the more settlements will occur at the face’s projection. The unsupported tunnel length and the lining installation delay influence the settlements ahead of the face. Behind the face, settlements are influenced by the distance in which the permanent lining is installed from the face. Finally, the liner itself will be subjected to deformations which implies in surface movements. In shield tunneling, settlements ahead and above the face are conditioned by the confinement at the face, the ground conditions and hydraulic conditions. Over-excavations due to peripheral cutters, driving difficulties and roughness of the cutting wheel may be a cause of settlements along and ahead of the shield. At the shield tail, the gap between the shield and the lining may be another cause, together with the grout injection pressure (or lack of) and clear­ ance on the interface soil-lining after the shield pas­ sage. Behind the shield, the lining deformation due to soil loading is other cause of settlements. Hydro-geological conditions and groundwater movements are another source of settlements. Lowering the groundwater table (by drainage sys­ tems or by the tunnel itself) may cause the consoli­ dation of soft soil layers inside of its radius of influence, thus generating settlements. Hydraulic gradients towards the tunnel face may reduce the soil’s mechanics properties, favoring deformations to occur. In the case of excavations in karstic soils or rocks, voids can be created by the water percolation. Working conditions and workmanship quality is another cause of settlements. Vibrations due to blast­ ing may imply in near surface ground compaction, as observed by Bilfinger et al (2016). Settlement’s magnitude and time-dependent behavior is closely related to soil’s geotechnical aspects such as compressibility, strength, permeabil­ ity, anisotropy and the spatial distribution of those properties. For a given ground condition, the work quality and a design focused on displacement control will play an important role. A very important aspect related to soil settlements induced by tunneling is its consequence. In densely occupied urban environments, ground movements could result in damage to existing buildings and/or utilities. This very complex soil-structure problem depends on many variables, such as the building’s stiffness and its strength, which imply on its capacity to resist to differential settlements. Tunnel design in such areas should be oriented to minimize or mitigate damage, while tunneling in non-occupied areas often can be optimized regardless of surface settlements.

Settlement trough

Tunneling is mainly a three-dimensional problem and so do the settlement trough induced by its construc­ tion. The settlement trough evolves as the tunnel advances and it is supposed to stabilize to its max­ imum value far behind from the tunnel face (except if time-dependent behavior takes place). Settlements can be observed ahead of the tunnel face. Field measurement observations have shown that the transverse settlement shape can be well described by a Gaussian distribution curve (Schmidt (1969), Leca & New (2007)). This very simple formulation has been used in the past years and it is wide spread amongst practical engineers due to its simplicity, which lead to a first estimation of surface settlements requiring very few parameters to be used. Other methods are also available. Celestino et al (2000), for example, propose yield-density functions to fit the settlement trough, stat­ ing that it represents better ground distortions. Numer­ ical modeling is another tool to estimate the settlement magnitude and its spatial distribution. It enables includ­ ing other geotechnical parameters to estimate more complex effects, such as soil consolidation. Its use, however, is more time consuming. This Gaussian curve is mathematically described by the Equation 1:

where Sv ð yÞ is the soil settlement at a distance y perpendicular to the tunnel axis. Sv;max is the max­ imum settlement measured above the tunnel axis, and i is the horizontal distance from the inflexion point of the Gaussian curve to the tunnel axis. This idealized curve is presented in Figure 1. The volume of the settlement trough per unit length of the tunnel can be calculated by integrating Equation 1, resulting in Equation 2:

By considering the soil mass as incompressible, neglecting dilatation and swelling, the volume of the settlement trough observed in the surface should be equal to the volume loss (VL ), which is the volume of the ground that has deformed into the tunnel after it has been constructed (see Figure 2 for definition). Based on these assumptions, the ground loss ratio is equal to the ratio between the volume loss (or the volume of the settlement trough) and the excavated volume (At), thus following:

27

(2012)): pre-cambrian embasement, paleogen to neogen sediments (also known as tertiary sediments) and quaternary deposits. The city’s rock embasement, in a simplified way, is mainly composed by metamorphic and igneous rocks, besides masses of intrusive granit­ oids. The physical or chemical weathering of those rocks gave origin to the residual soils (lat­ eritic soils) which also compose the city’s geo­ logical background. São Paulo’s sedimentary basin is formed by ancient sediment deposits, which is mainly filled by the so-called “São Paulo” and “Resende” formations. Resende formation covers the most part of the city’s sedimentary basin and it is composed by an intercal­ ation of stiff over-consolidated clays and loose to compact, fine to coarse sands. Alluvial deposits occur mainly near the city’s rivers, in layers not thicker than ten meters. Very soft organic clays, peat and loose sands characterize this formation. Manmade ground with the most varied compos­ itions was a result of the human interventions in the past centuries as an effort of urbanization. A detailed description of Sao Paulo’s geo­ logical background can be found in Monteiro et al (2012) and Gurgueira (2013). Geotechnical background regarding the city’s subsoil can be found in Pinto et al. (1993), Negro et al. (1992), Massad (2012) and Futai et al (2012). Tables 1 and 2 summarize some typical geomechanical parameters of local soils. Sedimentary sandy soils have variable fine con­ tent and, consequently, variable shear strength

Figure 1. Idealized transverse settlement trough (O´Reilly & New, 1982).

Figure 2. Definitions of ground loss and the volume of the settlement trough.

The width of the settlement trough is related to the distance of the inflexion point (i) to the tunnel axis. O’Reilly & New (1982) have found strong evi­ dence that the distance of the inflexion point varies linearly with the tunnel’s depth (z0), by means of the Equation 4:

Table 1. Sedimentary clayey soils - geomechanical parameters from literature (in parenthesis: lower and upper boundaries). Red porous clay

in which K is the so-called “trough width param­ eter” and depends on the soil type. According to Leca & New (2007) the excavation methodology does not affect K. In the following chapters, historical registers of soil settlements due to tunneling in the city of São Paulo are analyzed. Equation 3 is used to estimate the ground loss ratio resulting from excavations. By fitting Equation 1 to the settlement measurement data available from Metro line 4 it was possible to scatter values of i and z0, thus leading to an estima­ tion of K values as a function of local geology.

γn (kN/m3) 15.2 (13.5-16.5) w (%) 41 (33-47) e 1.54 (1.38-1.85) c´ (kPa) 32 (10-70) ϕ´ (º) 29 (26-33) Ei (MPa) Su (kPa) Ei/Su 600 (300-1000) 480 E50/Su (200-1000)

3 SÃO PAULO’S GEOLOGICAL ASPECTS 3.1

Generalities

The city of São Paulo’s geology can be basically div­ ided into three compartments (Monteiro et al

28

Red stiff clay

Variegated soil

Grey clays

17.2 (16.2-18.0) 42 (35-47) 1.28 (1.17-1.46) 69 (50-90) 26

15.5-21.4

18.8-19.9

16-43

22-29

0.48-1.55

0.61-0.87

10-90

50-150

15-35

23

21 x NSPT 520 (340-740) 420 (120-600)

32 x NSPT 400 (300-600) 290 (140-600)

19 x NSPT 14 x NSPT 1200 (750-1700) 230

4 HISTORIC REGISTERS

Table 2. Residual soils - geomechanical parameters com­ piled from literature.

γn (kN/m3) w (%) e k (m/s) c´ (kPa) ϕ´ (º) G0 (MPa)

Gnaisse

Granite

16.1 19.8 1.05 5.7x10-5 14 (0-50) 30.5 (28-32) 127 + 1.4 x NSPT

15.5 24.2 1.18 7x10-6 10

4.1

Many works have been published in the past 25 years containing data and analysis related to tunnel­ ing in the city of São Paulo. A vast bibliographic review about this subject can be found in Bilfinger et al (2012). Among the available data, there are settlements measured during the metro line 2 construction, those published by Negro (1992), Parreira & Levada (1998), Sozio et al (1998) and Celestino & Ferreira (1996). Registers covers both NATM and TBM tun­ nels, excavated in quaternary and tertiary sediment­ ary soils and gneiss residual soils. The historical registers reveal that surface settle­ ment magnitude tends to be higher in tunnels exca­ vated in residual soils. This same tendency was observed in the metro line 4.

30 (27-31)

(32° < ϕ 1), defined as the ratio of the bearing capacity to the net vertical pressure at the base of the soil, can be expressed as

45

Figure 4. Pillar numbers and excavation condition.

excavation depth become H+T (T is the thickness of the improvement). Thus the Fs are expressed as

in which, B' is the extension depth of the failure surface below the excavation, γ is the unit weight of the saturated clay, sua is the average undrained shear strength within the excavation depth and sub is the average undrained shear strength on the failure pffiffisur­ ffi face below the base. The B' is equal to B/ 2 or depth to hard layer below formation level which is smaller. For an excavation of limited length (L), a modification factor λs = (1 + 0.2B/L) can be applied to Equation 1 (Chang, 2000) to consider the geometry effect. The factor of safety (Fs) can be expressed as

The calculated factor of safety refer to Fs3. The results are shown in Table 2 for standard sec­ tion and Table 3 for south section respectively. Fs2 are approximately 1.2 - 1.4 times that of Fs1, indicat­ ing that the 3 m jet grouting increases the safety factor by 20% - 40%. Because of geometry differ­ ence, the Fs of south section is about 1.15 times that of standard section. Fs2 decrease to below 1.5 when H of the standard section exceeds 15 m and H of the south section exceeds 17 m. Due to yielding and nonlinear behavior of soil, base heave will become critical when the factor of safety is below 1.5 (Terza­ ghi et.al., 1996).

In the two proposed methods, wall penetration below the formation level is ignored and the effect of wall stiffness is not considered. Therefore, the cal­ culated Fs is not the real safety factor, but it can be used as a reference to evaluate the displacement of the retaining system and ground movement (Terza­ ghi et al., 1996). The Fs of standard section (L/B = 176.2/22 = 8, B/Hmax = 22/17.64 = 1.25) were calculated with Equation 1 and that of the end section (L/B = 26.8/ 16.3 = 1.64, B/Hmax = 16.3/19.5 = 0.84) were calcu­ lated pffiffiffi with Equation 2. The B' is determined by B' = B/ 2 because the hard layer of layer ⑦ was much deeper. Two average undrained shear strengths of Sub, Sub1 and Sub2, were used and the corresponding results were denoted by Fs1 and Fs2 respectively. The Sub1 was determined by the undrained shear strength profile given in Figure 1, which neglected the effect of jet grouting; the Sub2 considered the effect of jet grouting by using undrained shear strength 175 kPa for the 3 m thick improved soil. In addition, the effect of ground improvement on base heave stability at the final excavation stage was analyzed using another method which treats the improved soil as an overburden load and the

Table 2.

Safety factor Fs of standard section.

H (m) Sub (kPa) Fs

Table 3.

Sub1 Sub2 Fs1 Fs2 Fs3

Fs

46

9.5

12

15

17.6

27.4 39 1.66 2.37 /

34 42 1.31 1.62 /

39 49.5 1.19 1.51 /

45 55 1.11 1.35 /

48.5 59 1.02 1.24 1.15

Safety factor Fs of end section.

H (m) Sub (kPa)

6

Sub1 Sub2 Fs1 Fs2 Fs3

6

9.5

12

15

17.6

27.4 41 1.89 2.69 /

22 34 1.49 1.84 /

25 37 1.36 1.72 /

29 40 1.31 1.60 /

35 46 1.14 1.35 1.32

4 UPLIFT OF THE PILLARS AND ITS EFFECT 4.1

Uplift of the pillars

Figure 5 shows the variations of cumulative uplift δhv (positive for up toward) with time of the two pil­ lars (LZ1 in south section and LZ2 in standard sec­ tion) in the layered excavation of Sa1 to Sa5. The two pillars uplifted continuously during excavation. Even at the early stage of the slab construction and no excavation was conducted nearby, they still con­ tinuously uplifted in a short period, which indicates that the basal heave and pillar uplift was time related. Then the uplift slightly reduced and almost maintained constant during the slab construction ­ a gradually increased load applied on the base. The late stepped excavations (stages of S2-S5) in stand­ ard section do not seem to affect the uplift of the two excavated pillars. The maximum cumulative uplift of the two pillars were very different, 30 mm at LZ1 (end section) and 80 mm at LZ2 (standard section). The very different observations in the end section and the standard section were likely related to the geometry effect: the base heave in the end part of a long excavation was three-dimensional and had large safety. Figure 6 shows the variations of cumulative uplift with time for pillars LZ3 to LZ11 in the standard section excavated by the stepped procedure. The par­ tial excavation near the south section in the layered excavation made the pillars LZ3-6 uplift 7-41 mm before the stepped excavation, while the pillars LZ7­ 11 uplifted from the beginning of the stepped exca­ vation of S2. All these pillars continuously uplifted during the stepped excavation and then maintained stable during the slab construction. However, LZ3 uplift very little in the excavation stage S2. Nearby constructed concrete slab in south end section seems to reduce the base heave around the LZ3 in the excavation. The maximum cumulative uplifts of the thirteen pillar (LZ1-13) are shown in Figure 7, which were in

Figure 6. Variation of uplift with time in stepped excavation.

Figure 7. Measured maximum cumulative uplifts.

the range of 20 mm – 87 mm. Due to the geometry effect, the maximum cumulative uplifts of two pil­ lars of LZ1 and LZ13 in the end sections were below 30 mm. The maximum cumulative uplift of pillars in the standard section tended to decrease from south to north, i.e. the direction of the stepped excavation: pillars in the early excavation area had larger uplift than those in late excavation area. LZ3 and LZ4 recorded unexpectedly small uplift, which may be attributed to the constructed concrete slab in the south end section and other factors related to the complex excavation and construction procedure in field. The monitoring data of LZ6 and LZ10 shown in Figure 8 are selected to investigate the potential influence of construction procedure on the uplift. It can be seen that the uplift increments of the two pil­ lars in the last excavation stage (S4 for LZ6 and S6 for LZ10) were very different, 25 mm for LZ6 and 11 mm for LZ10. Smaller excavation slot (see Figure 4) therefore quicker slab construction may be the reason of the smaller uplift increments of the late excavated LZ10 than the early excavated LZ6 in the last excavation.

Figure 5. Variations of uplift with time in layered excavation.

47

Figure 8. Uplift of LZ6 and LZ10 with excavation.

Figure 10. Variation of the uplift with safety factor.

Figure 9 shows the variation of the δhv with excava­ tion depth (H) for four selected pillars. It should be noted that LZ1 and LZ2 are layered excavated so the excavation depth can be easily determined; LZ6 and LZ10 were stepped excavated and the excavation depth was determined by the elevation of the excava­ tion surface where the pillar locates. It can be seen that the uplift of LZ1 in the south section almost increased linearly with the excavation depth. For the three pillars (LZ2, LZ6 and LZ10) in the standard sections, uplift increased with a larger rate than the end section and accelerated significantly at the final stage where exca­ vation depth exceeded 15 m. Relationships between pillar uplift and the factor of safety (Fs2) of the four pillars were shown in Figure 10. Because the Fs versus H relationship is not a linear one (see Equation 1), these δhv versus Fs curves were lightly different from the δhv versus H curves shown in Figure 9. For the three pillars (LZ2, LZ6 and LZ10) in the standard section, the first turning point appears around Fs = 1.6 and the second turning point appears around Fs = 1.3. The pillar LZ1 in the end section presents much smaller uplift in the last excavation stage because of larger Fs. According to Figure 10, there is a tendency that the Equation (2) underestimated the

Fs of the end section in the last excavation stage where the B/H below 1.0. 4.2

Effects on the reinforced concrete strut

Pillar uplift had significant impact on the reinforced concrete strut of the standard section. As an example shown in Figure 11, transverse cracks occurred on the top surface of some reinforced concrete struts because of significant uplift. Figure 12 shows a typical struts force pro­ files measured at the standard section near pillar Q6 in different excavation stages. The axial force of the reinforced concrete strut (i.e. the first strut in the figure) decreased in the last three stages. Although there is some uncertainty in the accur­ acy of the measured axial force of the concrete strut, it may be concluded that the reinforced concrete strut changed from compression state to bending state under the progressively uplift of the support pillar. Treating the reinforced concrete strut as a beam with fixed ends and the support uplift acting on its

Figure 9. Variation of the uplift with excavation depth.

Figure 11. Cracks on top of the reinforced concrete strut.

48

5 RELATIONSHIPS WITH THE DEFLECTION OF THE DIAPHRAGM WALL Basal heave results from the inward movement of the soft clay around the excavation under undrained condition. Therefore, basal heave is often accompan­ ied by large deflection of the retaining wall. In the standard section, most of the observed maximum deflection of the diaphragm wall δhhm were in the range of 45 mm - 72 mm and the non-dimensional deflection δhhm/H were in the range of 0.25% ­ 0.40%. They were much smaller than the value of about 0.7% estimated by the δhhm/H versus Fs rela­ tionship provided by the Mana and Clough (1981) with Fs = 1.2. The maximum deflections δhhm in the end section were in 40 mm and the δhhm/H was in 0.2%, smaller than the standard section because of geometry effect. Figure 13 shows a typical result measured at the point Q9 in standard section. At the final excavation step (S5, H = 17.6 m) when the Fs was around 1.2, the deflection accelerated to 60 mm and the toe of the diagram wall suddenly moves inward 18 mm. The retaining system and the construction procedure successfully controlled the deflection of the dia­ phragm wall, but not necessarily the inward move­ ment of the deep soil under low factor of safety. Therefore, it can be concluded that the soil move­ ment under the toe of the diagram wall also contrib­ uted to the uplift of the pillar. The deflection of Q28 in south section shown in Figure 13 were very differ­ ent from that of Q9. The toe of the diagram wall was almost fixed and the inward movement of the soil was limited within the depth of the diaphragm wall. Figure 14 shows the relationships of pillar uplift δhv with the maximum deflection of the diaphragm wall

Figure 12. Typical measured axial force at standard section.

middle, the relationship between the mid-span moment (M) of the strut and the support uplift δhv can be formulated by

in which, M, Mq and MN are the moment caused by the pillar uplift δhv , the weight of the concrete strut in term of a uniform load q, and the axial force N acting on the concrete strut, respectively; l0 is the half the length of the strut; EI is the bending rigidity of the strut. The concrete strut was 20.4 m long, 700 mm wide, 900 mm high and it was reinforced with 8Ф25 rebar at both top and bottom. The ultimate bending capacity of the strut (Mu) was 1113 kN·m. Based on monitoring data, the axial force N of the concrete strut was within 1000 kN·m, while the uplift of the pillar was within 10 cm, so the MN was within 100 kN·m. The q was 25.3 kN/m and the Mq was about 110 kN·m. Thus the concrete strut had great safety under the MN and Mq, and an increasing Mδ led to the cracking. According to Equation (4), the Mδ is strongly related to the bending rigidity EI. An equivalent bending rigidity EI = 3:44 x 1011 kN · mm2 which is much smaller than the elastic bending rigidity (EI ¼ 1:29 x 1012 kN · mm2 ) was used to consider the non-linear characteristics of the reinforced con­ crete strut caused by the cracking. Taking the M in Equation 4 by the Mu of 1113 kN·m and l' = L/2 = 10.2 m, the calculated critical uplift corresponding to bending capacity of the concrete strut was 64mm. Therefore, the middle of the concrete strut seems to reach the ultimate state when uplift reached about 64mm. However, field performance indicated that the concrete strut didn’t failure even the uplift over 80 mm. It was very likely that the ends moments were smaller than the mid-span moment and still in safety.

Figure 13. The deflections of the diaphragm wall.

49

The retaining system and construction procedure successfully controlled the maximum non-dimensional deflection δhhm/H below 0.5%, much smaller than that estimated with the Terzaghi’s safety factor by the rela­ tionship provided by Mana and Clough (1996). The final δhv/δhhm of the standard section and end section were 1.33 and 0.70 respectively. Inward movement of the soil under the toe of the retaining wall contribute to large δhv and large δhv/δhhm in the standard section.

ACKNOWLEDGEMENT The authors appreciate the financial supports pro­ vided by the Natural Science Foundation of China (NSFC) (Grant No. 41972273) for this research. The authors also thank for Henry Zhang’s review for this technical paper and the improvement of the writing.

Figure 14. The relationship between the pillar uplift δhv an the maximum deflection of the diaphragm wall δhhm.

δhhm near the pillar LZ6 in the standard section and LZ1 in the south end section. The ratio of δhv/δhhm were around 0.5 - 0.7 when the δhhm was below 40 mm and the toe of the diaphragm wall was almost fixed. When the δhhm exceeded 40 mm, accompanied by the inward movement of soil under the toe of the diaphragm wall, the ratio of δhv/δhhm of the standard section sharply increased to 1.33 at the end of excavation. 6

REFERENCES Chang, M.F. 2000. Basal Stability analysis of braced cuts in clay. Journal of Geotechnical and Geoenvironmental Engineering 126(3): 276–279. Chen R.P., Li Z.C., Chen Y.M. et al. 2015. Failure investi­ gation at a collapsed deep excavation in very sensitive organic soft clay. Journal of Performance of Con­ structed Facilities–ASCE 29: 04014078–1. Feng H., Liu Y.H. & Xu C.L. 2014. Effect of column uplift to the stability of steel supporting system in metro deep excavation. China Science Paper 9(11): 1301–1305. (in Chinese) Mana, A.I. & Clough, G.W. 1981. Prediction of movements for braced cuts in clay. Journal of Geotechnical Engin­ eering Division, ASCE 107(6): 759–777. Shirlaw, J.N, Tan, T.S. & Wong, K.S. 2006. Deep excava­ tions in Singapore marine clay. In Bakker, et al (ed.), Geotechnical Aspects of Underground Construction in Soft Ground; Proc. intern. symp., Amsterdam, 15-17 June 2006. London: Taylor & Francis Group. Sun, Y.F. 1988. The ancient river channel buried in the Quaternary sedimentary layers of Shanghai and the Yangtze River estuary. Shanghai Land Resources (4): 9–16. (in Chinese) Tanaka, H. 1993. Behavior of braced excavation stabilized by deep mixing method. Japanese Society of Soil mech­ anics and foundations Engineering, 33(2): 105–115. Terzaghi, K, Peck, R.B. & Mesri, G. 1996. Soil Mechanics in Engineering Practice. New York: John Wiley & Sons.

CONCLUSIONS

Significant geometry effect on the uplift of the pil­ lars was observed in this case: the standard section with an excavation depth of 17.6 m presented much larger uplift than the end section with an excavation depth of 19.5 m. Significant pillar uplift in the standard section damaged the safety of the concrete struts and made the mid-span moment reach its cap­ acity. Pillar uplift in the standard section was roughly divided into three stages: the early lowspeed stage, the acceleration stage with low factor of safety (Fs) around 1.3, and stable stage during slab construction. The maximum uplift in the stand­ ard section tended to decrease with the direction of excavation. Small excavation slob and quick con­ struction of the concrete slab can help to reduce the uplift of the pillars.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Solution implementation for a tunnel collapse in a weak embankment soil: The case of the Ali Boumendjel Metro station, Algiers, Algeria R. Hebib Department of Geology, Ferhat Abbas University, Setif, Algeria

Z. Derriche Laboratory of TransportEnvironment Engineering Public Works, National high School of Public Works, Algiers, Algeria

B. Alloul & D. Belhai University of Sciences and Technology Houari Boumediene, Algiers, Algeria

ABSTRACT: This study illustrates the utmost importance of applying the ‘‘Learn-as-you-go ‘‘philoso­ phy in the design and construction of underground structures in densely populated urban areas. It describes the speaking case of a tunnel section that forms a part of the Ali Boumendjel station of the Algiers metro development. The construction works of this tunnel section led to the collapse of a part of the vault provoking ground subsidence and surface settlements which could expose to danger the overlying buildings and also the main waste water collector of the region. After the project description and the initial design details presentation, the study describes the instability events that occurred, focus­ ing on the unexpected geological conditions disclosed by the excavation and the construction difficulties experienced. The study analyses then the ground movements recorded by the monitoring system imple­ mented, and details the technical solutions and interventions that made it possible to finalize the works with the required safety measures.

work projects. The present study illustrates a real case where the observational method made it pos­ sible to avoid the occurrence of disastrous instabil­ ities during the excavation of a subway gallery in the old portion of the Algiers city center.

1 INTRODUCTION Tunnel projects in densely populated urban areas are among the most difficult constructions to carry out (Guglielmetti et al. 2008), particularly because of the difficulties to achieve sufficiently accurate descrip­ tion of the project real conditions before starting the works (geotechnical, building foundation, presence of underground utility networks). Excavation in urban areas imposes specific chal­ lenges to address which are related to the impact of the induced ground and structural movements on the integrity of the urban environment (buildings, roads, utility networks, etc.). Despite their relative develop­ ments, the tunnel design methods still cannot provide real-time response to the stability problems that may arise during excavation work. This is due to the com­ plexity of the phenomena that occur and the great uncertainties related to the real conditions that cause them. For these reasons, the implementation of the observational type of approach (Terzaghi & Peck 1967, Peck 1969, Nicholson et al. 1999) is necessary to ensure the good execution of the underground

2 PROJECT DESCRIPTION The “Ali Boumendjel” Algiers metro station is part of the extension of the first line of the Algiers Metro that links “Place des Martyrs” in the West to “AinNaadja” and El-Harrach “in the East of the city. It is located in the old center of Algiers (Figure 1a) and can be reached from several locations (Figure 1b) (Larbi Ben Mhidi Street, Ali Boumendjel Street, the Ibn Badiss Mosque, and the Algerian National Theater). The excavation of the southern access (at Larbi Ben Mhidi Street) of the station is carried out in two phases; the excavation of a well from the ground surface followed by the excavation of a 17% sloped tunnel branch that must lead to the metro station proper location (Figure 1c). This pro­ ject presents quite specific conditions: The site is

DOI: 10.1201/9780429321559-7

51

Figure 1. Situation of the southern access to Ali Boumendjel metro station. (a): Algiers old center (b): Ali Boumendjel metro station; (c): location of the study area.

difficult to access, characterized by narrow roads and very dense circulation of vehicles, old and vulner­ able buildings. The subsoil contains several under­ ground utility networks that are in the zone of influence of the metro station excavation. This is to say that any instability during excavation will have a disastrous impact on the safety and human activ­ ities in the area.

altered and highly fractured schist, with intercalations of clay levels with small scattered shale fragments; ­ and between 15m and 40m: a moderately fractured slightly altered schist bearing some circulations of water. The results of the geotechnical characterization of these three horizons are presented in Table 1. The results were obtained on the basis of labora­ tory tests and parameterized with RockLab software for schist (following the standards given in the refer­ ences column of Table 1).

2.1 Geological and geotechnical conditions As part of the metamorphic basement of the Algiers massif (Aymé 1963, Saadallah 1992), the geological unit concerned by the excavation of the southern Ali Boumedjel metro station access is represented by a light gray to dark gray schist formation (TEIXCO 2009). More specifically, the borehole logs allowed to distinguish three horizons (Figure 2) within the ground: - between 0 and 4.15 m: an anthropogenic embankment soil; - between 4.15 and 15 m: a highly

2.2 Initial design of excavation work and temporary support Given the estimated geotechnical conditions and the gallery situation, the upper part of the gallery is dug within the weathered schist, and the lower part is dug within the slightly weathered schist. The designer (CENOR 2011) has planned (Figure 3) to excavate the gallery in full section (46 m2). Consid­ ering the shallow depth of the tunnel (07 m),

Figure 2. Geological and geotechnical context of the southern access to Ali Boumendjel metro station.

52

Table 1.

Geotechnical characterization.

Geotechnical characterization

Units References

Embankment soil Weathered schist Unweathered schist

Wet Density Plasticity index Porosity Point Load Test Uniaxial Compressive Strength Standard Penetration Test Pressiometer limit Pressure Pressiometer Modulus Lugeon Test Rock Mass Rating Geological strength index Cohesion Friction angle Deformation modulus

% MPa MPa MPa MPa UL KPa ° MPa

1,9 15,6 6 1,6 40 5 30 15

EN 17892-2 (1991) EN 17892-12 (1991) NF P 94-410-3 (2001) XP P 94–429 (2002) NF P 94-420 (2000) NF P 94-116 (1991) NF P 94-110-1 (2000) NF P 94-110-1 (2000) NF P 94–131 (1994) Bieniawski (1989) Hoek et al. (1998) Hoek et al. (2002) Hoek et al. (2002) Hoek et al (2002)

2,5 40 - 60 10 - 25 >20 60 minutes on the ground. In gen­ eral, the zone in front of the TBM cutter face to be grouted covered an area of 15.5 m high and 12.8 m wide; the grouted zone around the TBM covered the 3 m thick area outside the TBM (Figure 6). The layout of grout holes is shown in Figure 7a. Grout holes were at 1.4 m centre-to-centre on a triangular grid. The amount of MK grout injec­ tion was 20% of ground volume for the zone in front and around the TBM. But the volume of grout injec­ tion was increased to 25% of ground volume for the grouting under the TBM. Since the grout hole layout

Figure 4. Worn-out condition of cutter bits on 300° spoke of cutter head (after Chang et al., 2016).

was dug out from the manhole on the cutter head for the visual inspection of the cutter head. The prelim­ inary visual examination had disclosed the seriously worn out situation of the cutter bits on the cutter head (Figure 4). A quick fix of the cutter head was not possible. So it was decided to carry out a thorough overhaul of the cutter head. A longer time for repairing was required. So the excavated space in front of the cutter head was backfilled with boulders and the rescue work moved on to the next stage. 4 TAM GROUTING PROGRAM Tube-a-Manchette (TAM) grouting method (Figure 5) was used for the waterproofing work around the TBMs. The grout holes were drilled with a duplex drilling system. The outer casing of the drilling system had an outside diameter of

Table 1. Mixing proportions and grouting parameters of the grouts used in this project. Cement-bentonite grout (CB grout) Cement (kg) Bentonite (kg) Water (litre) ∑ = (litre) Gel time (min) Injection rate (litre/min/port)

Figure 5. Schematic of TAM grouting method.

93

Silicate grout (MK grout)

150 - 250 Sodium silicate (litre) 40 – 80 MK reagent (kg) The rest Water (litre) 1000 ∑ = (litre) Gel time (grout itself)(min) 8 ~ 15 Injection rate (litre/min/port)

250 35 – 65 685 – 715 1000 60 8 ~ 12

Figure 6. Soil grouting zone for the in-place TBM repairing (Courtesy of SanShin Corporation).

and grouting plan for both TBMs are basically the same, this paper will only use the up-track TBM as the study subject here. The area under the TBM was unable to be directly grouted using the vertically drilled grout holes. So, 7 holes (black dots in Figure 7a) right next to the TBM was used for the under TBM grouting through lateral permeation method. The under-the-TBM grouting was carried out after the outer row grouting had been done. So the flow of MK grout could be restricted and dir­ ected to the area under the TBM. During TAM grouting, grout was injected from the bottom of M-tube first and then proceeded step­ by-step (0.33 m per step) upward. The grout volume control method was adopted here as the quality assurance measure for this TAM grouting work. While injecting the grout into the ground, the responding grouting pressure at each injection port was recorded until a designate amount of grout had been injected. The grout injection rate from the port was kept at 10±2 litre/min/port. Once the injection volume of a port reached the designated value (= 150 litre/port in front of TBM and 350 litre/port under the TBM), grouting was terminated and the grouting pressure response per port was recorded.

TAM grouting was usually conducted in a multipleport injection manner: 5 to 10 grout holes were injected as a group. 5 GROUTING PRESSURE DISTRIBUTION Since each injection port on the M-tube has its own spatial coordinates, the injection pressure gathered from each port can be used to establish the spatial distribution of grouting pressure response in the grouting zone (Liao et al. 2020). Basically, the grouting pressure response from the ground is dir­ ectly related to the effectiveness of TAM grouting. So the pressure response can be used as an index for checking the soundness of TAM grouting. Established with the GRAPHER software, Figure 10 shows the 2D grouting pressure distribu­ tion contours of the MK grouting at the depth of TBM centre and 1 m under/above the TBM. It can be seen that at the depth of TBM centre, the grout­ ing pressure on the right hand side of TBM (i.e., upside in the Figure 7) and in front of TBM was around 10 kg/cm2; while the grouting pressure on the left hand side of TBM (i.e., down-side in the

Figure 7. (a) Layout of grout holes (top view) for the main grouting; (b) drill holes from inside of the TBM for water leak test and supplementary grouting (Courtesy of SanShin Corporation).

94

the required value of 1x10-5 cm/sec. In fact, only a trace of water flow was detected from the water test holes on the cutter face and no groundwater leak was observed when the manhole on the cutter face was opened later. To further verify the effectiveness of grouting under the TBM, 10 holes (30 mm – ϕ) were drilled 10 – 20 cm into the grouted zone from inside the TBM for the water leak test (Figure 7b). The maximum water leak per hole was 3 ml/min; 8 holes (up-track TBM) and 4 holes (down-track TBM), respectively, had no water leaked out. The coefficient of permeability (k, cm/sec) of water leak tested soil (Figure 9) was cal­ culated from the following equation:

Figure 7) was around 18 kg/cm2. The grouting pres­ sure distribution at 1 m below the TBM showed a similar distribution pattern; but the grouting pres­ sure is higher than that at the centre of TBM. Higher grouting pressure could be the result of extra grout volume (350 litre/port) being injected; compared to 150 litre/port injected in front of the TBM. For the grouting pressure under the TBM, the area around the front end of TBM had a grouting pressure of > 10 kg/cm2; but a lower pressure response at the rear end of TBM where no confinement was provided for the grout flow. The grouting pressure of >10 kg/cm2 at the front end of TBM was considered to be an indication of good water tightness. In comparison, the grouting pres­ sure response 1 m above the TBM (Figure 10a) does not look as good as that below or in front of the TBM (Figure 10c). In addition, there is a transverse low pressure band (< 10 kg/cm2) 1 m above the TBM (GL -12.83~-13.83 m) going through the whole grouted zone. The water tight­ ness of this area was not certain. It needed to be checked by the in-situ permeability test. Two vertical holes were drilled to check the per­ meability of TAM grouted soil in front of the TBM using the falling head method. The locations of per­ meability test are marked in Figure 7a. One (T2) located from GL -18.2 to -19.2 m, was about at the depth of the TBM centre; the other (T1) located from GL -13.6 to -14.6 m was about 1 m above the TBM (Figure 8b). Equation 1 was used to calculate the coefficient of permeability (k):

where Q = discharge rate (ml/sec); H = water head at test location (cm); r0 ¼radius of drill hole (cm); L0 = length of drill hole (cm). Substituting the following water leak test data of one drill hole to Eq. 2 (L0 = 20 cm, r0 = 1.5 cm, H = 192 cm, Q = 3 ml/min), yields a k value of 5.26x10-6 cm/sec which is smaller than the criter­ ion (< 1x10-5 cm/sec). It showed that the water tight­ ness of the grouted zone met the design requirement. So only a small amount of quick setting silicate grout (gel time = 3 ~ 5 sec) was actually injected for the sup­ plementary grouting purpose. After the water tightness of grouting work had been checked by water leak tests, workers were sent out of TBM to dig out a space in front of the cutter face for cutter bits repairing and reinforcing the

where h1= water level in observation pipe at time t1; h2= water level at time t2; L= length of test hole; r = radius of test hole; R: inner radius of casing (Figure 8a). The calculated k values at T1 and T2 locations are equal to 1.34x10-6 cm/sec and 2.05x10-7 cm/sec respectively. Both were lower than

Figure 8. In-situ permeability test on grouted soil (a) fall­ ing head test; (b) locations of water test holes (Courtesy of SanShin Corporation).

Figure 9. Water leak test of grouted zone (Rapid jet Asso­ ciation, 2019).

95

Figure 10. Grouting pressure distribution contours (top view) at the depths of TBM center and 1 m above/below the up-track TBM.

Figure 11. Dug-out working space in front of TBM for the in-place repairing of cutter head and cutter bits (Courtesy of DORT, Taipei City Government).

Figure 12. Exposed grouted boulder layer and steel frame + timber lagging for the dug-out space support in front of TBM.

spokes on the cutter head (Figure 2). To minimize the exposed area and maintain the stability of the grouted boulder layer, only the upper half of the cutter head was dug out (Figure 11). In fact, by redu­ cing the height of the excavated space, it can not only increase the stability of the dug-out space but also reduce the thickness of grouting zone and save the grouting cost. By rotating the cutter head, the spokes and cutter bits below the open space could be brought up and repaired. Since most of the cutter bits were seriously worn out (Figure 4) and needed to be replaced and additional bits were added, the entire repairing process might take months. So, the dug-out space was protected by a frame of 100×100 steel H-sections and timber lagging to ensure the safety of workers (Figure 12). In the meantime, the structure of the cutter head was also reinforced by installing extra support to the spokes (Figure 2). After the cutter bits replacement and spoke reinforcement had been done, TBM was able to launch again and drove through the boulder layer without further problems. The TBM rescue project caused a programme delay of 210 days for the up-track tunnel and 171 days for the down-track tunnel respectively. But the rescue of the two stuck TBMs was successful.

6 CONCLUSIONS Usually, the design of TBMs is based on the ground conditions revealed from desk study and the site investigation work. However, it is practically not possible to obtain all the critical information needed for the design of TBM in the conglomerate boulder layer, such as the boulder size and drifted wood. But, the original cutter head design of the TBMs used in the Ding-Pu project was obviously unable to deal with the in-situ condition of con­ glomerate boulder layer. So the TBMs got stuck shortly after launching. To facilitate the in-place repairing work of the TBM, Tube-a-manchette (TAM) grouting was adopted to construct a waterproof zone in front and around the TBM, including the area below the TBM. By careful execution of TAM grouting and systematically moni­ toring and analyzing the grouting pressure response, a good watertight grouted zone was constructed (GL -11.6 m ~ -24.4 m) under high groundwater level. Workers were safely sent out of the TBM to dig out

96

a working space in front of TBM for the repair of worn out cutter bits and the reinforcement of cutter head. For the double protection of workers working outside of TBM, steel frame and timber lagging were used to support the dug-out space (Figure 12). After the repair and structure reinforcement, TBM was able to launch and drove through the boulder layer again. Both TBMs managed to finish the tunneling work with no further problems. The TBM rescue mission was successfully accomplished.

paper. Also, the comments made by Mr. Albert Ho on the contents of this paper are also appreciated.

REFERENCES Chang, H. K., Chou, C. R., Chang, R. F. & Hsieh, Y. H. 2016. Geotechnical Aspects of MRT Construction in Mixed Ground Conditions – The Dingpu Extension Pro­ ject, Magazine of Civil and Hydraulic Engineering, Vol. 43, 2, pp. 34–41. (in Chinese) Liao, Hung-Jiun., Cheng, Shih-Hao., Wong, Ricky K. N. & Shao-Jie Weng 2020. Base Grouting against Uplifting Water for a Deep Excavation in Silty Soil”, Proc. of 10th Int. Symposium on Geotechnical Aspects of Underground Construction in Soft Ground (ISCambridge 2020). Rapid Jet Association, 2019. Rapidjet method – Jet grout­ ing guidelines.

ACKNOWLEDGEMENTS The Authors would like to thank the Department of Rapid Transit (DORT), Taipei City Government and Taipei Branch, SanShin Corporation for providing valuable site information and grouting data of this

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Tunnelling in urban areas and pile interception challenges – a case study: Bank station upgrade project (BSCU) A. Nasekhian, C. Anthony & B. Haig Designer, Dr Sauer & Partners Ltd. London, UK

M. Dewhirst Client, TfL, London, UK

J. Ares Contractor, Dragados SA, London, UK

C. Barker Building Owner’s Engineering Representative, Arup, London, UK

ABSTRACT: This paper gives a detailed account of the approach taken at the Bank Station Capacity Upgrade project, where existing pile foundations required load transfer structures to be installed after fully intercepting 4 out of 25 end-bearing unreinforced concrete under-ream piles (shaft diameters range from 1.5 to 1.8m) and partially intercepting 7 others. The article addresses early stage pile interception options through to the detailed design. Elements presented include the stage 3 damage assessment of the building, which was based on the results of a complex three-dimensional finite element model considering the soil-structure inter­ action, and the extensive instrumentation and monitoring measures that were employed to verify the predic­ tions made by the 3D FE model.

In accordance with planning processes, the project sought to minimise the impact of tunnelling on the piles and building structure, while also retaining cap­ acity within piles for future development of the building. Thus, understanding the responses of all building piles was crucial during and after tunnel construction. This paper addresses design of the transfer structure required for the intercepted piles and damage assessment of the building from design stage to construction phase.

1 INTRODUCTION Frequently described as a rabbit warren beneath the City, the Bank- Monument complex has presented challenges for tunnellers for over a century. Bank Station Capacity Upgrade (BSCU) aims to match projected ridership, improve journey experience within the station and provide step free access to the Northern line and DLR. The project threads even more tunnels into the complex layers of existing assets: above, below and through many of the exist­ ing operational structures. As part of the upgrade a new platform for the Northern line has been constructed parallel with the existing platforms, necessitating a new section of running tunnel that connects into the existing align­ ment to the north and south of the station. Where possible underground obstructions were avoided; however, to optimise the track alignment, the new route inevitably deviates from the footprint of the streets above and, in one location, intercepts the existing end-bearing piles of Princes Court; a commercial building built in the early 1970s (Figure 1).

2 EXISTING BUILDING AND GROUND CONDITIONS The new southbound running tunnel passes under Princes Court, an 8-storey building with concrete frames and a 2 level basement on Princes Street, London. The Princes Court basement raft is a 1.2 metre thick slab and is unusually supported by 25 large diameter deep under-reamed piles (Figure 2). The building façade is glass and masonry. The alignment of the new tunnel could not avoid clashing with the building’s existing under-ream

DOI: 10.1201/9780429321559-13

98

Throughout the project, the building remained occupied with all the piles under load and designed to support between 6 and 10MN. While there are many cases of tunnels intercepting and cutting fric­ tion piles, as far as the authors are aware, this project is first time end-bearing piles have been intercepted and permanently supported around a tunnel in the UK and likely worldwide. The pile loads used to design the load transfer structures (Figure 3) are based on an increase on the historical design; these loads were agreed between London Underground (LU) and the building owners to allow for future redevelopment of the building. The current pile loads, calculated by Robert Bird Group (RBG) using STRAND7 (a 3D FE analysis model of the superstructure), were considered for the building damage assessment during pile interception and tunnelling. These were more appropriate than the historical estimation of building loads as there was a significant conservatism included at the time of their construction. The pile load to be transferred at the interception level was reduced by pile skin friction, applying a factor of 3.0 over the length of the pile left after interception, based on the ground strength profile (Cu) adopted for the BSCU project, see Table 1. The new running tunnel lies within London Clay (LC) underlying nearly eight metres of River Terrace Deposits and Made Ground. The track level of the new running tunnel is at +81.9mTD (Tunnel Datum). The groundwater table is assumed 8m below the ground surface. The base of the building’s basement slab is nearly at the interface of the River Terrace Deposits and London Clay strata. It should be noted that the stiffness of the London Clay assumed for design and analysis was based on back-analysis of ground movement. A ground stiffness of 200Cu for the London Clay aligned with the observed ground and adjacent underground structure move­ ments from tunnelling undertaken earlier on the project. The undrained shear strength and stiffness of the LC layers are increased linearly with

Figure 1. The Northern section of the BSCU scheme.

Figure 2. A perspective view of the building, and the new running tunnel through the end bearing piles.

piles without significantly increasing the length of the tunnels and impacts to further structures. The intercepted piles are illustrated in Figure 3. The clashes included full interceptions, where the whole under-ream and shaft would need to be removed, and partial interceptions, where - depending on asbuilt location - only part of the piles would need to be removed to allow the tunnel profile to be maintained.

Figure 3. Pile design loads agreed with building owner for future development.

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Table 1. Pile loads used for the transfer structure design and building damage assessment.

Pile No

Shaft Diameter m

Bell Diameter m

Current Load kN

Future Capacity kN

4 14 24 25

1.52 1.83 1.83 1.83

3.81 4.72 4.72 3.66

4000 4090 5525 5290

7038 10500 10500 6589

depth. More details associated with the ground conditions and geotechnical parameters of subsoil layers in the region of the BSCU scheme can be found in Nasekhian & Spyridis (2017). 3 INITIAL OPTIONS AND RISK ASSESSMENT DURING DESIGN DEVELOPMENT A comprehensive design process was undertaken to gain acceptance from the building owner for the pro­ posed works. The LU-Dragados team worked on refining the solution from when they started the con­ cept design stage post tender award in August 2013 until the detailed damage assessment and design of the final selected option was submitted in April 2018. Initially, an active load transfer structure had been considered. There was limited information available on the building, so a conservative approach was ref­ erenced, in case it should be required. This included the construction of handworks tunnels a few metres above the running tunnel structure to provide space for a jacking system (Figure 4). During the Concept Design phase (equivalent to RIBA 3 see LU S1050) a review process was estab­ lished to choose the most appropriate solution for pile interceptions. • The first stage was to determine all solutions that were considered feasible. • The second stage was to work through a weighted comparison matrix for each case and to identify practical and acceptable options to meet the requirements. • The third stage was to score the options against a set of criteria including health and safety, CDM, buildability, noise and vibration and conformance with requirements, including limiting damage to the building and maintaining separation of LU and private assets. Options considered included intercepting and cut­ ting the existing piles while strengthening the build­ ing, but this option would be intrusive for the occupants. Other options minimised the profile of the tunnels to minimise ground movement and inter­ ception, but these did not provide the building owner

Figure 4. Two options considered in the conceptual design stage.

with the capacity required for future development of the site. The jacking system considered during tender did not minimise the risk to the workforce to as low as reasonably practicable (ALARP), neither during excavation and installation nor during the life and maintenance of the structure. In line with risk assessments carried out for the whole of the project it was concluded that preference should be given to designs that allowed mechanical excavation and sup­ port methods to be used during construction. The decision as to what design solution could be taken forward was linked to the assessment of the impact of various cutting and supporting sequences that could be employed. The damage assessment car­ ried out at this phase suggested that cutting two piles in a row prior to provision of permanent support was a viable option. It was accepted that immediately after cutting a pile there is no load, but that load may be later applied through changes to the distribution of loads within the building through ground move­ ment or changes to the loading (such as continuing the tunnel to cut through further piles).

100

The chosen option for detailed design was there­ fore to support the intercepted piles around the tunnel lining without active transfer of load through jacking or dowelling. As the cut intercepted piles would be temporarily unsupported before construc­ tion of the load transfer structure, further analysis of this solution using a coupled 3-dimensional struc­ tural and geotechnical FE models was required to demonstrate to the building owner and their tech­ nical advisor Arup, the details of this option. Figure 5. Load path for pile load transfer structure.

4 DETAILED DESIGN An initial step during the detailed design stage was to optimise the proposed track alignment (with con­ sideration of train speed and track maintenance), which saved one pile from being fully intercepted This left 4 full interceptions and a need to support piles 4 and 14 (on a combined transfer structure), and 24 and 25 (on individual structures). The profile of the tunnel was reduced back to a standard profile as quickly as practicable between transfer structures to minimise interceptions and ground movements. However, a further 3 piles (3, 15 and 23) were iden­ tified as potentially being partially intercepted and a requirement for load transfer would depend on the as-built locations and amount of interception. As usual in the design process, the geometry started with the most statically efficient shape (e.g. near to circular) but in this case there was a limitation on the depth that meant that this shape was not workable. A circular profile was likely to undermine surrounding piles, reducing their bearing capacity and thus, a closed horse-shoe shaped cross section was developed. 4.1

Load transfer mechanism

The structural mechanism utilised can be explained as follows: the bottom of the cut pile is socketed into an arched reinforced concrete structure around the new running tunnel, (passive approach). The structure bridges over the new running tunnel so that when load is applied in the future it will be delivered to a reinforced concrete invert. Those loads are then transferred to the ground in bearing. Figure 5 shows the final composition of the lin­ ings for the transfer structure. The transfer structure, a bar reinforced cast-in-place concrete arch with a series of straight edges on the inner surface to aid construction. Between the transfer structure and the running tunnel lining a compressible separation material was installed to allow for independent movement of the running tunnel and building struc­ ture, so potential future demolition unloading and subsequent reloading of the piles would not impact the tunnel. A waterproof membrane is applied at the intrados of the concrete arch to ensure that the run­ ning tunnel is water-tight. The running tunnel

secondary lining is designed to take the hydrostatic pressure and any required fixing loads. It is noted that a low vibration, high attenuation Sonneville LVT HA track form was included in the design to mitigate any noise and vibration generated by the relationship between the tunnel and the build­ ing foundation. 4.2

Finite element modelling

In the design of the structures, both closed form analyt­ ical solution and finite element analyses were employed. First, the initial dimensioning of the cross section was designed in accordance with Eurocode 2 using simple arch beam formulae. After establishing the geometry by ensuring that the arched structure can bear the loads of the pile as a footing, the final geom­ etry was numerically modelled. To simplify construc­ tion it was decided to use a single cross section for all intercepted piles, varying the length of the structure based on each pile load. The transfer structure of each intercepted pile was modelled using 3D FE analysis to evaluate the struc­ tural behaviour, understand realistic load transfer mechanism from pile to the structure and the under­ lying soil layer, and soil-structure interaction (Figure 6). Also, the impact of uncertainty of the pile loca­ tion, due to the construction tolerances, on the reinforced concrete design was investigated. The envelope of steel reinforcement was further enhanced by assessing the stress distribution in the volume elements representing the transfer structure in various pile locations. The transfer structure was designed so that minimal alteration was required on site. Slight tweaks were applied by the site supervi­ sion team for bars that interfaced with the actual pile cut line. Ease of rebar installation was considered at the design stage in collaboration with the Dragados construction team. Various options, for instance bringing in prefabricated reinforcement cages, were considered but not chosen after reviewing the risks to health and safety on site. The modelling assumptions and consider­ ations are: 1. A slice of the ground aligned with the length of the transfer structure has been modelled. For

101

Figure 6. 3D FE model of the transfer structure in ABAQUS software; a) mesh, b) extremes of pile location assuming given pile tolerances.

instance the ground model for a 5m long transfer structure is 5m thick. 2. To consider a realistic soil structure interaction between the transfer structure and the surrounding soil, contact elements have been employed. The contact element at the interface of the soil and transfer structure has been set up to consider com­ pressive interaction and allows no tension to be transferred, i.e. the associated elements will be detached in the event of tension at the interface (See Figure 6- a). 3. Pile load is applied on a squared surface as a pressure in the middle of the transfer struc­ ture where the pile is cut. The area of the patch is equivalent to the pile shaft diameter. For each pile interception, two extremes of the pile location are analysed to allow for the uncertainty of the pile locations due to the construction tolerances (See Figure 6- b). The horizontal tolerance allowance for piles

included in the design is 525mm relative to the best estimate location of the piles. 4. The ground is modelled using the Mohr-coulomb material model using a total stress analysis with undrained. The concrete material model is elastic. 5. A 40% live load and 60% dead load with partial factors applied, as appropriate, was considered. Furthermore, this assumption was later reviewed during the building damage assessment which revealed it is closer to 20% live load and 80% dead load, hence the loads used in the structural design are more conservative. Figure 7 illustrates the maximum compression stress developed in the structure for best estimate pile location. 5 BUILDING DAMAGE ASSESSMENT The building damage assessment was undertaken in accordance with LU Civil Engineering – Common Requirements S1050 (2013) which requires a staged based assessment. In Stage 2 the “Greenfield” ground movement was analysed in terms of settle­ ment and horizontal displacement assuming a volume loss of 1.5% for SCL tunnelling. Subsur­ face tunnelling induced ground movement profiles were determined in accordance with the method­ ology described by Mair et. al. (1993 &1996). The Stage 2 assessment concluded that: 1. The maximum settlement of the building at foun­ dation level at the end of construction would be 31mm (Figure 8).

Figure 7. Principle compression stress in concrete mass for piles P4, P14 &P24 and P25. Units in N, negative values repre­ sent compression.

102

Figure 9. Soil-Structure interaction model.

Figure 8. Stage 2 damage assessment results based on greenfield ground movement.

2. The maximum tensile strain within the build­ ing based on these movements falls within damage Category 2 based on the Burland damage category definition (Burland, 1995). It is noted that this categorisation is not strictly applicable to a non-masonry structure such as this building. 3. Damage Category 2 (DC2) corresponds to “Slight” damage and no mitigation measures were identified as required, however, given the complexity of the foundations it was recom­ mended that a Stage 3 assessment validate the assumptions at the detailed design stage. Part of the initial Stage 3 damage assessment was to determine an appropriate construction sequence that would not result in damage to the building that exceeded the agreed parameter; DC2. As mentioned previously initial calculations, using a combination of methods including empirical analysis and simplified FE modelling, established the principal of cutting the piles; redistributing the current building loads amongst the rest of the foundation system and then supporting them to allow for future loading. It was proposed that the fully-intercepted piles could be addressed in a sequence wherein the first two piles (4 and 14) would be cut before installation of a load transfer structure, then the sequence would be repeated for the second two fully intercepted piles (24 and 25). To further demonstrate this, a comprehensive soilstructure interaction model was undertaken using three dimensional Finite Element (FE) modelling (Figure 9). Due to the extent of the problem and the level of detail to be included in the analysis it was

impractical to model all the ground and the structure in one FE model. Therefore, two separate FE models using appropriate software have been developed. The ABAQUS ground model developed by Dr Sauer & Partners (DSP) includes the pile-raft foundation, building basement and all tunnel excavation sequences. It was coupled with a STRAND7 super­ structure model created by Robert Bird Group (RBG) to assess the building response due to the tunnelling induced settlements given by the substructure model. A full soil-structure interaction was assumed in this assessment, wherein two separate models for the super- and sub-structures equilibrate load and dis­ placement across a common interface. The modelling assumptions and features in the sub-structure model are: • The sub-structure model developed is comprised of approximately 1.5 million tetrahedral volume elements for soil/piles concrete and 14,500 shell elements for slabs, the SCL tunnels and transfer structures with averaged thickness. • All construction sequences of the SCL tunnelling have been simulated (For instance Figure 10 depicts extraction from the model related with Pile 4 and 14). • The ground model is based on Mohr-Coulomb model with total stress analysis and undrained parameters of the London Clay. Stress/strain dependency of the ground stiffness have been considered in the analysis by means of increasing the stiffness with depth and based on the appro­ priate stiffness associated with the expected strain around the excavation zone. • No information about the pore water pressure is obtained during the analysis. Pile interception and tunnelling underneath the building leads to excess pore water pressure during the construction phase. Over a long period of time the water pres­ sure will be dissipated which will result in some additional movement. • Stress history in the model due to the building current load and pile installation was established prior to the tunnelling stages. Contact elements were employed to model the interface between the piles/foundation slab and the ground.

103

existing piles, including the intercepted ones, will be capable of supporting the future loads. The super-structure analysis concluded that the proposed construction sequence should not cause damage beyond DC2. Negligible damage to the façade or internal walls of the building was expected. Very slight changes were observed between the ini­ tial and final internal forces in the building structure including floor slabs, beams and columns. In accord­ ance with these findings no explicit mitigation meas­ ures were proposed for the building. Figure 10. Construction sequence for Piles 4 and 14.

6 CONSTRUCTION • Ground deformations and pile movements have been extracted at various construction stages, selected to align with the tunnelling programme, and to allow ground movements for the building foundation to be assessed at different stages of the excavation. • Sensitivity analysis was undertaken to ensure a robust assessment that includes the impact of ground stiffness, pile stiffness and creep effect, modelling non-linear and plasticity behaviour of unreinforced concrete of the existing piles (using Concrete Damage Plasticity model available in ABAQUS) and finally variation in the building loads. The following conclusions could be drawn in regards to the sub-structure modelling results: • The maximum ground movement at the founda­ tion level was predicted in the North East corner of the building approximately 10 mm. This was lower than the Stage 2 damage assessment but this was to be expected as the building stiffness, soil-structure interaction and site specific param­ eters had been introduced. The study agreed with the conclusion made in the previous stages that tunnelling through the pile foundation can be undertaken without impacting the building beyond a level described as damage Category 2 by Burland. • It was found that the partially intercepted piles have negligible impact on the building slab deformation (practically zero), as long as the piles are found not to exceed the assumed max­ imum horizontal construction tolerances of 525mm from their assumed locations. • The pile/raft foundation of the building is stable during tunnel construction and after pile interception. • The induced tensile strains/stresses due to pile interception are reversed to compressive strains/ stresses in the majority of the intercepted piles when the future development loads are applied. The vertical compressive stresses do not exceed the allowable concrete stress limit and the

Tunnelling was on the project’s critical path and it was paramount to gain acceptance from the building owner when considering tunnelling resources and potential delays to other work. Using the modelling described above, the construction sequence and the method of the pile load transfer was presented to the building owner and their engineering representatives, Arup to allow tunnelling to proceed under the build­ ing in July 2018. Acceptance was granted and the work was completed on time by the end of November 2018. The transfer structures were built in just under five months with over 600m3 of concrete and 35 tons of steel. Through a combination of careful excava­ tion, detailed monitoring and full comprehension of the design by all involved in the construction, the transfer structures reached successful completion with minimal impact on the building above. One challenge from a construction perspective was working out how best to cut through the piles. After considering various methods, and in consult­ ation with the building’s owner, the site team chose stitch drilling. To access the pile for drilling, the excavation was carried out as per design with the pile exposed in the tunnel face Figure 11a. As-built surveys were taken of the exposed pile to validate the design. The existing piles locations were within the expected tolerances. Stitch drilling was carried out from an access platform, which resulted in the structural separation of the top and the base of the pile. The shaft was then removed by mechanical breaking. The reinforcement was installed closely around and supporting the base of the pile shaft. Once the cages were installed concrete was poured into bespoke shutters built by the construc­ tion team. A compressible material was installed to allow movement of the transfer structure. The thick­ ness and composition of the material was such that it allowed movement matching that predicted during the detailed design. A sheet waterproofing mem­ brane was installed to complete the structural separ­ ation. The LU tunnel lining was installed within this membrane providing the envelope for the running tunnel.

104

construction, local adjustments to the tunnel lining and the as-built geometry meant that no partial removal of the pile shafts was necessary. As a result, partial pile interception load transfer structures were not required. However, a separation material was installed to allow the foundation (transfer structure) and tunnel lining to behave independently. 7 MONITORING Monitoring considers the use of instruments to record movements, stresses or strains. It was decided to monitor to allow i) the assessment results (and inher­ ent assumptions) to be validated, ii) decisions to be made with regard to the construction process through trigger levels assigned to the monitoring points. An extensive monitoring scheme was applied including: • 19 no. levelling sockets installed in the basement to monitor pile movements at slab level to be read manually. • 14 no. levelling sockets installed at low level around the entire perimeter that measured manu­ ally. These sockets used to monitor differential settlement across party walls. • 21 no. ground surface levelling points installed around the entire perimeter. • Total of 25 no. 3D prisms installed on the façade (10no. read automatically & 15 no. manually) at low and high level. These were used to monitor differential movement across party walls. • Fibre optic implemented at the centre of two piles (Pile 23 partially intercepted at the underream only, and the fully intercepted Pile 14)

Figure 11. Photos from the construction of the transfer structures.

Several piles had the potential to be partially inter­ cepted. A contingency measure had been designed in case the encountered shaft of ones of those piles was out of assumed tolerance and encroached to the sec­ ondary lining space of the running tunnel. During

Presenting all monitoring results is beyond the scope of this paper. Fibre optic pile monitoring is presented in Barker et al (2020). The results of levelling sockets at foundation slab level, targets on pile’s shaft above the cut level are discussed herein. As soon as the pile shaft was exposed two monitor­ ing targets were installed on the pile shafts above the cut level. It took nearly a couple of working shifts to stitch drill the pile. Monitoring of the piles prior to cutting and after removal of the piles showed rela­ tively consistent 3 to 4 mm vertical settlement (i.e. immediate settlement) at the cut level. The observed movements are in a very good agreement with those predicted by FE analysis which showed a range of 2.5 to 3.5mm. The sockets on the foundation slab right above the intercepted piles moved slightly (i.e. 1 to 2mm) less. The final vertical settlements (at the end of tunnelling) of the piles at the cut level for piles 4, 14, 24 and 25 are 9, 8, 12 and 12 mm, respectively. The predicted values from the numerical modelling are half for P4, P14 and P24, but for P25 nearly 12mm.

105

have extended the tunnel programme and not reduced risks to ALARP. In summary, the project took an approach in the delivery phase that satisfied all party interests while providing an optimised track alignment, which result in whole life benefits including reduced operating costs and noise pollution as well as improved maintainability and safety. The method used at Bank and the results obtained will become increasingly valuable for geotechnical engineering, especially as subsurface infrastructure expands in increasingly congested urban ground. It will also allow building owners to have more confidence with redevelopment in the future. The building owner requested to maintain the pile load capacity after interception (N.B. even higher than originally designed for future development including demolition and building a new structure). If this had not been the case a viable solution would have been to remove the base of the piles and separ­ ate them (e.g. placing flexible material) above the tunnel lining. Once demonstrated that the building performed satisfactorily without the end bearing cap­ acity of the cut piles this would have been the sim­ plest and the safest solution and obviously, the smaller the tunnel the less impact on the building from settlement. However, engineering solutions need to balance multiple requirements, and while a simple solution might meet some, all must be met.

Figure 12. End of tunneling stage, FEA results (contour) vs levelling sockets (individual values) at top of foundation slab; Unit mm, recorded in June 2019.

Figure 12 compares the FE modelling predictions (contour plot) versus the levelling sockets results on top of the foundation slab at the end of tunnelling underneath the building and beyond. The observa­ tion represents the building movement 9 months after intercepting the last pile. The magnitude of the measured settlement on the building foundation slab showed a similar range with respect to the predictions given in the damage assessment. However the pattern of the movement shows that the settlement trough under the building is wider than expected and the build­ ing foundation has mostly moved 1 to 4.5mm more than predictions except the pile 25 which moved 3 to 4mm less. Since the actual settlement trough is wider (i.e. differential settlement is less than prediction) the overall building damage assessment outcome remained valid. As it can be seen in Figure 2, there are two slabs in the southern and northern of the building that have elevation higher than the main slab. The northern slab is supported by 3 pedestals sit­ ting on piles 7, 16 and 25. The archive drawings are not clear about the connection detail at the perimeter of this slab to the adjacent slabs or party walls. Thus, the worst case assumptions applied in the modelling to address the concerns raised over the pile 25 interception during the damage assessment. The reason for over predic­ tion around pile 25 could mainly be attributed to this uncertainty. In reality the slab over pile 25 moved 6mm, while 10mm was expected. 8

CONCLUSIONS

As a result of the complex finite element model and monitoring, the Bank Station Capacity Upgrade project was able to proceed with its preferred approach of cut­ ting the piles prior to permanently supporting them around the tunnel profile confident in the predicted impact this would have on the structure and the build­ ing owner’s future use of the assets. This allowed the project to avoid difficult and potentially dangerous alternative options including pile jacking, which would

ACKNOWLEDGEMENT The authors would like to thank all individuals from the companies & organisations that contributed to the development and construction of this highly challenging and successful pile interception includ­ ing: Transport for London (TfL), The Worshipful Company of Grocers, Dragados, Dr Sauer & Part­ ners (DSP), Robert Bird Group (RBG), Ove Arup & Partners (Arup), Geocisa, Cambridge Centre for Smart Infrastructure and Construction (CSIC) and Geotechnical Consulting Group (GCG).

REFERENCES Barker C.A., Kechavarzi C., Xu X., Nasekhian A., (2020), Fibre Optic Monitoring of Existing Under Ream Piles for Tunnelling Interception, British Geotechnical Asso­ ciation Piling 2020 Conference, Durham, 15-16 Septem­ ber 2020. Burland J. B. (1995), Assessment of risk of damage to buildings due to tunnelling and excavation. Proceedings: 1st International Conference of Earthquake Geotech­ nical Engineering, IS Tokyo. London Underground CAT1 Standard (2013), S1050, Civil Engineering - Common Requirements; Issue No. A7,. Mair R. J., Taylor R. N. & Bracegirdle A. (1993), Subsur­ face settlement profiles above tunnels in clays. Géotech­ nique 43, No. 2, pp. 315–320.

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Mair R.J., Taylor R. & Burland J. (1996), Prediction of ground movements and assessment of risk of building damage due to bored tunnelling, Conference on Geotechnical Aspects of Underground Construction in Soft Ground, London, pp 713–718.

Nasekhian, A. & Spyridis, P. (2017), Finite Element modelling for the London Underground Bank Station Capacity Upgrade SCL design and deep tube tunnels assessment. Proc. of IV Int. Conf. on Computational Methods in Tunnelling and Subsurface Eng. (EURO:TUN).

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Impacts of new development on existing underground assets using greenfield model K. Neaupane & Y. He Atkins SNC-Lavalin, Ground Engineering, UK

ABSTRACT: This paper presents two case studies involving new development and their impacts on existing underground sewers and a live tunnel. To assess the effects of the construction of proposed new structures on the existing underground assets, two- and three-dimensional Finite Element (FE) greenfield models were developed in Plaxis 2D and 3D. Radius of curvature and bending strains were estimated from diameter and curvature of the pipe using the method proposed by Bracegirdle et al. (1996). The analyses undertaken capture the movements of the ground as a direct response to proposed constructions and provides a basis for an assessment of the likelihood of damage to the underground assets.

1 INTRODUCTION 1.1 Background When planning new development, it is important to control and minimise the risk of damage to existing underground assets. During the planning stage, the information available on the current condition of the existing underground assets and jointing properties is often limited. In addition, it adds complications to the numerical modelling process if the underground assets are made of composite materials such as masonry. This paper presents two case studies involving application of a relatively simple design approach to rapidly assess the potential damage to existing con­ crete and masonry buried structures due to proposed constructions. 2 CASE STUDIES 2.1 Case A: Sewers under bridge piers using Plaxis 3D 2.1.1 Project overview Two existing branches of a trunk sewer pass adjacent to a pier of a proposed two-span bridge. The bridge pier is supported by piled foundations. The piles are connected via a reinforced concrete pile cap and the exterior surfaces of the piles are at a minimum dis­ tance of 1750 mm from the sewer branches. The pier is supported by 8 No. 1500m diameter bored piles at 3750 mm c/c spacing. Details of the sewer lines are given below.

– Masonry branch: Approximately 1372 mm internal diameter with 225 mm wall thickness, passing at a minimum clear horizontal distance of 1750 mm from the surface of the piles. – Concrete branch: Approximately 1372 mm internal diameter with a 114 mm wall thickness, passing at a minimum clear distance of approxi­ mately 6.5 m from the outer piles. The cross section of the sewers and pile founda­ tion is given in Figure 1. 2.1.2 Ground conditions Based on available ground investigation, the sequence of strata encountered at the site comprises engineered embankment fill, old fill over the wea­ thered and un-weathered London clay over Lambeth Group. The existing ground level within the model boundary varies between 41.50 m and 50 m AOD (above ordnance datum). The plastic index of the London Clay from the site correlates to a range of effective friction angle φ' values which generally agree with the φ' values for the London Clay. A correlation of E'/cu = 700 has been assumed to determine the soil stiffness param­ eters. The hydrostatic water pressure profile with depth corresponds to a groundwater pressure head at the existing ground level. It is expected the site would maintain dry during construction. Hence, the groundwater table is taken to be at the finished ground level for the numerical analysis. 2.1.3 Constitutive model and input properties To assess the effects of the construction of the pro­ posed structure above the two sewers, a finite

DOI: 10.1201/9780429321559-14

108

Table 1.

Figure 1. Cross sectional view of central pier and pipes.

Design ground profile and input soil parameters. 0

γ

c'

Stratum

mAOD

kN/m3

kPa °

Embankment Fill Old Fill Weathered London Clay* London Clay*

49.2 – 47.9

19.0

0

35

0.43

41 33.50

18.5 19.5

0 5

30 23

0.50 1.00

27

20.0

5

23

1.00

υ

cu kPa

Stratum

element analysis was carried out using Plaxis 3D (Brinkgreve et al., 2015). A hardening soil model (HS model) has been used to represent the soil constitutive behaviour. In the HS model, soil stiffness is described more accurately by using three different input stiffnesses: the triaxial stiff­ ness E50, the triaxial unloading stiffness (Eur) and the oedometer loading stiffness (Eoed). The HS model also accounts for stress-dependency of stiffness moduli. This means that all stiffnesses increase with pressure. Hence, all three input stiffnesses relate to a reference stress. Besides the model parameters mentioned above, initial soil condition, such as pre-consolidation pres­ sure, plays an essential role in most soil deformation problems and can be taken into account in the initial stress generation. A summary of the ground model parameters adopted in the analysis is given in Table 1. The degree of stress dependency in the Hardening Soil model is controlled by the power function parameter m. This parameter is often derived by curve fitting against measured values. Figure 2 illus­ trates the comparison between Mohr-Coulomb (EMC =E50) and Hardening Soil stiffness (EHS) from which reference pressure (Pref) and constant (m) are taken for each soil type.

’

Top level

Embankment Fill Old Fill Weathered London Clay* London Clay*

K0

Pref kPa

E50ref Eurref MPa MPa

m

10

30

90

0.01 0.3 60

10 263

30 65

90 195

0.01 0.3 45 0.90 0.2 90+4z

337

90

270

0.90 0.2 120 +4z

* z (depth in meter) is measured from the top of the stratum.

2.1.4 Key assumptions and numerical model The following key assumptions have been made for the analysis: – Piling induced displacements is minimal as bored piling techniques will be utilised. Hence it has been omitted in the modelling exercise. – The undrained condition was used to simulate short term behaviour at the construction stages. Excess pore-water pressure was allowed to dissi­ pate during the consolidation stage to simulate the long-term behaviour. – The piles were taken to have rigid connections with the pile cap (i.e. fixed pile heads). – Loads and moments were applied at each pier base. – No partial factors have been applied to the ana­ lysis as this is a serviceability check.

Figure 2. Comparison between Mohr-Coulomb (EMC) and Hardening Soil (EHS) Stiffnesses.

The model contains 8 No. 30m long concrete bored piles connected at the top via a rectangular pile cap and a brick and concrete sewer at a horizontal clear

109

Table 2.

Applied loadings in the analysis*. Forces (kN)

Column Column 1 (Inside the pile group) Column 2 (Outside)

Moment (kNm)

Fx

Fy

Fz

Mx

My

614

-457 8003

716

Mz

-4121

-6530 40

-405 91242 -4363

-6328 21

* No partial factor was applied

Figure 3. Structures and applied loadings. Table 3.

distance of approx. 1.75m and 6.5m respectively from the exterior surface of the piles. The horizontal extent of the model is 180 m × 140 m. The vertical boundary is taken as -10 m AOD. The piles are modelled using embedded beam elements and the pile cap is modelled using plate elements (see Figure 3). The skin friction is assumed to increase linearly with depth (Tskin start =100 kN/m and Tskin end = 180 kN/m). The base resistance of each pile in the London Clay was taken as 1100kN. To understand the maximum possible strains that may develop at the sewer location in the absence of the sewer, a ‘greenfield’ model was developed. This model aims to give the worst-case movement and assumes that the sewer offers no resistance/support to the soils at that location. The current construction effects are considered in the analysis, neglecting the effects of historical load­ ing. Based on above understanding, the analysis was carried out in stages as given below:

Thames Water suggested assessment criteria.

Allowable strain, ε

Max. Rotation, θ

Pipe Type

Dia. (mm) Δεmax ðμεÞ

Dia.(mm) θmax ðo Þ

Brick

N/A

N/A

N/A

Concrete (unreinforced)*

>750

>1400

0.3

500 (Tensile) 25% of allow­ able Stress (Compressive) 60 (Tensile) 400 (Com­ presssive)

* The above concrete sewer limits are applicable to unre­ inforced concrete pipes. It is understood that reinforced concrete pipes were installed via pipe-jacking. Hence, a conservative assumption is made regarding the con­ crete sewer limits.

The maximum phase displacement results at the sewer locations close to the foundation from the model are summarised in Table 4. The signs are with respect to the global axis directions. The x- and y-axes lie on a horizontal plane, x- being sewer axis and y- perpendicular to the sewer. The z-axis is ver­ tical. The maximum displacements are recorded in the z-direction, a positive sign indicates heaving and a negative sign indicates settlement. Details on the greenfield vertical movements along the centreline of the sewers are presented in Figures

– Initial phase – Establish existing condition – Consolidation and reset displacement – Excavation to proposed formation level of 39.50m AOD (heaving effect due to the excavation influ­ ences both brick and concrete sewers) – Install foundation and apply bridge loads – Long-term settlement For the purpose of this study, the most critical combination of three-dimensional forces and moments was considered. Given that the bridge is skewed, the loadings acting on the two pier columns are different. A summary of the loadings applied to the two columns is given in Table 2.

Table 4. Phase displacement results at sewer locations from the greenfield analysis.

Location Stages

Ux

Uy

Uz

2.1.5 Displacement results and discussion The results of the analyses were compared with the suggested assessment criteria for existing Thames Water pipeline and assets given in the Thames Water document ‘Guidance on piling, heavy loads, excava­ tions, tunnelling and dewatering’. The suggested assessment criteria relevant to the assets being assessed are given in Table 3.

Brick sewer

2.2 2.4 0.6

4.7 4.8 1.1

6.2 4.4 23.5

1.7 1.9 0.4

4.3 4.4 0.7

6.4 5.4 26.4

Max. phase disp. at sewer locations near bridge foundation (mm)

Concrete sewer

110

Excavate Apply bridge load Long-term consolidation Excavate Apply bridge load Long-term consolidation

longitudinal strain reduction factor was not applicable; hence a value of 1 was used. Combined strains were calculated as the sum of the bending and longitudinal effects. It is envisaged that the joints in the concrete sewer will accommodate some rotation, and the induced tensile axial strain may be compensated for by the joint opening. It has been conservatively assumed that the concrete pipe will follow the move­ ment profile of the ground and that the rotation (α) of the joints may be estimated using Equation 2 below:

Figure 4. Computed vertical displacement at the brick sewer location.

where k is the curvature of the deformed soil pro­ file and L is the pipe segment length. For the purpose of this analysis, length of concrete segment was assumed to be 2.5m as the true length of the seg­ ments is unknown. This approach takes no account of the actual location of joints; it is assumed that joints may be located at the most severe position, and that the joints are free to rotate. Where joint rotation is combined with tensile strain, the joints will open. This combined effect has also been con­ sidered to ensure that the joints are likely to remain water-tight. Computed profiles of tensile and compressive strain in the sewers, and rotation of the joints are presented in Figures 6 to 9. Peak values are summar­ ised in Table 5.

Figure 5. Computed vertical displacement at the concrete sewer location.

4 and 5, respectively. The maximum heave due to unloading at the excavation stage slightly reduces due to the foundation load in the following stage.

2.2 Case B: A concrete lined tunnel under new building development using Plaxis 2D

2.1.6 Effect of ground movements The ‘greenfield’ ground movements computed along the sewer axis in the finite element analyses have been assessed using spreadsheets to calculate poten­ tial strains in the brick and concrete sewer and rota­ tion of the joints for the concrete sewer. The general approach taken follows the semi-empirical method­ ology described by Bracegirdle et al. (1996). For the calculation of bending strain in the brick and concrete pipes, it is assumed that the pipe follows the movement profile of the ground, and the bending strain was estimated using Equation 1 below:

where D is the outside diameter of a pipe, and ρ is the curvature. For the calculation of longitudinal strains in the pipe, the strains in the soil were modified by a strain reduction factor (RF). This reduction factor is to allow for potential differential movement between the pipe and the soil. For concrete pipes, a strain reduction factor (RF) of 0.2 has been taken based on discussion by Bracegirdle et al. (1996). For brick work, the

2.2.1 Project overview A multi-storey residential building founded on pile group was being constructed directly on top of an operating underground line. The piles are connected via a reinforced concrete slab among which, two rows of piles are located about 3m away from both side of the tunnel. A crosssectional view of the scheme is provided in Figure 10.

Figure 6. Computed axial strain in the brick sewer.

111

Figure 7. Computed flexural strain in the brick sewer. Figure 10. Section of proposed building foundation.3.

2.2.2 Ground conditions The ground profile in the vicinity of the existing tunnel and the proposed building location predomin­ ately comprises Fill, Alluvium, River Terrace Gravel (RTG), London Clay, Lambeth Group (mainly granular) and Thanet Sand. The pile group is founded in the Lambeth Group. The existing ground level is taken as +5.80 m AOD. In the preliminary settlement calculation, a correlation of Eu/cu = 1000 was assumed to deter­ mine the soil parameters for London Clay and Lam­ beth Group. However, back analysis based on the monitoring data showed that the initial assumption was too optimistic for vertically loaded pile groups and Eu/cu = 700 was founded to be more appropriate. The hardening soil model, which accounts for stressdependency of stiffness modulus was used for the simulation of soil behaviour in the current analysis. A summary of the ground model and soil parameters adopted in the analysis is given in Table 6.

Figure 8. Computed axial strain in the concrete sewer.

Table 6.

Ground profile and input soil parameters. c' kPa

φ' °

40 60+12z 225 -

0 1 0 5 5 0

20 20 35 20 27 36

E50 MPa

Eur MPa

m

ν

10 9.6 30 45 110 250

30 28.8 90 135 330 750

0.01 0.01 0.01 0.9 0.01 0.01

0.3 0.2 0.3 0.2 0.2 0.3

Stratum

Thickness m

γ cu kN/m3 kPa

Figure 9. Computed flexural strain in the concrete sewer.

Fill Alluvium RTG London Clay* Lambeth Group Thanet Sand

4.5 6.25 3.5 11 16.5 Not proven

18 17 20 20 20 21

Table 5. joints.

Stratum

Pref kPa

Fill Alluvium RTG London Clay Lambeth Group Thanet Sand

110 110 110 175 364 524

Computed peak strains in sewers and rotation of

Results

Concrete sewer

Brick sewer

Peak compressive strain (με) Peak tensile strain (με) Peak rotation of joint (°) Peak joint opening (mm)

35.8 47.0 0.005 0.183

265 393 -

* The depth z is measured from the top of the stratum.

112

Water strikes recorded groundwater at variable depths within the Fill and Alluvium. Groundwater level used in the analysis is selected as that is thought to represent the most unfavourable conditions which could realistically occur during the design lifetime of the proposed building, which is at the base slab level (+3.60 m AOD). 2.2.3 Structural modelling The existing circular tunnel has segmental concrete lin­ ings and the lining thickness is 250mm. The outer diam­ eter of the tunnel is 4.9m. Based on a segment number of 5 and one key segment, the second moment of inertia is reduced according to the formula as in: Ie=Ij+I(4/n2), where I and Ij are the moment of inertia of the intact lining and segmental joint and n is the number of equal segments in the lining (Wood, 1975). A reduction factor of 0.65 was applied to the longitudinal bending stiffness (EI) of the tunnel to account for the structural behaviour of joints between segments (Hoefsloot, 2009). The con­ crete linings are modelled as elastic materials with a Young’s modulus of 20 GPa, which is about 70% of the uncracked modulus of concrete. The bored piles are 900 mm in diameter and ~25 m in length. The two rows closest to the tunnel have centre to centre pile spacings of 1.74m (left) and 1.24m (right). The pile cap is 2 m thick. They are both modelled as elastic materials with a Young’s modulus of 30 GPa.

Figure 12. Configuration of the 2D longitudinal model.

2) Longitudinal (Greenfield) model: To assess the radius of curvature of the tunnel, a simplified longitudinal model was set up. Although the interaction between the tunnel and the building above is a three-dimensional problem, the plane strain simplified longitudinal greenfield model can provide conservative tunnel movement results. The longitudinal section considered is taken along the neutral axis of the tunnel. The piles and the pile cap are not modelled but a uniformly distributed pressure of 200kPa was applied at a depth of 2/3 pile lengths as per the equivalent raft method (Tomlinson, 2008). The model geometry is shown in Figure 12. The analyses have been carried out in a phased sequence to represent broadly the construction activ­ ities on site. The analysis phases are described below: – Initial phase to establish initial stress condition – Installation of the tunnel – Consolidation to model existing tunnel – Excavation to the base slab level – Installation of the structures (piles and pile cap) and application of 75% of the total loads – Consolidation for 6 months after 75% loading – Application of 100% of the total building loads – Consolidation to model long-term condition

2.2.4 Numerical model To assess the impact of the proposed works on the existing tunnel. two plane strain (2D) finite element models were set up using Plaxis 2D. Both models use the same ground model and input parameters as in Table 4. The models developed are described further below: 1) Transverse model: The section considered in the analysis is chosen to represent the most loaded section of the tunnel, where the pile number is the largest. There is one row of piles on the left side of the tunnel. The model geometry is shown in Figure 11. The tunnel lining is modelled as plate elements. A volume loss of 1.5% is con­ sidered by applying a tunnel contraction on the lining. The skin friction is assumed to increase linearly with depth.

In the transverse model, the tunnel is assumed wished-in-place and no installation effect has been considered. In the consolidation phases, coupled flow deformation analysis was performed. 2.2.5 Computed displacement results The computed tunnel movement from the two models are summarized in Table 7: Table 7.

Computed maximum tunnel vertical displacement.

Stage 75% loading 2nd consolidation (6 months) 100% loading 3rd consolidation (long-term) Figure 11. Configuration of the 2D transverse model.

* From monitoring data

113

Transverse mm 6.8 16.5 (15.1)* 19.0 24.0

Longitudinal (Greenfield) mm 5.8 20.4 22.9 30.3

3 CONCLUSIONS The analyses undertaken are based on a ‘greenfield’ approach; this results in a relatively simple conserva­ tive analysis designed to capture the movements of the ground as a direct response to the loading and construction of the proposed structures. Applications to assessments on underground services and tunnels are demonstrated. The strain and radius of curvature results can be derived from the greenfield ground movement results, though the strain results can only be considered relative to the underground assets in their current states. These analyses are therefore not designed to capture the gross strains at the end of construction. This, however, provides a basis for a risk assessment of the likelihood of damage to the underground assets as a result of the proposed works in line with relevant damage assessment criteria. If the criteria are violated, more detailed analysis including modelling of the underground structures and back analysis based on monitoring data may be required.

Figure 13. Typical radius of curvature results calculated from longitudinal settlements.

The maximum tunnel settlements calculated from the transverse and longitudinal models are compar­ able. The longitudinal model gives higher settlement results as expected. Maximum settlement of circa 15mm was observed and results from the transverse model indicated the maximum tunnel settlement of 16.48 mm 6 months after applying 75% of the load­ ing. This is comparable with the monitoring data.

REFERENCES Bracegirdle, A. Mair, R.J. Nyren, R.J. & Taylor, R.N. 1996. A methodology for evaluating potential damage to cast iron pipes induced by tunnelling. In Mair & Taylor (eds), Geotechnical Aspects of Underground Construc­ tion in Soft Ground: 659–664. Rotterdam: Balkema. Brinkgreve R.B.J. Kumarswamy, S. & Swolfs, W.M. 2018. Plaxis 3D Manuals. Netherlands: Plaxis BV. Hoefsloot, F.J.M. 2009. Analytical solution of longitudinal behaviour of tunnel lining. Geotechnical Aspects of Underground Construction in Soft Ground: 775–780. London: Taylor & Francis Group. Muir Wood, A.M. 1975. The circular tunnel in elastic ground. Géotechnique 25(1): 115–127. Thames Water. Guidance on piling, heavy loads, excava­ tions, tunnelling and dewatering. Available on develop ers.thameswater.co.uk. Tomlinson, 2008. Pile design and construction practice. 6th ed. Boca Raton: Taylor & Francis Group.

2.2.6 Calculation of radius of curvature One of the damage assessment criteria for the exist­ ing tunnel is that the minimum longitudinal radius of curvature needs to satisfy an allowable limit. Hence, we need derive the radius of curvature along the lon­ gitudinal direction. The radius of curvature of the tunnel is estimated from the computed displacement results from the greenfield model. It is calculated by taking the second derivatives of vertical displacement (ρ = d2u/dx2). A typical radius of curvature plot is shown in Figure 13. The radius of curvature was calculated for differ­ ent stages and the minimum value was compared with the allowable limit to determine whether miti­ gation measures are needed.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Cementitious systems with carbon nanomaterials for underground infrastructure I. Papanikolaou & A. Al-Tabbaa Department of Engineering, University of Cambridge, Cambridge, UK

M. Goisis Italcementi HeidelbergCement Group, Global Product Innovation, Bergamo, Italy

T. Embley Costain Group, Maidenhead, UK

ABSTRACT: Cementitious materials are widely used for underground infrastructure; however, the durabil­ ity of underground structures is a challenge and frequent inspections and repairs are often needed. Over the last years, biomimetic materials have been developed that could self-diagnose their condition, adapt to their environment, develop immunity and self-heal. This research investigates experimentally the effect of carbon nanomaterials on the mechanical performance of cement mortars to be used in underground construction with an outlook that they can instigate additional functionalities such as self-sensing. These additives include carbon nanotubes (CNTs), graphene oxide (GO) and graphene nanoplatelets (GNPs). Their addition was found to reduce the fluidity of the mixes, whilst they had a limited impact on mechanical strength. The microstructure remained largely unaltered whilst porosity was improved slightly by some of the additives. This research paves the way in a preliminary understanding of the effect of different carbon nanomaterials on the properties of mortars for underground structures.

1 INTRODUCTION

In addition, some forms of underground construc­ tion, such as sprayed concrete linings (SCL), present further material requirements. An industry survey, car­ ried out in the UK in 2016, showed that bonding between layers; high early strength gain; good flexural strength and ductility are key requirements to ensure that the required performance of SCL is achieved (Papanikolaou et al., 2018). Despite the widespread use of SCL in the UK, some notable failures have taken place, such as the collapse of three SCL tunnels in 1994 as part of the Heathrow Express Rail link pro­ ject (HSE, 1996) and also the partial collapse of a tunnel crown at the Crossrail project in 2014 that resulted in a fatality (Smith, 2015). To meet some of these unique underground infra­ structure demands, the concept of biomimetic mater­ ials in cementitious systems has been developed. These biomimetic materials, which are inspired by nature, could potentially and ultimately selfdiagnose their condition, adapt to their environment, develop immunity against harmful agents and selfheal if they are damaged (Al-Tabbaa et al., 2017), (Davies et al., 2018). Such biomimetic materials would therefore significantly reduce the burden of inspecting and maintaining buried infrastructure and

Cementitious materials are widely used for under­ ground infrastructure such as tunnel linings and shafts for water and transport networks, geotech­ nical structures and grouting applications. In the UK, over £600 billion of infrastructure will be delivered in the next decade (HM Infrastructure and Projects Authority, 2018). Underground construc­ tion plays a prominent role in the infrastructure investment pipeline, with projects such as Thames Tideway Tunnel, HS2 and Crossrail2 in London, along with many other utility projects that take place across the country. Despite the national significance of underground infrastructure, the condition of these structures is often unknown due to difficulties experienced when accessing such assets for inspection and repair and the associated prohibiting costs in their regular upkeep. Ensuring that underground structures per­ form as expected is critical, since any structural fail­ ure would lead to great disruption for the public. Therefore, durability of underground infrastructure is a fundamental priority for the civil engineering industry.

DOI: 10.1201/9780429321559-15

115

and also in the generation of a self-sensing mechanism (Papanikolaou et al., 2019). This section reviews the effect of different carbon nanomaterials in mortar.

the asset reliability would be enhanced. In the UK, the £4.9M EPSRC-funded Resilient Materials for Life (RM4L) Programme Grant is conducting worldleading research on the development of biomimetic cementitious materials (Resilient Materials 4 Life, 2019). A range of self-healing systems has so far been developed and field trials as well as a number of commercial applications have been carried out (Al-Tabbaa et al., 2018), (Al-Tabbaa et al., 2019). As part of the RM4L Programme Grant, carbon nanomaterials have been investigated as potential addi­ tives to reinforce the cement composites at nanoscale. These include carbon nanotubes (CNTs) and graphene derivatives such as graphene oxide (GO) and gra­ phene/graphite nanoplatelets (GNPs). These carbonbased additions have been found in some cases to improve the mechanical performance and durability of cement composites due to their barrier properties, but they also allow for additional functionalities to be introduced such as self-sensing. However, graphene was only isolated in 2004 and therefore, the literature is still limited and there are some discrepancies between the authors. This research has mainly focused on the influence of low carbon nanomaterial dosages on the mechanical performance of mortars, also giving a limited look at porosity refinement and microstruc­ ture evolution. These are important technical focus areas for underground construction. Mechanical strength gain is necessary for applications such as sprayed concrete linings, whilst a reduction in porosity is also critical for ensuring the durability of concrete structures. 2

2.1

Carbon nanotubes

Carbon nanotubes (CNTs) are allotropes of carbon with a cylindrical structure which have been success­ fully produced since 1991 (Iijima, 1991) and can be categorized as single-walled nanotubes (SWNT) and multi-walled nanotubes (MWNT). CNTs show an remarkable elastic behavior, have a Young’s modulus of 1 TPa, possess excellent thermal and electrical con­ ductivities along with an extremely high aspect ratio (length/diameter) that can help in improving the mech­ anical properties of cement composites (Singh, Kalra and Saxena, 2017) provided that effective dispersion is achieved. CNTs are usually tested at a dosage of 0.1% - 1% by weight of cement. In terms of mechanical properties almost all authors have found higher mechanical strengths (Paul et al., 2018), however, poor dispersion and agglomeration were key challenges that sometimes compromised the performance. Due to their hydropho­ bic nature and high specific surface area, CNTs agglomerate in aqueous solution (Figure 1) and reduce the fluidity of the mixes, so ensuring homogeneous dispersion is key. CNTs can also improve the microstructure - small amounts of CNTs (up to 0.4% by weight of cement) were beneficial for the transport properties and they resulted in up to 50% reduction of sorptivity, due to the refinement of pores in the mortars (Alafogianni et al., 2019). However, another study found that when the CNT content was raised, the corrosion potential and corrosion rate of steel reinforcement increased as CNTs lowered the resistivity of the cement matrix (Paul et al., 2018) fostering electrical current conduct­ ivity. Therefore, these two contradicting mechanisms

CARBON NANOMATERIALS IN CONCRETE

Nanotechnology in construction could allow for the development of more advanced structural materials since it provides reinforcement at nanoscale. Nanoma­ terials are defined as those that have particles with at least one dimension less than 100 nm (Jones et al., 2016). Nanomaterials are primarily used for enhancing the concrete performance in terms of mechanical and durability properties but most recently they are also used for advanced functionalities such as electrical conductivity (self-sensing concrete) and photocatalysis (self-cleaning concrete). Examples of nanomaterials that are already in use include nano-silica, nano­ alumina, polycarboxylates (in superplasticiser admix­ tures), nano-clay and titanium dioxide (for photocata­ lytic concrete) (Norhasri et al., 2017). Carbon based nanomaterials have recently emerged as a potential concrete addition due to their excellent electrical con­ ductivity that could allow for a self-sensing mechan­ ism to be instigated. Self-sensing is of interest in underground construction as it reduces the need for inspections and allows for a more proactive infrastruc­ ture management approach. In a recent industry survey carried out in the UK, it was highlighted that the biggest opportunity for graphene-based nanomater­ ials lies in the improvement in flexural/tensile strength

Figure 1. CNTs in aqueous solution with a polycarboxylate superplasticiser, prior to any mechanical mixing. CNT aggregation can be observed.

116

of reduced electrical resistivity and reduced porosity need to be carefully balanced in reinforced concrete to ensure that the durability performance is maintained. 2.2

Walls interactions between them (Goisis, 2017). A way to overcome the challenge of hydrophobicity is to introduce some functional oxygen groups on the graphene surface. Graphene oxide (GO) is a monolayer material functionalised with oxygenbearing groups such as hydroxyl and carboxyl types that allow it to mix better with water. Therefore, the hydrophilic islands facilitate better dispersion com­ pared to that of the more hydrophobic CNTs or GNPs. GO is usually tested at much lower dosages com­ pared to CNTs – at a scale of 10 times less (~0.01% ­ 0.1% by weight of cement). Despite these low con­ centrations and the partially hydrophilic nature of GO, its addition has been found to lower the fluidity of the mix significantly. This is because the large sur­ face area of GO absorbs water molecules and also due to the flocculation of GO-cement grains and the chemical cross-linking of de-protonated GO that forms agglomerates and results in water entrapment (Shamsaei et al., 2018). A 34% reduction in fluidity was found with 0.03% GO (Gong et al., 2014) whilst another researcher found a fluidity reduction of 42% with 0.05% GO addition (Pan et al., 2015). Both flexural and compressive strengths have been found to increase with the addition of GO, however there are many discrepancies between the authors. As an example, the addition of 0.03% GO in a cement paste has been found to result in a 46% increase in compressive strength at 28 days (Gong et al., 2014), whilst another author found only 6% increase for the same GO content (Li et al., 2017). The influence of gra­ phene on cement composites strength is still not fully understood since it may depend on the type of graphenic structure, cement composition, mix design and dispersion methodology. Far less studies have explored the durability per­ formance of GO-reinforced cement composites and therefore it is difficult to reach a conclusion at this stage. The addition of 0.04% GO was found to reduce the sorptivity coefficient by 29% (Qureshi et al., 2019) and another study also found similar reductions (Alharbi et al., 2018). The improved impermeability of cement composites with GO add­ ition, is attributed to the improved microstructure and accelerated hydration with GO addition.

Graphene-derivatives

Scientists have theorised about graphene for dec­ ades. It was originally observed in electron micro­ scopes in 1962 while supported on metal surfaces. The material was later rediscovered, isolated and characterised in 2004 by Andre Geim and Konstan­ tin Novoselov at the University of Manchester (Novoselov et al., 2012). Research was informed by existing theoretical descriptions of its composition, structure and properties. High-quality graphene proved to be surprisingly easy to isolate, making more research possible. This work resulted in the two scientists winning the Nobel Prize in Physics in 2010. Graphene was obtained through exfoliation of graphite which is composed of numerous stacked graphene layers, each of them in the form of twodimensional hexagonal lattice having a carbon atom at each vertex (Madhuri and Maheshwar, 2015). The interest aroused by the potential applica­ tions of graphene has caused a proliferation of graphene-type structures with varying number of graphene layers, average lateral sizes and carbon to oxygen (C/O) atomic ratios. Therefore, gra­ phene should be considered a class of materials, rather than a well-defined chemical structure, within the common meaning of the term (Goisis, 2017). Standardisation of graphene nomenclature is still ongoing and an attempt to a rational nomenclature based on three parameters - number of graphene layers, average lateral size and carbon-to oxygen (C/O) atomic ratio - has been issued (Wick et al., 2014). Although the rapid progress on graphene-based cementitious composites has benefited from the rela­ tively mature research on other carbon-based nano­ materials, especially CNTs, the use of graphene in concrete has some intrinsic advantage over them. Compared to CNTs, for example, graphene is: - 2-dimensional, presenting larger specific surface area of ~2600 m2/g, more than twice that of CNT (Peigney et al., 2001); - easier to produce in large quantities with accept­ able reproducible properties; - lower potential risks for health and environment (IOSH, 2017); - more easily dispersible in media, therefore more attractive for various applications in construction materials (Shamsaei et al., 2018).

2.2.2 Graphene nanoplatelets Graphene nanoplatelets (GNPs) is a sheet-like mater­ ial that consists of a small number of well-defined graphene layers (typically 95%

14 nm = 40 layers

CNTs Black colour, powder >93% purity -

30 μm (D50 = 25 μm) 30 (m2/g)

< 10μm

-

-

-

-

-

26-30kg/ m3

GNPs were supplied in paste form containing approximately 5% active matter and 95% water. GO was supplied in water dispersion with 0.4 wt% con­ centration. CNTs were an experimental product. The key properties of the carbon nanomaterials used in this study are illustrated in Table 1. The chosen quantities for each nanomaterial are based on the averages used in the literature for mechanical and durability properties improvement.

ideal and some oxygen content (2-6%) could also be present (Kovtun et al., 2019). Despite this, the integ­ rity of the structure permits GNPs to enhance dra­ matically the electrical properties of the cement composite (and generate self-sensing behaviour). GNPs behave in a similar manner to CNTs - high hydrophobicity, tendency to agglomerate and a reduction of fluidity are present in both cases. Most authors have tested GNPs at dosages of 0.1% ­ 1% by weight of cement. However, as with CNTs and GO, the effect of GNPs on the mechanical and durability properties is still unclear. Whilst most authors report performance improvements with GNPs addition, the percentage improvement is not always clear and sometimes the performance is com­ promised as illustrated in Figure 2 for compressive strength. In terms of durability, several authors have found that GNP addition reduces the porosity (mainly tested with Mercury Intrusion Porosimetry (MIP)) and that pore structure shifts from larger to finer pores (Du and Pang, 2015), (Wang et al., 2018). However, as with the other carbon nanomaterials, research on the effect of GNPs on the durability per­ formance remains limited.

3.2

Sample preparation

Initially, the carbon nanomaterials were added to a solution containing water and polycarboxylate superplasticiser. All the dispersions were mixed fol­ lowing the same protocol to isolate the effects of nanomaterial addition. A QSonica Q700 tip sonica­ tor (with a 12mm tip) was used; for a 15 min sonic­ ation that was paused every 5 mins for 1 min to control and measure the temperature. The power ranged from 55-60W and the total energy output in the 15 mins ranged from 48000–52000 J. A coldwater bath was applied to control the temperature. Standard mortars were then prepared as per BS EN 196-1:2016, adding a polycarboxylate-comb polymer which is typically used in advanced con­ crete for underground construction and adjusting the water-to-cement ratio to 0.48 to account for the improved fluidity with the superplasticiser. The sam­ ples were cast in two layers (no vibration to avoid segregation) and cured at 95%RH and 20°C for 24 hours and then in water at 20°C until testing. Close attention was paid to the mortar preparation procedure on which the measured strength values depend. The weight of the dispersion before and after sonication was measured to ensure that the loss of water is compensated. The increasing temperature during sonication can result in a slight but important

3 RESEARCH METHODOLOGY 3.1

GNPs GO (G2NanPaste) (Graphenea)

Materials

Portland cement CEM I 52.5R by Italcementi, EN sand by a German supplier, G2Nan graphite nano­ platelets (GNPs) by Nanesa, graphene oxide (GO) by Graphenea, experimental carbon nanotubes (CNTs), and Driver 31 polycarboxylate superplasti­ ciser by Axim (SP) were used in the study.

118

loss of water that impacts the w/c ratio of the mix­ ture and hence on its strength. 3.3

Experimental procedures

3.3.1 Fresh properties Immediately after mixing, the fluidity and fresh density of the samples were determined. The fluidity was measured as per BS EN 1015-3:1999 with a flow table test. The fresh density was measured using a fixed volume container. 3.3.2 Strength testing Flexural and compressive strength tests were carried out at 2, 7 and 28 days as per BS EN 196-1-2016 using a 300kN Controls testing machine. After strength testing, small pieces were kept in ethanol and acetone (5 minutes in each) to stop hydration and then in the oven at 30°C for 24 hrs.

Figure 3. Consistency (mm) of CEMI mortars.

careful selection of grading curves and the use of comb-polymer superplasticiser permits high fluidity mixtures, up to S5 slump class, to be developed. The fresh density of the samples (Table 2) remained stable, indicating that carbon nanomater­ ials did not result in any air entrapment in the mix.

3.3.3 Durability performance Porosity is a key indicator of durability performance as lower porosity prevents harmful agents from penetrating the concrete. The porosity was assessed by Mercury Intrusion Porosimetry (MIP).

4.2

The effect of carbon nanomaterials on the flexural strength is illustrated in Figure 4. A slight enhance­ ment was found at 2-days with all nanomaterials. GNPs improved the 2-day flexural strength by 5.5%, CNTs improved it by 3% whilst GO improved it by 7.5% and 1% at 0.03 and 0.06 wt.% dosages, respectively. The indication that the optimum GO content in mortar is in the range 0.02-0.06% is also present in the literature. At higher GO dosages, par­ ticle agglomeration takes place and mechanical strength is negatively impacted. At 7 and 28 days the addition of GNPs resulted in a slight enhancement of 2.9% and 0.7% respectively, however, the addition of GO was found to compromise the flexural strength. The addition of 0.03% GO led to a decrease in strength of 12.5% (7 days) and 9.1% (28 days). The compressive strength shows a similar pattern to the flexural strength (Figure 5). The addition of GNPs showed almost no effect at 2 and 7 days but resulted in a 7.6% reduction at 28 days. The CNTs also showed an insignificant change in strength at 2 and 7 days. On the other hand, GO had an accelera­ tory effect at 2 days as it increased the compressive strength by 7.4% and 4% at 0.03% and 0.06 wt.% dosages, respectively. However, at 7 days this effect disappeared and at 28 days a reduction in strength was observed from 67.2MPa (plain mortar) to 62.8MPa (0.03% GO) and 66.4MPa (0.06% GO). The strength results of the CEMI mortars show that the effect of carbon nanomaterial addition does not follow a clear relationship with dosages and curing time. An early positive effect was sometimes observed at 2 days that resulted in an enhancement of the mechanical strength, which

3.3.4 Microstructural observation A ZEISS EVO LS 15 scanning electron microscope (SEM) permitted observations at micro level. Small chipped pieces were extracted from the cracked faces of the mortar specimens. The samples were gold-coated before testing. 4 EXPERIMENTAL RESULTS AND ANALYSIS This section presents and discusses the effect of carbon nanomaterials on the fresh, mechanical and durability properties of cement mortars. 4.1

Strength testing

Fresh properties

The addition of all nanomaterials resulted in a significant decrease in the fluidity of the mortars (Figure 3). The increasing GO dosage linearly decreased the fluidity. GNPs (G2NanPaste) and CNTs (both hydrophobic) reduced the fluidity by approximately 30%. Not only their high specific sur­ face area played an important role but also their dosage which was 10 times higher than that of GO. Due to the low fluidity of the CNTs mix, the speci­ mens were vibrated for 10 seconds after mixing. This dramatic reduction in fluidity with nanoma­ terial addition, could pose a key challenge for under­ ground construction, especially for sprayed concrete. The reduction in fluidity can be prevented by func­ tionalising or treating the nanomaterials with chem­ ical admixtures, however, this could increase the cost and require specialist equipment. Tests on con­ crete (by Italcementi in 2019) showed that the

119

Table 2.

Several papers have confirmed the increase of strength of pastes and mortars containing graphene at dosages in the range R, thus when both the gradient at the tunnel face, Equa­ tion 2 and the pore velocity, Equation 3 are less than for the unconfined aquifer. 3 PLASTERING 3.1

where k is the permeability of the aquifer, k’ perme­ ability of the less permeable layer on top of the aqui­ fer, and H and S the thickness of the aquifer and less permeable layer, see also Figure 1. The pressure distribution, as sketched in Figure 1, becomes a more complex function (see Bezuijen and Xu, 2018). However, of interest for this paper is whether the drilling is in ‘permeable’ or ‘impermeable’ sand and what will be the gradient and the pore velocity. For that Equations (2) and (3) can be used, however in those Equations, R has to be replaced with Rs (Bezuijen and Xu, 2018):

3.2

Figure 1. Cross-section and top view, sketches of equal-potential lines and flow lines in aquifer (Bezuijen and Xu 2018).

Importance of plastering

For the interaction between the groundwater flow and the penetration of slurry, the period of high permeabil­ ity is of importance, because when the tunnel face has a low permeability all pressure drop will be over that tunnel face and there will be hardly any tunneling induced groundwater flow anymore. It is therefore important how long it will take before the plastering of the slurry has caused the situation of low permeability. In literature there is often reference to the experi­ ments of Krause (1987) to have an indication how fast a slurry will penetrate the soil and cause plaster­ ing. However, the results of Krause are based on infil­ tration tests that were performed with a much higher hydraulic gradient than the gradient that can be expected in tunneling (Bezuijen et al. 2016, Xu 2018). Since slurry or foam can only penetrate the soil when it can replace the groundwater, the penetration vel­ ocity of the slurry and foam depends on the hydraulic gradient at the tunnel face as calculated in Section 2. Furthermore, Xu & Bezuijen (2019) have shown that the velocity and the extend of infiltration in sat­ urated sand depends on the sand concentration in the slurry, see Figure 2. The sand in the slurry in these tests was the same sand as the sand to be excavated by the TBM, to simu­ late the excavation process where the slurry will also be loaded with sand. The results show that there is a filter cake formation for pure slurry infiltrating sand but not for sand loaded slurry. This results in an ongoing penetration of the slurry in the sand propor­ tional with the square root of time (Xu & Bezuijen, 2019). Penetration velocity in front of a slurry TBM

To calculate the penetration velocity of the slurry in front of a TBM, the method outlined in Bezuijen et al. (2016) is used. However, this method is extended to include the situation that no filter cake is formed and that the tunnel is drilled in a semi-confined aquifer. Assume the situation that the sand has a ‘low per­ meability’, thus there is hardly any slurry penetration

235

where x is the amount of penetration of the slurry or muck and t the time. It should be realized that these equations are only valid in case there is a linear rela­ tion between the penetration depth L and the head drop over the infiltrated zone and a constant permeability of the sand for the slurry (kb). In case of the creation of a filter cake, these conditions are not necessarily ful­ filled, and these equations will only be an approximation. Equation 7 can be solved for the boundary condi­ tion x = 0 as t = 0, leading to:

Figure 2. Discharged volume against square root of time for tests on slurries containing sand (Xu & Bezuijen, 2019).

during drilling and the TBM is drilling in an uncon­ fined aquifer. When drilling stops, the slurry starts to penetrate the soil, see Figure 3. The maximum penetration depth is L. The pres­ sure drop over the slurry or muck is caused by a combination of permeability of the sand for the slurry and a term that is caused by the yield stress between the and the sand or by cake formation. Assume a piezometric head that is zero far from the tunnel. In that case Equation 1 presents the relation between the piezometric head in the soil just in front of the slurry. The piezometric head in the mixing chamber can be written as:

where kb is the permeability of the soil for the slurry. Since vp=dx/dt and using Equation 3, this leads for flow of slurry through an unconfined aqui­ fer to the equation:

The equation can also be used for semi-confined flow but then R has to be replaced with Rs as given in Equation 5. Equation 7 and 8 show that the pene­ tration velocity is a function of the permeability of the sand in front of the TBM. Since the radius of the TBM (R) is normally much larger than the penetra­ tion depth (x), the groundwater flow normally dom­ inates the penetration velocity, unless very thick bentonite is used with a orders lower permeability through the sand than the permeability of water. In case there is no maximum penetration depth (This for example the case for slurry with soil with a density of 1500 kg/m3), the last term of Equation (7 can be omitted and the equation can be written as:

Solving this last equation for the boundary condi­ tion that t = 0 and x = 0 results in:

According to this formula x will increase linearly with time as long as:

Figure 3. Definition sketch, penetration of slurry during standstill.

and with the square root of time when this term is much larger than 1. This result agrees with the results of the infiltration experiments with the extra flow resistance, see Figure 2 and the slurry with soil with

236

higher densities. In this figure the penetration is plotted against the square root of time resulting in a quadratic increase of x for small times and a linear increase for larger times. The presented solutions are two extremes of what can happen during slurry or foam penetrate on. These two solutions are capable to simulate the slurry penetration for the pure slurry and the 1500 kg/m3 slurry with sand. Using the param­ eters given in Table 1, the result shown in Figure 4 was obtained. The calculations were run for a real tunnel, but since the gradients in the infiltration tests are approximately the same as the ones to be expected for a real tunnel (this was realized by installing a constriction the tube used for the infiltration tests, Xu & Bezuijen, 2019), the results are directly comparable. In case of pure slurry, Equation. 7 is used and Equation 9 for the 1500 kg/m3 slurry with sand. Integra­ tion of the dx/dt found leads to the penetration depth (x). dx/dt is equal to the pore velocity (vp), therefore, Equation 3 can be used to calculate the pore water pressures just before the infiltrated Table 1 .

Parameters used in calculations.

Parameter

slurry

dim

Excess pressure face Diameter tunnel Permeability soil, water Perm. soil bentonite Porosity sand Drilling velocity (drilling) Max. penetr. pure slurry

50 10 4.e-4 1.e-6 0.4 6.7.e-4 0.08

kPa m m/s m/s m/s m

14.85 24 4e-4 8e-6 1.5e-6

m m m/s m/s m/s

GHT calculations Sect. 6.3 Diameter tunnel App. diameter tunnel Permeability soil, water Perm. soil bentonite Permeability top layers

Figure 4. Results of simulations with different models, see also text. Slurry penetration can be compared with meas­ ured values (see Figure 2). Excess p means excess pore pressure.

zone in front of the TBM. This result is also shown in Figure 4. It can be seen as the course of the excess porewater in front of the TBM when drilling stops. It appears that in both cases the excess porewater pressure decreases rapidly. However, for the pure slurry, where the flow is blocked by a filter cake, the excess porewater pressure goes to zero. While some excess pore pressure remains in case of high-density slurry with sand and bentonite. 4 CALCULATION EXAMPLES 4.1

Drilling

The first example is a slurry shield in low permeable sand. Parameters are shown in Table 1. The slurry has taken some soil due to drilling and the density of the soil is 1050 kg/m3. Parameters are the same as for the infiltration experiments and the maximum infiltration of the slurry for one-hour penetration is approximately 0.08 m. During drilling, the cutter head rotates with 1.7 revolutions per minute. On average 2 teeth will pass at the same position in one revolution, thus that means that slurry can penetrate for a bit less than 18 s before it is scraped away due to the next teeth passage. The excess pore water pressure is 50 kPa thus 5 m water column, the diameter of the tunnel is 10 m. For this situation, the slurry penetration is com­ pletely dominated by the groundwater flow before the tunnel. Solving Equation (7 numerically it was found that the slurry penetrated only 0.012 m into the soil before the next teeth is coming. Using Equa­ tion 3, a penetration of 0.014 m is found, thus the influence of both the slurry flow through the sand and the maximum penetration depth of 0.08 m is very small and indeed the groundwater flow domin­ ated the slurry penetration. Since an average velocity of a slurry TBM during drilling is around 40 mm/ min, or 0.67 mm/s (see Table 1), the TBM will have moved forward 0.012 m in 18 s and thus nearly all penetrated slurry will be removed by the next teeth that passes that same area of the front face. For a bit lower permeability, all slurry will be removed. In case the sand has a higher permeability, not all pene­ trated slurry will be removed and a slurry penetrated sand layer forms in front of the TBM. The thickness of this layer will increase until the penetration vel­ ocity (given in Equation (7 is equal to the drilling velocity. Consequently, there is hardly any cake formation at the tunnel face during drilling in this example and excess pore pressures close to the applied over pres­ sure in the mixing chamber can be expected during drilling. Another consequence is that it will not make much difference during drilling what is the exact density and permeability of the slurry, since this is scraped away before it can form a cake.

237

4.2

Standstill in semi-confined aquifer

Infiltration will go on to larger depths in the sand in front of the TBM during standstill. The density of the slurry and the permeabilities will determine how far the slurry will penetrate and also how long it will take before excess pore pressures are reduced in the soil. If the slurry in the mixing chamber still has a density of 1050 kg/m3, it will not penetrate more than 0.08 m, see Figure 2. It is also shown that for a higher slurry density (1500 kg/m3) the penetration will be more, up to 0.27 m, if ring building can be accomplished within one hour. The field measurements are not conclusive what densities can be expected. The pressure gradient at the 2nd Heinenoord Tunnel correspond to densities between 1260 and 1450 kg/m3 (COB, 1999) and for the Green Hart Tunnel (GHT) in the Netherlands values of around 1500 kg/m3 were found. In the GHT machine, the density was also measured in the slurry pipelines coming from the mixing chamber. Here a density of only 1260 kg/m3 was found during drilling. Although not conclusive, looking with these results to Figure 2, only limited cake formation can be expected and there will be a more or less ongoing infiltration into the soil. The model described in Section 5.2 can be used to simulate the pore pressure decay when drilling stops. In this example parameters for the GHT were used. Since the measurements for this tunnel were obtained in a semi-confined aquifer, Equation 5 has to be used to calculate the effective radius. This appeared to be 12 m. The permeability of the sand is 4×10-4 m/s according to Autouri and Mimec (2005) and they also mention the excess pressure in the mixing chamber, 50 kPa above the pore water pressure in the soil at the tunnel axis. The excess pore pressure measured is less (see Figure 5 and Figure 6), but this was measured not in front of the TBM but to the side. To simulate the infiltration, the pressure at the tunnel face has to be used as input. Assuming Darcian flow the result can then be scaled to the measured pressure. This is done in Figure 5. It is clear from this figure that the decay predicted assuming clean slurry is too fast. The measured value for a slurry loaded with sand up to a density of 1500 kg/m3 results in a better agreement, but now the measured decay is faster than according to the simulation. Possibly the slurry has a lower density then the assumed 1500 kg/m3 and there is still some cake formation. However, also slight variations in the permeability can cause these differences. Fur­ thermore, the permeability of the sand for bentonite is a difficult parameter. The process of slurry pene­ tration and cake formation is described here with two parameters, the permeability and the penetration depth. Measurements of Xu & Bezuijen (2019) have shown that this is a rather crude way to describe the processes. Cake formation will lead to very low per­ meabilities, but also when there is no cake formation as can be expected when a high-density slurry-sand

Figure 5. Measured and calculated pressure decay com­ pared during standstill. Measurements from the GHT.

Figure 6. Pore pressure measurements during construction of the GHT. WE2 is measured 1 m from the tunnel, WR1 at approximately 100 m away. perpendicular to the axis.

mixture is used in an infiltration test, the permeabil­ ity of the sand for the bentonite decreases over time because the bentonite concentration in the infiltrated sand increases. The simplified approach chosen in this paper allows to see the interactions between groundwater, slurry and TBM drilling, but will not result in perfect agreement. 4.3

Slurry shield, restarting

When, after ring building, the TBM is restarted to drill for the next ring in the lining, the filter cake (if present) and the penetrated slurry will be ‘eaten’ by the TBM. An external filter cake will disappear immediately when the cutter head starts moving and rotating because it is an external filter cake which grows from the soil at the front face into the mixing chamber. For the bentonite that has infiltrated into the soil during standstill, it will take longer before it is removed by the TBM. It therefore can be expected that when drilling starts, there is a sharp increase in pressure when the filter cake is removed followed by a slower increase as also the soil layer in which the slurry has penetrated becomes thinner.

238

This is what is found during measurements at the GHT, see Figure 6. A steep rise is followed by a slower rise during drilling and a decrease during standstill. Since the model as described here cannot handle cake formation, the model was run assuming only slurry penetration over 24 cm. The result of a comparison between the calculated pressure increase during drilling for the peak shown around 14:00 hour in Figure 6 is shown in 7. As could be expected, the agreement is not very good. Since the cake is not modelled, the calculated pressure increase is too slow in the beginning. Fur­ thermore, a maximum is reached in the calculation (as all ground in which the bentonite has penetrated is excavated). Such a maximum is not so clear from the measurements. A previous calculation (Bezuijen et al. 2016) using the same model, but not taking into account that the infiltrated zone could be thicker due to the sand loaded slurry and also not taken into account that the tunnel was drilled in a semi-confined aquifer resulted in a pressure rise that was even stee­ per than measured. This indicates that enough infor­ mation on the aquifer and on the slurry properties is necessary to make an accurate calculation. With some different assumptions, it is possible to make a better fit. If there is a filter cake, then the per­ meability of the sand for bentonite that was infiltrated before the filter cake was formed will be rather large, larger than the value in Table 1 (Xu, 2018). A value of 4x10-5 m/s is found in tests. Furthermore Broere (2001) and Hoefsloot (2001) have argued that there can be some elastic storage in the soil layers. That means that the semi-confined aquifer will start as a more or less unconfined aquifer and therefore the effective diameter of the tunnel is only 14.85 m instead of 24 m as mentioned in Table 1. With these assumptions a better fit between simulations and measurements is possible, see Figure 8. This can be seen as pure curve fitting, and in fact it is. However, it is interesting to study the difference between the calculated curves of Figure 7 and Figure 8

Figure 8. As Figure 7, but with different parameters, see also text.

a bit more in detail. In Figure 7 a semi-confined aquifer was assumed, that leads to a reduction of the hydraulic gradient in front of the tunnel. Therefore, the drilling velocity is higher than the penetration velocity and at t=6000 s in Figure 7 all infiltrated bentonite is removed due to the higher drilling velocity compared to the infiltration velocity. In Figure 8, it is the other way around, now the maximum infiltration velocity is a bit higher than the penetration velocity. When drill­ ing start and the slurry has penetrated over some dis­ tance into the sand, the penetration velocity is partly determined by the permeability of the sand for the slurry, the penetration velocity is still slow, and the infiltrated zone becomes thinner. However, when the infiltrated zone is rather thin, it is the higher perme­ ability of the sand for water that determines the infil­ tration velocity and there remains a thin infiltrated zone in this calculation. Looking at the measurement results and their conformity with the simulation, it is not unreasonable to assume that was also during the measurements still a small infiltrated zone during drilling. In this way such a ‘curve fitting’ can be useful to study the possible mechanism that occur. 4.4

Figure 7. Measured pressure increase during drilling com­ pared with calculation.

Determination of leakage length

The calculation with a semi-confined aquifer requires the leakage length, Equation 4. This is determined from the measured excess pore water distribution. As mentioned before, the excess pore water pressure around the tunnel is also measured 100 m away from the tunnel as shown in Figure 6. The excess pore water pressure can be measured at such a large distance because the tunnel was con­ structed in a semi-confined aquifer. According to the calculation method presented for an unconfined aqui­ fer, the excess pore water pressure at a certain dis­ tance from the tunnel is proportional with 1/R. The GHT is a large tunnel with a shield diameter of 14.85 m, thus R=7.425 m. WE2, see Figure 6, is placed 8.8 m from the axis of the tunnel and WR1 107.5 m. In an unconfined aquifer, the increase in

239

piezometric head when drilling would be at WR1 only approximately 10% of the increase in WE2. However, it appears that because the GHT is con­ structed in a semi-confined aquifer, the fluctuation in pore water pressure at WR1 is 30% of that of WE2. The leakage length, Equation 4 of that aquifer has an influence. To determine the leakage length (λ) of the semi-confined aquifer, it is necessary to know the permeabilities and the dimensions of the various layers. The permeability of the sand layer is 4×10-4 m/s. The thickness of the peat layer in the polder at the measurement location is 10 m and the thickness of the sand layer is approximately 30 m. The perme­ ability of the peat layers is not given. However, this can be calculated using the Equation 4, the equations for the piezometric head presented by Bezuijen & Xu (2018) and the measurement result that the amp­ litude in the measured pore pressure around 100 m from the tunnel axis is still 30% of the pore pressure measured 1 m outside de tunnel, 8.8 m form the axis. This corresponds with a leakage length of 300 m and a permeability of the peat layers of 1.5×10-6 m/s. This seems a reasonable value. With Equation 5, the effective radius can be determined to be 12 m, which means an effective diameter of 24 m as mentioned in Table 1. What is written above, is based on steady ground­ water flow as can be described with Darcy’s law. No elastic nor phreatic storage is included. Analytical solutions as presented by Broere (2001) and Hoef­ sloot (2001) do include an elastic storage term. This may be useful to take into account for some aspects of the groundwater flow. 5 CONCLUSIONS A calculation method is presented that allows to cal­ culate the excess pore pressures around a slurry shield for drilling both in an unconfined and semiconfined aquifer, assuming steady flow. This results in an indication how far slurry will penetrate during drilling and during standstill for ring building. The calculation method described in this paper shows the influence of tunnel diameter, permeability of the soil, the type of aquifer, the viscosity of the slurry and the sand content in the slurry on the infil­ tration velocity. All these aspects are not taken into account in empirical infiltration equations. As a consequence, these empirical relations determined by small scale tests can vastly overpredict the pene­ tration velocity of the slurry and the time is takes to form a filter cake. From the calculations presented in this paper it can be concluded that the influence of the parameters mentioned is as follows (assuming that only that parameter varies): -

A larger diameter tunnel leads to a lower gradi­ ent in the groundwater flow and thus will it take longer before a certain infiltration is reached, or a filter cake is formed.

-

The same can be mentioned for lower perme­ ability soil and for a semi-confined aquifer, com­ pared with an unconfined aquifer. More sand content in the slurry leads to more infiltration and therefore it will take longer before a low permeable layer is formed during standstill and slurry penetration stops. It also will take longer before a slurry infiltrated zone is ‘eaten’ by the TBM when drilling starts again after ring building.

Checking the consequences of excess pore water pressures in the design phase of a tunnel can further improve the tunneling practice and lead to a reduction of ‘unexpected’ settlements or blowouts.

REFERENCES Aime, R., Aristaghes, P., Autuori, P. and Minec, S. 2004. 15 m Diameter Tunneling under Netherlands Polders, Proc. Underground Space for Sustainable Urban Devel­ opment. Proc WTC Singapore, Elsevier. Autuori, P. and Minec, S. 2005. Large diameter tunnelling under polders. Proc. 5th Int. Symposium on Under­ ground Construction in soft Ground, IS-Amsterdam 2005, pp 181–186 Bezuijen, A. and Xu T. 2018. Excess pore water pressures in front of a tunnel face when drilling in a semiconfined aquifer. Proc. WTC 2018, Dubai. Bezuijen, A., Steeneken, S.P. and Ruigrok, J.A.T. 2016. Monitoring and Analysing Pressures around a TBM. Proc. 13th Int. Conf. Underground Construction Prague, Prague, Czech Republic. Bezuijen A. and Talmon A.M. 2008. Processes around a TBM. Proc. 6th Int. conf. Geotechnical aspects of under­ ground construction in soft ground, Shanghai, China pp: 19–30 Bezuijen, A. 2002, The influence of permeability on the properties of a foam mixture in a TBM. Proc. 4th Int. Symp. on Geotechnical Aspects of Underground Con­ struction in Soft Ground - IS Toulouse. Bezuijen A., Pruiksma J.P. and Meerten H.H. van 2001. Pore pressures in front of tunnel, measurements, calcula­ tions and consequences for stability of tunnel face. Proc. Int. Symp. on Modern Tunneling Science and Techn. Kyoto. Broere, W. 2001 Tunnel Face Stability & & New CPT Applications. PhD thesis. Delft: Delft University of Technology. COB committee K100. 1999, Monitoring during construc­ tion of the second Heinenoord Tunnel. Report of largescale research for bored tunnels. COB Dias, T. G. S. and Bezuijen, A. 2016. A different view on TBM face equilibrium in permeable ground. In Proceed­ ings of the ITA world tunnel congress 2016 – uniting an industry, vol. 1, pp. 94–103. Englewood, CO, USA: Society for Mining, Metallurgy and Exploration. Hoefsloot, F.J.M. 2001. Pore pressures in front of the bore front: A simple hydrological model (in Dutch). Geotechniek, October. pp. 26–33. Kaalberg, Nijs R.E.P. de, and Ruigrok J.A.T. 2014. TBM face stability & excess pore presssures in close proxim­ ity of piled bridge foundations controlled with 3D FEM. Proc. of 8th Int. Symp. on Geotechnical Aspects of Underground Construction in Soft Ground, IS-Seoul.

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Krause, T. 1987. Schildvortrieb flüssigkeitsgestuetzte Orts­ brust. (in German) Dissertation. Technische Universität, Braunschweig. Xu, T. 2018. Infiltration and excess pore water pressures in front of a TBM: experiments, mechanisms and com­ putational models. PhD Thesis, Ghent University, Belgium. Xu T. and Bezuijen A. 2019. Bentonite slurry infiltration into sand: filter cake formation under various conditions. Géotechnique [https://doi.org/10.1680/jgeot.18.P.094]

Zizka, Z. 2018. Stability of a slurry supported tunnel face considering the transient support mechanism during excavation in non-cohesive soil. PhD thesis, Bochum University, Germany. Zizka, Z., Schoesser, B., Popovic, I. and Thewes, M. 2017. Excess Pore Pressures in Front of the Tunnel Face During Slurry Shield Excavations due to Different Time Scales for Excavation Sequence of Cutting Tools and Penetration Time of Support Fluid. EURO: TUN 2017, Innsbruck University, Austria.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Face pressure and volume loss relationships for pressurized tunneling in granular soils S.J. Boone Golder Associates Ltd., London, Canada

J.N. Shirlaw Golder Associates (HK) Ltd., Hong Kong

ABSTRACT: Urban tunneling in sand using pressurized face tunnel boring machines must apply pressures sufficient to avoid large ground losses, comply with surface settlement performance requirements and also avoid ground heave. While empirical methods are commonly used to set performance requirements, these do not adequately address root causes of volume loss at tunnel level. Controlling ground movement at the source has become the primary means of protecting infrastructure through specification of very small settlement or volume loss tolerances. Until the contributions of each ground loss source are quantitatively examined, field control will be dominated by subjective opinion, disparate methodologies and trial and error adjustments. While centrifugal modelling of tunnels in sand has been completed, the results are of limited practical use. This paper summarizes centrifuge and field data in the context of face and annular slurry or grout pressures at ultimate failure and pressure-ground displacement relationships for tunnels driven in sand.

1 INTRODUCTION The typical behavior of unsupported sand in a tunnel face is to run back to a natural angle of repose (dry sand) or to flow (sand below the water table), Terza­ ghi (1950) with raveling behavior temporarily exist­ ing within a narrow range of water content. Traditionally, tunnels in sand have been, and some­ times still are, driven by means of close timbering, often in combination with dewatering, spiling/fore­ poling or ground treatment. The introduction of Pres­ surized Tunnel Boring Machines (PTBMs) over the last 50 years has allowed a support pressure to be applied to stabilize sand and allow rapid tunneling with generally small resulting ground movements in conditions that would have been considered very dif­ ficult or impossible prior to PTBM development. Design of PTBMs to handle various ground condi­ tions has continually improved with respect to slurry and muck conditioning materials and in testing and understanding of how best to apply the diversity of conditioning agents now available. The very large number of PTBMs used over the last two decades means that there is a pool of expertise in how best to plan for and operate the TBMs. As a result, the abil­ ity to achieve very low values of ground movement on a consistent basis has substantially improved and is being recognized in design and specification of new projects. However, what is required in practice

may now have started to extend beyond our empir­ ical understanding of the pressure-ground displace­ ment relationship for tunneling in sands. 2 DESIGN ISSUES Until recently, it was common practice for designers to specify values for maximum ground surface settlement trough unit volume (“volume loss”) over PTBM driven tunnels in the range of 1% to 2% of the tunnel volume. The potential for damage to nearby buildings and utilities would typically be assessed based on similar or higher values (to pro­ vide a margin for safety). Where a building or utility was identified as being at risk of unacceptable damage, building/utility protection measures were either specified in detail or outlined for the contrac­ tor to develop into a final design. This common prac­ tice has continued to evolve with lower and lower values for maximum volume loss being specified for PTBM tunneling as a means of controlling ground and infrastructure damage by controlling the source of the ground displacement, i.e. the tunneling. As a result, the need for conventional building or utility protection measures is often eliminated, based on the low specified values for volume loss. This approach realizes the full potential for PTBMs to control ground movements. It also avoids the need for

DOI: 10.1201/9780429321559-31

242

expensive building protection measures. These can cause as many problems as they solve, particularly for ground improvement around tunnels in sand, which is prone to disturbance by multiple drill holes needed for many treatment technologies. This current approach has many advantages but places responsibility on the contractor’s engineer and operating crews to control all sources of ground loss at the tunnel level. Sources of ground loss at tunnel level during PTBM tunneling include ingress of ground at the face related to support pressures, convergence around the PTBM, convergence around the lining and subsequent consolidation of soft clay soils. Of these, operating PTBM face pressure and grouting pressure and volume are the most readily controllable. Operating pressures for PTBMs include: • target, minimum and maximum values for the face pressure during PTBM advance; • fluid pressure in the annulus around the PTBM skin; • minimum and maximum values for tail void grouting pressure and volume; and • face pressure needed for interventions under com­ pressed air. Target operating pressures are normally estab­ lished by calculation before the commencement of the tunnel drive. As the operating pressures are a critical factor in avoiding damage to buildings and utilizes, it is now becoming common for the pressure calculations to be subject to an independ­ ent check, and to review by owners and regu­ lators. Franza et al. (2019b) record that on one project there were seven parties involved in the review of the operating pressure calculations. Where there are multiple parties involved in review and approval of operating pressures, this tends to restrict flexibility in terms of adjusting the calculated pressures during tunneling. There should be procedures to adjust these initially cal­ culated values for the target pressure, to allow for changes in the critical input parameters, such as the groundwater pressure. While 1% volume loss has traditionally been regarded as an excellent outcome for tunneling, we have experienced many project specifications have required the contractor to consistently achieve lower values, such as 0.5% or 0.75% volume loss (e.g., Kramer et al., 2015; Shirlaw et al., 2017). None of these specifications for very low volume loss was for tunneling in soft clay and many included some tun­ neling in sand, cement-ed sand or sand-like mater­ ials, such as completely weathered granite. Establishing operating pressures necessary to achieve a maximum value for volume loss of 1% in sand is not onerous. Golder (2009) suggest a simple method of calculation based on analogy with the design pressure on non-yielding tunnel linings. This has been used successfully in completely weathered

243

granite and alluvial and marine sand deposits in Hong Kong. Finite element analysis is also com­ monly used to establish the operating pressures necessary, as outlined in Kwong et al. (2019) or Ring and Comulada (2018). Ring and Comulada (2018) discuss some of the problems in modelling, particularly in establishing the actual pressure applied to the tail void grout. Vu et al. (2016) provide a summary of field data from five tunneling projects in sand, with a range of surface volume loss of 0.21 to 1.2%. Most of the results were less than 1%, but less than half of the results were lower than 0.5%. Kwong et al. (2019) provide field data indicating 0.61% to 1.3% volume loss over a very large, very shallow tunnel. The current trend of specifying maxi-mum values for volume loss below 1% imposes a significant potential constraint on those that set the target oper­ ating pressures. Design of instrumentation over and adjacent to the tunnel typically includes pre-set review levels that relate to the design and specifica­ tions. Exceeding specified maximum values for volume loss or settlement will typically trigger a review of the operating pressures and procedures, and may result in a more severe penalty, such as an imposed halt in the tunneling. Establishing the cor­ rect values for the operating pressures is therefore essential to successful tunneling, as failure to do so can adversely affect tunneling progress and risk damage to third parties. Normal variation in the operation of the PTBM and in ground properties result in a significant scatter of measured values for volume loss or settlement measured over tunnels, even with the same PTBM and tunneling crew in apparently consistent ground conditions. Shirlaw (2000) reported that over well controlled tunneling in apparently consistent ground conditions, the minimum and maximum settlement recorded over a tunnel could differ by a factor of 2.5, comparable with the ranges quoted above (Vu et al., 2016; Kwong et al., 2019). Much larger ranges could be measured in difficult ground conditions or for poorly controlled tunneling. The contractor’s engineer must establish appropri­ ate operating parameters for the PTBM at regular intervals along the tunnel alignment, suitable for the ground and groundwater conditions as they vary along the alignment. In this process, the designer must also allow for some degree of variation in the performance of the tunneling. As a result, the target volume loss should be lower than the maximum value specified, unless occasional exceedance of that value can be tolerated. So, if the owner specifies a maximum volume loss of 0.75%, or the equivalent value in terms of surface settlement, then the con­ tractor’s designer may decide to aim for 0.5% volume loss or less, based on theoretical calculations and an assumption of reasonably well controlled tun­ neling. Maintaining values of volume loss as low as this in sand is beyond typical practice and is trending

toward an environment within which statistical qual­ ity control concepts (e.g., Shewart, 1986) need to be applied. Very low values for volume loss can often be achieved in cemented sands or cemented sand-like materials, such as weathered granite. Shirlaw et al. (2017) record a maximum volume loss of 0.75% over two slurry PTBM drives in mixed grades of weathered granite. Weathered granite typically has some residual cementation, and like other cemented materials, conservatively low values for the cemen­ tation are adopted for design purposes. As a result, the actual volume loss values are typically signifi­ cantly below those predicted. However, tunneling in uncemented sands will not experience this benefit, and consistently achieving very low values of volume loss still extends beyond most current prac­ tical experience.

Commonly, 2D and 3D numerical modelling as used in practice does not relate the support pressure and the surface volume loss but relies on simple assump­ tions related to ground losses at the face and conver­ gence around the tunneling systems. This is no different to assuming a value for volume loss and does not provide the necessary linkage between the pressures applied at tunnel level and the resulting volume loss. We have used published data from centrifuge test­ ing of model tunnels in clay (e.g., Mair, Gunn and O’Reilly, 1981) to calculate face pressures required to limit ground displacements on many tunneling projects. Subsequent field monitoring data has shown the robustness of this approach. Currently, similar charts that relate tunnel and surface volume loss to the Load Factor do not exist for tunneling in sand.

3 PTBM OPERATION PLANNING

4 CENTRIFUGE MODELLING OF GROUND RESPONSES TO TUNNELLING IN SAND

Currently, there are three common methods for assessing the pressures needed to control surface volume loss over PTBM tunnels in sand:

While there have been numerous papers published on the results of centrifuge tests on model tunnels in sand, as discussed by Franza et al. (2019), the appli­ cation of these results is limited since:

1. a Factor of Safety against failure (or partial fac­ tors on geotechnical parameters) is chosen under the assumption that the resulting support pressure will control both the surface volume loss and the risk of failure, in which the surface volume loss estimate is based largely on empirical case his­ tory data; 2. a simplistic Serviceability Limit state analysis that uses an analogy to the design load for rigid tunnel linings to assess the support pressure required at small values for ground movement (e.g., Golder Associates, 2009 and CEDD, 2014); and 3. numerical methods, such as finite element or finite difference analysis (2D or 3D).

1. Most of the testing has been carried out in dry sand. From the limited testing carried out in sat­ urated sand, the results from testing in dry sand may not be directly applicable to saturated sand, even after allowing for the water pressure. 2. Almost all the testing has been either of move­ ment of the face only or 2D plane-strain conver­ gence, neither of which fully represents the scope for movement and arching that occur during PTBM tunneling.

Methods 1 and 2 have provided a reasonable basis for assessment on numerous projects where empir­ ical performance data are available and are suffi­ ciently comparable and relevant. There is no theoretical foundation to these methods that allows adjustment to cover new or atypical situations. Methods 1 or 2 are commonly used for routine ana­ lyses. It is now common to provide values for target pressures at intervals of 10 m to 50 m along a tunnel, requiring 20 to 100 sets of calculations per kilometer of tunnel, and spread-sheet calculations provide a simple means of rapidly and transparently provid­ ing the necessary values. Numerical analyses (Method 3) are commonly used to assess ground responses in critical locations (Golder, 2009) and many case histories and research papers. Detailed, non-linear 3D analysis required to relate support pressure to surface volume loss requires specialized numerical modelling expertise.

Recent centrifuge testing of model tunnels in sand provides some potentially useful, if limited, guidance to designers on aspects of the design problem related to very low values of specified volume loss. Particu­ lar issues that relate directly to volume loss include: the width of the settlement trough; dilation or con­ traction of the sand; and the relationship between support pressures (at face and around the PTBM) and their consequent influences on volume losses at tunnel level and ground surface. 4.1

Settlement trough width

The width of the settlement trough is a major fac-tor in assessing the potential for damage to buildings and other infrastructure during tunneling. As the settlement trough becomes wider: • more buildings and utilities are potentially affected by the tunneling; however • for a given value of volume loss, settlement and horizontal strain values are smaller.

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The distance to the point of settlement trough slope inflection, i, is typically represented as a function of the depth to the axis level of the tunnel (Zo), such that:

4.3 Face and radial support pressure and volume loss relationships

A commonly quoted value for K is 0.35 for tunneling in sand, based on Mair and Taylor (1997). However, Franza et al. (2019) have shown that the value for K in sand varies with the cover to diameter ratio (C/D) and the volume loss. At low values of volume loss, the value for K is significantly greater than 0.35. For 0.5% volume loss, the value for K may be as high as 0.5 for shallow tunnels (C/D = 1.3) and greater than one for deeper tunnels (C/D = 6.3). This wider settlement trough, through a higher K factor, can represent a significant difference between tunnels in sand with very low values for volume loss and the higher values that have gen­ erally been the case in the past. 4.2 Dilative or contractive ground behavior Franza et al. (2019) distinguish between the volume loss recorded at the level of the tunnel, Vlt, and the volume loss recorded on the surface, Vls, in keeping with other researchers’ concepts of volume losses propagating from the source to final surface expression (e.g., Atkinson and Potts, 1977; Lee et al. 1992, Loganathan and Poulos, 1998) rather than focusing on only empirical cor­ relations of surface volume loss case history data to ground types. The value of Vls is not necessar­ ily the same as the value of Vlt. If the sand is exhibiting dilatant behavior, Vls will be smaller than Vlt. If the sand is exhibiting contractive behavior, Vls will be larger than Vlt. Dilatent ground is generally beneficial for design and con­ struction, since the effect on the surface is less than that at tunnel level. Conversely, if the ground behavior is contractive, volume loss at the ground surface is greater than the volume loss at tunnel level and adverse to both design and con­ struction expectations. Given that volume losses at tunnel level are controlled by the operating pressures (at the face, annular gap slurry and grouting), contractive behavior requires excep­ tionally close control over Vlt since, tunnellers have to achieve even lower values for Vlt to achieve a given value of Vls. Franza et al. (2019a) show that loose and medium dense sands are contractive at low values of Vlt ( 5%). In both Figures 1 and 2, the illus­ trated volume losses are those at tunnel level.

245

Figure 1. Summary of centrifuge experiment data for plane strain (2D) conditions. Centrifuge data for dry conditions at Dr=0.3 and 0.9 from Franza et al. (2019a), saturated conditions from Lee et al. (2004). Field data taken from Orfila et al. (2007), Mignini et al. (2008), Hsiung and Lu (2008), Antiga and Chiorboli (2009) and Gens et al. (2012). Field data for completely decomposed granite (CDG) taken from Shirlaw, et al. (2017).

Figure 2. Summary of centrifuge experiment data for face support conditions. Centrifuge data for dry conditions at Dr=0.65 from Chambon and Corte (1994), Dr=0.3 from Idinger et al. (2011), Dr=0.26 and 0.83 from Kirsch (2009) and saturated conditions from Plekkenpol et al. (2006). Field data taken from Orfila et al. (2007), Mignini et al. (2008), Hsiung and Lu (2008), Antiga and Chiorboli (2009) and Gens et al. (2012). Field data for completely decomposed granite (CDG) taken from Shirlaw, et al. (2017).

Figure 1 can be used to establish some general relationships: • saturating the ground significantly worsens the Vlt performance, with a higher volume loss for a given value of Load Factor, compared with dry ground; and

• as C/D increases, the volume loss reduces for a given value of Load Factor • as the relative density increases, the volume loss reduces for a given value of Load Factor A summary of centrifuge data relating pressure and movement at the face is presented in Figure 2.

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Compared with the number of plane strain tests, there are few published tests that relate the support pressure to the movement at the face of the tunnel, and Figure 2 includes data from 1g as well as centrifuge tests. There are also very few published tests of model tun­ nels in sand under saturated conditions; an exception is a study by Plekkenpol et al. (2006). Most of the testing is for stationary model tunnels in dry sand. However, the majority of actual tunnels in sand will likely be below groundwater levels. As demonstrated in Figures 1 and 2, what evidence there is shows a significant difference between the ground behavior under dry and saturated conditions. What can be seen in Figure 2 is a very abrupt change from generally small values of minimal volume loss up to a Load Factor of 0.8 to 0.9, to the rapid increase in volume loss as the face approaches failure in dry sand. This almost “brittle” ground behavior of tunnels in dry sand can be contrasted with the much more “ductile” behavior of clays seen in the data presented by Mair, Gunn and O’Reilly (1981). It is likely that the rela­ tionship between tunnel volume loss and load factor for PTBM tunneling will lie somewhere between the values for movement at the face only (Figure 2) and the plane strain tests summarized in Figure 1. 5 FIELD PERFORMANCE OF PTBM TUNNELS IN SAND Field data published by Antiga and Chiorboli (2009), Hsiung and Lu (2008), Orfila et al. (2007), Mignini et al. (2008), Gens et al. (2012) and Shirlaw et al. (2017) are also included in Figures 1 and 2 where information was available to discern the appropriate parameters. Field data is shown in both Figures 1 and 2 since separation of face losses and convergence was not fully possible based on the available information. As a comparison to the dry sand model data, data is shown for a tunnel in Milan completed above groundwater in dense granular soils using an earth pressure balance (EPB) TBM (Antiga and Chiorboli, 2009). Cover to diameter ratios ranged from about 0.8 to 2.3. In this case, Antiga and Chiorboli (2009) pre­ sent face pressures (EPB chamber pressures) versus surface settlement performance, approximately con­ verted to surface volume losses for the purpose of this paper. Almost all data is for surface volume loss of less than 1%. Based on the work of Franza et al. (2019a), the volume losses at surface level would be approximately equal to those at tunnel level. As a result, the field data can be roughly compared with the data for Vlt obtained from the centrifuge testing. Support pressures at collapse were assumed to be con­ sistent with those calculated using the approach of Anagnostou and Kovari (1996). Since an EPB TBM was utilized, there would have been little to no slurry pressure within the cut-to-shield gap and closure of the shield to lining gap would have been controlled by the effectiveness of the tail void grouting. Antiga

and Chiorboli (2009) indicated that approximately 25% of the volume loss (or settlement) occurred at or ahead of the TBM face, with the remainder due to closure of the gap around the TBM and lining, neg­ lecting, however, the time delay associated with verti­ cal propagation of the ground losses in relatively dry ground. Field data from tunneling in saturated mediumdense silty sand, below the groundwater table, in Kao­ hsiung, are illustrated in Figures 1 and 2 (Hsiung and Lu, 2008). An EPB TBM was used for this project. Hsiung and Lu (2008) reported that annular grout vol­ umes ranged between 100 to 150%, indicating that little gap closure occurred and most of the ground losses were associated with face pressure control. Some data from Kaohsiung are above 1% surface volume loss and are not consistent with the general trend seen in the other points. It is possible that these results show a greater than normal contribution from ground losses due to TBM steering and/or less than ideal tail void grouting. In assessing the Kaohsiung data, the methods of Anagnostou and Kovari (1996) were used for the purposes of defining the pressure at (face) collapse and it was assumed that volume losses at tunnel level were approximately equal to volume losses at the ground surface. Limited data associated with EPB tunneling below groundwater levels in Bar­ celona’s sandy soils were discerned from Orfila et al. (2007), Mignini et al. (2008) and Gens et al. (2012) and are also shown in Figures 1 and 2. Shirlaw et al. (2017) documented a case of slurry tunneling in com­ plex weathered rock profile where target pressures were developed using the methods of Golder (2009). Data illustrated in Figures 1 and 2 are for areas where tunneling passed through saturated and completely decomposed sand-like granite. Comparing field and centrifuge test data: • field data show that the effect of saturation is to increase the volume loss for a given value of Load Factor, compared with dry sand, consistent with the interpretation of the model test results; and • field data are typically between the ‘face only’ results and the 2D results, but closer to the ‘face only’ results. Our experience is consistent with the general trends shown in Figure 2. Provided that an adequate face pressure is applied, and there is effective tail void grouting, the measured values of surface volume loss over tunnels in sand are typically very low. However, if the face pressure falls below an appropriate value, even for a very short time period, the result will be rapidly increasing settlement and/ or a sinkhole, depending on the magnitude of the drop in the pressure. For tunnels in saturated sand, the data suggests that maintaining the face pressure so that the load factor to less than about 0.8 is essen­ tial for controlling volume losses at tunnel level to less than 1%, under the assumption that gap grouting is fully controlled.

247

have become increasingly sophisticated, the rigor and reliability of analysis, displacement estimation and planning has lagged.

6 CURRENT STATE OF KNOWLEDGE ­ PRACTICAL CONSEQUENCES The consequences of not applying correct operating pressures during PTBM tunneling are potentially severe: excessive settlement, damage to buildings and utilities, and/or a sinkhole. Occasional sinkholes, areas of large local settlement or loss of ground were recorded over many of the EPB driven tunnels for the North East Line in Singapore (Shirlaw et al., 2003), and in many cases were due to an inadequate face pressure being applied. The authors have observed similar cases of sinkholes over slurry micro-tunneling and large diameter slurry machine tunneling else­ where in Asia and North America. There are many reasons why an inadequate face pressure might be applied including, based on our experience: • no calculations or estimates of the required face pressures were completed prior to tunneling; • errors were made in the calculations for target face pressures; • target face pressure calculations were based on an incorrect geological or hydrogeological model; often far more effort is devoted to assessing the relevant geotechnical parameters for the soil than is devoted to defining the water pressure, which is typically the dominant term in the equations; • available geological and hydrogeological data were dismissed in favour of subjectively chosen and often optimistically interpreted values; • failing to communicate target face pressures to the TBM operator; • failure to monitor and adjust face pressures to keep within the required range; • failure to achieve the target face pressure by, for example, incorrect conditioning, imbalance between screw conveyor rotation, discharge rates and advance rates in EPB tunneling or inappropri­ ate slurry properties for slurry TBMs; • mechanical failure of the TBM, such as excessive wear of the screw conveyor, or the failure of the main bearing seals or pressure bulkhead; • sudden loss of face pressure, due to encountering a well, deep instrumentation or other open path to the ground surface, or flow back into the shaft on launching the TBM; • operational personnel being unconcerned about short term drops in the face pressure; and/or • using an arbitrary basis to develop the allowable range within which face pressures may vary, without assessing the effect on volume loss of tunneling at the lower end of the range. As urban projects are increasingly demanding low tolerances for surface displacement and potential damage to infrastructure and as tunnels get larger and/ or shallower, the real risks to owners, designers, con­ tractors and their engineers become more severe and result in more frequent disputes and litigation. While PTMB pressure control and data collection systems

7 CONCLUSIONS Developing the appropriate target values for face, slurry and grouting pressures is an essential step in successful PTBM tunneling. For the planning, design and construction of tunnels in sand, charts that relate Load Factor to volume loss, such as have been available for clays for over 30 years, could potentially provide a valuable basis for assessing target values for face and grout pressures. Establish­ ing the relationship between pressure and volume loss is also a tool for focusing risk management on the processes that control ground movements at the tunnel level. Recognizing that defining the support pressure at collapse before construction is yet subject to a wide range of opinion (e.g., Vermeer et al. 2002, Kirsch 2009; Iglesia et al. 2013), further centrifuge testing and metanalysis of existing data should prove beneficial. Tests that are representative of tunnel heading geometry and for saturated granular soil conditions, would potentially be of great value. While modelling of complex PTMB construction conditions in saturated sand is challenging, such work may yet lend valuable insights for probabilistic analysis, pre-construction planning, development of risk mitigation strategies and, finally, during con­ struction, permit a more robust approach to produc­ tion quality (displacement) control. Data from parametric centrifuge testing could pro­ vide a relatively simple basis for understanding crit­ ical mechanisms of propagation of ground loss at tunnel level, its surface expression and influences of PTBM pressure controls on surface displacement estimates. In the meantime, Figures 1 and 2 may assist with providing insight into relative influences and magnitudes of face pressures and their effects on ground surface displacement in a form that permits some clarity for design, construction planning, tun­ neling quality control and risk management.

REFERENCES Anagnostu, G. & Kovari, K. 1996. Face stability in slurry and EPB shield tunnelling. Proceedings of the Sympo­ sium on Geotechnical Aspects of Underground Con­ struction in Soft Ground, London, 379–384. Antiga, A. and Chiorboli, M. 2009. Tunnel face stability and settlement control using earth pressure balance shield in cohesionless soil. Geotechnical Aspects of Underground Construction in Soft Ground, Balkema, 365–371. Atkinson, J.H. and Potts, D.M. 1977. Subsidence above shallow tunnels in soft ground. J. of the Geotech. Div., ASCE, 103(4), 307–325. CEDD (Civil Engineering and Development Department) 2014. Ground control for EPB TBM tunnelling, GEO Report No. 298. Hong Kong, Hong Kong SAR: Civil

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Engineering and Development Department, The Govern­ ment of the Hong Kong Special Administrative Region. Chambon, P. and Corte, J.-F. 1994. Shallow Tunnels in Cohesionless Soil: Stability of Tunnel Face. Journal of Geotechnical Engineering, 120(7),1148–1165. Chen, R., Li, J., Kong, L. and Tang, L. 2013. Experimental study on face instability of shield tunnel in sand. Tunnelling and Underground Space Technology, Elsevier, 33, 12–21. Franza, A., Marshall, A. M. & Zhou, B. 2019a. Greenfield tunnelling in sands: the effects of soil density and rela­ tive depth. Géotechnique 69, No. 4, 297–307, doi: 10.1680/jgeot.17.p.091. Franza, A., Marshall, A. M., Zhou, B., Shirlaw, N. and Boone, S. 2019b. Discussion: Greenfield tunnelling in sands: the effects of soil density and relative depth. Géo­ technique, https://doi.org/10.1680/jgeot.19.d.002 Gens, A., DiMariano, A. and Yubero, M.T. 2012. EPB tunnelling in deltaic deposits: Observations of ground move­ ments. Geotechnical Aspects of Underground Construction in Soft Ground, Taylor & Francis, London, 987–993. Giardina, G., DeJong, M. J. & Mair, R. J. (2015). Inter­ action between surface structures and tunnelling in sand: centrifuge and computational modelling. Tunnelling and Underground Space Technol. 50, 465–478. Golder Associates (HK) Ltd. 2009. Ground control for slurry TBM tunnelling, GEO Report No. 249. Hong Kong, Hong Kong SAR: Civil Engineering and Development Department, The Government of the Hong Kong Special Administrative Region. Hsiung, B.B-C. and Lu, K-L. 2008. A Bored Tunnel on Kaohsiung Rapid Transit System, Contract CR2. Journal of GeoEngineering, Taiwan Geotechnical Society, Vol. 3, No. 1, 33–40. Idinger, G., Aklik, P., Wu, W. and Borja, R.I. 2011. Centri­ fuge model test on the face stability of shallow tunnel. Acta Geotechnica (2011) 6:105–117 Iglesia, G.R., Einstein, H.H. and Whitman, R.V. 2013. Investigation of Soil Arching with Centrifuge Tests. Journal of Geotechnical and Geoenvironmental Engin­ eering, ASCE, 04013005-1 – 04013005-13. Kirsch, A. 2009. Experimental investigation of the face sta­ bility of shallow tunnels in sand. Proceedings, World Tunnel Congress 2009, Budapest. Kramer, G., Walters, D., Cording, E., Poon, TYSSE Tunneling Test Section and Excavation beneath Schulich Building. Rapid Excavation and Tunneling Conference, 2015. Kwong, A. K. L., Ng, C. C. W. and Schwob, A. 2019. Con­ trol of settlement and volume loss induced by tunneling under recently reclaimed land. Underground Space, https://doi.org/10.1016/j.undsp.2019.03.005 Lee, C.-J., Chiang, K.-H. and Kuo, C.-M. 2004. Ground Movement and Tunnel Stability when Tunneling in Sandy Ground. Journal of the Chinese Institute of Engineers, Vol. 27(7), 1021–1032. Lee, K. M., Rowe, R.K. and Lo, K.Y. 1992. Subsidence owing to tunnelling. I. Estimating the gap parameter. Canadian Geotechnical Journal, 1992, 29(6): 929–940, https://doi.org/10.1139/t92-104. Loganathan N and Poulos H.G. 1998. “Analytical Solutions to Predict Tunnelling induced Ground Movements”, Journal of Geotechnical and Geoenvironmental Engin­ eering, ASCE. Vol. 124, No. 9, September 1998. Mair, R. J. and Taylor, R. N. (1997). Bored tunnelling in the urban environment. Proceed. 14th international con­ ference on soil mechanics and foundation engineering, volume 4, 2353–2385. Balkema, Hamburg

Mair, R. J., Gunn, M.T and O’Reilly, M.P. 1981. Ground movements around shallow tunnels in soft clay. Pro­ ceedings, Xth International Conference on Soil Mechan­ ics & Foundation Engineering. Balkema, Rotterdam. Vol. 1, pp323–328 Marshall, A. M., Farrell, R., Klar, A. & Mair, R. (2012). Tunnels in sands: the effect of size, depth and volume loss on greenfield displacements. Géotechnique 62, No. 5, 385–399, https:/doi.org/10.1680/geot.10.P.047. Mignini, A., Orfila, T., Colomer, M. and Bertagnolio, F. 2008. Surface settlement minimization in soft soil when excavat­ ing with an earth pressure balance shield. Barcelona metro Line 9 Mas Blau – San Cosme Station. Tunnelling work­ ing procedure. Jornada Tecnica de Tuneles con EPB. Simulacion y Control de la Tuneladora, Barcelona. O’Reilly, M. P. & New, B. M. 1982. Settlements above tun­ nels in the United Kingdom – their magnitude and pre­ diction. In Tunnelling ’82: papers presented at the third international symposium (ed. M. J. Jones), pp. 173–181. London, UK: Institution of Mining and Metallurgy. Orfila, T., Moya, N., and Della Valle, N. 2007. Optimizing Line 9’s EPBM parameters. Tunnels and Tunnelling, April 2007, 40–45. Plekkenpol, J.W., van der Schrier, J.S.and Hergarden, H.J. 2006. Shield tunnelling in saturated sand - face support pressure and soil deformations. In Bezuijen, A. and van Lottum, H. (eds.), Tunnelling: A Decade of Progress, Geo-Delft 1995-2005, Taylor & Francis, London. Ring, B. and Comulada, M. 2018. Practical numerical simulation of the effect of TBM process pressures on soil displacements through 3D shift iteration. Under­ ground Space 3, 297–309 Shewart, W.A. 1986. Statistical Method from the Viewpoint of Quality Control. Edited by W.E. Deming. Dover Publications, Inc. New York. 155 pgs Shirlaw, J.N. 2000. Discussion: Can settlements over tun­ nels be accurately predicted using advanced numerical methods? Proc. Geotechncial Aspects of Underground Construction in Soft Ground, Tokyo. Kusakabe, Fujita, Miyazaki, eds, publ Balkema 471–472 Shirlaw, J.N., Ong, J.C.W. Rosser, H.B., Tan, C.G, Osborne, N.H. and Heslop P.J.E. 2003. Local settle­ ments and sinkholes due to EPB tunnelling. Proc. ICE, Geotechnical Engineering, 56(GE4), 193–211. Shirlaw, J.N., Saw, A.L., Dudouit, F. and Salisbury, D. 2017. Operating Pressures for Twin Slurry TBM Drives, SCL1103, Hong Kong. 16th Australasian Tunnelling Conference 2017, Sydney, Australia. Australasian Tun­ nelling Society Terzaghi, K. 1950. Geologic Aspects of Soft Ground Tun­ nelling, Chapter 2. Applied Sedimentation. P.D. Trask (ed.), John Wiley and Sons, NY. Tsang A.C.M., Salisbury C.D. and Yeung 2012. Confine­ ment pressure for face stability of Tunnel Boring Machine (TBM) tunnel excavation under Hong Kong‟s Western District. Proc. HKIE Geotechnical Division, Geotechnical Aspects of tunnelling for infrastructure development. 2012 May; p. 147–158 Vermeer, P.A., Ruse, N.M. and Marcher, T. 2002. Tunnel heading stability in drained ground. Felsbau, 20(6), 8–18. Vu, M. N., Broere, W. & Bosch, J. 2016. Volume loss in shallow tunnelling. Tunnelling and Underground Space Technology 59, 77–90, doi: 10.1016/j.tust.2016.06.011. Wan, M.S.P., Standing, J.R., Potts D.M. and Burland J.B. 2018. Pore water pressure and total horizontal stress response to EPBM tunnelling in London Clay, Geotechnique, doi: 10.1680/jgeot.17.p.309

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Albert Embankment: Design of deep excavations in the River Thames foreshore O. Brown, P. Stewart, S. Thomson, B. Patel & A.M. Waller AECOM, London, UK

S. Sismondi & F. Quesada FLO, London, UK

ABSTRACT: London’s 150-year-old Victorian sewerage network was originally designed to service a population half the size of its current 8.7 million inhabitants. The system is in good working order but is operating over-capacity. Each year, millions of tonnes of untreated sewage and stormwater is discharged into the River Thames at the overflow points. This poses considerable health risk to river users and the environ­ ment. A £3.8 billion Tideway tunnel is currently underway to capture and divert overflows to be treated before being discharged. At Vauxhall Bridge, an interception chamber on the tidal foreshore of the River Thames directs two existing overflows to the new tunnel via a 50m deep drop shaft. This paper describes numerical modelling carried out for the design of the shaft and field monitoring of the shaft excavation and modelling for the interception chamber. The results provide valuable information for the design of deep exca­ vations in typical London basin strata.

1 INTRODUCTION 1.1

General

London’s 150-year-old Victorian sewerage network includes a series of control points, known as Com­ bined Sewer Overflows (CSOs). The CSOs direct excess flows into the River Thames to prevent flood­ ing of streets and homes when it rains. However, the London sewerage network was designed to serve a population of 4 million people, but the population of London is currently 8.7 million - and this is esti­ mated to rise to 16 million by 2160, Tideway (2020). Consequently, millions of tonnes of excess untreated sewage and stormwater, annually spills into the River Thames via the CSOs. This volume is increas­ ing and is a great health risk to river users and the environment. 1.2

CSOs that currently discharge 18 Mm3 of untreated sewage per year into the river. Once complete, the Tideway tunnel will connect to the Lee Tunnel and transfer sewage for treatment at Beckton Sewage Treatment Works before it is discharged (Figure 1). The works, which are expected to be completed in 2024, will protect the river for at least the next 100 years and create a cleaner and healthier environment.

Tideway super sewer

A new £3.8 billion Tideway Super Sewer is being con­ structed to direct London’s sewer overflows into a new 7.2m diameter and 25km long tunnel beneath the River Thames. The project is currently the largest water infrastructure scheme in London - a plan view of the route across London is shown in Figure 1. The depth of the new tunnel ranges from 30m below ground level (bgl) to 65m bgl. It will inter­ cept and/or control thirty-four of the worst polluting

DOI: 10.1201/9780429321559-32

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1.3

Central section

In September 2015, Ferrovial - Laing O’Rourke Joint Venture (FLO JV) were selected to design and construct the central 12.6km section of the Tideway scheme. This section connects eleven CSOs at eight separate drop shaft sites. FLO JV appointed AECOM as their main designer for the permanent works, which include the main tunnel, shallow and deep connection tunnels, drop shafts, interception chambers, and associated river and civil engineering works. This paper presents details for the Albert Embank­ ment Foreshore shaft (ALBEF) located 100m north­ east of Vauxhall Bridge, as shown in Figure 2. Details are also provided on the geological conditions and parameters as well as selected monitoring data for the shaft and interception chamber. The findings provide valuable information for the design of future intercep­ tion chambers and deep shafts in London.

Figure 1. Plan view of the tideway super sewer.

2 THE ALBERT EMBANKMENT FORESHORE 2.1

Overview

At ALBEF (see Figure 2) a new interception cham­ ber was built to capture sewage and stormwater overflows from two existing CSOs Clapham and Brixton. The discharge points for these two CSOs are located within the embankment walls either side of Vauxhall Bridge on the foreshore of the tidal Thames. The flows are transferred from an Intercep­ tion chamber to a 50m deep drop shaft via a connection culvert tunnel (CCT). At the base of the drop shaft flows pass via a short connection tunnel into the main Tideway tunnel. The new struc­ tures are all located under and within c.100m down­ stream of Vauxhall bridge. The shaft is also located directly in front of Camelford House, leased by Thames Water, the Project Sponsor. Figure 3 shows a schematic of the new shaft and connection struc­ tures. Figure 4 provides a BIM model view of the interception chamber.

Figure 3. Schematic of the ALBEF drop shaft.

2.2

Geological conditions

The ALBEF site is situated within the London Basin on the southern limb of an open syncline. The ground conditions are typical London Basin strata, comprising Made Ground, Alluvium and River Ter­ race Deposits (Kempton Park Gravel) overlying London Clay Formation, Harwich Formation, Lam­ beth Group, Thanet Formation and Chalk. A marine seismic survey and borehole logs reviewed for the desk study indicated the presence

Figure 2. Plan view of ALBEF, Tideway monthly report (2019).

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impact its construction, as well as the interception chamber, connection tunnel and other shallower infrastructure. 2.4

Dewatering of the Lambeth Group sand channels and the lower aquifer (Thanet Formation and Chalk) was required to mitigate the risks associ­ ated with hydraulic failure and inundation of the ground during excavation in or near to confined aquifers under pressure. Dewatering also miti­ gated the higher risks of shaft instability during shaft excavation associated with Sprayed Con­ crete lined (SCL) construction, and concerns for collapse and impacts on operatives and third par­ ties. Groundwater control at the ALBEF site was carried out by WJ Groundwater Limited. The primary dewatering network comprised 15 no. surface wells. Eleven of the wells targeted the Thanet Formation and Chalk in order to reduce the groundwater level in the lower aquifer to 52.8mATD (1m below shaft formation). Four surface wells targeted the Lambeth Group and associated sand channels, the piezo­ metric pressure was reduced from 80mATD to 61mATD. The surface wells were supplemented with 22 in-shaft vacuum well points installed from 63mATD. The aim was to achieve full drawdown of a granular/cohesive layer interface at 60.5mATD and ensuring all pockets were intercepted. This addressed spacing constraints on the surface wells. Figure 5 indicates the well points.

Figure 4. BIM model for the ALBEF interception chamber. Table 1.

Assumed ground profile. Top of stratum

Strata

Made Ground Alluvium River Terrace Deposits London Clay Formation Harwich Formation Lambeth Group Thanet Formation Chalk

m bgl

m ATD

Thickness m

0 0.5

100.03 99.53

0.5 ­ 2.10

2.6

97.43

28.1

30.75 31.35 49.25 58.53

69.28 68.68 50.78 41.50

0.6 17.9 9.28 Not proven

of geological faults, one being close to the shaft. As a precautionary measure, additional marine site investigations were undertaken and spatial mapping of strata levels across the ALBEF shaft footprint during its excavation. The assumed ground profile based on interpret­ ation of historical and Tideway exploratory holes, located adjacent to the shaft is given in Table 1. 2.3

Groundwater control

3 GEOTECHNICAL PARAMETERS 3.1

General

Geotechnical design and FEA modelling parameters, notwithstanding sensitivity ranges, are shown in Table 2.

Hydrogeological conditions

There are two aquifers within the London Basin that are present at the ALBEF site, an upper aquifer and a lower aquifer. The upper aquifer is located in the River Terrace Deposits, and locally in the Made Ground and Alluvium. The lower aquifer is in the Chalk. The lower aquifer is in hydraulic continuity with the Thanet Formation and more permeable units of the lower Lambeth Group. Groundwater is also present in granular units in the upper part of the Lambeth Group, in particular the channel sands, and is termed the ‘Intermediate’ Aqui­ fer. The thickness, lateral variation and interconnec­ tivity of the units in this aquifer varies significantly. The distribution of groundwater is significant to the design of the drop shaft and also has the potential to

Figure 5. In-shaft wellpoints in Lambeth Group.

252

Table 2.

Geotechnical and modelling parameters.

Parameter

Units

Name

Kempton Park Gravel

London Clay Formation

Lambeth Group

Material model Type of material behaviour Dry/Wet soil unit weight Secant stiffness

[–] [–]

Model Type

Hardening Soil Drained

Hardening Soil Undrained (A)

Hardening Soil Undrained (A)

[kN m-3] [kN m-3] [kN m-2]

γunsat γsat Eref 50

19 20 60.0E3

20 21 48.0E3

21 22 76.0E3

[kN m-2]

Eref oed

51.4E3

48.0E3

76.0E3

[kN m-2]

Eref ur

180.0E3

144.0E3

228.0E3

[–]

m

0.1E-3

0.75

0.75

[kN m-2] [°] [–]

cref ’ φ’ νur’

0.1 34.0 0.2

5.0 22.0* 0.2*

5.0 22.0* 0.2*

[–]

K0

0.44

[–]

OCR

1.0

0.7 1.5 3.0

0.7 1.5 4.0

(drained triaxial)

Tangent stiffness (primary ed meter ° ° l ading) ° Loading/reloading stiffness Power (f°r stress-level dependency f stiffness) ° Cohesion Friction angle Poisson’s ratio (unl ad/rel ad) ° ° Lateral At Rest Earth Pressure Overconsolidation Ratio

* Soil model is ‘Undrained A’ and uses effective stress parameters

4 THE ALBEF SHAFT 4.1

inundation of the site. Construction of the shaft is shown in Figure 6. The shaft incorporates tem­ porary and permanent linings constructed using Support Before Excavation (SBE)construction in the top 26m and an Excavation Before Support (EBS) construction in the bottom section, as defined by Faustin et al (2018).

Geometry

The ALBEF circular drop shaft is approximately 50m deep with a minimum internal diameter of 16m. 4.2

Shaft construction

The shaft was constructed from foreshore level of the River Thames within a temporary twin-wall cofferdam in order to provide dry conditions for construction and protection against tidal

4.2.1 SBE shaft construction For the SBE section through the superficial deposits and London Clay, the shaft was supported by 1.3m diameter and 26m long secant piles. The secant piles comprise ‘hard/hard’ secant piles of concrete grade C40/50 with a general spacing of 903mm between piles (66 no. in total). A ‘hard/hard’ secant piled wall was required in order to achieve the required pile depths and to limit ground movements. Only the male piles were reinforced, and a reinforced concrete capping beam was required to tie the pile system together within a requirement for compliance with 1:200 verticality criteria. 4.2.2 SBE excavation Shaft excavation commenced once the secant piled wall and associated capping beam were installed. When excavation reached below the level of the CCT, a temporary working platform was placed and the CCT eye broken out by stitch drilling of the secant piled wall. Following con­ struction of the CCT, shaft excavation continued

Figure 6. The ALBEF shaft.

253

to the toe of the piles together with secondary lining installation. A 5m high shaft heading beam was positioned above the secant piles and partially constructed early to mitigate inundation risks. Both the shaft heading beam and internal shaft lining will be cast-in-place reinforced concrete. 4.2.3 EBS shaft construction For the EBS section, in London Clay, Harwich Forma­ tion and Lambeth Group, the shaft was supported by a sprayed concrete lining (SCL) as excavated. The pri­ mary SCL was followed by installation of a 3m thick reinforced concrete base slab. The short connection tunnel was then excavated wholly within the Lambeth Group to join the shaft to the main TBM tunnel. A combination of a pipe arch and dewatering was used to manage the temporary stability of the excava­ tion and monitoring was installed inside the main tunnel to record convergence (or divergence) of the main tunnel lining. The SCL section comprised 600mm thick steel fibre reinforced concrete (SFRC) of concrete grade C32/40 with a 75mm sacrificial sealing layer. The SCL was constructed in 1m high advances in a quadrants, halves or full circumference excavation sequence. The SCL thickness doubled around the short con­ nection tunnel portal at the bottom of the shaft to withstand the increased lining stresses that occur due to the opening for the tunnel. Furthermore, trad­ itional steel reinforcement was used instead of steel fibre reinforcement in the portal thickening to resist the concentrated flexural and tensile stresses.

Figure 7. FEA 2D axisymmetric model.

5 SHAFT DESIGN (FEA) 5.1

Figure 8. 3D FEA shaft model.

General

A circular shaft is essentially a self-supporting struc­ ture in that the external pressures acting on the shaft are resisted by the compressive ‘hoop force’ in the wall. No extra support is required to balance the external forces and the circular wall is inherently stable provided the hoop force does not exceed the limits of the material properties. Additionally, bend­ ing moments and shear forces will be introduced due to changes in ground and loading conditions with depth. The primary lining is required to resist the short-term ground and water loads and will share the long-term ground loads (consolidation) and other long-term loads with the cast in place secondary lining. Other long-term loads include internal surge pressures, future development and ship impact load­ ing. A two-dimensional axisymmetric finite element analysis (FEA) of the soil-structure interaction (SSI), as in Figure 7, was carried out using PLAXIS soft­ ware to analyse the behaviour of the secant piled and SCL shaft in the short-term and long-term. The corresponding ground pressures from the SSI analysis were then used as input to a three-

dimensional structural-shell model carried out using LUSAS software, as in Figure 8, to calculate the design structural forces for each element. This model also incorporated the openings to the CCT and short connection tunnel. This paper will focus on the SSI analysis using PLAXIS. Verification of ultimate limit states was car­ ried out in accordance with Eurocode Design Approach 1. This was applied within the numerical analysis using the material factor approach for Com­ bination 2, and by applying partial factors to action effects, i.e. factoring the resulting lining forces, for Combination 1. Combination 1 proved to be critical to the lining design. In order to best represent likely soil behaviour, a hardening soil constitutive model was adopted. This elastoplastic model accounts for shear and com­ pression hardening and stress-dependency of the soil stiffness, including an increased stiffness in unload­ ing and reloading. The model provides a more

254

realistic ground response to shaft excavation than simpler constitutive models and allowed for a more refined and therefore economical solution to be developed with reduced volumes of concrete and reinforcement. Reduced groundwater pressures are used to repre­ sent the effects of dewatering and depressurisation, and enhanced material strength is used to represent the effects of any ground treatment. For the London Clay and cohesive units of the Lambeth Group, undrained behavior was modelled with the undrained (A) approach, an effective stress analysis in which the stiffness of water is added to the stiffness matrix to distinguish between effective stresses and (excess) pore pressures. After construc­ tion of the secondary lining, a consolidation phase was used to dissipate excess pore pressures and determine long-term ground loading on the shaft. One drawback to using the hardening soil model is that it does not distinguish between large stiffness at small strains and reduced stiffness at engineering strain levels (Plaxis, 2015). Therefore, input stiffness parameters were specified in accordance with the dominant strain levels expected. A parametric study of the soil strain level was carried out to determine the sensitivity of the design to this parameter and an appropriate level to use in design; this led to adop­ tion of a critical soil strain level of 0.1%. A parametric study of the at-rest coefficient of lateral earth pressure, K0, was also carried out and it was concluded that using the upper bound value of K0 was the critical case for shaft design as this increases lateral pressures on the shaft lining, increasing bend­ ing moments and shear forces. The secant pile wall was characterised using linear elastic two-dimensional plate elements with axial, shear and flexural resistances. SSI was defined using interface elements between the plate elements and the ground. These interface elements have a Mohr Cou­ lomb failure criteria which limits the degree of shear force transmitted. As the walls are con­ structed by the installation of individual piles, it is likely there will be joints between the piles. The presence of joints between the piles can be represented by reducing the bending stiffness in the hoop direction, Zdravkovic et al (2005). This was accommodated in the elastic modulus of the plate element in the hoop direction, where Eh was varied between 10% & 100% of the elastic modulus in the vertical direction, Ev. The case with no reduction in Eh (isotropic stiffness) results in smaller bending, larger hoop stresses and smaller shaft and ground movements. Fur­ thermore, a coefficient of lateral earth pressure, K0, of 1.0 was adopted for the London Clay to account for the effects of secant pile installation. Excavation and spraying of each SCL ring were simulated step by step assuming an advancement rate of approximately 0.5-1m per day. The SCL was also characterised using

linear elastic two-dimensional plate elements which were assigned an isotropic stiffness, updated to reflect the time-dependent behaviour of the SCL, Chang & Stille (1993). The structural forces in the SCL are highly dependent on the excavation sequence and the stiffness of the SCL and this method takes full advantage of the arch effect in the ground making the soil a constitutive part of the support, reducing the thickness of the lining to a minimum, Dias et al (2014) to provide a more economical design. Accompanying the analysis of the shaft, the design also considered the influence of the shaft on the neighbouring piled foundations for Camelford House, a multi-storey building adjacent to the ALBEF drop shaft site. The influence of these foundations on the shaft was also con­ sidered by predicting the load taken down through the building piles based on available asbuilt information and assuming this as a surface load applied at the toe level for of the building piles, which is closer to the level for SCL shaft construction.

255

6 SHAFT MONITORING 6.1

Instrumentation

Instrumentation and monitoring of the shaft were carried out to validate the design process and to con­ trol the construction works. For the piled section of the drop shaft, this comprises inclinometers within 4 no. piles to monitor lateral movement and optical equipment on the capping beam to monitor vertical movement. For the SCL section of the drop shaft, it comprises monitoring bolts to monitor convergence. A system of trigger limits and zones for the monitoring system aim to either confirm that the design predictions and assumptions are satisfied on-site or, otherwise, enable the timely introduc­ tion of preventative/contingency support meas­ ures as required. The system employed trigger values (mm) of Clear (no problem), Green (0.75 - early warning of potential problem), Amber (1.0 - design predictions exceeded), Red (1.25) and Black (1.5 - stop works as necessary). 6.2

Data analysis

Figures 9 and 10 show a comparison of the monitored and predicted movements for the piled and SCL shaft elements respectively. On Figure 10 movements correspond to the stage at which the base slab is cast, prior to breakout for the short connection tunnel construction. 6.3

Discussion

For the secant piles, and from the sensitivity ana­ lyses there are two design predictions which

between the piles may have limited impact on the axial stiffness in the hoop direction, possibly related to high joint-quality with minimal gaps between piles. For the SCL, all recorded move­ ments sit within design predictions, albeit the monitored values for SCL may be lower due to a difference between actual and assumed param­ eters such as the SCL stiffness, excavation sequence (this varied between quadrants and full ring) and timing of monitoring installation. 7 THE ALBEF INTERCEPTION CHAMBER 7.1

Geometry

The ALBEF interception chamber (IC) measures 19.5m by 11.4m internally at the widest point, with two different levels, the deepest almost 20m below ground and the smaller, shallower area, around 11m. 7.2

Interception chamber construction

IC construction is from foreshore level within a temporary cofferdam to provide dry conditions and protection against tidal inundation. The IC comprises a 750mm diameter secant piled box with a capping beam. Internally the IC incorporates a series of cas­ cade slabs to direct flows down and into the CCT. During design several construction methods were assessed. The adopted solution comprises topdown construction, with permanent concrete waling beams cast into the piled walls as the excavation des­ cends the design mitigates against collapse risk of a 17 m deep excavation protected only by a single skin cofferdam installed 1 m from the works. The IC tem­ porary shoring solution by FLO, uses a combination of pre-stressed hydraulic props at capping beam level and below this permanent reinforced concrete waling beams. Figure 11 indicates the pile layout, upper props and inclinometer positions. Figure 12 indicates the initial construction stages. In the early design rectangular steel diaphragms were incorpor­ ated at each cascade slab level, and later as design

Figure 9. Piled shaft displacements.

Figure 10. SCL shaft displacements.

correspond to the two cases of elastic modulus considered, i.e. Eh=Ev and Eh=0.1Ev. Recorded movements generally sit between the two predic­ tions but more closely match the isotropic stiff­ ness condition (Eh=Ev). This suggests the joints

Figure 11. IC pile layout, inclinometer and support positions.

256

from the Tideway project and from comparison with published data, Hight et al (2007) from other central London projects. There are differ­ ences in permeability, both horizontal and verti­ cal for the London Clay A3 and A2 sub­ divisions present at ALBEF. Both contain thin horizontal partings of coarse silt and very fine sand, interlaminated with a silty clay matrix, which reduces the vertical permeability. From data reviewed a design ratio of horizontal to vertical permeability of 5 was adopted. Sensitiv­ ity analyses were then carried out to assess both longer construction periods and changes in key parameters such as the London Clay permeability. Applied loads were modelled in the analyses, with a surface construction surcharge load of 60kN/m2 applied throughout the construction stages around the IC, beyond a 3m exclusion zone. SLS and ULS-C1 and C2 design load cases were considered, and also, a ULS acciden­ tal over-dig of 0.5m before construction of the upper and lower chamber slabs. Shear forces and bending moments were analysed for the dif­ ferent design cases and sensitivity runs. With the refinements in construction sequence and sup­ port, acceptable structural forces were obtained for the perimeter walls and support system.

Figure 12. IC upper level excavation and temporary support.

7.4 Figure 13. IC FEA modelling section A-A’.

developed most were replaced with solid corners and permanent reinforced concrete struts. 7.3

Interception chamber design (2D FEA)ii

For design of the IC secant piled wall and proposed support systems, three Plaxis 2D models were devel­ oped to provide representative cross-sections of the interception chambers. Figure 13 shows Sec­ tion A-A’. As shown support elements are modelled as either node-to-node or fixed-end anchors. Secant pile walls and base slabs are modelled as plate elements. Interface elements are used for soil structure interaction. The construction sequence in the Plaxis 2D models is aligned with the construction programme activities and durations. Calculation type, i.e. plastic/undrained or coupled/time­ dependent, uses the construction timescales. In total 128 days were allowed to excavate and construct the lower chamber base slab, and at this stage fully drained conditions were assessed. The permeability of the London Clay has been assessed using site specific and route-wide data

Interception chamber monitoring

Monitoring at ALBEF IC involves multiple tech­ niques, both for structure movements and forces and also for the surrounding areas, including the river walls, Vauxhall Bridge, adjacent buildings and public spaces. These include inclinometers in the IC secant piles and strain gauges on the props. Results from inclinometer IC05, located in the mid-point of the deeper chamber piles, are potted on Figure 14. On 9th January 2020 the IC was at 9.5m depth, and movements were all within desig­ nated trigger levels.

Figure 14. IC Results from pile inclinometer 05.

257

7.5

ACKNOWLEDGEMENTS

Discussion

The IC chamber design, incorporating top-down construction and structural support elements at each cascade slab level, delivered benefits as follows: - Increase in the structural capacity of the permanent concrete waling beam at each cascade slab - Increase in the overall support system stiffness, reducing predicted wall deflections, and associated ground movements and their wider effects - Adequate support to resist residual Accidental Ship­ ping Impact forces during construction once the dissipation effects of a barge are considered, and - More realistic consolidation timescales through adoption of anisotropic hydraulic properties of the London Clay. 8 SUMMARY/CONCLUSIONS Design of the drop shaft posed several challenges including dealing with challenging ground and groundwater conditions to allow programme-efficient piled and SCL methods to be used safely and success­ fully. The design methodology and complex SSI ana­ lysis for the shaft used staged construction and an advanced soil model which helped predict more real­ istic shaft behavior and enabled several design improvements. These include taking full advantage of the arch effect and unloading stiffness in the ground which in turn reduced the thickness of the lining, resulting in a more economic design and reduced con­ struction time. Design of the IC posed ground condi­ tion challenges, including accurate prediction of movements, forces and consolidation timescales. The adoption of London Clay permeability anisotropy enabled an efficient design to be undertaken.

The authors wish to thank Tideway and FLO JV for making information available to use within this paper and their input during the design process.

REFERENCES British Standards Institution. (2004). BS EN 1997-1:2004 +A1:2013 Eurocode 7: Geotechnical Design – Part 1: General rules. London: British Standards Institution. British Standards Institution. (2010). BS EN 1536: 2010 Execution of special geotechnical works – Bored piles. London: British Standards Institution. Chang, Y. and Stille, H. (1993). Influence of early-age proper­ ties of shotcrete on tunnel construction sequences. Shot­ crete for Underground Support VI, ASCE: 110–117. Dias C.C., Hirata F.P. and Kuwajima F.M. (2014). Settle­ ments due to the Excavation of Shafts in São Paulo. Proceedings of the World Tunnel Congress 2014 – Tunnels for a better Life. Foz do Iguaçu, Brazil. Faustin N.E., Elshafie M.Z.E.B & Mair R.J. (2018). Case studies of circular shaft construction in London. Pro­ ceedings of the Institution of Civil Engineers – Geotechnical Engineering Vol 171:5, 391–404 Gaba, A., Hardy, S., Doughty, L., Powrie, W., Selemetas, D. (2017). CIRIA C760 Guidance on embedded retaining wall design. London: CIRIA. Hight, D.W., Gasparre, A., Nishimura, S., Jardine, R.J., and Coop, M.J. (2007). Characteristics of the London Clay from the Terminal 5 Site at Heathrow Airport. Geotech­ nique 57, (1): 3–18. Thames Tideway (2020). URL: https://www.tideway. london/supersewer/Date accessed: 24th January 2020 Tomlinson, M. and Woodward, J. (2008). Pile Design and Construction Practice, 5th edition. London: Taylor & Francis. Zdravkovic, L., Potts, D.M. & St John, H.D. (2005). Mod­ elling of a 3D excavation in finite element analysis. Geotechnique 55 (7): 497–513.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Centrifuge and numerical modelling of the influence of structural stiffness on basement heave in over-consolidated clay D.Y.K. Chan & S.P.G. Madabhushi Department of Engineering, University of Cambridge, Cambridge, UK

ABSTRACT: Long-term basement heave is a pertinent problem in the construction of underground spaces in over-consolidated clay strata, notably the clays of southeastern England. When a basement is constructed, the permanent removal of soil above formation level inevitably causes upward movement of the remaining soil, or heave pressures on the base slab. In over-consolidated clay strata such as London Clay, this process of heave continues over many years after the basements structural completion. The designer must predict these future heave pressures and movements when designing the basement structure. However, there has been much conservatism in the methods of design due to the scarcity of site monitoring data to calibrate the methods of heave prediction. There is a need for further physical data to improve the methods of design. This paper presents a research project that seeks to fill this research gap through the technique of geotech­ nical centrifuge modelling. The model simulates the construction and long-term heave of a 15 m deep base­ ment underlain by stiff clay. The model basement is fitted with extensive instrumentation and this is the first research project to provide simultaneous measurements of the vertical movement of the base slab and the distribution of slab-soil contact pressure in a centrifuge model of basement heave. The experimental results are validated against finite element simulations of the same prototypes. The results show that the prediction of high heave pressures is a self-fulfilling prophecy: the assumption of high heave pressures by the engineer leads to the specification of strong structures to sustain the load, and the stiffness of these structures in turn restrain the soil, generating high heave pressures. This paper advocates an alternative design approach so that leaner basement structures can be specified, and thus urban underground spaces can be provided more economically.

1 INTRODUCTION When a basement structure is excavated, the perman­ ent removal of soil overburden leads to a reduction in vertical effective stress, causing the remaining soil to swell. In over-consolidated clay strata, such as London Clay and Gault Clay, this process of swelling continues after the completion of the basement struc­ ture, generating upward displacement and heave pres­ sures on the base slab as the clay re-consolidates. This process is known as “long-term heave” and engineers must design the base slab to restrain or allow for these gradual movements and pressure changes that often continue for over a decade beyond structural completion (Chan et al. 2018). One recent example of this issue was the Liverpool Street Crossrail site in London (Figure 1), where uncertainties in predictions of long-term heave effects on the station concourse basement had neces­ sitated much conservatism in the substructure design. Some designers seek to model the effect of long­ term heave as an upward load at formation level,

typically quoting heave pressures of 50% to 65% of the pre-existing effective overburden. There is much uncertainty surrounding these estimates and the soilstructure interaction that generates the heave pres­ sures and movements, so there is a desire for further research to fill this gap. This paper presents a research project that investi­ gates the phenomenon of long-term heave, using both geotechnical centrifuge modelling and finite element simulations to quantify the effects of long­ term heave on a rectangular basement underlain by over-consolidated clay. 2 CENTRIFUGE MODEL A 100 g geotechnical centrifuge model was used to simulate a 15 m deep basement (model scale: 150 mm) with plan area 30 m x 15 m (model: 300 mm x 150 mm). Each centrifuge test involved two layers of soil: a 16 m (model: 160 mm) layer of Speswhite kaolin at the bottom, and a 15 m (model: 150 mm) layer of dry Hostun sand on top (Figure 2).

DOI: 10.1201/9780429321559-33

259

took for the clay to consolidate to equilibrium during an experiment and models the common construction practice of using an under-slab drainage layer to minimise buoyancy loads. The clay was pre-consolidated to a maximum vertical stress of 800 kPa. In-flight cone pene­ trometer tests established that the undrained shear strength of the clay was about 100 kPa, consist­ ent with Vardanega et al. (2012). The Hostun sand was poured into the centrifuge model using an automatic sand pourer, at a dry density of 1600 kg/m3 . The properties of the model soils are given in Table 1 and their representative stiff­ ness parameters are given in Table 4. Two different basement models were used in the experiments to investigate the influence of basement slab and wall stiffness on heave behaviour. The stiff basement model was made from 3.25 mm thick stainless steel plates, which matched the bending stiffness of 1 m thick reinforced concrete retaining walls and base slabs (Table 2). The flexible basement used 1.22 mm thick brass plates to create an Figure 1. Photograph of liverpool street crossrail construc­ tion site. Table 1.

The water table of the centrifuge package was set at the level of the sand-clay interface and a thin layer of sand separated the base slab and the clay to ensure adequate drainage at formation level. The use of these sand layers reduced the amount of time it

Properties of soils used in centrifuge model.

Dry density of sand Saturated density of clay Swelling index of clay (CS ) Pre-consolidation stress of clay

Figure 2. Cross-section drawing of centrifuge model.

260

1600 kg/m3 1750 kg/m3 0.1155 800 kPa

Table 2.

3 FINITE ELEMENTS MODEL

Properties of stiff basement model.

Basement footprint (model scale) Basement footprint (prototype) Wall and slab stiffness (model) Wall and slab stiffness (prototype)

150 mm x 300 mm 15 m x 30 m 533 Nm2 /m 533 MNm2 /m

arbitrarily flexible structure which would allow large heave movements (Table 3). Before spin-up, the basement cavity was filled with a heavy fluid (sodium polytungstate solu­ tion) of the same density as the dry sand outside the basement. This heavy fluid was removed from the basement cavity during centrifuge flight by draining it through a network of pipes and valves, to model the effect of excavation. Subse­ quently, an electrical actuator would apply a vertical load onto the top of the basement walls to model the construction of a three-storey under­ ground building. The centrifuge model included instrumentation to record pore pressures, vertical displacements, structural bending curvatures, prop forces, and slab-soil contact pressures. In particu­ lar, the use of a Tekscan tactile sensing mat to measure the profile of slab-soil contact pressure (Figure 3) meant that this research project is the first centrifuge modelling study to provide simul­ taneous measurements of the profiles of heave displacement and swell pressures. Further details about the preparation and the oper­ ation of the centrifuge model, and the validation of experimental results against site data, can be found in Chan and Madabhushi (2018) and Chan et al. (2019). Table 3.

Properties of flexible basement model.

Basement footprint (model scale) Basement footprint (prototype) Wall and slab stiffness (model) Wall and slab stiffness (prototype)

The finite element models in this research project were performed using PLAXIS 2D 2017-01. The plane of symmetry of the basement was taken as the representative cross-section, giving plane-strain models of 85 m overall width with a 15 m wide, 15 m deep basement cavity. The over-consolidated clay was represented by a small-strain hardening (HSS) model (Obrzud & Truty 2011) to capture the non-linear stiffness of the clay, and the constitutive parameters were calibrated using triaxial test data from Vardanega et al. (2012) and one-dimensional compression data from the preparation process of the clay samples used in the experiments reported in this paper. The sand was represented by a MohrCoulomb model with parameters obtained from the experiments reported in Heron (2013) and Deng & Haigh (2018). Table 4 summarises the constitutive parameters used in the finite element simulations. The software package and soil constitutive models were chosen to match current practices in industry where finite element models of basement heave in over-consolidated clays are needed. To match the conditions in the centrifuge experi­ ments, the left and right boundaries of each finite elements model were normally fixed and imperme­ able, while the bottom boundary was vertically fixed and fully permeable. Each basement model was repre­ sented by linear-elastic plate elements whose stiffness matched the prototype of the corresponding centrifuge test (Tables 2 and 3). Fully coupled consolidation analyses were performed, with undrained behaviour derived from effective stress parameters and excess pore pressures. Figure 4 shows the finite elements

Table 4.

150 mm x 300 mm 15 m x 30 m 14 Nm2 /m 14 MNm2 /m

Figure 3. Plan view of basement model, showing locations of instrumentation.

Properties of soils used in finite elements model.

Model

Hostun sand Mohr-Coulomb

Speswhite kaolin HSS

Dry density Saturated density einit Permeability (m/s)

1600 kg/m3 2000 kg/m3 0.64 1 x 10-7

1750 kg/m3 1.00 5 x 10-10

Poisson ratio l E (kPa) G0 (kPa) γref Eur (kPa) E50 (kPa) Eoed (kPa) pref (kPa) m

0.20 47500 -

0.12 45000 2.5 x 10-4 16800 5600 4800 250 0.65

¢0 ψ c’ref (kPa) Rf

33� 20� 1 0.9

20� 0� 0 0.8

261

Figure 4. Finite element model of stiff basement, showing excess pore pressures immediately after excavation.

model for the stiff basement excavation, with con­ tours of excess pore pressures caused by excavation. Each simulation was initialised with greenfield conditions. Then, the basement model was added and the soil inside the basement cavity was replaced by line loads representing the hydrostatic loads of the heavy fluid in the centrifuge experiments. Excess pore pressures were then allowed to dissi­ pate and displacements were zeroed after the simu­ lation had reached equilibrium. The hydrostatic loads were turned off to simu­ late excavation, and then point loads were imposed onto the top of the walls to simulate construction. The same prototype lengths of time were used for the finite element models as the centrifuge models, to capture the concurrent effects of construction and consolidation.

4 RESULTS AND DISCUSSION The aim of the centrifuge and numerical models pre­ sented in this paper is to measure the magnitude of heave displacement and swell pressures caused by long-term heave, so Figure 5 and 6 will focus on the equilibrium slab-soil contact pressures and the total displacements caused by excavation, construction, and consolidation. The differences in behaviour between the flexible basement models and the stiff basement models will be highlighted, and the goodness of fit between numerical data and experimental data will be discussed. The profiles of equilibrium slab-soil contact pres­ sures along the centre-line of the base slab are shown on Figure 5. There was good agreement between the experimental results and finite element

Figure 5. Graph of slab-soil contact pressures in long-term equilibrium after excavation and construction.

262

Figure 6. Graph of heave displacement against swell pressure in long-term equilibrium after excavation and construction.

results in terms of the profiles of pressure, giving confidence to the validity of the numerical model. The pressure profiles of the both the stiff base­ ment slab and the flexible basement slab showed concentrations of pressure near the toes of the walls and relaxations of pressure towards the centre of the slab, compared to the pressures observed before excavation (about 240 kPa). However, the effect of relaxation was much more profound in the case of the flexible slab, with almost complete relaxation of pressure at the centre. In contrast, the centre of the stiff slab attracted about 150 kPa of swell pressure, which agrees with the oft-quoted design rule of thumb that 50 - 65% of the pre-existing effective overburden would manifest itself as a long-term heave load after the construction of a basement of typical stiffness in over-consolidated clay. The difference between the stiff basement and the flexible basement’s behaviours can be explained by soil-structure interaction. This is best illustrated using a plot of vertical displacement versus slab-soil contact pressure for various positions along the centre-lines of the basement slabs (Figure 6). The low stiffness of the flexible base slab allowed it to undergo some 200 mm of differential heave, let­ ting the underlying clay expand and relax. In con­ trast, the high stiffness of the stiff base slab restrained the clay, permitting little upward move­ ment and therefore attracting large swell pressures. For the stiff basement, the finite elements model appeared to over-predict the magnitude of heave by about 30 mm compared to the experimental results. On closer inspection, this discrepancy arose from an over-estimation of the undrained stiffness of the clay when the building load was imposed onto the base­ ment structure. Otherwise, there was good agreement between experimental and numerical data in terms of the changes in vertical displacement between short-

term and long-term conditions. Further research will be needed to refine the constitutive model so that the short-term interactions between stiff slabs and over-consolidated clays can be modelled more accurately. 5 IMPLICATIONS FOR BASE SLAB DESIGN The two basement structures presented in this paper had the same overall dimensions and were subject to the same soil conditions and construction sequence. They only differed in terms of structural stiffness and self-weight. Nevertheless, they exhibited very different heave pressures and movements. The stiff basement suppressed post-construction heave movement at the expense of generating high heave pressures. The flex­ ible basement allowed drastic relaxation of swell pres­ sures at the expense of permitting large heave movements. The magnitudes of swell pressures and heave displacements depend strongly on the stiffness of the basement slab. The practical implication of this finding is that the prediction of high heave pressures is a self-fulfilling prophecy that should be avoided in design. If a design engineer attempted to predict heave pres­ sures before specifying the sub-structure, the assumption of high heave pressures would lead to the specification of stiff and heavy slabs to carry the assumed loads (Figure 7). The high stiffness of such slabs would then constrain the vertical movement of the clay and generate high swell pressures. It would be preferable to specify a permissible movement limit based on serviceability requirements, then design a slab using soil-structure interaction methods to provide the appropriate flexibility to accommodate the specified displacement at the expected swell pressures (Figure 8).

263

Figure 9. Illustration of serviceability-based heave design using relaxation ratio method.

Figure 7. Flowchart showing the “self-fulfilling prophecy’’ of large heave load predictions in design.

2. Set a movement limit based on serviceability requirements (δ1 þ δ2 þ δ3 ) and mark the amount of heave that is predicted to occur before the con­ struction of the slab (δ1 ) and that due to buoy­ ancy (δ2 ); 3. From the intersection between the movement limit line and the soil curve, draw a straight line that represents the stiffness of the struc­ ture. The gradient of this line is the required structural stiffness. Design a slab with this stiffness. 4. Verify that the slab can carry the estimated heave pressure that is associated with the chosen move­ ment limit. Simpson (2018) demonstrated that, as long as there is a net reduction of soil stiffness due to clay swelling, the actual heave pressure and movement will not exceed this methods predictions. Furthermore, it will be greatly beneficial for the calibration of soilstructure interaction design methods if more basement construction sites in over-consolidated clay can be monitored to obtain data of actual heave pressure. 6 CONCLUSIONS

Figure 8. Flowchart showing the serviceability-driven approach of design.

preferable,

One common used soil-structure interaction method is the “non-FE method’’, sometimes known as the “relaxation ratio method’’ (Chan & Madab­ hushi 2017; Simpson 2018). The procedure for ser­ viceability design with the relaxation ratio method is illustrated in Figure 9: 1. Plot a soil curve to represent the non-linear stiff­ ness of the clay stratum;

This research project used both centrifuge modelling and numerical modelling to shed light on the influence of basement slab stiffness on the phenomenon of long­ term heave. It is the first published study to provide simultaneous measurements of the vertical movement of the base slab and the distribution of slab-soil contact pressure in a centrifuge model of basement heave. For both the stiff basement and the flexible base­ ment, there was good agreement between the two methods of investigation in terms of their estima­ tions of slab-soil contact pressure. Further research is required to refine the finite element model’s pre­ dictions of short-term displacements. The results show that designers should avoid making guesses of the heave pressure before

264

designing the sub-structure, because the heave pres­ sure is strongly dependent on the basement slab’s stiffness. It would be preferable to decide on a serviceability limit first and then use soil-structure interaction methods to design a suitable sub-structure. Site monitoring data of actual heave pressure will be beneficial for the calibration of these design methods.

ACKNOWLEDGEMENTS The authors would like to thank the EPSRC Centre for Doctoral Training in Future Infrastruc­ ture and Built Environment at the University of Cambridge (EPSRC grant reference number EP/ L016095/1) and Mott MacDonald Geotechnics for supporting this research project.

REFERENCES Chan, D. Y. K. & S. P. G. Madabhushi (2017, September). Designing urban deep basements in south east england for future ground movement: Progress and opportunities for experimental simulation of long-term heave. In Pro­ ceedings of the International Symposia for Next Gener­ ation Infrastructure. Chan, D. Y. K. & S. P. G. Madabhushi (2018, July). Centri­ fuge simulation of heave behaviour of deep basement slabs in over-consolidated clay. In Proceedings of the 9th International Conference on Physical Modelling in Geotechnics, London. Taylor & Francis.

Chan, D. Y. K., S. P. G. Madabhushi, Y. S. Hsu, A. S. O’Brien, S. A. Solera, & M. Williamson (2019, September). Experimental study of structural move­ ments and swelling pressures on deep basements caused by long-term heave in over-consolidated clay. In Proceedings of the XVII European Conference on Soil Mechanics and Geotechnical Engineering, Reykjavik. Chan, D. Y. K., S. P. G. Madabhushi, D. P. Nicholson, T. J. P. Chapman, & S. A. Solera (2018, November). Twenty-one years of heave monitoring in London Clay at Horseferry Road basement. Ground Engineering 51 (11), 28–33. Deng, C. & S. K. Haigh (2018, July). Soil movement mobilised with retaining wall rotation in loose sand. In Proceedings of the 9th International Conference on Physical Modelling in Geotechnics, London. Taylor & Francis. Heron, C. M. (2013, August). The dynamic soil structure interaction of shallow foundations on dry sand beds. PhD thesis, University of Cambridge. Obrzud, R. & A. Truty (2011, March). The hardening soil model - a practical guidebook. Technical Report 100701, Zace Services Ltd, Lausanne, Switzerland. Simpson, B. (2018, February). Effective heave pressures beneath restrained basement slabs. Proceedings of the Institution of Civil Engineers - Geotechnical Engineering 171(1), 28–36. Vardanega, P. J., B. H. Lau, S. Y. Lam, S. K. Haigh, S. P. G. Madabhushi, & M. D. Bolton (2012). Laboratory measurement of strength mobilisation in kaolin: Link to stress history. Geotechnique Letters 21), 9–15.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Influence of TBM geometry on lining loads of deep tunnels

V. De Gori Geotechnical Design Group, Rome, Italy

A. de Lillis & S. Miliziano Department of Structural and Geotechnical Engineering, Sapienza University of Rome, Rome, Italy

ABSTRACT: The paper presents a numerical study aimed at investigating the influence of the geometrical design of Tunnel Boring Machines (TBMs) on the forces acting on the lining. The study was conducted using a 3D numerical model that accounts for the major features of the excavation such as front pressure, overcut, shield conicity, tail void grouting, grout hardening over time and installation of the lining. To accurately reproduce the soil-shield-lining interaction, a simple procedure was adopted to overcome the modelling inac­ curacies associated with the misidentification of the excavation profile due to the development of pre­ convergences ahead of the excavation front. A parametric study was conducted to evaluate the influence of the shield’s conicity and length on the lining forces. The results remark the great influence of the analysed parameters and show that the developed model can provide novel and significant insights on the interaction process.

1 INTRODUCTION The structural design of tunnel linings must account for the stress release in the soil induced by the exca­ vation and for the overall soil-shield-lining inter­ action process. This is particularly true for deep tunnels, in which the lining is subjected to high loads and high convergences are usually induced by design to reduce the stress state and minimize the shield’s jamming risk. Given the complexity of the soil behaviour and excavation process, empirical and analytical methods cannot provide solutions as accurate as those result­ ing from numerical modelling. 2D plane-strain numerical analyses though, suffer a series of short­ comings mainly related to the intrinsic threedimensionality of the excavation process (Karakus, 2007). 3D modelling, on the other hand, allows to simulate realistically the main features of mechan­ ized tunnelling and can yield satisfactory results (Kasper & Meschke, 2004; Litsas et al., 2018; Losacco & Viggiani, 2019; Miliziano & de Lillis, 2019). Although recent improvements in computational power have made 3D modelling significantly more affordable, the time and skills required to develop and fine-tune the models and the actual runtime can still bottleneck the design process. Among the main factors to be accounted for, the following require special consideration: i) the face pressure applied by the front of the TBM (and by the muck inside

DOI: 10.1201/9780429321559-34

266

the excavation chamber in case of Earth Pressure Balance TBMs); ii) the geometry of the TBM, including the overcut and the conicity of the shield, which are still often overlooked; iii) the annular void behind the tail of the machine; iv) the grouting of the tail void, its mechanical properties and their evolution over time. The structural system of the segmental lining has been studied by several authors with somewhat mixed results (e.g. Arnau & Molins, 2012; Do et al., 2014). Kavvadas et al. (2017) proposed a comparison of lining forces obtained simulating the lining in three different ways (i.e. continuous shell, shells with aligned joints and shells with staggered joints) and found the resulting differences to be quite small. This study focuses on the influence of the geomet­ rical design of the TBM on the soil-shield-lining interaction process and, ultimately, on the lining forces. Specifically, an advanced 3D numerical model was developed accounting for the main factors influ­ encing the problem and used to investigate the influ­ ence of the shield’s conicity and length. The model is founded on a very accurate geometric and mechanical representation and the excavation is simulated in detail. The model, which was tested satisfactorily against monitoring data by De Gori et al. (2019), is fine-tuned for deep tunnelling in soft ground. Also, the model adopts a simple numerical procedure, pro­ posed by de Lillis et al. (2018), to overcome modelling issues associated with the misidentification of the actual excavation boundary induced by the

development of pre-convergences (radial soil dis­ placements ahead of the excavation front). The analyses are performed in effective stresses and, assuming clayey soils, the tunnel excavation is simulated in undrained conditions. Once a stationary solution is reached, drained conditions are simulated. The effects of gravity (and also the weight of the shield and the lining) are neglected. In the following, after a brief description of the numerical model and the mesh design technique, the main results are illustrated and discussed, focussing on the decisive influence of the soil-shield inter­ action on the forces arising in the lining. 2 NUMERICAL MODEL

Figure 2. Numerical model.

The numerical model, developed with the finite differ­ ence code FLAC3D, simulates the main factors affect­ ing the soil-shield-lining interaction, namely: front pressure at the excavation face, overcut, conicity (tapering) of the shield, annular tail void behind the shield, void grouting and grout hardening over time, installation of the lining and application of the jacks thrust (Figure 1). The model is fine-tuned for deep mechanized tunnels; as such, the effects of gravity can be neglected and, taking advantage of the resulting plane of symmetry, just one quarter of the problem can be modelled. The mesh size and density (Figure 2) were thor­ oughly checked to ensure negligible boundary effects and a proper accuracy of the simulation of the TBM’s geometry and the interaction process, especially close to the tunnel. Furthermore, the model works in largestrain mode, thus, the position of the grid nodes is updated at the end of each calculation step. This enhances the simulation of the contacts between moving grids due to lower geometrical tolerances. TBMs for deep tunnels are usually designed to min­ imise the risk of jamming and the lining forces. For this reason, the overcut and the conicity are wilfully increased to magnify the steering gap and the stress relaxation. Clearly, in such cases, higher care must be placed in the accurate identification of the excavation boundary. The shield of the TBM is simulated via very stiff elastic continuum elements, whose geometrical

properties are reported in Table 1 together with those of the lining. The lining is also simulated using elastic continuum elements (Augarde & Burd, 2001), with a Young’s modulus of 37.2 GPa, neglecting the pres­ ence of joints. Both the lining and the shield are assumed to be weightless. On the external surface of the shield, an elastic interface is applied. The stiffness of the interface was calibrated checking the system’s behaviour during contacts and ensuring the absence of overlapping between soil and shield grid points. The forces exerted by the hydraulic jacks on the lining are simulated as a uniform longitudinal pressure. The soil mechanical behaviour is described adopt­ ing the CYsoil model; an elastic-plastic constitutive model with hardening, both deviatoric and volumet­ ric, characterized by an elliptic volumetric cap and a frictional Mohr-Coulomb shear envelope (Itasca, 2012). It is a model able to describe with sufficient accuracy the non-linearity of the soil behaviour. The main soil parameters are listed in Table 2 where, with reference to effective stresses, φ is the friction angle, c the cohesion, υ the Poisson’s ratio, ψ the dilatancy angle, E°ref the reference Young’s modu­ lus, pref the reference mean pressure, Rf the failure ratio and β a calibration factor. The initial stress state is: total vertical stress σv = 1800 kPa, total horizontal

Table 1.

TBM and lining geometry.

Cutterhead radius (m)

Shield front radius (m)

tail radius (m)

length (m)

Lining extrados radius (m)

2.5

2.48

2.43

8

2.38

Table 2.

Main soil parameters.

φ (°) c (kPa) υ (-) ψ (°) E0ref (MPa) pref (kPa) Rf (-) β (-) 22

Figure 1. Main features of the simulation scheme.

267

0

0.3

0

15

100

0.85

5

stress σh = 1300 kPa, pore pressure u = 500 kPa. The soil is normally consolidated and the at-rest coeffi­ cient of lateral earth pressure K0 is 0.615, adopting Jaky’s law. At this stage of the study, the soil is assumed to be able to withstand any value of suction. The excavation is simulated adopting a step-by­ step approach, which involves the following substeps: 1) the excavation advances one ring (1.2 m) and the corresponding soil slice is removed; 2) the shield moves forward and applies the front pressure to the new excavation face; 3) the lining ring installed in the previous phase is now outside the shield; a new ring is generated and the jacks thrust is applied; 4) the algorithm reads the current position of the soil and the tail void is injected with grout; previ­ ously injected grout is hardened following the law proposed by Kasper & Meschke (2006) and assuming an advancement rate of the excavation of 12 m/day. The analyses are carried out in effective stresses and the excavation is simulated in undrained condi­ tions, imposing total volumetric deformations equal to zero everywhere in the domain, assuming a lowpermeability fine-grained soil and thus a negligible dissipation of excess pore pressure in the surround­ ing soil. Upon reaching a stationary solution (about 4 diameters, D, behind the tail), long-term drained conditions are simulated re-imposing the initial pore pressure in the entire calculation domain. 2.1

Mesh design

Most numerical models adopt a mesh initially designed to have grid nodes located on the excava­ tion boundary. This allows to instruct the excavation algorithm to remove the slice of soil inside those nodes once the TBM passes through a generic sec­ tion. Since the stress release induced by the excava­ tion propagates beyond the tunnel face, though, pre­ convergences (radial displacements ahead of the excavation front) develop and the above-mentioned grid nodes move inside the excavation boundary

(Figure 3a). This means that the algorithm will “excavate” a portion of soil which is smaller than the actual excavation, dictated by the diameter of the cutting wheel. Furthermore, this entails that the subsequent interaction process will start from a configuration in which the soil nodes are at the wrong distance from the shield. Also, if the starting stress state is anisotropic, the soil nodes will not be equidistant from the tunnel axis due to the pre­ convergences being anisotropic as well. To tackle this issue, a simple technique proposed by de Lillis et al. (2018) was adopted. The proced­ ure consists in designing a starting mesh that cor­ rectly reproduces the actual excavation profile after the development of the pre-convergences. To this aim, a trial analysis is performed adopting a standard circular mesh and recording the pre­ convergences along the tunnel wall. Then, the recorded values are added radially to the starting location of the nodes obtaining an elliptical mesh (being the pre-convergences asymmetrical given the anisotropic starting stress state) that will coin­ cide with the excavation diameter after the pre­ convergences (Figure 3b). Even though it is an iterative procedure, one iteration usually provides satisfactory results. 3 NUMERICAL RESULTS In this paragraph, some of the main results are described, focusing on the development of radial dis­ placements (convergences) in the soil, the soil-shield­ lining interaction and the resulting lining forces induced by the passage of TBM in undrained condi­ tions, at first, and then in long-term conditions. Figure 4 shows the longitudinal profile of the radial position of two soil points located at the crown and at the springline. The oscillations in the results are due to the excavation advancing step-by-step 1.2 m at a time. When the excavation moves forward the soil nodes in the newly excavated stretch are at

Figure 3. Mesh design.

268

Figure 4. Longitudinal profile of radial displacements.

different distances from the excavation front and thus respond differently. The starting position of the nodes is different because the pre-convergences, given the initial asymmetrical stress state, will be asymmetrical (smaller at the springline and higher at the crown). As anticipated, the mesh is designed in such a way that, when a given section is to be excavated, the location of the soil nodes coincides with that of the actual excavation boundary. After the excavation, more displacements develop and the soil tends to close on the machine’s shield, following its conicity. At the tail of the machine, the radial distance from the tunnel axis is smaller at the springline due to a more plastic behaviour. Roughly 4 diameters behind the excavation face, three-dimensional effects fade and stationary undrained conditioned are attained after the injection of the grout and its hardening. The described behaviour can be further illustrated by looking at the longitudinal profile of the radial stres­ ses in the soil (Figure 5). About 1.5 diameters ahead of the excavation front, a slight rise in radial stresses, due to the development of a 3D arch, can be observed. Then, the stresses decrease rapidly and become zero at the excavation face, where the overcut generates a void (there is no contact between soil and shield).

Behind the front face, because of the high initial stress state and its poor mechanical properties, the soil closes abruptly onto the shield and the radial stresses increase again once contact is estab­ lished. Due to the stress release induced by the excavation, the radial stress decreases and the cir­ cumferential stress increases around the tunnel, causing a stronger increase of the deviatoric stress near the springline. This redistribution of stresses is associated with a more plastic behaviour near the side of the tunnel, as shown in Figure 6, where the mobilised friction angle at different distances from the excavation face is reported. As the excavation face gets further (Figure 5), the conicity of the shield allows further displacements in the soil and the radial stresses progressively decrease. Near the tail, the stress is zero at the crown, where there is no contact with the shield. Behind the TBM, the annular tail void is injected with grout and the radial stresses increase. Overall, the changes in the stress field induced by the excavation and the soil-shield interaction are such that at the tail of the machine the stress state is the opposite of the starting one, with the radial stress being higher at the springline. The forces induced in the

Figure 5. Longitudinal profile of radial stresses.

269

Figure 6. Evolution of the mobilised friction angle: a) 1D ahead of the excavation front; b) at the excavation front; c) after the grout injection; d) undrained stationary conditions.

lining are generated by the interaction with this, new, profoundly changed, stress distribution in the soil. The normal force N and the bending moment M in stationary undrained conditions and in drained condi­ tions are shown in Figure 7. The axial forces range between 1850 kN and 2000 kN in undrained condition (Figure 7a); the minimum value being at the springline and the maximum at the crown. The bending moment, quite small in absolute values, is negative (internal fibers elongated) at the springline and positive at the crown (Figure 7b). The maximum compressive stress in the lining (σc), calculated assuming a homogeneous concrete section, is higher at the crown and lower at an angle of about 45° along the tunnel wall (Figure 7c). The internal forces can be related with the longitu­ dinal profile of the radial stresses seen in Figure 5. In fact, the higher normal force at the crown is associated with the higher radial stress near the springline, and vice versa. The small difference between the radial stresses induces small variability of the axial forces along the tunnel wall and small values of bending moments. In long-term drained conditions, the dissipation of the negative excess pore pressure induced by the excavation induces an appreciable increase of the normal forces, which reach values between 2100 kN and 2250 kN; the bending moment changes are very

small; the maximum compressive stress rises almost uniformly due to the increase of N. A set of parametric analyses was carried out to investigate the influence of changes in the TBM geometry on the forces induced in the lining. In par­ ticular, the study focused on the influence of the conicity and length of the shield. 3.1

Influence of the shield’s conicity

At first, two analyses were performed increasing the shield’s conicity, c, from 5 cm to 7.5 cm (c7.5­ analysis) and 10 cm (c10-analysis); all the other inputs of the analyses were kept the same. As seen in the previous paragraph, the lining forces are closely related to the radial displacement and the evolution of stresses, especially radial ones, in the surrounding soil. The results of the analyses, reported in Figure 8 in terms of radial position of the soil nodes located at the crown and at the springline, show that, as the conicity of the shield increases, the pre-convergences increase (appreciably going from c5 to c7.5, much less from c7.5 to c10) and the soil closure onto the shield decreases. The increase in conicity forces the soil to displace more to reach a contact point. Near the tail of the shield, the contact does not materialize along a portion of the shield that

Figure 7. Stress state in the lining: a) normal force; b) bending moment; c) maximum compressive stress.

270

Figure 8. Influence of the shield’s conicity on the radial displacements.

grows larger as the conicity increases, thanks to 3D effects allowing a stable configuration even though the soil-shield contact is incomplete. In particular, this is possible because of the contact points located near the excavation face and behind the tail of the machine, where grout has been injected. At the rear end of the shield, the soil is in contact with the shield at the side of the tunnel wall, in the springline area, with the notable exception of the c10­ analysis. From a stress point of view, this means that near the tail of the TBM, the radial stresses are zero at the crown for a longitudinal stretch that increases with c, while they are greater than zero in the springline area. In the c10-analysis, instead, since there is a residual void near the springline too, the radial stresses are zero all along the tunnel wall.

As discussed in the previous paragraph, the stress-displacement field at the tail of the machine, has a decisive influence on the lining forces. The greater the shield’s conicity, the greater the stress relaxation in the surrounding soils. Thus, as the conicity increases, the normal force in the lining decreases (Figure 9a). The distribu­ tion of N resulting from the c10-analysis, differs qualitatively from the others: in this case the soil does not touch the shield’s tail at any point (the radial stresses are zero both at the crown and at the springline), and this leads to a different inter­ action process. As c increases, the variability of N along the tunnel wall decreases, inducing smaller bending moments (Figure 9b). These dis­ tributions result in the maximum compressive stress decreasing as c increases. Particularly, σc is

Figure 9. Influence of the shield’s conicity on the lining loads: a) normal force; b) bending moment; c) maximum compres­ sive stress.

271

20% and 40% smaller than the reference case, respectively in the c7.5 and the c10 analyses (Figure 9c). It is worth to point out that greater displacements are obviously associated with a more plastic behav­ iour and an increase of the plastic radius around the tunnel, which could, in some cases, induce local fail­ ures in the surrounding soil. In drained conditions, the normal forces increase uniformly while the bending moments do not vary sig­ nificantly. In particular, a greater increase in N is observed in the analyses with larger conicities. This is clearly related to the amount of negative excess pore pressure developed during the excavation, which grow larger with the conicity and the associated stress relaxation. 3.2

Influence of the shield’s length

Two further analyses were performed investigat­ ing the influence of the shield’s length on the longitudinal displacements and the lining forces. The shield’s length, L, was assumed to be 4 m (L4-analysis) and 12 m (L12-analysis) respect­ ively, while keeping constant all the other inputs. Figure 10 shows the longitudinal displacements profiles. The results show that as L increases, both the pre-convergences and the portion of the shield along which the soil touches the TBM increases. Assuming a 4m-long shield, the soil does not close onto the shield because a stable configuration can be achieved thanks to the short distance between the excavation face and the injected annular void (strong 3D effects). Assuming a 12m-long shield,

instead, the soil displaces more gradually, as the stabilizing 3D effects induced by the closeness of the excavation front fade moving further along the shield. In terms of lining forces (Figure 11), the results show that the stress relaxation is directly related to the shield’s length; thus, N decreases as L increases. Con­ versely, the variability of N along the tunnel wall, and thus M, increases appreciably with the length of the shield. The similarity between the results obtained from the L8 and the L12 analyses can be explained looking at the longitudinal displace­ ment profile: in both cases the shield is long enough to ensure weak 3D effects, resulting in similar behaviours. In both analyses, in fact, the soil touches the shield along the majority of its length and only at its tail, in the crown area, a gap persists. As seen in the c10-analysis, the distribution of N resulting from the L4-analysis is different from the others. Once again, this is due to the radial stresses being zero all along the tunnel wall at the tail of the shield (while in the other cases σr = 0 just near the crown), leading to a strongly different soil-lining interaction. In drained conditions, the normal forces increase homogeneously in all the analyses, while the bending moments does not vary sig­ nificantly. In particular, the increase in N due to the consolidation process increases slightly with L. In terms of maximum compressive stress, the case of L = 4 m, provides maximum values which are about 30% lower than the case with the longer shield.

Figure 10. Influence of the shield’s length on the radial displacements.

272

Figure 11. Influence of the shield’s length on the lining loads: a) normal force; b) bending moment; c) maximum compres­ sive stress.

The length of the shield has a significant influence on the development of 3D effects; if the machine is short enough, a strong longitudinal arch can develop between the soil in contact with the injected grout (behind the tail) and the soil just ahead of the excava­ tion front, profoundly altering the entire interaction. Such phenomena can only be observed performing advanced 3D analyses, as the ones presented herein. The results provide novel insights on the soilshield-lining interaction process and further develop­ ments shall have significant implications on both the geometrical design of TBMs and the structural design of deep tunnel linings.

4 CONCLUSIONS To accurately predict the lining loads of deep tun­ nels, the stress release induced in the soil by the excavation must be accounted for. This paper pre­ sented an advanced 3D numerical model that simu­ lates the main features of mechanized tunnelling with remarkable geometrical accuracy. Moreover, a numerical procedure to avoid the misidentification of the excavation boundary is adopted. Under the assumptions made in this study (grav­ ity is neglected, as is the weight of the shield and the lining, undrained soil-shield-lining interaction during the excavation, final drained condition achieved without studying the consolidation pro­ cess), the numerical results show that the soilshield-lining interaction causes a great stress redis­ tribution around the tunnel, primarily associated with 3D effects and soil-shield contacts. The result­ ing lining loads depend on the soil stress state at the rear end of the shield, which in most cases was found to be the opposite of the initial K0 stress state (horizontal stress at the springline higher than the vertical stress at the crown). This phenomenon is due to a complex interaction with the shield of the TBM, characterized by a localized contact at the springline and a residual gap near the crown. The parametric analyses show that the geometry of the TBM has a great influence on the stress state of the lining. As the conicity increases the stress relaxation increases and the resulting stress state in the lining is smaller and more homogeneous. In the investigated cases, the maximum compressive stress decreases by 25% and 40% as the conicity increases by 50% and 100%, respectively. Of course, as the conicity increases, larger plastic zones will develop and the extension of the areas where the shear strength is fully mobilized will increase.

REFERENCES Arnau, O. & Molins, C. 2012. Three dimensional structural response of segmental tunnel lining. Engineering Struc­ tures 44: 210–221. Augarde, C.E. & Burd, H.J. 2001. Three-dimensional finite element analysis of lined tunnels. International Journal of Numerical and Analytical Methods in Geomechanics 25(3):243–262. De Gori, V., de Lillis, A. & Miliziano, S. 2019. Lining stresses in a TBM-driven tunnel: a comparison between numerical results and monitoring data. In D. Peila, G. Viggiani & T. Celestino (eds.), Tunnels and Under­ ground Cities: Engineering and Innovation meet Archaeology, Architecture and Art; Proceedings of the WTC 2019 ITA-AITES World Tunnel Congress, Naples, 2019. London: CRC Press. de Lillis, A., De Gori, V. & Miliziano, S. 2018. Numerical modelling strategy to accurately assess lining stresses in mechanized tunnelling. In A.S. Cardoso, J.L. Borges, P. A. Costa, A.T. Gomes, J.C. Marques & C.S. Vieira (eds.), Proceedings of the 9th European Conference on Numerical Methods in Geotechnical Engineering, Porto, 2018. London: CRC Press. Do, N.A., Dias, D., Oreste, P. & Djeran-Maigre, I. 2014. Three-dimensional numerical simulation for mechanized

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tunnelling in soft ground: the influence of the joint pattern. Acta Geotechnica 9(4): 673–694. Itasca Consulting Group, Inc. 2012. FLAC3D Version 5.0, Fast Lagrangian Analyses of Continua in Three-Dimensions, User’s manual. Minneapolis. Karakus, M. 2007. Appraising the methods accounting for 3D tunnelling effects in 2D plane strain FE analysis. Tun­ nelling and Underground Space Technology 22(1): 47–56. Kasper, T. & Meschke, G. 2004. A 3D finite element simu­ lation model for TBM tunnelling in soft ground. Inter­ national Journal for Numerical and Analytical Methods in Geomechanics 28(14): 1441–1460. Kasper, T. & Meschke, G. 2006. On the influence of face pressure, grouting pressure and TBM design in soft ground tunnelling. Tunnelling and Underground Space Technology 21(2): 161–171.

Kavvadas, M., Litsas, D., Vazaios, I. & Fortsakis, P. 2017. Development of a 3D finite element model for shield EPD tunnelling. Tunnelling and Underground Space Technology 65: 22–34. Litsas, D., Sitarenios, P. & Kavvadas, M. 2018. Parametric investigation of tunnelling-induced ground movement due to geometrical and operational TBM complexities. Italian Geotechnical Journal 51(4): 22–34. Losacco, N. & Viggiani, G.M.B. 2019. Class A prediction of mechanised tunnelling in Rome. Tunnelling and Underground Space Technology 87:160–173. Miliziano, S. & de Lillis, A. 2019. Predicted and observed settlements induced by the mechanized tunnel excava­ tion of metro line C near S. Giovanni station in Rome. Tunnelling and Underground Space Technology 86: 236–246.

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Development of a new excavation technique for centrifuge testing in sand N.E. Faustin & R.J. Mair University of Cambridge, UK

M.Z.E.B. Elshafie Qatar University, Qatar

ABSTRACT: Even though shaft excavation is a key part of tunnelling projects in large cities, many uncertainties about their actual performance still exist in practice and only very few case studies have been reported in the literature. In this research work, centrifuge testing was used to investigate the performance of circular shafts and the adjacent ground movements during excavation. The most widely used excavation technique in centrifuge model-ling, employs a hydrostatic stress state which may not be the case in practice. This paper describes novel apparatus and centrifuge testing proced­ ures that were developed to simulate shaft excavation in dry sand. The dry sand was carefully removed from the centre of a circular shaft under a centrifugal acceleration of up to 125 times the earth’s gravity. The new in-flight excavation system allows more realistic stress changes to be induced on the shaft lining, compared with existing techniques. Detailed centrifuge test measurements of longitudinal bending and hoop strains in the circular shaft lining and displacement of the adjacent ground during excavation, reported in this paper, provide invaluable insights for future shaft construction.

1 INTRODUCTION 1.1

1.2

General

Centrifuge model testing is useful to investigate complex engineering problems that cannot be resolved using routine design procedures. In many cases, this is due to limited knowledge of the prob­ lem itself. One such example is circular shafts. There are few well-documented case studies of cir­ cular shafts despite these structures being a key com­ ponent of tunnelling schemes in urban environments. This makes it difficult to accurately predict the per­ formance of circular shafts or estimate the magni­ tude and extent of the surrounding ground movement. Consequently, the design of circular shafts is conservative, resulting in thick, heavily reinforced shaft linings and protective measures for adjacent infrastructure that may not necessarily be required. Centrifuge testing of circular shafts, coupled with field observation from actual case studies in practice, can lead to more efficient designs and ultimately could reduce the cost of circular shaft construction. An important part of the centrifuge test is to model excavation of the shaft under an enhanced gravita­ tional field i.e. in-flight.

Modelling excavations in centrifuge testing

Three different excavation methods are reported in the literature for modelling excavations in a geotechnical centrifuge. The most common method replaces the soil at the centre of the shaft with a fluid of similar density (Lade et al. 1981, Kusakabe 1982, Britto & Kusa­ kabe 1984, Elshafie et al. 2013 and Divall & Goodey 2016). The excavation is subsequently carried out by draining. the fluid during the centrifuge test. This technique can only replicate a hydrostatic stress state in which there are equal horizontal and vertical stres­ ses at a given depth (Ko=1). This condition may not actually simulate field conditions. A slight variation to this technique was adopted by Divall and Goodey 2016 to excavate a semi-circular shaft in clay at 100g. A 70 mm diameter and 200 mm long solid semi-circular shaft lining wrapped in a latex membrane was placed in a slightly larger pre­ excavated hole. The 3 mm wide annulus that formed between the shaft and the clay was filled with a heavy fluid, sodium polytungstate, which was drained inflight. The presence of the solid shaft at the centre of the excavation reduced the amount of heavy fluid required for the centrifuge test. However, this still only replicates a hydrostatic stress state.

DOI: 10.1201/9780429321559-35

275

The second excavation method is to force a displacement of a semi-circular shaft lining that may occur during shaft excavation (Fujii et al. 1994, Hagiwara et al. 1998, Imamura et al. 1999 and Jeong & Kim 2014). This method was generally adopted for laboratory and centrifuge tests that modelled active conditions in order to examine lateral earth pressures. The third method is to remove sand at the centre of the shaft during the centrifuge test. Azevedo 1983 achieved this by removing flexible bags of sand from the centre of the shaft and Ueno et al. 1996 developed an in-flight excavation device. The device comprised a perforated tube located at the centre of the shaft and a negative pneumatic pressure was applied to the bottom of the perforated tube to draw dry sand from the centre of the shaft to a collection point. However, the apparatus included a bulky vacuum system at the bottom of the shaft that may have interfered with the soil displacements. 1.3

Circular shaft construction

In practice, a circular shaft is constructed using either a Support Before-Excavation (SBE) or Excavation Before Support (EBS) method (Faustin et al. 2018). For the first category, the shaft is supported by pre­ installed walls that are either diaphragm walls, secant bored piles or sheet piles. Excavation is carried out after the pre-installed walls are built. For the second category, the shaft is progressively excavated in sec­ tions to expose the ground, typically in one metre height increments, followed by the erection of either pre-cast segments or a sprayed concrete lining. When a complete ring is formed the sequence is repeated for the underlining ring until the desired shaft depth is reached. This paper describes new centrifuge apparatus and testing procedures that were developed to excavate circular shafts in-flight in dry sand. The tests were conducted in the 10m diameter beam centrifuge at the Schofield Centre, Cambridge Uni­ versity. A detailed description of this centrifuge is given by Schofield 1980. The excavation proced­ ures allowed the actual stress changes on the shaft lining to be modelled and were analogous to a SBE shaft construction in which the soil is sup­ ported by pre-installed walls before the excavation is carried out. Further details of the centrifuge tests are provided in Faustin et al. 2017.

acceleration of 125 times the earth’s gravity. The model shaft and the surrounding sand were wellinstrumented with strain gauges and miniature dis­ placement transducers to measure longitudinal bending and hoop (compressive) strains in the shaft lining as well as deformation of the adjacent sand during excavation. Figure 1 is a typical cross section through the sand centrifuge model. It shows that container filled with dry Leighton Buzzard Fraction E sand, an instru­ mented model shaft and an in-flight excavation system at the centre. Excavation of the shaft was carried out in-flight. 2.2

Leighton Buzzard Fraction E sand

Leighton Buzzard Fraction E sand was used for the centrifuge tests. This sand has been used exten­ sively for centrifuge testing. Its particle size (D10, D50 & D60), minimum and maximum void ratio (emin & emax), Specific Gravity (Gs) and critical state angle of shearing resistance (ϕcs), are summar­ ised in Table 1 (Tan 1990). A uniform dense sand bed with an average rela­ tive density of 80% was prepared using an auto­ matic sand pourer (Zhao et al. 2006). A uniform density could not be obtained around the circular shaft with the rectangular sand pour pattern rou­ tinely adopted for tests. Therefore, the automatic sand pourer was programmed to pluviate the sand in a spiral pattern. The sand pour set up is shown in Figure 2.

Figure 1. Schematic of the centrifuge model.

2 CENTRIFUGE MODELLING OF SHAFT EXCAVATION IN SAND

Table 1.

2.1

Properties of Leighton Buzzard Sand.

Overview

A completely new apparatus was designed and built for the 1:80 scale centrifuge tests in dry dense sand. All the components were proof tested for safe use in the centrifuge, at a centrifugal

Particle size (mm) D10 0.095

276

D50 0.14

Void ratio D60 0.15

emin 0.613

emax 1.014

Gs 2.65

ϕcs 32°

Figure 2. Preparation of centrifuge model – sand pouring.

2.3

Main apparatus

The sand centrifuge tests were carried out in a 700 mm diameter and 620 mm deep aluminium model container. A 2 mm thick circular shaft was machined from aluminium alloy 6082 having a Young’s Modulus of 70 GPa. The model shaft had an internal diameter of 140 mm and a length of 290 mm. It was instrumented with strain gauges on both the internal and external surfaces and carefully positioned at the centre of the sand during prepar­ ation of the centrifuge model at 1g. During the centrifuge test, the shaft was excavated using a novel sand excavation system comprising a lifting device (similar to the one used by Silva 2005), stacked trays and a cylinder with several openings. The excavation was analogous to a shaft made up of secant bored piles or diaphragm wall panels, installed in a circular arrangement, that are excavated after all the piles or diaphragm wall panels have been installed. A maximum excavation depth of 198 mm achieved was equivalent to a prototype shaft 11 m in diameter and approximately 16 m deep. 2.3.1 In-flight excavation system The new in-flight excavation system comprised a series of nine stacked trays and a cylinder with openings, as shown in Figure 1. The trays were loosely connected to each other using tie rods. The excavation system was placed at the centre of the model shaft and was supported entirely from above

so that no additional stresses were induced in the underlying soil. This created a narrow 18 mm wide annulus between the excavation system and the shaft. A 10m thick compressible foam layer was also placed along the perimeter of the slotted cylinder to prevent sand from flowing inwards and to absorb any deflections from the overlying excavation system. Figure 3 shows a schematic of the excavation system using the bottom two trays. At the start of the centrifuge test, the shaft was embedded in the sand with the slotted cylinder and stacked trays at the centre. The stacked trays covered openings in the slotted cylinder thereby preventing sand from flowing from the 18 mm wide annulus into the trays (Figure 3(a)) During the centrifuge test, the trays were lifted sequentially, to expose the openings, thereby allow­ ing the sand to flow into from the annulus into the tray, Figure 3(b). This staged excavation continued until all the trays were lifted. Each lift lowered the excavation level by approxi­ mately 20 mm. Figure 4 shows the trays filled with sand after a centrifuge test. 2.3.2 Excavation system verification A simple finite element analysis was conducted using PLAXIS 2D to verify that the in-flight shaft excavation device did not change the stress field around the shaft or affect the behaviour of the shaft lining. The analyses showed that the presence of the

277

Figure 4. Sand-filled trays after a centrifuge test. Figure 3. Schematic of the sand excavation system.

excavation system, at the centre of the shaft, has a negligible effect on the behaviour of the shaft and the surrounding soil during excavation. 2.4

Monitoring instrumentation

Encapsulated 350Ω strain gauges were secured to the internal and external surface of the shaft to meas­ ure strains in the shaft lining. The strain gauges were configured in a Wheatstone full bridge and half bridge configurations to measure bending strains and hoop strains respectively. Ground surface displacements were also meas­ ured at varying distances from the shaft, using a combination of miniature linear variable dis­ placement transducers (LVDTs) and lasers. The CVDTS are labelled A to H on Figure 1 and the lasers are labelled L1 to L4. There was at least one level of redundancy between the LVDTs and lasers to verify that the correct measurements were being recorded. Excavation of the model shaft was expected to generate only small ground movements. Therefore, the LVDTs were positioned so that at the start of excavation the LVDT voltage output was very close

to zero and linear readings were obtained during excavation of the shaft. Two sets of instruments were used to monitor the shaft excavation progress. The first instrument was a load cell, mounted on the excavation system, to measure the cumulative mass of the sand-filled trays as they were lifted. The second instrument was a laser, with a measuring range of 50 mm to 350 mm, that was focussed at the centre of the narrow sand annulus. The laser measure­ ment represented an average depth across the annulus. An accelerometer monitored the variation in cen­ trifugal acceleration throughout the centrifuge test and a camera and an endoscope provided good visual inspection of the excavation. 2.5

Model preparation

The instrumented model shaft was placed at the centre of the model container after 250 mm of sand had been poured (Figure 5). This arrangement pro­ vided a clearance of approximately one shaft depth from the bottom of the shaft to the base of the model container and approximately two shaft diameters to the model container walls. Finite element analyses showed these clearances are enough to avoid bound­ ary effects. A further 290 mm of sand was then

278

was then excavated by lifting the stacked trays in turn to expose the adjacent openings and allow sand to flow from the annulus into the trays. On average it took one hour for eight excavation steps to be carried out. 3 TYPICAL CENTRIFUGE TEST RESULTS 3.1

Figure 5. Placement of instrumented model shaft and exca­ vation system during sand pouring.

poured around the shaft and within the 18 mm annu­ lus between the shaft excavation system. Once the sand was poured, the monitoring instrumentation was placed, and the rest of the model built. It took 10 days to prepare the centri­ fuge model and a photograph of the completed centrifuge model being lowered onto the beam centrifuge is shown in Figure 6. 2.6

Centrifuge test procedure

Nine centrifuge tests were conducted between Octo­ ber 2014 and May 2016. The centrifuge model was swung up to 80 g in approximately hour. The shaft

Figure 6. Lowering the 750kg model onto the beam centrifuge.

Shaft excavation progress

Figure 7 shows the typical shaft excavation progress in terms of shaft excavation depth recorded by the laser and cumulative mass measured by the load cell. As expected, each tray lift increased the shaft exca­ vation depth and caused a tensile load increase in the load cell reading. On average, each tray lift increased the shaft excavation depth by approximately 20mm Spikes in the load cell measurement at the start of each excavation step correspond to the lifting device having to overcome some initial friction to lift the sand-filled tray. 3.2

Ground surface movements

Figure 8 shows surface settlement observed during swing up to the test centrifugal acceleration. The sand experienced a maximum settlement of 2 mm to under a centrifugal acceleration of 80g. At the end of the test, the thickness of the sand model was permanently reduced by an average of 1.2mm. This irrecoverable deformation of the sand surface equates to an average increase in relative density of approximately 1.2 %. The distribution of surface settlement adjacent to the shaft is shown in terms of settlement plotted against distance from the shaft in Figure 9. Gener­ ally, the magnitude of settlement was highest close to the shaft and decreased with increasing distance away from the shaft. Relatively smaller settlements were observed within approximately 20 mm from the shaft wall. Minimal variation in the excavated sand level, at the end of the test, confirmed that the new shaft excavation system produced a relatively uniform (symmetric) excavation.

Figure 7. Shaft excavation progress at 80 g.

279

Figure 8. Measured ground surface settlement during swing up to 80g.

Figure 9. Measured ground surface settlement during excavation at 80g.

280

3.3

Figure 11 shows the hoop stresses, derived from the strain readings, as the excavation progressed. Posi­ tive stresses indicate tensile hoop stresses and nega­ tive values compressive hoop stresses. In general, the hoop stresses increase as the excavation progresses. The measured longitudinal bending strains were considerably smaller than the hoop strains and are therefore not presented.

Strains in the circular shaft lining

Results are presented for two arrays of hoop strain gauges, Array A and Array B, shown in Figure 10, used to monitor circumferential or hoop stresses in the shaft lining. The convention adopted for strain gauge identification is [Array identifier, level of the strain gauge]. For example, A1 represents a strain gauge in Array A at level 1 (at a depth of 40 mm) and B1 repre­ sents a strain gauge in Array B at level 1 (at a depth of 40 mm).

4 CONCLUSIONS

Figure 10. Schematic of hoop strain gauges in Arrays A & B.

Circular shafts are a key part of tunnel schemes in urban environments but there are few welldocumented case study shafts. As a result, there is limited knowledge and understanding of the perform­ ance of circular shafts and the adjacent ground behav­ iour during construction. With recent construction of major tunnelling projects in the UK, like Crossrail, and other planned or underway, there is a pressing need to improve industry’s understanding of these structures. To obtain realistic centrifuge testing results it is import­ ant to accurately model the in-situ horizontal stress changes on the shaft lining as well as the vertical stress changes at the excavation level during shaft construction. The most common method of carrying out excava­ tions in centrifuge testing involves replacing the soil to be excavated with a heavy fluid of similar density. The heavy fluid is then drained in-flight. This technique can only replicate a hydrostatic stress state having

Figure 11. Derived hoop stresses in the model shaft lining during excavation.

281

equal horizontal and vertical stresses at a given depth. This condition may not actually be the case in the field. The centrifuge apparatus and testing procedures reported in this paper were developed to create more realistic stress changes on the shaft lining by remov­ ing sand from the centre of the shaft during excava­ tion at high centrifugal accelerations. The measurements showed the distribution of ground movements adjacent to the shaft due to excavation. Strain measurements showed that most of the horizontal soil stresses acting on circular shafts are translated into hoop stresses with min­ imal bending stresses in the shaft lining. Hoop strains gradually increased as the excavation progressed.

ACKNOWLEDGEMENTS The authors are grateful to the Engineering and Physical Sciences Research Council (Award Refer­ ence 1220514) and Geotechnical Consulting Group for financial support. Special thanks are given to Neil Houghton for his contribution to the detailed design of the new apparatus and to Cambridge Insitu Limited for the strain gauging. The invalu­ able contribution from the technicians at the Scho­ field Centre at Cambridge University if gratefully acknowledged.

REFERENCES Azevedo, R. 1983. Centrifuge and analytical modelling of excavation in sand. PhD thesis, University of Colorado, Boulder, USA. Britto, A. and Kusakabe, O. 1984. On the stability of sup­ ported excavations. Canadian Geotechnical Journal, 21:338–348. Divall, S. and Goodey, R. J. 2016. An apparatus for centri­ fuge modelling of a shaft construction in clay. In Euro­ fuge2016, 3rd European Conf. on Physical Modelling in Geotechnics, Nantes, France. Elshafie, M.Z.E.B., Choy, C. K. C., and Mair, R. J. 2013. Centrifuge modeling of deep excavations and their inter­ action with adjacent buildings. Geotechnical Testing Journal, Vol. 36, No. 5, 2013, pp.1–12.

Faustin N.E. 2017. Performance of Circular Shafts and Ground Behaviour during Construction. PhD thesis, University of Cambridge, Cambridge, UK. Faustin N.E., Elshafie M.Z.E.B & Mair R.J. 2018. Case studies of circular shaft construction in London. Pro­ ceedings of the Institution of Civil Engineers Geotech­ nical Engineering Vol 171:5, 391–404 Fujii, T., Hagiwara, T., Ueno, K., and Taguchi, A. 1994. Experiment and analysis of earth pressure on axisym­ metric shaft in sand. In Proceedings of Centrifuge ’94, 791–796. Hagiwara, T., Imamura, S., Fujii, T., Nomoto, T., and Kusakabe, O. 1998. Earth pressure acting on a deep cir­ cular shaft and associated ground deformation. Pro­ ceedings of the International Conference on Centrifuge, 643–648, Tokyo, Japan. Imamura, S., Nomoto, T., Fujii, T., and Hagiwara, T. 1999. Earth pressures acting on a deep shaft and the move­ ments of adjacent ground in sand. Proceedings of the International Symposium on Geotechnical Aspects of Underground Construction in Soft Ground, 647–652, Tokyo, Japan. Jeong, S. S. and Kim, K. Y. 2014. Three-dimensional arch­ ing effect on vertical circular shafts. Proceedings of Geotechnical Aspects of Underground Construction in Soft Ground, 79–85, Seoul, Korea. Kusakabe, O. 1982. Stability of excavations in soft clay. PhD thesis, University of Cambridge. Lade, P. V., Jessberger, H., Makowski, E., and Jordan, P. 1981. Modeling of deep shafts in centrifuge tests. In Proc. 10th International Conf. on Soil Mechanics and FoundationEngineering, Stockholm, 683–691, Balkema, Rotterdam. Silva, M. F. 2005. Numerical and physical models of rate effects in soil penetration. PhD thesis, University of Cambridge, Cambridge UK. Schofield, A. N. 1980. Cambridge geotechnical centrifuge operations. Géotechnique, 30(3):227–268. Tan, F. S. C. 1990. Centrifuge and Theoretical Modelling of Conical Footings on Sand. PhD thesis, University of Cambridge, Cambridge, UK. Ueno, K., Yokoyama, Y., A, O., and T, F. 1996. Earth pres­ sures acting on flexible circular shafts in sand. Proceed­ ings of Geotechnical Aspects of Underground Construction in Soft Ground, 237–242, Balkema, Rotterdam. Zhao, Y., Gafar, K., Elshafie, M. Z. E. B., Deeks, a. D., Knappett, J. a., and Madabhushi, S. P. G. 2006. Calibra­ tion and use of a new automatic sand pourer. 6th Inter­ national Conference of Physical Modelling in Geotechnics, 65:265–270.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Interpretation of soil parameters used for numerical analysis with small strain model for deep excavation in loose to medium dense sand Hsiung Benson Bin-Chen & Phan Khac Hai Department of Civil Engineering, National Kaohsiung University of Science and Technology, Kaohsiung, Taiwan

Ching Hung Department of Civil Engineering, Tainan, Taiwan

ABSTRACT: In this paper, a Hardening Soil- Small strain (HSS) model, capable of characterizing the small strain behavior of the soils, is utilized to evaluate the performances of deep excavations in loose to medium dense sands. First, the stiffnesses of the sandy soils are collected mainly from the results of in- situ and labora­ tory tests from previous studies, and additional two small strain stiffness parameters, the shear modulus at very small strain level, Go, and the threshold shear strain, γ0.7, are also interpreted through laboratory tests. Second, the sandy soil parameters are validated to examine the behaviour of excavation in terms of lateral wall movements and ground surface settlements. The results of the numerical analyses show that the Go can be defined as a function of the depth and the threshold of 1 times 10-4 of shear strain γ0.7 can be revealed. Those input parameters are recommended for analyses of deep excavation in loose to medium dense sand if the HSS model is adopted. Also, the stress-controlled method is followed in the current test standard of one of in-situ tests, Pressuremeter test which may have to be re-considered in order to fit the reality on site.

1 INTRODUCTION Numerical analyses play an important role in predicting the performance of underground con­ struction, especially for deep excavation behav­ iors due to the complexity of soil- structure interactions. It is also important to have compara­ tively accurate soil parameters as inputs before any prediction is undertaken. Indicated by several research works, the small strain stiffness shall be considered in the constitutive model used to pre­ dict the ground responses caused by deep excava­ tion (Atkinson & Sallfors, 1991; Mair, 1993). Nikolinakou et al. (2011) used the advanced MIT-S1 model to evaluate the performance of deep excavation in Berlin sand. This model is relatively complex and 13 input soil parameters were required to govern the small strain behavior in sand. Thus, to simply capture the small strain behaviour, the study utilized the “Hardening soil model with the small strain (HSS)” (Brinkgreve et al., 2006). This constitutive model is an exten­ sion of the Hardening Soil (HS) model with two additional soil input parameters to express the small strain stiffness behavior, namely the small strain/initial shear modulus, G0, and the reference threshold shear strain, γ0.7. Dao (2015) expressed the use of HSS is eligible to provide

comparatively better estimations in terms of ground surface settlements and horizontal wall movements when compared to the results obtained by Mohr-Coulomb (MC) model and HS model. Evaluations of the HSS model in the simulation of deep excavation for both clayey and sandy soil can be found in Lim et al. (2010) and Hsiung & Yang (2017). Those results have shown that the analysis using this model provided a more realistic displacement when compared to the field monitoring data. Besides, application of this model in clayey soil is seen to be reasonable in deep excavation as suggested by Likitlersuang et al. (2013). In this study, the small strain stiffness of loose to medium dense sand was examined based on the laboratory and in-situ tests. The tests were carried out for grounds in Kaohsiung city, Taiwan and Kuala Lumpur city, Malaysia. The tests include dilatometer tests (DMTs), downhole tests, and pressuremeter tests (PMTs) on project sites and resonant columns, as well as dynamic axial tests in laboratory. Finite element simulations, utilizing the HSS model, are also performed in the study. The additional two soil parameters, G0 and γ0.7, will be interpreted from insitu and laboratory tests. At last, the method of prop­ erly control of PMT used in sand will be examined and discussed.

DOI: 10.1201/9780429321559-36

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2.2

2 PROJECT BACKGROUND 2.1

Case KHH

A deep excavation in Kaohsiung city, Taiwan was selected for the first case study as research back­ ground of this study. This excavation is used as an underground car-park basement and the excavation shape is rectangular with 70 m in length and 20 m in width. An 0.9 m thick and 32 m deep diaphragm wall was adopted to retain the excavation. The maximum excavation depth was 16.8 m and was constructed by bottom-up method in five excavation stages with 4 levels of steel struts. Figure 1 illustrates the crosssection and subsurface soil profile as the excavation. Although there were three thin clayey layers, the ground condition was almost fully rested on the loose to medium dense sand due to the dense thickness of sandy layers. The groundwater level slightly fluctu­ ated from 2 m to 2.3 m below ground-surface level (bgl). More details of background of this excavation can be found in Hsiung et al. (2016). Furthermore, Hsiung et al. (2016) indicated the empirical equation to obtain the soil stiffness in loose to medium dense sand and initial elastic modulus of soils, E, measured from the in-situ DMT tests could also be interpreted by linear regression analysis. The function of linear increases (E) which increases the ground depth (Z) was shown as follows:

Case KLCC

The second case of deep excavation in sand was selected from Kuala Lumpur, Malaysia. This project was also used as an underground car-park and the shape of excavation is rectangular with a length of 78 m and a width of 43 m. The excavation was also performed by the bottom-up method and retained by a 0.6 m thick and 20 m deep diaphragm wall. The maximum excavation depth is up to 13.2 m and the pit is only supported by a single level steel H-beam strut (400 x 400 mm) at 4 m below the ground sur­ face. Figure 2 presents the cross-section of the exca­ vation and ground profile in KLCC. As shown in Figure 2, the ground profiles can be possibly cat­ egorized into three ground formations (Phan et al., 2019) such as (i) Upper Kuala Lumpur sand (UKLS), which is the loose to medium dense sand ranged from 0 to 13.5 m bgl (SPT-N < 30); (ii) Lower Kuala Lumpur sand (LKLS), which is dense to very dense sand ranged from 13.5 to 25.5 m bgl (30 < SPT-N < 100); and (iii) Kuala Lumpur sand/ silt-stone (KLSS), which is the sandstone and silt­ stone layer beyond 25.5 m bgl (SPT-N > 100). The bulk density of the residual soil layer generally ranges from 19 kN/m3 to 22 kN/m3 with depth. The groundwater table is observed at approximately 2.5m below the ground surface. More details of this project background can be found in the related litera­ tures by Law et al. (2014); Law et al. (2016); Ang et al. (2018); Phan et al. (2019). As indicated before, the soil stiffness at the site is also possible to be categorized into three ground for­ mations accordingly with the characteristics of soil based on in-situ PMT tests as well as similar ground profiles in Berlin, Germany and Kaohsiung sand. It is also found that initial elastic modulus of soils of all three ground formations can also become func­ tion of ground depth by applying linear regression

Figure 1. Cross-section of excavation in Case KHH (Hsiung

et al., 2016).

284

Figure 2. Cross-section of excavation in Case KLCC.

analyses (Phan et al., 2019). Following are the details of interpreted initial elastic modulus of soils: UKLS EUPKL ¼ 1250ðZ þ 9:7ÞðkPaÞ

ð2Þ

ELKLS ¼ 10396Z þ 28900ðkPaÞ

ð3Þ

LKLS

KLSS EKLSS ¼ 1588740Z - 37083380ðkPaÞ ð4Þ

3 THE SOIL SMALL STRAIN PROPERTIES The small shear modulus (G0) and the reference threshold shear strain (γ0.7) are the two required input parameters in the finite analyses using HSS model. Therefore, the laboratory tests and in-situ tests were conducted in the central area of Kaohsiung city in order to obtain said parameters for those loose to medium dense sand. It is noted that the Go parameter was carried out in both laboratory and field tests. Meanwhile, the parameter of γ0.7 was interpreted by the stiffness degradation curve. The following section will describe interpretation of small strain parameters of the soil tests. 3.1

Field data test

The most common approaches to obtain the stiffness at small strain level is by the measurement of the shear wave velocity in soil materials (Viggiani & Atkinson, 1995; Kramer, 1996; Likitlersuang et al. 2013; Mu & Huang, 2016). Thereby conducting the downhole seis­ mic test and s-wave velocity (Vs) measured from three various locations close to the site using said testing are presented in the same figure too. It was tested once for every meter from surface level to 10 m below surface level but changed to 2 meters of an interval for ground deeper than 10 m below surface level. Shown in Figure 3a, Vs increases from approximately 100 m/sec at the depth near surface level to around 400 m/sec at 65 m below surface level. More importantly, Figure 3b shows the G0 values calculated as illustrated in Equa­ tion 5 from the measured shear wave velocity (Vs) and the soil density.

where ρ is the density of soil. All data points for G0 at various depths are plotted in Figure 3b. G0 likely increased in proportion to the ground depth regardless of the material, and this is also consistent with results reported by Hsiung et al. (2016) in terms of initial elastic modulus of soils. Therefore, a similar regression analysis was conducted so G0 was defined as a function of the depth (Z). In which Z is in units of meters, and G0 is in units of MPa.

Figure 3. Interpretation of G0 used for analyses using the HSS model: (a) Shear wave velocity of ground (Vs) measured on site; (b) Interpretation G0 from linear regression of downhole test data and comparison with G0 from the lab test.

This relation can thus be expressed by following Equation 6 below:

In addition, the G0 interpreted from the laboratory test was plotted in Figure 3b together for a comparison purpose. It is likely to be smaller in comparison with results from in- situ tests which might be con­ nected with disturbance of sampling and testing.

285

3.2

Laboratory test

The additional laboratory tests, such as dynamic tri­ axial and resonant column test were conducted at the

depth of 12 m and 10 m bgl to obtain γ0.7 for sandy soil by measuring the stiffness reduction curve. For the dynamic triaxial test, a 70 mm of diameter, 140 mm of height cylinder sample was used and 100 kPa of effective vertical stress was applied for the testing. Cyclic loading was adopted, and axial strain level was controlled in the range of 0.0001 % to 10 %. In addition, the extra bender elements were installed at two ends of the sample in order to gener­ ate shear wave at different strain levels. Regarding resonant column test, the sand sample was taken from 10 m bgl and the axial strain level was con­ trolled in the range of 0.0001% to 1%. Figure 4 shows the results of the testing which plotted the degradation curve of shear modulus of sand and it shows γ0.7 is around 0.0001. 4 NUMERICAL MODELS 4.1

Finite element models

The commercially available PLAXIS 3D version 2017 was adopted in this study for the finite element simulations. Figure 5 illustrates the boundary conditions of two deep excavations from Case KHH and Case KLCC. The roller

Figure 4. Degradation curve of shear modulus of sand with shear strain level.

boundaries as lateral displacement constraints were set vertically at 4 sides; whereas the bottom of the model was placed hinge boundaries. The distances of mesh boundary each model to the excavation boundary were set approximately to seven times the maximum depth of excavation, He as recommended by Finno et al. (2007). A ten nodded tetrahedron element elements were used for the soil volumes in the analyses (PLAXIS BV 2017). 4.2

Input parameters

As stated above, the HSS model was adopted to govern the soil small behavior in this study. There are three soil input moduli in this model at the reference pressure pref as taken 100 kPa (Schanz et al., 1999) such as E50ref, Eoedref and Eurref, those parameters stand for tri­ axial loading secant stiffness, oedometer loading tangent stiffness, and triaxial unloading/reloading stiffness, respectively. Besides, E50ref can be used to estimate the value of Eoedref and Eurref with more accurate prediction subjected the unloading-reloading feature (Schanz et al., 1999; Schweiger, 2009; Hsiung et al., 2017; Ang et al., 2018). Hsiung et al. (2017) and Ang et al. (2018) suggested that E is approximately 1.5 ref ref times of the E50, Eur and Eoed were set to 3 ref ref times of the E50 and E50 . Furthermore, the shear modulus at small strain in dense to very dense sand (LKLS layer), Go, is directly adopted from the relationship with E0 through Equation 7, and E0 can be 10 times E50 as suggested from RTRI (2010). Therefore, the rest of soil small strain input parameters using HSS model for those cases are listed in Table 1.

where vur is the unloading/reloading Poisson's ratio, and it is assumed a constant with vur=0.2 (PLAXIS BV 2017)

Figure 5. The cross-section and boundary condition of the 3D models. (a) Case KHH; (b) Case KLCC.

286

Table 1.

Basic soil properties adopted for the validation analyses.

(a) Kaohsiung City Layer

Depth

Soil type

E50ref (MPa)

Eoedref (MPa)

Eurref (MPa)

γ0.7

Goref (MPa)

2 4 5 6 8 9

2.0- 6.5 8.0- 17.0 17.0- 23.5 23.5- 28.5 30.5- 42.0 42.0- 60.0

SM SM SM SM SM SM

26.2 21.9 24.1 26.3 29.6 33.1

26.2 21.9 24.1 26.3 29.6 33.1

78.7 65.7 72.2 78.8 88.7 99.4

0.0001 0.0001 0.0001 0.0001 0.0001 0.0001

101.3 112.7 127.8 141.5 162.1 184.2

Noted: Additional parameters not listed above please refer to paper written by Hsiung et al. 2016 (b) Kuala Lumpur City Group layer UPKLS LKLS

Depth

Soil type

E50ref (MPa)

Eoedref (MPa)

Eurref (MPa)

γ0.7

Goref (MPa)

0.0- 7.5 7.5- 10.5 10.5- 13.5 13.5- 18.0 18.0- 25.5

SM SM SM SM SM

20.3 21.2 22.5 145.9 172.9

20.3 21.2 22.5 145.9 172.9

60.8 63.7 67.5 437.7 518.6

0.0001 0.0001 0.0001 0.0001 0.0001

95.8 106.7 115.4 608.0 720.3

Noted: Additional parameters not listed above please refer to paper written by Phan et al. 2019

For simulations of structures, the temporary strut systems were modeled by the anchor-to-anchor elements and the diaphragm wall was simulated by plate element. These elements were assumed as a linear elastic material. The interface elements were also used in these analyses to represent the inter­ action between the soil and diaphragm wall. The node pairs at the interface of the structure and soil are used in this model. For details of structural input parameters in Case KHH and Case KLCC can respectively be found in the papers written by Hsiung et al. (2016) and Phan et al. (2019).

5 DISCUSSIONS 5.1

Wall and ground surface displacements

Figure 6 shows the performance of the Case KHH through the stages of constructions by observation of the wall deflections and ground surface settlements. The second and final excavation results are chosen to compare with the ones in field measurements which aim to explore the small strain behavior in the shal­ low and deep excavation in sand. The inclinometer reading of SID3 and SID4, which respectively located

Figure 6. Comparison of observations and predictions of wall delection and surface settlement in case A Wall deflection: (a) Inclinometer 3 (SID3); (b) Inclinometer 4 (SID4); Surface settlement: (c) Section 4.

287

in the center of the long side and the short side of excavation, are adopted to figure the horizontal wall movements, while the ground surface settlements are observed at Section 4 (please refer to Hsiung et al. 2016). The comparisons show that the predictions of wall deflection in the long side of excavation are well agreed with the observations rather than those of in the short side. The predicted maximum wall deflec­ tion in section of SID3 increases from 20 mm in the 2nd excavation stage to 54 mm in the final excavation stage (refer to Figure 6a). As indicated previously, the short side wall deflections, however, seem to be overpredicted (refer to Figure 6b). Hsiung et al. (2016) have already explained the heavy corner diagonal bra­ cing of each strut stiffened the wall stiffness system which led to reduce some amount of wall displace­ ments on the short side. Unfortunately, the heavy corner diagonal bracing struts were excluded in this simulation owing to reducing the simulation effort and the calculation time. Figure 6c shows the comparisons of ground sur­ face settlement in Section 4 between simulation and monitoring data. Despite limited data of observa­ tion of ground settlements up to 3 m away from the excavation, it is obviously seen that a close agree­ ment between field measurements and results from numerical analyses. So far, the largest ground sur­ face settlements had been measured from the site roundly 16 mm and 26 mm at the second and final excavation stages. Those values are pretty fitted with the initial value of the ground surface settle­ ment trough produced by using HSS model in cor­ responding the excavation stages at the second and the final, approximately 14 mm and 28 mm, respectively. Figure 7 shows the comparisons of the wall deflections between predictions and field data meas­ urements at the first and final excavation stages

among various cross-sections at the long and short side of the excavation for Case KLCC. The two inclinometers in the middle and quarter of the long side of the excavation, SID8, and SID9, are adopted to examine the horizontal wall movements. Mean­ while, the inclinometer reading of SID6 is selected to observe the wall deflection on the short side of excavation. As shown in the Figure 7, the diaphragm wall was experienced rapidly in the changes of its movement modes, i.e. the wall behaved first in canti­ lever-mode and then changed to prop-mode after the struts were installed in both measured data and prediction. The observations show that the very close agreements were found in predictions at the very early stage of excavation as the shallow excava­ tion in both long and short side of excavation. The maximum wall deflections in prediction were almost the same with observations roundly 45 mm, and 18 mm in the SID8 and SID6. How­ ever, a slight over-predicted in section of SID9 was found in this stage, approximately less than 10 mm in difference for maximum wall deflec­ tions. The small soil movement in shallow exca­ vation induced by the construction stage was obviously well-predicted by using the HSS simu­ lation. Similarly, the deeper excavation as the final excavation stage are also given the good results of prediction in SID9 and SID6, but the exception for some slight underestimations in the SID8. To start with, almost 10 mm of the max­ imum wall deflection in predicting was greater than measurements of SID8 and such difference might be connected with ground variance on site or measurement error. Since the field observation of ground settlement was unavailable in this project, the in situ lateral wall movements were only used to compared with HSS

Figure 7. Comparison of observations and predictions of wall delection in case KLCC (a) Inclinometer 3 (SID8); (b) Inclinometer 4 (SID9); (c) Inclinometer 6 (SID6).

288

model. Fortunately, these predictions came along a comparatively good consistency in both shallow and deep excavation. Against, the underestimated wall deflections such as SID8 is likely related to either the strutting system, which it was re-arranged in this model as suggested by Ang et al. (2018) due to lack of its details of coordinates and spacing, or the soil stiffness of LKLS below, which the soil properties appeared to reflect higher strength in this group layer. 5.2

The ranges of strain level

Figure 8 shows the ranges of strain level from the finite analyses for deep excavations in KHH and KLCC induced by excavation construction sequences. By examining the reduction of secant shear modulus of the soil in the certain depths of 12 m bgl (this depth is equaled to the soil sample was taken in the laboratory test), the shear strain values significantly increase from the first excavation stage to the final excavation stage (i.e. 0.014 to 0.054 % in case KHH, and 0.064 to 0.66 % in case KLCC). However, there was only Case KHH is lied on the range of strain level for retaining wall during construction sequences as suggested by Mair (1993) as 0.01 % to 0.1 %. Meanwhile, Case KLCC was out of this strain range when the excavation reached its final exca­ vation stage. The larger horizontal wall displace­ ment at its final stage is likely to the fact that the excavation would produce larger shear strain. In addition, it is obviously seen that the labora­ tory degradation curve of sandy soil at 10 and 12 m bgl is able to represent the soil behavior in simulations subjected by HSS model. Therefore, this degradation curve can be obtained to inter­ pret the small strain behavior in similar ground condition. 5.3

Strain level in PMT tests

It is widely known that the PMT can be carried out by either using the stress-controlled or strain-controlled procedures. In Case KLCC, several pre-bored

Figure 9. Strain level exploration in PMT tests.

pressuremeter (PBP) tests have been conducted and those results were used to estimate the soil stiffness as the input parameters for geotechnical analyses and foundation design. This test is produced by applying stress-controlled method which the changes in soil volume were measured after pressure application at equal intervals of time, the soil stiffness is obtained from the initial elastic segment of the pressure-volume change curve. This measured stiffness, in other words, is obtained at relatively large shear strains (Shanaid, 2009). Figure 9 shows the large range of strain level amongst the three PMT tests at various depths in sand which the soil stiffness was interpreted by measured the changes in volume of the probe during the process of pressure supply as following the Equation 8 below:

where, ε is the strain level of the PMT tests, ΔV is the increments of probe volume, and Vo is the initial volume of probe. It is clearly seen that the range of strain level to interpret the soil stiffness is compara­ tively large, roundly 18 to 35 % in loose to medium dense sand and 44 to 50 % in dense to very dense sand, respectively. As shown the range of strain level from two laboratory tests (Figure 4), the small strain level of these stress-controlled PMTs can be recognized in order to predict the lateral wall move­ ment more reasonable if the ratio E/Eo is roundly equal to 0.15. Based on the stiffness degradation curve above, the ranges of strain level for loose to medium dense sand, thus, should be able to control in the ranging about from 1 % to 2 %. 6 CONCLUSION The following conclusions can be drawn from the results of this study:

Figure 8. Strain level exploration in simulation.

1. It is suggested that a linear regression analysis is applicable to represent the relationship between

289

2.

3.

4.

5.

shear module at small strain, G0, versus the depth of loose to medium dense sand, Z for in such ground. The reference threshold shear strain, γ0.7, of 0.0001 can be used as an input parameter for the HSS model to employ the sandy soil small strain behavior. The degradation curve of shear modu­ lus of loose to medium dense sand can be drawn by either conducting the dynamic triaxial or column resonant tests because those tests results converge at γ0.7 around 0.0001 (Figure 4). A comparatively well consistency is reached between the predictions and field measurements by using the HSS model by examining the ground and structural behavior of deep excava­ tion in sand, particularly for loose and medium dense sand. The key soil small strain input parameters such as Go and γ0.7 have been inter­ preted in this study those values play a vital role to predict well in prediction of excavation behav­ ior by using HSS model. The ranges of strain level of retaining wall in soft soil are generally ranged from 0.01 to 0.1 %, however, it could be out of this range when excessive large horizontal displacement was gen­ erated due to the launching of the final excavation stage. In addition, the degradation curve in laboratory test can be obtained to interpret the small strain behavior of the excavation within its construction sequences. It is suggested that the stress-controlled method in PMT should be re-considered to address the small strain level; hence, the strain level could be ranged about from 1% to 2 % in loose to medium dense sand. This range of small strain level aims to achieve the reasonable results of soil stiffness once the stress-controlled method has been adopted on sites.

REFERENCES Ang J. S., Hsiung B. C. B. and Ching Hung (2018). Plane strain ratio and waling size evaluation of deep excavation of Kuala Lumpur using 3d finite element analysis. Proceeding 20th SEAGC – 3rd AGSSEA Conference in conjunction with 22nd Annual Indo­ nesian National Conference on Geotechnical Engin­ eering. Jarkata – Indonesia, 6 – November 2018. ISSN 978-602-17221-9. Pp. 193–201 Atkinson, J.H., Sallfors, G., (1991). Experimental deter­ mination of soil properties. In: Proceedings of the10th ECSMFE, vol.3, Florence, pp.915–956. Brinkgreve, R.B.J., Bakker, K.J., Bonnier, P.G., (2006). The relevance of small-strain soil stiffness in numer­ ical simulation of excavation and tunneling projects. In: Numerical Methods in Geotechnical Engineering. 6th European Conference in Geotechnical Engineer­ ing. Taylor and Francis, Graz, Austria, pp. 133–139. Finno, R. J., Blackburn, J. T., & Roboski, J. F. (2007). Three-dimensional effects for supported excavations in

clay. Journal of Geotechnical and Geoenvironmental Engineering, 133(1), 30–36. Fernando Schnaid (2009). In Situ Testing in Geomecha­ nics- The Main Tests. Taylor & Francis, London and Newyork. Hsiung, B. C. B., Yang, K. H. (2017). Displacement of ground and diaphragm wall induced by deep excava­ tions in loose to medium dense sand. Proceedings of the 19th International Conference on Soil Mechanics and Geotechnical Engineering, Seoul, 2017 Hsiung, B. C. B., Yang, K. H., Aila, W., and Hung, C. (2016). Three-dimensional effects of a deep excavation on wall deflections in loose to medium dense sands. Computers and Geotechnics 80: 138–151. Likitlersuang, S, Teachavorasinskun, S, Surarak, C, Oh, E and Balasubramaniam, A (2013). Small strain stiffness and stiffness degradation curve of Bangkok Clays. Soils and Foundations 53(4): 498–509. Law K. H., Ismail Z. and Roslan H. (2016). 3D Finite Element Analysis of a Deep Excavation Considering the Effect of Anisotropic Wall Stiffness. 19th Southeast Asian Geotech­ nical Conference & 2nd AGSSEA Conference(19SEAGC & 2AGSSEA)Kuala Lumpur, pp659–664. Law K. H., Siti Z. O., Roslan H and Zubaidah I. (2014). Determination of Soil Stiffness Parameters at a Deep Excavation Construction Site in Kenny Hill Formation. Measurement, Vol. 47, pp. 645–650. Mu, L. and Hunag, M. (2016), Small strain based method for prediction three- dimensional soil displacements induced by braced excavation, Tunnelling and Under­ ground Space Technology, Volume 52, 12–22 Mair, R.J. (1993). Developments in geotechnical engineer­ ing research: application to tunnels and deep excavations. Proceedings of Institution of Civil Engin­ eers: Civil Engineering, 93. pp. 27–41. Nikolinakou, M. A., A. J. Whittle, S. Savidis& U. Schran (2011). Prediction and interpretation of the performance of a deep excavation in Berlin sand. Journal of Geotechnical Geoenvironmental Engineer­ ing, 137(11): 1047–1061 Ou, C. Y. (2006). Deep Excavation: Theory and Practice. Taylor & Francis, Netherlands Phan K.H., Hsiung B.C.B., Giang H.N. (2019). Re-categorized soil layers and Performance Predic­ tion of Deep Excavation in Kuala Lumpur. The 3rd international conference on transport infrastructure with sustainable development. Construction Pub­ lisher, Hanoi, Vietnam. PLAXIS BV., 2016. PLAXIS 3D User’s Menu. Delft, The Netherlands. RTRI. 2012. Design guideline of cut-and-cover tunnel for railway structure. Tokyo, Japan (in Japanese) S.L. Kramer (1996). Geotechnical Earthquake Engineering. Prentice Hall, Upper Saddle River, New Jersey; pp. 653. Schanz, T., Vermeer, P. A., and Bonnier, P. G. (1999). The hardening soil model: formulation and verification. Beyond 2000 in Computational Geotechnics, Balkema, Rotterdam, pp281–296. Schweiger HF. (2009). Influence of constitutive model and EC7 design approach in FEM analysis of deep excava­ tions. PLAXIS, ISSMGE International Seminar on Deep Excavations and Retaining Structures, vol. 3D. Buda­ pest, pp. 99–114, 2016. Viggiani, G., Atkinson, J.H., (1995). Stiffness of fine-grained soil at very small strains. Géotechnique 45(2), 249–265.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Prediction of damage intensity of reinforced concrete tunnels and soil against blast loading K. Senthil & S. Rupali Department of Civil Engineering, Dr. BR Ambedkar NIT Jalandhar, India

L. Pelecanos Department of Architecture and Civil Engineering, University of Bath, UK Department of Civil Engineering, Dr BR Ambedkar NIT Jalandhar, India

ABSTRACT: In the present study, a finite element technique was used to study the dynamic response of tunnels against underground explosions using finite element software ABAQUS, Explicit Analysis. The reinforced concrete tunnel of 0.8 × 0.8 × 4.0 m size along with the boundary of soils were mod­ elled using FE three dimensional analysis. The inelastic behavior of concrete and steel bar has been incorporated through concrete damage plasticity model and Johnson-cook models available in ABAQUS. The Drucker-Prager model as well as acoustic infinite medium have been used in order to model the damage behavior of soil and tunnel. The simulated results thus obtained from the present study were compared with the experimental results available in literature and found in close agreement. Further, the simulations were carried out on the effect of important parameters such as standoff dis­ tance of detonation and mass of TNT in order to predict the damage intensity in tunnel as well as soil strata.

Keywords: Tunnels, Reinforced concrete, Blast load, Dynamic analysis, Finite element method

1 INTRODUCTION Underground tunnels used for roadway and railway, utility lines and water pipelines are indivisible part of the modern civil infrastructure. In the recent dec­ ades, explosion incidents caused by terrorist activ­ ities have proved to be a growing threat to the human civilization and the civil infrastructure. Among the various schemes that terrorists may use, bombing is one prime option, is called explosion. Explosion inside a underground structure could dir­ ectly threaten the lives of people inside and it might also damage the structure and cause further loss of lives and properties. Preventive measures should therefore be implemented not only to significantly reduce the possibility of terrorist attacks, but also to protect existing subway structures from collapse under internal blast loading. Hence, in order to safeguard the tunnels, it is necessary to understand the response of these structures under blast load. Lu (2005) concludes that the neural network approach proves to be successful and able to predict the unseen test data with satisfactory accuracy. The results also provide new insight into the signifi­ cance of some influencing variables such as the

path orientation on the ground shock parameters. Lu et al. (2005) found that the 2D model predict reasonably accurate results in terms of the crater size, blast loading on the structure, and the critical response in the front wall. Otherwise, the response in the remaining part of the structure shows notice­ able differences between two dimensional and three dimensional models. Choi et al. (2006) performed three- dimensional finite element (FE) analyses in order to develop the chart to assess the response of tunnel against blast loading. The tunnel damage assessment charts were developed from the numer­ ical analysis and provided a methodology for simple and practical application to vulnerability assessment of tunnels. Gui and Chien (2006) con­ cluded that a designer may adopt dynamic soil parameters which are obtained from the good ground investigations as well as soil testing may result in a more economical analysis. Ngo et al., (2007) presents a comprehensive overview of the effects of explosion on structures and they were introduced different methods to estimate blast loads and structural response. Liu (2009) concluded that Grouting to improve the stiffness of soil around subway tunnel could be an effective mitigation

DOI: 10.1201/9780429321559-37

291

measure to increase blast resistance. Chen et al (2014) performed experiment in order to provide believable results of dynamic responses of largespan structures, including the dynamic loads, deflections, strains and failure modes. It was con­ cluded that subjected to close-in explosions, the buried arch deforms at a dominant flexural mode accompanying with compression mode. Yang et al. (2010) observed that the upper part of the tunnel lining cross-section with directions ranging from 0° to 22.5° and horizontal distances 0 to 7 m away from the explosive center are the vulnerable areas, and the metro tunnel might be safe when tunnel depth is more than 7 m. Soheyli et al (2016) per­ formed large scale experiment as well as simula­ tions on the underground tunnels and predicted the response in terms of acceleration at 1.69 and 2.76 kg mass of TNT and this study is helpful for engineers and future theoretical and numerical stud­ ies. Senthil et al. (2018a and b) and Singhal et al. (2018) performed numerical analysis on the build­ ing frame in order to understand the influence of mass of TNT, Standoff distance, boundary condi­ tions etc. and it was demonstrated very well the accuracy and effectiveness of the continuum finite element based numerical models in the chosen problems. In addition to that, Di Murro (2019) and Koon Lok and Pelecanos (2019) have performed numerical investigations on the behavior of tunnels through experiments as well as simulations. Based on the detailed literature survey, it is observed that effectiveness and accuracy of continuum level finite element model on the behavior of underground tun­ nels subjected to external blast loading is not obtained due to complex nature of loading and soilstructure interactions. Therefore, the present work is focused on the behavior of underground tunnels by external blast loading using the 3D nonlinear finite element analysis. The response of the tunnel against external blast loading was studied using ABAQUS/Explicit and the findings are compared with the available experimental results. In addition to that, the influence of standoff distance and mass of TNT on the behavior of reinforced concrete tunnel is studied in order to verify the robustness and effectiveness of the continuum level finite element models in light of predicting the problems. 2

CONSTITUTIVE MODELLING

2.1

Johnson-Cook model for reinforcement

The flow and fracture behavior of reinforcement material was predicted employing the JohnsonCook (1985) elasto-viscoplastic material model available in ABAQUS finite element code. The material model is based on the von Mises yield cri­ terion and associated flow rule. It includes the effect of linear thermo-elasticity, yielding, plastic flow, isotropic strain hardening, strain rate harden­ ing, softening due to adiabatic heating and damage. Johnson and Cook (1985) extended the failure cri­ terion such as strain path, strain rate and tempera­ ture in the fracture strain expression, in addition to stress triaxiality. The section of the reinforcement was assigned 460 MPa and the strength and failure properties proposed by Borvik et al., (2001), have considered in the present study, however the yield strength of the steel reinforcement was 340 MPa by Soheyli et al (2016). 2.2 Concrete damaged plasticity model for concrete In finite element modelling, inelastic behaviour of concrete was defined by using concrete damaged plasticity model (CDP) providing a general capabil­ ity for modelling concrete and other quasi-brittle materials. The model assumes that the two main failure mechanisms are tensile cracking and com­ pressive crushing of the concrete material. The evo­ lution of the yield surface is controlled by two hardening variables which are linked to failure mechanisms under tension and compression load­ ing, namely εcpl and εtpl are compressive and tensile equivalent plastic strains, respectively. The damage variables can take values from zero to one, where zero represents the undamaged material and one represents total loss of strength. The stress strain relations under uniaxial compression and tension loading are given by the following equations where Eo is the initial (undamaged) elastic stiffness of the material: σt = (1-dt)Eo(εt-εtpl) and σc = (1-dc)Eo(εc -εcpl), where dt and dc are tension damage variable and compression damage variable respectively. The concrete damaged plasticity model parameters such as (dt) tension and (dc) compression damage vari­ ables are shown in Tables 1 - 5.

Table 1.

The inelastic behavior of concrete has been incorp­ orated through concrete damage plasticity model and the model includes compressive and tensile behavior. The elastic and plastic behavior of steel reinforcement bar has been incorporated using Johnson-cook model includes the effect of state of stress, temperature and strain rate and discussed in this section. The plastic behavior of soil, acoustic infinite elements as well as the characteristics of the TNT was defined using

Material constants for concrete material.

Description

Numerical Value

Density (kg/m3) Young’s Modulus (N/mm2) Poisson’s ratio Dilation angle Eccentricity(m) K σb0/σc0

2400 19700 0.2 35° 0.1 0.66 1.16

292

Table 2.

Concrete compressive behavior.

Yield stress (N/mm2)

Inelastic strain

Yield stress (N/m2)

Inelastic strain

20.0 19.8 19.6 19.4 19.1 18.8 18.5 18.1 17.7 17.4 17.0 16.6

0 0.00015 0.00025 0.00035 0.00045 0.00055 0.00065 0.00075 0.00085 0.00095 0.00105 0.00115

16.3 15.9 15.5 15.2 14.9 14.5 14.2 13.9 13.6 13.3 13.0

0.00125 0.00135 0.00145 0.00155 0.00165 0.00175 0.00185 0.00195 0.00205 0.00215 0.00225

Table 6.

Material constant for soil.

Density (kg/m3)

Elastic modulus Poisson (N/mm2) ratio

Dilatation angle

Friction angle

Flow stress ratio

1850

29

1

31

0.778

Table 6. The Drucker-Prager model shear criterion was considered as linear and Drucker-Prager harden­ ing behavior was defined as compression which was having yield stress versus absolute plastic strain. The cohesive behavior such as normal stiffness, shear stiffness and tensile strength was 315, 82 and 1 MPa respectively, also defined between tunnel and soil as contact property. 2.3

Table 3.

Concrete tensile behavior. 2

Yield stress (N/m )

Cracking strain

1.80 1.50 0.60 0.10 0.05

0 0.000127097 0.000246433 0.000651263 0.000805343

Drucker-Prager model for soil

The simplification of Mohr-Coulomb model where the hexagonal shape of the failure cone was replaced by a simple cone was known as the Drucker-Prager model (1952). Generally, it shares the same advantages and disadvantages with the Mohr-Coulomb model. In Drucker Prager Model the yield is circular, from the centre to the yield surface it is equidistance and the properties of soil are shown in 3 FINITE ELEMENT MODELLING

Table 4. Concrete compression damage. Damage parameter dc

Inelastic strain

Damage parameter dc

Inelastic strain

0 0.006 0.015 0.027 0.041 0.057 0.074 0.092 0.110 0.129 0.148 0.166

0 0.00015 0.00025 0.00035 0.00045 0.00055 0.00065 0.00075 0.00085 0.00095 0.00105 0.00115

0.18 0.20 0.22 0.23 0.25 0.27 0.28 0.30 0.31 0.33 0.34

0.0012 0.0013 0.0014 0.0015 0.0016 0.0017 0.0018 0.0019 0.0020 0.0021 0.0022

Table 5.

0.36

Concrete tensile damage.

Damage Parameter

Cracking Strain

0 0.40 0.69 0.92

0 0.00012 0.00024 0.00065

The finite element model of the soil, reinforcement, concrete and acoustic infinite element were made using ABAQUS/CAE. The length, width and height of the model was 12.0, 4.0 and 4.5 m respectively, considered in the present study. The thickness of tunnel wall was 100 mm and clear cover is 50 mm on both the side of wall and the size of the tunnel was 0.8 × 0.8 m internal clear square as exactly pro­ posed by Soheyli et al., (2016), see Figure 1. The single layer of main as well as transverse steel reinforcement of 8 mm diameter placed at 100 mm center to centre distance, see Figure 1(a). The geometry of the soil, concrete and steel reinforce­ ment were modelled as solid deformable body, Figure 1(a)-(e). The interaction between concrete and steel was modelled using the tie constraint option available in ABAQUS/CAE wherein the concrete was assumed as host region and the steel as embedded region. The constitutive and fracture behavior of steel and concrete have been predicted using Johnson-Cook and Concrete damaged plasti­ city model respectively available in ABAQUS (2018). The origin of blast considered against 4 m from the exterior top surface of slab and 1.69 kg mass of TNT using CONWEP model with AIR BLAST definition.

293

Figure 1. Tunnel (a) reinforcement (b) concrete (c) left side view (d) front view (e) right side view (f) acoustic infinite element and (g) isometric view of combined finite element model.

The TNT blast material interacts as an incident wave with tunnel lining and surrounding soil. The acoustic infinite element was modelled as acoustic medium having bulk density of 1500 MPa and density of 110 kg/m3 was defined in order to define the exterior boundary for the soil strata, see Figure 1(f). In Abaqus/ Explicit the possibility of using acoustic infinite elem­ ents to model the effect of the exterior fluid is explored. The use of acoustic infinite elements removes the need of impedance-type absorbing bound­ ary conditions on the outer boundary. Acoustic infinite elements ACIN3D4 were defined on the outer bound­ ary of this row. The fake contact was defined between the soil exterior surface and acoustic infinite elements. The surface-to-surface contact was assigned between the tunnel lining and surrounding soil using contact algorithm available in ABAQUS. The tunnel was con­ sidered as master and the contact surface of soil as slave surface and the complete model of the soilstructure is shown in Figure 1(f). The results thus obtained were compared with experimental data by Soheyli et al (2016), after detailed mesh convergence study. A detailed mesh sensitivity analysis has been car­ ried out to understand the influence of mesh size. The size of transverse reinforcement was 5 mm and considered as linear line element type T3D2. The size of main reinforcement was also 5 mm and linear hexahedral element of C3D8R type and the size of the reinforcement was kept constant for all the simu­ lations in light of mesh convergence study. The size of soil element at the right side was considered as 150 mm whereas the size of element at right size was 400 mm however, the mesh size near the

explosion zone was kept 125 mm. The size of acous­ tic infinite element was also considered similar to soil strata. The mesh sensitivity was studied by vary­ ing the mesh size only on the concrete tunnel. The size of concrete element was varied as 0.05 × 0.05 × 0.05 m, 0.04 × 0.04 × 0.04 m, 0.03 × 0.03 × 0.03 m and 0.02 × 0.02 × 0.02 m. Total number of elements in case of 50, 40, 30 and 20 mm mesh size of concrete were 11520, 26800, 63973 and 191400, respectively. The predicted acceleration and vonMises stresses in the concrete tunnel were compared corresponding to varying mesh size, see Figure 2 and 3. The maximum predicted acceleration at the tunnel was found to be almost same, i.e. 2.3 g against varying mesh size, see Figure 2, whereas the measured acceleration through experiment was 2.37 g in the front wall of at points positioned 125 mm away from the central axe of the tunnel and 400 mm lower than the roof, see Figure 4. It is observed that the acceleration in the tunnel corres­ ponding element size of 50 mm was found in good agreement with the experimental results considering the fact that computational cost. The lowest compu­ tational time in case of 50 mm mesh was 34 hours. It is also observed that the Mises stresses in concrete was found to be 1.09, 1.44, 1.42 and 1.42 MPa against 50, 40, 30 and 20 mm mesh size respectively, see Figure 3. In light of the von-Mises stresses, the difference in the results between 40 and 50 mm seems insignificant. Therefore, it is concluded that the mesh size of 50 mm for concrete was found to be suitable for further analysis considering the less computational time and cost. On the basis of detailed mesh convergence study, 50 × 50 × 50 mm size was considered for further simulations. Hence, the total number of linear hexahedral element of C3D8R having 402888, linear line element type T3D2 having 28080, linear quadrilateral elements of type ACIN3D4 having 1918, quadratic tetrahedral elem­ ents of type C3D10M having 5735 and the total number of elements for the standard simulations was 438621.

Figure 2. Acceleration in concrete tunnel at (a) 50mm (b) 40mm (c) 30mm and (d) 20 mm mesh size.

294

Figure 3. von-Mises stresses in concrete tunnel at (a) 50mm (b) 40mm (c) 30mm and (d) 20 mm mesh size

Figure 5. Predicted acceleration (m/s2) in the soil (longitu­ dinal section) at (a) 0.0012 (b) 0.016 (c) 0.0228 and (d) 0.0264 Sec.

4 VALIDATION OF FINITE ELEMENT RESULTS The simulations were carried out against 1.69 kg mass of TNT at a distance of 4.0 m from the surface of the front wall. Theh concrete damaged plasticity model has been employed for predicting the material behavior of the concrete, whereas the Johnson-Cook model has been used for predicting the material behavior of steel reinforcement. The Drucker-Prager model has been employed for predicting the behav­ iour of soil. In addition to that, the acoustic infinite elements were used in order to remove the need of impedance-type absorbing boundary conditions on the outer boundary. The three dimensional finite element modelling of the tunnel and soil were discussed in Section 3. The simulated results thus obtained have been compared with the experiments carried out by Soheyli et al (2016) and are discussed in this Section. The actual and predicted acceleration of the tunnel has been compared in Figure 4. The acceleration was predicted from the node 14533 which was almost near to the location where the same was measured from the experiment. Overall, the predicted numerical results

accurately predicted the pattern of acceleration throughout the time step and found in good agreement with the experiments. In general, a maximum deviation of almost 20% has been found between the actual and predicted peak acceleration in the tunnel. Based on this observations, it is concluded that the present study suc­ cessfully demonstrates the accuracy and effectiveness of the finite element models of the tunnels. The acceleration in the soil strata was also studied and the acceleration path at 0.0012, 0.016, 0.0228 and 0.0264 Sec, were shown in Figure 5. It is observed that the maximum acceleration generated at the time of detonation, 1931 g, whereas the accel­ eration of 10.5 g moves radially and reaches the tunnel front wall at 0.02 sec, see Figure 5(c). 5 IMPACT OF VARYING STANDOFF DISTANCE AND MASS OF TNT The simulations were carried out in light of import­ ant parameters such as varying standoff distance and mass of TNT. The response of tunnels was observed in terms of deflection, velocity, acceleration and vonMises stresses therein were presented and discussed. 5.1

Varying standoff distance

The simulations were carried out against varying standoff distance such as 4, 4.5 and 5 m from the surface of the tunnel front wall at constant mass of TNT, i.e. 1.69 kg, please see Figure 6. The behaviour

Figure 4. Comparison of Experimental and numerical simulations with 1.69 Kg mass TNT at the distance of 4 m from the explosion charges.

Figure 6. Finite element modelling of tunnel location at (a) 4.0 (b) 4.5 and (c) 5.0 m.

295

of reinforced concrete tunnel in terms of acceleration, velocity, deflection and Mises stresses at varying standoff distance is shown in Figure 7 - 10. The acceleration in slab against blast load of 1.69 kg mass TNT originated at 4, 4.5 and 5 m from

Figure 7. Effect of Varying standoff distance from the explosion charges.

Figure 8. Velocity in the tunnel at 0.12 Sec of (a) 4, (b) 4.5 and (c) 5.0 m standoff distance. Figure 10. Mises stress in the tunnel front wall at (a) 0.024 (b) 0.036 and (c) 0.048 Sec of (i) 4, (ii) 4.5 and (iii) 5.0 m standoff distance.

Figure 9. Displacement in the tunnel at 0.12 Sec of (a) 4, (b) 4.5 and (c) 5.0 m standoff distance.

the central axis of the tunnel front wall is shown in Figure 7. The maximum acceleration was found to be 1.9, 1.46 and 1.02 g against 4, 4.5 and 5 m standoff distance respectively. It is also clearly seen that the acceleration reaches its peak value within 0.022 Sec i.e., from the time of detonation in case of 4 m, see Figure 7. However, the peak acceleration was observed at 0.025 and 0.027 Sec in case of 4.5 and 5 m standoff distance. The impulse velocity on the tunnel due to the blast load of varying standoff distance at 0.12 Sec is shown in Figure 8(a)-(d). The velocity was found to be 0.019, 0.016 and 0.014 m/s against 4, 4.5 and 5.0 m standoff distance, respectively.

296

It is observed that the decrement of velocity is almost consistent with increase of standoff dis­ tance, see Figure 8. Maximum displacement in the tunnel predicted at 0.12 Sec by 4, 4.5 and 5.0 m standoff distance, see Figure 9. The maximum displacement of tunnel against blast load originated at 1.5, 2, 5 and 8 m from the sur­ face is shown in Figure 9(a)-(c). The maximum deflec­ tion was found to be 2.49, 1.94 and 1.65 mm against 4, 4.5 and 5.0 m standoff distance respectively. The max­ imum deflection on tunnel was about 2.49 mm against 4 m distance and it is found to be more vulnerable among the chosen case. The deflection of slab against 4 m standoff distance was found decreased to 28 and 50% as compared to the deflection of slab by 4.5 and 5.0 m respectively. However, overall it was observed that the systems doesn’t experience damage against any of the case. The von-Mises stresses in concrete against blast load of 1.69 kg mass of TNT at varying standoff dis­ tance is shown in Figure 10(a-i)-(c-iii). At 4 m standoff distance, the stresses in concrete are 1.28, 0.18 and 0.14 MPa at 0.024, 0.036 and 0.048 Sec respectively. At 4.5 m standoff distance, the stresses in concrete are 0.345, 0.088 and 0.11 MPa at 0.024, 0.036 and 0.048 Sec respectively. Similarly, at 4.5 m standoff distance, the stresses in concrete are 0.026, 0.18 and 0.11 MPa at 0.024, 0.036 and 0.048 Sec respectively. However, it is observed that the stress in concrete was found to be almost in the range of 1.5-0.1 MPa. The tunnel experienced high­ est stress i.e. 1.28 MPa at 0.024 second for 4.0 m standoff distance. Also, it is observed that the stress in concrete was found to be decreased when the standoff distance increase from 4.5 to 5.0 m. Therefore, it is concluded that The maximum stress in the concrete was found to be in the range of 1 to 1.5 MPa and is almost constant for varying standoff distance, however, the stress in concrete was found decreasing with increasing of standoff distance. 5.2

Figure 11. Acceleration function of time in the tunnel against varying mass of TNT at node 14533.

increased by 45 and 126% compared to the 1.69 kg mass of TNT. However, the acceleration at time 0.016 Sec was found to be 18.3, 27.1 and 43.8 g against 1.69, 2.76 and 5 kg mass TNT respect­ ively. In case of the acceleration at time 0.0228 Sec, it was found to be 1.05, 1.57, 2.57 g due to 1.69, 2.76 and 5 kg mass TNT respectively. It is observed that the intensity of acceleration was decreased by almost one hundred times within 0.016 Sec and decreased by one thousand times when it reaches the target point. Also, it is observed that the acceleration was found to be almost dissi­ pated by tunnel may be due to the distance, see Figure 12(c). However, it is concluded that the acceleration of tunnel was found to increase with increase of mass of TNT. 6 CONCLUSIONS

Varying mass of TNT

The simulations were carried out against varying mass of TNT such as 1.69, 2.76 and 5 kg at constant standoff of distance, i.e. 4.0 m. The behaviour of reinforced concrete tunnel in terms of acceleration against varying mass of TNT is shown in Figure 11. The maximum positive acceleration was found to be 1.9, 2.85 and 4.62 g against 1.69, 2.76 and 5 kg mass TNT respectively at the time 0.022 Sec. The acceleration in the soil against blast load of varying mass TNT originated at 4.0 m from the sur­ face of the wall is shown in Figure 12(a-i)-(d-iii). The unit of the acceleration contours was “m/s2”. The maximum acceleration was found to be 1931, 2816 and 4380 g against 1.69, 2.76 and 5 kg mass TNT respectively, at the time of detonation, i.e. 0.0012 Sec. It was observed that the acceleration by 2.76 and 5 kg mass of TNT was found to be

297

The present study focuses on the finite element investigation on the behavior of underground tun­ nels against external blast load. The simulations were carried out against 1.69 kg mass of TNT at a distance of 4.0 m from the surface of the front wall. The concrete damaged plasticity model has been employed for predicting the material behav­ ior of the concrete, whereas the Johnson-Cook model has been used for predicting the material behavior of steel reinforcement. The DruckerPrager model has been employed for predicting the behaviour of soil. In addition to that, the acoustic infinite elements were used in order to removes the need of impedance-type absorbing boundary conditions on the outer boundary. From the investigations, the findings were compared with the available experimental results and further, the simulations were carried out in light of influ­ ence of standoff distance and mass of TNT and the following conclusions have been drawn. • The maximum acceleration in the tunnel obtained from simulation was 1.9 mm which is very close

within 0.016 Sec and decreased by one thousand times when it reaches the target point, i.e. 0.022 Sec. However, it is concluded that the acceler­ ation of tunnel was found to increase with increase of mass of TNT.

ACKNOWLEDGEMENT This work is part of the research project “Safeguard­ ing civil underground tunnels against terrorist blast and explosion shocks”. The authors highly acknow­ ledge the financial support provided by The Royal Society, UK, the research grant No. IES\R2\181054 for carrying out the present study.

REFERENCES

Figure 12. Predicted acceleration in the soil of (i) 1.69 (ii) 2.76 and (iii) 5 kg mass of TNT at (a) 0.0012 (b) 0.016 (c) 0.0228 and (d) 0.0264 Sec.

to the experimental results of 2.3 g. The actual and predicted pattern of the acceleration-time history as a result of blast load has been compared and almost exact pattern of acceleration-time history has been predicted through numerical simulations. • The maximum stress in the concrete was found to be in the range of 1 to 1.5 MPa and is almost con­ stant for varying standoff distance, however, the stress in concrete was found decreasing with increase of standoff distance. • It is observed that the intensity of acceleration was decreased by almost one hundred times

ABAQUS/CAE User’s Manual, SIMULIA. Version 6.14. Borvik, T., Hopperstad, O.S., Berstad, T. & Langseth, M. (2001) A computational model of viscoplasticity and ductile damage for impact and penetration. European Journal of Mechanics and Solids 20: 685–712. Choi, S., Wang, J., Munfakh, G. & Dwyre, E. (2006) 3D Nonlinear blast model analysis for underground structures, Geocongress 2006. Chen, H., Zhou, J., Fan, H., Jin, F., Xu, Y., Qiu, Y., Wang, P. & Xie, W. (2014) Dynamic responses of buried arch structure subjected to subsurface localized impulsive loading: Experimental study. International Journal of Impact Engineering 65: 89–101. Di Murro V, Pelecanos L, Soga K, Kechavarzi C, Morton R.F. & Scibile L. (2019) Long-term deformation monitoring of CERN concrete-lined tunnels using dis­ tributed fibre-optic sensing, Geotechnical Engineering Journal of the SEAGS & AGSSEA, 50, 1–7. Gui, M. & Chien M. (2006) Blast-resistant analysis for a tunnel passing beneath Taipei Songshan airport–a parametric study. Geotechnical and Geological Engin­ eering 24: 227–48 Johnson, G.R. & Cook, W.H. (1985) Fracture characteris­ tics of three metals subjected to various strains, strain rates, temperatures, and pressures”, Eng. Fracture Mech., 21, 31–48. Koon Lok Nip & L. Pelecanos, (2019) Simplified numer­ ical load-transfer finite element modelling of tunnelling effects on piles, Nature, Tunnels and Underground Space, 21(2) 117–129. Liu, H. (2009) Dynamic Analysis of Subway Structures under Blast Loading. Geotechnical and Geological Engineering 27(6): 699–711. Lu, Y. (2005) Underground blast induced ground shock and its modelling using artificial neural network. Computers and Geotechnics 32:164–178. Lu. Y., Wang. Z. & Chong. K., (2005) A comparative study of buried structure in soil subjected to blast load using 2D and 3D numerical simulations. Soil Dynamics and Earthquake Engineering 25: 275–288. Ngo, T. & Mendis, P. (2009) Modelling the dynamic response and failure modes of reinforced concrete struc­ tures subjected to blast and impact loading. Structural Engineering and Mechanics 32(2): 269–282.

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Senthil. K., Rupali, S. & Kaur, N. (2018a). “The performance of monolithic reinforced concrete structure includes slab, beam and column against blast load”, Journal of Materials and Engineering Structures 5(2): 137–151. Senthil, K., Singhal, A., & Shailja, B. (2019) Damage mechanism and stress response of reinforced concrete slab under blast loading, Coupled Systems Mechanics, Vol. 8, No. 4, 315–338. Singhal, A., Senthil, K., & Shailja, B. (2018) Influence of boundary condition and mass of TNT on the behaviour of concrete slab under blast loading, Proc. Nat. Con. Advanced Structures, Materials and

Methodology in Civil Engineering, 3-4, November 2018, NIT Jalandhar, India, pp. 179–186. Soheyli, M.R., Akhaveissy, A.H., & Mirhosseini, S.M. (2016) Large-scale experimental and numerical study of blast acceleration created by close-in buried explosion on underground tunnel lining. Shock and Vibration 2016, Article ID 8918050. Yang, Y., Xie, X. & Wang, R. (2010) Numerical simu­ lation of dynamic response of operating metro tunnel induced by ground explosion. Journal of Rock Mechanics andeotechnical Engineering 2(4): 373–384.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

EPB-TBM tunnel under internal pressure: Assessment of serviceability

N.A. Labanda & A.O. Sfriso SRK Consulting, Argentina

D. Tsingas & R. Aradas Jacobs, Argentina

M. Martini Salini-Impregilo S.p.A., Italy

ABSTRACT: The Mantanza-Riachuelo basin recovery is one of the most ambitious environmental projects under construction in Argentina. In this context, the sanitary bureau of the metropolitan area of Buenos Aires (AySA) is building a sewage collection network to transport the waste water of the population in the southern area of the city, composed by almost five million people. The most complex tunnel in this big project is named Lot 3, an outfall EPB-TBM tunnel starting at a shaft located at the Rio de la Plata margin and running under the river 12 km to a discharge area. The tunnel runs through soft clay belonging to the post-pampeano formation and dense sands of the Puelchese formation. In operation, it will be pressurized by a pumping station which will produce a piezometer head that, in the first 2000 m, might be eventually higher than the confining pressure around the tunnel. This paper presents the numerical analysis of the structural forces acting on the tunnel rings using a riskoriented approach that considers the stochastic nature of materials, stratigraphy and tunnel-ground inter­ action. The compression of the lining is evaluated and compared with field measurements in order to predict the structural forces and the risk of the rings going into tension beyond the structural capacity of the system.

1 INTRODUCTION The Matanza-Riachuelo river flows along the ripar­ ian lands between the City and the Province of Buenos Aires (Argentina). It is a water stream 64 km long that flows into the Rio de la Plata. The river is heavily contaminated with industrial and residential wastewater discharged from both mar­ gins with limited -if any- pre-discharge water treat­ ment process. The Matanza-Riachuelo basin recovery project is an ambitious plan to manage the wastewater from the left margin of the Rio de la Plata through a sewerage interceptor tunnel, a treatment plant and a discharge tunnel into the Rio de la Plata. The project is divided into several contracts. One of them, named Lot 3, holds the outfall tunnel, an EPB-TBM tunnel running 35 meters below the riverbed of the Rio de la Plata. A global scheme of the outfall, which is the sub­ ject of this paper, is presented in Figure 1. The tunnel has an internal diameter of 4.3 m and a total length of 12 km, including a 1.5 km

DOI: 10.1201/9780429321559-38

300

diffusor zone with 34 standpipe risers, 28 m long, that daylight in the riverbed. It is being driven through soft clays of the Postpampeano formation and dense sands from the Puelchense formation. Water flows by gravity from a pumping station onshore built into the access shaft. Figure 2 shows the longitudinal elevation of the tunnel proposed in pre-feasibility stage. The tunnel will be subjected to a maximum internal pressure of 550 kPa and an external water pressure from 330 to 420 kPa, resulting in a net outwards pressure 130 to 220 kPa. Further infor­ mation about the tunnel and its design can be found in Aradas et al. (2019). EPB-TBMs employ muck to balance water and soil pressure at the tunnel face. By managing the excavation param­ eters i.e., screw conveyor speed, advance rate, thrust force, face support pressure, etc., the ground relaxation and, consequently, structural forces in the lining can be somewhat controlled. This aspect, usually not of high interest, is important in the case of segmental pressurized tunnels working at high internal pressure.

2 CONSTITUTIVE MODELING FOR SOIL AND LINING CONCRETE

Figure 1. Location of Matanza-Riachuelo outfall (Buenos Aires, Argentina). The thin red line is the border between the city of buenos aires and the province of buenos aires. The Riachuelo river is along the south border.

A common feature found when reviewing case histories of pressurized tunnels in soft ground is the recommendation to neglect the contribution of the external ground confinement as a safety redun­ dancy. This hypothesis, despite being conservative from the perspective of confinement, produces an unrealistic estimate of the deformation of the lining ring and therefore of the resulting bending moments. On the other hand, the adoption of a nonzero effective external ground pressure is challenging as it is largely dependent on the TBMsoil interaction during excavation and on the long­ term creep behavior of the soil. An assessment of the compression induced in the segmental rings by ground pressure via sto­ chastic numerical models and calibrated using back-analysis after field measurements is pre­ sented in this paper. Two rings are studied in this proposal: N � 237 and N � 497, embedded in soft clays and dense sands, respectively.

The tunnel was designed in pre-feasibility stage to be driven through dense sands, as shown in Figure 2, where yellow layers represent the Puelchense forma­ tion. However, field conditions departed from this assumption in several places. For instance, highplasticity clays belonging to the Paranaense forma­ tion were detected in the conveyor belt (Figure 3) betwen chainages PK 0+294 to PK 0+586. Ring N � 237 placed at PK 0+316 and consequently, embedded in clay, and Ring N � 497 placed at PK 0 +681, embedded in dense sand, were instrumented with strain gauges whose results allow for evaluating the compression load of the tunnel lining in two dif­ ferent grounds. The Hardening soil small (HSsmall) constitu­ tive model was used to simulate the mechanical behavior of soils and calibrated by in situ and lab tests and local experience. Statistical variabil­ ity of the friction angle was estimated using the 6-sigma method. A correlation between the com­ pression index Cc and the liquid limit ωL for Postpampeano clays was first reported by (Sfriso 1997) with a COV ¼ 0:30, and was later updated by(Ledesma 2008) as presented in Figure 4, with a COV ¼ 0:12. Swelling index Cs , relevant for

Figure 3. High-plasticity clay detected in the TBM’s con­ veyor belt.

Figure 2. Matanza-Riachuelo outfall: longitudinal elevation of the tunnel proposed in pre-feasibility stage.

301

Table 1.

Figure 4. Compression index Cc versus liquid limit for the post-pampeano formation.

the estimation of the interaction forces, correlates with Cc by:

HSsmall requires the definition of elastic and hardref ref ref , E50 and Gref ening parameters Eoed , Eur 0 , among other parameters. They were computed as

Model Drainage γ ¢0 c0 ψ Gref 0

γ0:7 Eref ur Eref 50 Eref oed m lur OCR K0nc Rinter k

Constitutive models and parameters used. Unit

Soft clay

Sandy soils

kN=m3

HSsmall Undrained (A) 16.0 P.D. 0 0 P.D. 10-3 P.D. P.D. P.D. 1.0 0.20 1.00 1 - sin ¢0 0.90 0.005

HSsmall Drained 20.5 32 0 0 200 10-4

120 40 40 0.5 0.20 1.00 1 - sin ¢0 0.70 1.00

�

kPa �

MPa MPa MPa MPa 10-6 m/s

where Ired is the reduced moment of inertia, Ireal is the real moment of inertia and n is the number of lining segments in each ring. 3 STATISTICAL CHARACTERIZATION 3.1

being lur the Poisson modulus, K0 the at rest earth pressure coefficient, pref the reference pressure and e0 the initial void ratio. Uncertainties in the properties of the clay where the ring N � 237 is placed were dealt with by consid­ ering a stochastic approach to bracket the prediction of the structural forces acting on the lining. The parameters employed are presented in Table 1, where some properties are defined using probability distributions that will be described in the next sec­ tion. A deterministic set of parameters were used for dense sand because its influence is negligible for the ring under analysis. The structural stiffness of the lining was reduced to account for the segment joints using the MuirWood expression (Muir Wood 1975)

Soil parameters

The proposed risk analysis requires that the variability of soil parameters be defined together with the bound­ aries presented in Equation (4). A statistical character­ ization for the effective friction angle ¢0 and the compression index Cc is presented in Figure 5. Prob­ ability distributions were obtained using the experi­ mental data shown in Figure 4 and classical correlations for the friction angle. Considering values a ± σa for parameters Cc , ¢0 and boundary multipliers in Equations (1) and (4), 24 ¼ 16 permutations can be performed to reproduce drained triaxial tests, where a is the mean value and σa is the standard deviation of the considered parameter. Results are presented in Figure 6 and compared with three experimental tests for this project. Good fitting is obtained for consolida­ tion mean pressures of p ¼ 50kPa, p ¼ 100kPa and p ¼ 200kPa. 3.2

Lining segment concrete

The Young’s modulus of concrete is required to com­ pute stresses and forces out of microstrains measured by the monitoring system. A histogram of 480 simple compression tests, a normal probability function and

302

Figure 8. Water level in the Rio de la Plata. Probability dis­ tribution and cumulative distribution.

Figure 5. Probability distribution for friction angle and compressibility index of the Postpampeano formation.

acting on the tunnel lining, making its statistical characterization a critical aspect of the analysis. The Argentine Naval Hydrographic Service provided the time-history evolution of the river over the last few years. Taking the data corresponding to the first semester of 2018, where the rings under analysis were installed, a normal probability function is fitted and presented in Figure 8, and compared with field measurements. A mean elevation 0:945 m and a standard deviation 0:547 m were considered in the analyses.

Figure 6. Drained triaxial test, comparisons between laboratory tests and HSsmall simulations.

4 NUMERICAL RESULTS AND COMPARISON WITH FIELD MEASUREMENTS

a cumulative probability function are shown in Figure 7. In order to compute the concrete stiffness at low strains Eci , the following expression is used (CEB­ FIP 1993)

4.1 Ground relaxation estimate by 3D modeling in ring N � 237

with Eco ¼ 21500MPa, fcmo ¼ 10MPa and fcm unconfined compression resistance of the sample. Considering a mean value fcm ¼ 59:6MPa and a standard deviation σfcm ¼ 4:94MPa, deformations measured by strain gauges can be translated into structural forces. 3.3 Water level in the river Variable water level in the river is one of the key contributors to uncertainty of the compression forces

A three dimensional model of the excavation was developed to estimate the ground relaxation induced by the TBM drive in ring N � 237. The TBM shield was simulated considering a face contraction cref ¼ 0:4% and a tail-to-face contraction increment cinc;axial ¼ 0:03571%/m. A grout pressure equal to the total field stress plus 0.5 bar was applied in the TBMs tail. The face pressure was calibrated with data provided by the sensors in the TBM, imposing a normal surface load equal to the measured value presented in Figure 9. The 3D model is presented in Figure 10. Due to the high computational effort required by this kind of simulations, mean values for all materials were used and a single model is ana­ lyzed. The excavation sequence was repeated until a reasonably stabilized zone was obtained. Settlements in the tunnel crown and effective verti­ cal stresses in the stabilized section are plotted in Figure 11. The mean value for settlement is close to 30 mm, while the effective vertical stress is around 145 kPa. These values were used to fit a ground relax­ ation factor in the two dimensional models presented in the next section. 4.2 Structural forces in ring N � 237

Figure 7. Histogram of unconfined compression test results for 480 samples of lining concrete. Probability distribution and cumulative distribution.

In order to evaluate the risk of tension forces being developed in the ring, a 2D model was developed and the influence of uncertainties in input data in

303

Equations (1) and (4), river water level and concrete quality were studied. The numerical model and construction stages are summarized in Figure 12. The excavation sequence was simulated in 2D by partial stress relaxation using the β-method, i.e. applying EMstage ¼ 1 - β51. Grout pressure in TBMs tail was simulated as a distributed load, the (impervious) lining was activated and consolidation until 99% dissipation of excess pore pressure -the expected condition by the time the tunnel starts operating- was allowed for. In operation, an internal head of 15.5 m, 14.0 to 15.1 meters above the river level was applied. 25 ¼ 32 models were gener­ ated by performing all permutations of data presented in Table 2.

Figure 9. Sensor data from the TBM face.

Figure 12. 2D model of ring N � 237. (a) Excavation using EM stage51 (b) grout pressure (c) Activation of lining and consolidation (d) Internal pressure in operation.

Table 2. Parameter ranges used to compute ground relax­ ation curves.

Figure 10. 3D model of settlements after excavation.

Cc Cs (eq. 1) Eref 50 (eq. 4) ¢0 River Elevation

Unit

Lower bound

Upper Bound

-

0.367 0.1 1.25 24 0.398

0.497 0.2 1.90 26 1.492

�

m

Figure 11. Settlements of tunnel crown and effective vertical stress in stabilized section.

304

Figure 13 shows the obtained ground relaxation curves and both displacement and effective vertical stress measured in the tunnel crown, plotted in terms of EM stage. The purpose of these curves is to obtain, in a 2D model, a stress state similar to the stabilized sec­ tion of a 3D model. By getting into the graph with the mean crown dis­ placement obtained in the 3D model, and intersecting curves corresponding to the 2D crown displacement (in blue), a range of EM stage values were obtained. The range of effective contact stresses obtained rea­ sonably contains the mean value obtained in the 3D model. Values adopted for the analysis are EM stage ¼ ½0:35; 0:40; 0:45; 0:50; 0:55; 0:60]. Tables 2 and 3 show values employed to calculate structural forces in the lining. Together with the EM stage values, a set of 6 x 26 ¼ 384 numerical models were obtained. Figure 14 plots translation and ovalization compo­ nents of the tunnel lining obtained with all permuta­ tions, during construction. The model shows that, for low relaxation values, a little rebound of the tunnel is observed and an ovalization pattern where the vertical stress is larger than the horizontal stress is obtained. While the ground relaxation increases, a tunnel settle­ ment and an invertion in the tunnel ovalization is obtained. Figure 15 shows the obtained structural forces for all models in the construction stage, compar­ ing those with field data measured in ring N � 237 using strain gauges. The actual concrete stiffness was used to translate deformations into forces, as indicated with markers for the mean value and lines for the standard deviation band.

Figure 14. Translation and ovalization of ring N � 237 during construction.

Figure 15. Structural forces of ring N � 237 in construction stage and comparison with field measurements.

It can be seen that the proposed procedure prodcues a good fitting with field data for both bending moments and normal forces. For low EM stage values, hihger normal forces and lower bending moments are obtained, being the last in agreement with the ovalization pattern. If EM stage is higher, normal forces decrease and bending moments increase. Figure 16 shows a comparison of the mobilized shear stress τmob for a 2D model and a stabilized section of the 3D model. For sake of simplicity, a single represen­ tative 2D model is shown but similar stress patterns

Figure 13. Ground relaxation curves for 25 ¼ 32 permuta­ tions in soil parameters and river water level. Ring N � 237.

Table 3. Concrete parameters used in permutations to compute structural forces.

fcm

Unit

Lower bound

Upper Bound

MPa

54.6

64.5

Figure 16. Comparison of τmob between the 2D model and a stabilized section of the 3D model.

305

are obtained in all permutations. It is interesting to see that the mobilized shear, a measure of the tunnel arch­ ing effect, is similar in both cases. After ensuring the validation of the proposed methodology, a study of the tunnel lining in operation was performed and the risk of tension forces being developed was analyzed. 4.3

Structural forces in operation in ring N � 237

Lining forces in operation in ring N � 237 are evalu­ ated in this section. Figure 17 shows the translation and ovalization for the tunnel under internal pres­ sure. Results are similar than those obtained for the construction stage, with a small increment in tunnel settlements due to the waste water weight. Ovaliza­ tion remains almost equal by virtue of the hydro­ static not generating higher deformations compared with the construction stage. Figure 18 plots the structural forces for all considered permutations. Results show that bending moments and shear forces are almost the same than during construction. Normal compression forces, however, decrease dra­ matically from values -1900 ± 100kN before the internal pressure to -200 ± 150kN in service. Des­ pite this fact, even for high relaxation values i.e. EM stage ¼ 0:60, lining is still under compression, and tension in connecting bolts is avoided.

4.4

Structural forces in ring N � 497

Structural forces and deformations in the most adverse chainage of the tunnel, represented by ring N � 497, is presented in this section. The procedure for the back analysis used in this case is different than the used in the previous case, avoiding 3D calculations and fixing the ground relaxation by means of EM stage and field measurements. The simulation procedure is the same than in the previous case. Soil proper­ ties for the soft clay and overlying materials were defined using the mean values presented in previous sections, while parameters for dense sand were considered to be permuted using bounds presented in Table 4. A two dimensional base model considering the stratigraphy corresponding to chainage PK 0 +681 is presented in Figure 19. The construction procedure is the same than in the previous case. The effective friction angle ¢0 was defined as the sum of the constant volume friction angle ¢cv ¼ 30� and the dilatancy angle ψ presented above. Ground relaxation curves for current stratig­ raphy are shown in Figure 20. Displacement are considerably lower than those observed for soft clays in Figure 13. Also, a lower ground mobil­ ization is required to achieve the same confine­ ment pressure on the lining. Concrete parameters presented in Table 3, together with relaxations EM stage ¼ ½0:35; 0:45; 0:55], produce 192 sets of paramteres that were employed to do the

Table 4. Parameters used to compute ground relaxation curves in dense sands.

Figure 17. Translation and ovalization of ring N � 237 in normal operation stage.

ψ Eref 50 Eref ur Gref 0 River Elevation

Unit

Lower bound

Upper Bound

�

8.0 82000 200000 280000 0.398

10.0 98000 240000 400000 1.492

[kPa] [kPa] [kPa] m

Figure 19. 2D base model for ring N � 497. (a) Excavaction with EM stage (b) grout pressure application (c) Lining construction (d) Tunnel under internal pressure in normal operation.

Figure 18. Structural forces of ring N � 237 in normal operation.

306

during construction are plotted in Figure 22, where vertical displacements are negligible while ovalization tends to increase with ground relaxation. 4.5

Figure 20. Ground relaxation curves for 25 ¼ 32 permuta­ tions in soil parameters and river water level. Ring N � 497.

stochastic analysis. Results are expressed in terms of bending moments, normal and shear forces during construction, and are shown together with field measurements for ring N � 497 in Figure 21. Good fitting is found between numerical and experimental results in bending moments, while predictions in normal forces tend to underestimate the compression forces in the lining. Tunnel translation and ovalization

�

Structural forces in operation, ring N � 497

Using the calibrated numerical model, an assess­ ment of serviceability in operation (under internal pressure) is presented in this section. Bending moments, normal and shear forces in the lining are presented in Figure 23, while translation and ovalization are shown in Figure 24. Similar to ring N � 237, internal pressure in the tunnel does not change much the bending moments, shear forces and lining deformations. However, normal forces were affected, being the internal pressure produced by the waste water transportation similar or even higher than the ground and water pressure applied in the tunnel lining. In this aspect, the behavior is different than the section embedded in clay; dense sands produce less pre-compression in the lining, thus increasing the probability to experience tensile forces in the joint bolts.

Figure 23. Structural forces of ring N � 497 in operation.

Figure 21. Structural forces of ring N 497 in construction stage and comparisons with field measurements.

Figure 24. Translation and ovalization of ring N � 497 in operation.

Figure 22. Translation and ovalization of ring N � 497 during construction.

307

5 CONCLUSIONS Pressurized TBM tunnels are traditionally designed neglecting the pre-compression induced by the soil-structure interaction, largely due to the uncertainties involved in their estimation. Despite enormous progress in the monitoring equipments and technology, and also in the numerical model­ ing tools available, it is still a challenging task to estimate a lower-bound contact pressure induced in the tunnel lining, reliable enough for being employed in the joint bolt design. A procedure for the risk assessment and the computation of the probability to have tension forces in the tunnel lining during the operation were presented and discussed in this paper. The stochastic nature of some relevant parameters was acknowledged, and the calibration of probability density distributions was done based on experience, published data and usual correlations for selected crit­ ical parameters. In the same way, a statistical analysis of the river level and the properties of the concrete used to manufacture lining segments was detailed and used in the simulations. A first stage, consisting in a convergence analysis using 3D simulations of the excavation sequence was briefly discussed, and mean values for vertical crown displacement and effective stress were shown and used to define an equivalent ground relaxation for the 2D models. After defining a reasonable range for EM stage, a set of 6 x 26 ¼ 384 permutations were used in the 2D numerical model to estimate a range for the struc­ tural loads. A comparison with results obtained by the strain gauges during the tunnel construction in two rings was presented, showing a good agreement

between the predictions and field measurements. Using this validation, the structural forces in service were estimated, concluding that the probability to support tension forces in this stage are negilgible for rings embedded in soft clay, while it increases for rings embedded in dense sands if the ground relaxation is not properly controlled.

ACKNOWLEDGMENTS The authors would like to acknowledge the Argentine Naval Hydrographic Service (www. hidro.gov.ar) for providing the water level timehistory of the Rio de la Plata and Salini­ Impregilo-Chediak for permission to publish the data contained in this paper.

REFERENCES Aradas, R., D. Tsingas, & M. Martini (2019). The design of a segmentally lined tunnel for a large sewer outfall. lote 3 - emisario planta riachuelo, argentina. In World Tunnel Congress 2019, Naples - Italy. International Tunnel Asociation. CEB-FIP (1993). MODEL CODE 1990. Thomas Telford Publishing. Ledesma, O. (2008). Calibración del Cam Clay para suelos del post-pampeano. Buenos Aires, Argentina: Tesis de grado de Ingeniería Civil - Universidad de Buenos Aires. Muir Wood, A. (1975). The circular tunnel in elastic ground. Géotechnique 25(1), 115–127. Sfriso, A. (1997). Formación post-pampeano: predicción de su comportamiento mecánico. CLICJ - Caracas, Vene­ zuela, 1–10.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

A risk assessment of downdrag induced by reconsolidation of clays after upwards pipe jacking N.A. Labanda & A.O. Sfriso SRK Consulting, Argentina Universidad de Buenos Aires, Argentina

D. Tsingas & R. Aradas Jacobs, Argentina

M. Martini Salini-Impregilo S.p.A., Italy

ABSTRACT: Salini-Impregilo is building part of the largest sanitary sewer system in the history of Argen­ tina in the suburbs of Buenos Aires City, to serve a population of almost five million people. The project is an outfall TBM tunnel 12 km long, starting from a reception shaft in the river margin, and transporting the sewage 35 meters below the Rio de la Plata riverbed to the point of discharge. Within the final kilometer of the tunnel, a set of 36 standing pipes so-called risers are constructed by driving steel tubes upwards and pass­ ing through dense sands, sandy clays and soft clays. Risers are linked-up with the launching lining segment using flange unions. Driving of risers upwards will generate excess pore pressure and disturbance in fine soils and, once the pipe is placed in its final position, negative skin friction due to reconsolidation and creep. A risk assessment of the downdrag is presented in this paper, based on the estimation of the force and/or displacement in the risertunnel union generated by this effect. The issues of whether it is desirable to instalock the riser-tunnel union at an early age after installation of the riser and the time lapse required to reduce negative skin friction effects are discussed. Results are validated by comparing the model results with field measurements in prototype models.

1 INTRODUCTION The Matanza-Riachuelo river flows along the ripar­ ian lands between the City and the Province of Buenos Aires (Argentina). It is a water stream 64 km long and 35 m wide in average that flows into the Rio de la Plata. The Riachuelo river is considered as one of the most polluted water courses in the world, ruined across the years by the discharge of several kind of untreated industrial and waste water. The Matanza-Riachuelo basin recovery project is an ambitious plan to manage the wastewater from the left margin of the Rio de la Plata through a sewerage interceptor tunnel, a treatment plant and a discharge tunnel into the Rio de la Plata. The pro­ ject is divided into several contracts. One of them, named Lot 3, holds the outfall tunnel, an EPB-TBM tunnel running 35 meters below the riverbed of the Rio de la Plata. A scheme of the plant together with the design of the system at conceptual level is shown in Figure 1.

The tunnel has an internal diameter of 4.3 m and a total length of 12 km including a 1.5 km diffusor zone with of 34, 28 m long standpipes named ‘risers’ that daylight in the riverbed. Risers are lifted from the interior of the tunnel by driving upwards steel tubes. A construction sequence of risers installation is shown in Figure 2. Driving the risers will generate excess pore pressure and dis­ turbance in the fine surrounding soils and, once the pipe is placed in its final position, negative skin friction due to reconsolidation and creep of soft clay layers. Due to an optimization of the design, the tunnel elevation was lifted and, consequently, the length of the risers reduced from 30.5 m to 28.0 m, thus reducing the thickness of sandy layers to be crossed. While this optimization is beneficial to reduce the jacking forces through dense sands, it does worsen the downdrag phe­ nomenon due to the loss of the positive skin fric­ tion provided by the sands. This could generate two scenarios:

DOI: 10.1201/9780429321559-39

309

Figure 1. Conceptual design of the Matanza-Riachuelo outfall. Detail of launching rings and risers in diffusor zone.

Figure 2. Risers construction stages.

� 30, σs0uv � 0:24j0:28 and fine content above the 85% (Sfriso 1997). The ‘Puelchense’ formation can be found below the post-pampeano, between 27 to 31 meters depth from the riverbed, composed by a dense clean sands with an in-situ relative density above 80%, an effect­ ive friction angle in constant volume ¢cv � 32� and a dilatancy angle ψ � 6� j8� . Between the puelchense and post-pampeano for­ mation, a transition of 5 m to 11 m thickness com­ posed by interleaved layers of soft clays and medium to dense silty sands is encountered. Figure 3 shows a detailed stratigraphy of the site, where the transition layer is represented as fundamentals soil layers classi­ fied by CPTu in-situ test and Robertson charts (Roberton 2009). It also shows the trace of the mentioned tunnel, the position of the CPTu tests used to calibrate the numerical models and, in red, the considered risers for our analysis. In order to illustrate the soil classification in the transition, Figure 4 plots the Robertson chart for the layer in CPTu-01A and CPTu-05A. It can be seen that the mentioned stratum is mainly composed by normally and under consoli­ dated soils, with an erratic distribution of soft clays and sandy soils. The underconsolidation is produced by a constant upflow coming from the puelchense for­ mation, a confined aquifer with an over pressure with respect to the river of, approximately, 2 m. It is clear

• Scenario N� 1: If the riser is not bolted to the tunnel, a vertical displacement is experienced by the pipe, slightly increasing the local hydraulic loss. • Scenario N� 2: If the riser is bolted to the tunnel, the negative skin friction will produce a nonnegligible force acting on the crown lining segment. A risk assessment using a simplified model to esti­ mate displacements (scenario N� 1) or forces (scenario N� 2) produced by negative skin friction in risers due to consolidation and creep is presented in this paper. A brief introduction to the geotechnical site character­ ization is presented in Section 2, while the constitutive models employed are presented in 3. The numerical model and its results about downdrag, structural forces in risers and comparisons with field test results are shown in 4. Finally, some conclusions are drawn in Section 5. 2

Figure 3. Site stratigraphy, risers location and crown elevation.

GEOTECHNICAL SITE CHARACTERIZATION

The mentioned project is placed below the Rio de la Plata, where first 20 to 25 meters depth from the riv­ erbed are composed by normally consolidated soft clays and silty clays stemmed by fluvial depositions, known as the ‘Post-pampeano’ formation. This for­ mation is characterized by a liquid limit ωL varying from � 30 to � 60, plasticity index from � 10 to

Figure 4. Robertson charts for the transition for CPTu-01A and CPTu-05A.

310

in the site profile that, as the crown elevation is lifted, the thickness of sandy soils across the risers decreases. This is crucial for our problem because, as is demonstrated in this paper, sandy soils provides a confinement pressure in the pipe diminishing dis­ placement or forces in the riser-tunnel union.

Table 1 .

Model Drainage γ ¢0 c0 ψ Cc Cs Cα Gref 0 γ0:7 ref Eur Eref 50 Eref oed m lur OCR K0nc Rinter k

3 CONSTITUTIVE MODELLING Soft clays in the riverbed of ther Rio de la Plata are anisotropically normally consolidated soils, suffi­ ciently ancient to assume that primary and secondary consolidation has been occurred. Nevertheless, during risers driving, an intensive soil remolding is carried out in the surroundings of the pipe, till a distance of 4 to 5 times its diameter. The soil disturbance, gener­ ated at constant humidity, produce excess pore pres­ sure erasing the stress history of the material. Once the driving is ended, excess pore pressure dissipation and reconsolidation in the influenced area, produce downdrag forces and, consequently, negative friction in the riser. Soft soil creep model is used for clays to take into account the time-dependent behavior pro­ duced by remolding. At the same time sandy layers lose densification, approaching the effective friction angle to the constant volume friction angle, reducing the dilatancy angle to 0� and, consequently, its con­ finement potential. Hardening soil small is used to simulate the mechanical behavior of this kind of soils. Figure 5 shows experimental results of compress­ ibility index as a function of the liquid limit, obtained by unidimensional consolidation test for the post­ pampeano formation, with a comparison with Terza­ ghi and Skempton correlations (Sfriso 1997). The 1 swelling index is considered as Cs ≈ 15 j 10 Cc and the secondary compressibility index for the mentioned formation is Cα ¼ 0:013 (Ledesma 2008). Param­ eters for sandy soils are obtained from classical cor­ relations based on the standard penetration test (SPT) and the transition layer is simulated as a composition of soft clays and sands layers. The geotechnical

Figure 5. Compressibility index Cc versus liquid limit for the post-pampeano formation. Comparisons with Terzaghi and Skempton expressions.

Constitutive models and parameters used. Unit

Soft clay

Sandy soils

kN=m3

Soft Soil Creep Undrained (A) 16.0 24 0 0 0.410 0.080 0.013 0.20 1.00 0.51 0.80 0.005

HSsmall Drained 19.0 32 0 0 200 10-4 120 40 40 0.5 0.20 1.00 0.60 0.70 1.00

�

kPa �

MPa MPa MPa MPa 10-6 m/s

profile for each riser is calibrated by comparing pore pressures calculated with the numerical model and CPTu measurements. Table 1 summarize constitutive models and parameters used for our simulations. Figure 6 shows a comparisons between consolidateddrained triaxial test of samples taken from the post­ pampeano formation and the constitutive model cali­ bration. Soft soil creep has a limitation to reproduce the mechanical behavior in low pressures in low strains. Nonetheless in this particular problem, the numerical model works in large strains and the pro­ posed constitutive model reproduce reasonably the maximum deviatoric resistance of soft clays. Figure 6 shows a comparisons between consoli­ dated-drained triaxial test of samples taken from the post-pampeano formation and the constitutive model calibration. Soft soil creep has a limitation to reproduce the mechanical behavior in low pressures in low strains. Nonetheless in this particular

Figure 6. Constitutive model calibration and comparisons with CD triaxial test.

311

problem, the numerical model works in large strains and the proposed constitutive model reproduce rea­ sonably the maximum deviatoric resistance of soft clays. The analyzed phenomena is not only depend­ ent on the geotechnical parameters, but also has a strongly dependence with the riser-soil inter­ action. In this sense, it is important to define a proper friction angle for the interface. Several research papers has been published regarding this matter for driven pile, combining different mater­ ials such as concrete, wood and steel with different confining pressures. It has been proved that the fric­ tion angle for the steel-soil interaction is independ­ ent with the contact tension and the saturation degree and, for smooth steel, ¢0 ¼ 24� (Potyondy 1961). This value is adopted for our numerical models. 4 NUMERICAL MODELLING OF DOWNDRAG 4.1 Soil stress state after pipe jacking Some authors have proposed theoretical and experi­ mental models to estimate the strain path in the sur­ rounding of a penetration test. Baligh (1985) provides a framework to study driven piles, assess­ ing the disturbance of soil and showing that the strain produced by a driving maneuver is comparable to a radial cavity expansion since 2 radii from the tip (see Figure. 7). Other researchers arrived to similar conclusions (Pestana et al. 2002, Chong 2013). An axial-symmetric model is considered for our ana­ lysis, where a displacement equals to 35.5 cm, the riser radius, is imposed from the rotation axis. After this, a plate to model the metal sheet of the riser is used to simulate the soil-structure interaction. Figure 8 shows a detail of the mentioned stages. Table 2 details characteristics of each analyzed riser and the closest CPTu test, taken as reference to val­ idate the geotechnical profile. It also summarize the sandy layers thickness crossed and riser length in both considered crown elevation. This data is used latter to correlate both, displacement and structural forces, with the presence of sands. The assessed geotechnical profiles in each riser is plotted in Figure 9 for the former elevation, where soft clays are represented with green and sandy soils are represented with pink. Remolded clays, with a noticeable creep effect, are represented with grey and it occupies 2 diameter in riser’s surrounding. After the radial cavity expansion, the riser is installed and an excess pore pressure is generated in soft clays. Figure 10 shows the total pore pressure field in the former crown elevation, that has to be dissipated in the consolidation procedure. It is important to note that models for the actual crown elevation considers the same geotechnical profile, but ends at 28.0 m depth.

Figure 7. Total Strain path in a driven pile surroundings and comparison with a radial cavity expansion (Baligh, 1985).

Figure 8. Construction stages of the finite element model.

4.2 Comparisons of numerical results with CPTu test Obtained total pore pressures in our numerical models are compared with CPTu reference tests per­ formed closest to the their future place, and the

312

Table 2 . Characteristics of analyzed risers and reference undrained Cone Penetration Tests (CPTu). New elevation

Former elevation

Sand layers Riser thickness CPTu [m] N�

Riser Sand layers length thickness [m] [m]

Riser length [m]

1 3 5 9 12 16 21 23 27 34

28.0 28.0 28.0 28.0 28.0 28.0 28.0 28.0 28.0 28.0

30.5 30.5 30.5 30.5 30.5 30.5 30.5 30.5 30.5 30.5

01 01A 01B 02 03 04 05 05A 06 07

2.7 1.0 2.7 3.5 2.8 1.9 6.0 5.0 4.2 1.3

5.3 1.0 5.2 6.0 5.3 4.4 8.5 7.5 6.7 3.8

Figure 9 . Geotechnical profiles proposed for each con­ sidered risers. Former crown elevation.

Figure 10. Numerical model results of total pore pressure in each riser. Former crown elevation.

results are presented in Figure 11. Informed pore pressures are surveyed in the soil-riser interface. It is interesting that our proposed model, des­ pite its simplicity, reproduce accurately the excess pore pressure measured with CPTu. Whether the cone penetration test is performed or the riser is driving, produce a relatively similar pore pressure profile due to the strain range at both cases are working. As is shown in Figure 6, after 12% of axial strain, the deviatoric stress reaches a plateau and, for an undrained test, the pore pressure is nearly stabilized.

Figure 11. Comparison between CPTu test (black line) and numerical results obtained after a radial cavity expansion (red line).

After riser collocation, a consolidation stage is calculated considering a lifetime of 100 years. As was described in the introduction, two scenarios are studied in the risk analysis: a non bolted riser is

313

analyzed in Scenario N� 1, allowing a vertical degree of freedom in the pipe to be displaced inside the tunnel. This case is presented in Section 4.3. Scen­ ario N� 2, presented in Section 4.4, considers a vertical fixity in riser base to calculate the applied force in the launching ring. 4.3

Scenario 1 - Time evolution of riser downdrag

Riser vertical displacement is conceptually given by

where δ is the total vertical displacement, δe is the elastic bouncing right after the driving maneuver, δc is the consolidation component and δcr is the creep component. It is well known that elastic bouncing displacement in driving piles is negligible. In this way, the δe component is not considered in our paper focusing our curiosity in time-dependent terms. Riser displacements over time due to primary con­ solidation δc (solid line) and due to primary and sec­ ondary δc þ δcr (dashed line), are plotted in Figure 12 for each considered riser in former crown elevation. Results obtained for the same analysis but in actual crown elevation, are presented in Figure 13. A hypothetical geotechnical profile where the riser does not cross any sandy layer, named ‘pure soft clay’, is included in both cases to estimate an upper bound for the results. When sand lenses are coarser, displacement are less than 3 mm after 100 years in all cases with the

exception of riser N� 3, where there are not sands over the tunnel crown, neither in actual or former elevation. In this case, displacements in the end of the lifetime is 35 mm and the 90% of the value is reached after 3 month since its installation. A completely different set­ ting is observed for the actual crown elevation, where risers N� 1, 3, 5, 9, 12, 16 and 34 have maximum dis­ placements highers than 10 mm, being a critical condi­ tion for the local hydraulic loss. Similarly to riser N� 3 in former elevation, 90% of maximum displacements are reached after 2 or 3 month since their installation. Collecting all maximum displacement obtained in both elevations and ordering them in terms of the sandy soil thickness crossed, it can be seen in Figure 14 that, while sands thickness increases over 4.0 m, values are stabilized between 1 and 2 mm and the creep effect is almost negligible because the confinement effect pro­ duced by coarser soils. The mentioned effect produced by sands generates an increment in structural forces in pipes. Figure 15 shows the maximum normal force reached in risers in terms of the sand soil thickness crossed, for both elevation. It is clear that, when the

Figure 14. Maximum displacements in terms of sand soil thickness crossed by riser in both elevations in the end of the lifetime.

Figure 12. Riser displacement due to primary and second­ ary consolidation in former crown elevation.

Figure 13. Riser displacement due to primary and second­ ary consolidation in actual crown elevation.

Figure 15. Maximum normal force in riser in terms of sand soil thickness crossed by riser in both elevations in the end of the lifetime.

314

equals, increasing in scenario N� 2 on the vicinity of the tunnel, being the maximum value is within the transition layer. If the geotechnical profile is composed mainly with soft clays, differences between both scenarios are notorious and the

displacement freedom is blocked by some reason, the structural element is more stressed. When sand layers are less than 3 m, the maximum normal force is almost constant between 900 and 1400 kN. From 3 m and up, the normal compression increase dramat­ ically, reaching values nearly to 2800 kN. Different from the displacements, the presence of sandy soils has not a blocking effect of creep in the structural forces. 4.4 Scenario 2 - Structural forces induced by consolidation When risers are bolted to the launching ring, the ver­ tical displacement is restricted and consolidation and creep of soft clays generates a concentrated force into the crown lining segment. This new scenario produce a rearrangement of internal stresses in pipes. Figure 16 shows maximum normal forces for riser-tunnel union (shaded in blue) and for all pipes (shaded in red), in the end of the lifetime, in terms of crossed sand layer thickness. When sand thickness is less than 3 m, the maximum normal force is placed in the riser-tunnel union and are almost three time higher than the maximum normal force reached in scenario N� 1. While the sand thickness increases, the force applied in the riser-tunnel union decreases to the half or less, but the maximum normal force in all over the riser is remains almost unalterable. In order to illustrate the internal distribution of normal and circunferential structural forces of two extrt4eme cases, diagrams for risers N� 1 and N� 3 are plotted in Figures 17 and 18 respectively. It can be seen in all cases that normal forces in the first day from the installation are related with its own weight and, while time elapses, normal compression increases considerably. In circumferential normal diagrams the opposite effect occurs, being higher right after the riser driving and decreasing with time. When presence of sandy soils are considerably, normal forces in both scenarios are almost

Figure 16. Maximum normal forces in risers in terms of sand soil thickness crossed in both elevations in the end of the lifetime.

Figure 17. Normal forces for risers N� 1 and N� 3 in both scenarios.

Figure 18. Circumferential forces for risers N� 1 and N� 3 in both scenarios.

315

maximum normal value is in the tunnel crown. Circumferential forces has their maximum values coincidentally with sand layers and differences between scenarios are negligible no matter the presence of sand layers, but being clear its contri­ bution to the riser confinement. 5 CONCLUSIONS The driving procedure of piles and tubes it is still a challenging task to scientist and engineers. In this paper, a simple but powerful finite element model has been proposed to estimate the downdrag phe­ nomena and the negative friction due to consolida­ tion and creep, calibrating the soil profile using CPTu test. Our proposal deals with the identification and modeling of a layered stratum, characterizing the transition as a composition of simple geotech­ nical units and calibrated comparing the pore pres­ sure obtained in our numerical models with those obtained in field tests, simulating the penetration strain path as a radial cavity expansion. After the riser driving, a negative skin friction in the soil-structure interface due to the consolidation and creep is generated due to clay remolding, moving the pipe inside of the tunnel in Scenario N� 1. In Scenario N� 2, where the riser is bolted in the launching ring, the mentioned phenomenon produce a non-negligible concentrated force in the crown tunnel. In both scenarios, displacements or the forces, has a strongly dependence with the sand layer crossed by the riser. In Scenario N� 1 a maximum displace­ ment of 3.5 cm is reached by Riser N� 3, and it was proved that a lifting in the elevation of the tunnel implies an increasing movement in the rest of pipes,

due to a less thickness in crossed sand layers. It was also showed that an increment in sand layers rebounds in a considerable reduction of the creep component and, after 60 to 90 days, the majority of the movement has been performed. When sand layer thickness is larger than 4 m, the riser displacement is negligible. If the pipe is bolted to the launching ring, the force expected in the union range from 2900 kN to 1000 kN, being also the maximum force within the riser if the sand layer thickness crossed is less than 2.5 m. While the thickness increases, the force in the tunnel crown decrease and the maximum normal force is moved to the transition layer.

REFERENCES Baligh, M. (1985). Strain path method. Journal of Geotech­ nical Engineering, 1108–1136. Chong, M. (2013). Soil movements due to displacement pile driving. Case Histories in Geotechnical Engineer­ ing, 1–13. Ledesma, O. (2008). Calibración del Cam Clay para suelos del post-pampeano. Buenos Aires, Argentina: Tesis de grado de Ingeniería Civil - Universidad de Buenos Aires. Pestana, J., E. Hunt, & J. Bray (2002). Soil deformation and excess pore pressure field around a closed-ended pile. Journal Of Geotechnical and Geoenvironmental Engineering 102, 1–12. Potyondy, J. (1961). Skin friction between various soils and construction materials. Geotechnique, 339–353. Roberton, P. (2009). Interpretation of cone penetration test - a unified approach. Canadian Geotechnical Journal 46, 1337–1355. Sfriso, A. (1997). Formación post-pampeano: predicción de su comportamiento mecánico. CLICJ - Caracas, Vene­ zuela, 1–10.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Small-scale modelling of pile drilling in sand – investigation of the influence on surrounding ground E.J. Lande & S. Ritter Norwegian Geotechnical Institute, Oslo, Norway

E.J. Lande, H. Tyvold & S. Nordal Department of Civil and Environmental Engineering, Norwegian University of Science and Technology, Trondheim, Norway

ABSTRACT: Overburden drilling of casings for tieback anchors and micro piles can cause considerable excess short-term ground displacements in the surroundings. The mechanisms of drilling effects on soil dis­ placements are not fully understood. In this context, a physical modelling approach was chosen to study the installation effects of overburden drilling. This paper presents a small-scale model test of overburden drilling in a medium dense saturated sand. A detailed description of the experimental setup and test procedure is described. Specifically, the modelling of the drilling procedure including penetration, rotation and flushing is discussed. Tests with water flushing show that there is a clear relation between the flow rate and penetration rate, and the ability to achieve fluidization of the soil in front of the drill bit. Increasing flow rates caused larger excess pore pressures and increased the influence area in the surrounding soil. Under this specific model conditions increased flow rate also generated more excess drill cuttings compared to the installed pile volume.

1 INTRODUCTION Sheet pile walls and tieback anchors that are drilled into bedrock are often used to support deep excava­ tions in soft soils with limited depth to bedrock. Moreover, micro piles and large diameter steel tube piles are frequently drilled from the final excavation level, and subsequently used as pile foundations for future structures. This type of drilling, which is char­ acterised by permanent casings that are drilled through soil and into bedrock, is often called over­ burden drilling. The influence on surrounding ground and neigh­ boring structures from deep excavations has been the subject of investigations for many decades. Results from Peck (1969), Mana & Clough (1981) and Karlsrud & Andresen (2008) show that ground settlements in areas surrounding a deep excavation in soft clay can range from about 0.5% to 2% of the final excavation depth. Langford et al. (2015) ana­ lyzed several deep excavations in Norway. The results showed that ground settlements caused by initial and secondary effects from installation of drilled tieback anchors and bored piles from inside deep excavations can be significantly larger than 2% of the excavation depth. The main causes for ground settlements were explained by three main factors:

• Horizontal displacements of support wall. • Erosion and loss of soil volume around casings for anchors. • Consolidation of clay due to reduced pore pres­ sures above bedrock. Recent case studies where overburden drilling has led to excessive ground settlements have been reported by Konstantakos et al. (2004), Küllingsjö (2007), Bredenberg et al. (2014) & Sandene et al. (2020). Sandene et al. (2021) presents results from a 9 m deep excavation in soft clay in Oslo, Norway. Excessive ground settlements occurred behind the sheet pile wall during and right after the tieback anchors were drilled into bedrock. Based on observa­ tions during drilling it is likely that the use of air driven down-the-hole (DTH) hammer have caused significant erosion and cavities around the anchor casings in a layer of silty sand above bedrock. Field data however, is inherently affected by vari­ ous uncertainties. For this reason, a conclusive inter­ pretation of the obtained results is often difficult, and the mechanisms of drilling effects on soil displace­ ments and pore pressures are not fully understood. While extensive research on driven piles and auger cast-in-place piles exists, there has been limited

DOI: 10.1201/9780429321559-40

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research on installation effects of overburden drill­ ing, e.g. Lande et al. (2020), Asplind (2017) and Ahlund & Ögren (2016). In this context, a physical modelling approach was chosen to study the installa­ tion effects of overburden drilling. This paper presents an experimental 1g model test that aims to replicate pile drilling with a continuous casing i.e. overburden drilling, in saturated sand. A novel test set-up made it possible to drill a smallscale pile with simultaneously penetration, rotation and flushing with water or air through the pile to transport the drill cuttings. This contribution investigates the influence of varying water-flushing parameters on the surrounding ground. 2 EXPERIMENTAL SET UP 2.1

Model tank and pile

Figure 1 presents a 3D drawing of the experimental set up. Model tests were carried out in a cube shaped steel tank (box) with a length and a width of 900 mm and a height of 840 mm. A linear actuator with a stroke length of 300 mm and a push capacity of 8000 N was connected to a reaction frame on the top of the model tank. Penetration force was measured by a load cell with a capacity of 5000 N connected between the actuator and a motor used to rotate the pile. The rotation motor had a swivel unit that made it possible to flush with water (or air) through the pile drill string and drill bit at the same time as it rotated

Figure 1. Experimental set up.

and penetrated. Both the penetration rate and rotation speed of the pile (rpm) was controlled by adjusting the voltage on the respective power supplies. Pile penetration was measured with an extensometer con­ nected to the frame and rotation motor. Figure 2 shows details of the model pile and drill bit. The pile had a length of 890 mm and consisted of a casing, i.e. steel tube with an outer diameter of 35 mm and 2 mm thickness. The pile diameter was adapted to the model tank to limit potential boundary effects. The mechanical design was based on a prototype concentric drill system, resulting in a scale ratio of about 1:3.2 between the diameter of the model pile (35 mm) and the prototype (114 mm). The dimensions (i.e. cross-sectional area) of the flush­ ing tube and flushing inlet channels in the drill bit as well as the annulus for the backflow were all based on the prototype to obtain representative flushing con­ ditions. The drill bit was connected to the bottom of the casing/pile. At the face/front of the drill bit, four flushing holes with a diameter of 4 mm were machined. The flushing medium (water or air) was applied from the swivel device on the top of the pile and continued through the flushing tube and accessed the pile through the flushing holes. Between the casing and the flushing tube was a middle steel tube that created an annulus against the outer casing. The flushing backflow went up through this annulus and out of the pile through holes in the casing at the pile top. A small catchpot was used to collect the backflow (water and soil) during drilling. The upper part of the drill bit has a packer system (i.e. rubber membrane) that enabled to “close” the annulus between the casing and middle tube, so that soil particles could be collected at the end of each test. The packer was activated by pumping air into it.

Figure 2. Model pile. Cross-section (left) and drill bit excluding casing (right).

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2.2

Instrumentation

Figure 3 provides an overview of the model test set-up including used instrumentation. Pore pres­ sure was measured with six sensors (PPs) con­ nected to separate standpipes (i.e. Ø4 mm plastic tubes) installed at two different soil depths (Zs = 170 and 370 mm) and with three radial distances from the pile center (r = 70, 140 and 210 mm) as can be seen in Figure 3b. The standpipes were sup­ ported and kept at the correct position by fastening them to three vertical steel rods (Ø10 mm) con­ nected to a steel plate on the bottom of the model tank. Vertical displacements of the soil surface (i.e. settlements) were measured with four linear vari­ able differential transformers (LVDTs) positioned at different distances from the pile (Figure 3c). A gantry (template for pile in Figure 3c) was used to keep the pile and the LVDTs in position during the tests. 2.3

Model sand and preparation

Baskarp sand No. 15 from Sibelco AB was used for all tests. This is a graded fine sand with welldocumented properties based on extensive laboratory investigations by among others Ibsen & Bødker (1994) and Ibsen et al. (2009). Table 1 show the main properties of the sand. Perforated plastic tubes were placed at the bottom of the model tank in a permeable filter layer of about 70 mm Leca marbles (Figure 3b and 3c). The filter layer and sand were separated with a geotextile. After filling the tank with dry sand water was pumped through the perforated tubes to saturate the sand. The water level was kept constant about 30 mm above the soil surface throughout the tests by using a weir at the top of the tank. An effective soil model preparation technique based on a procedure described by Foglia & Ibsen (2014) was adopted. This procedure did not require to remove and refill the sand between each test. The soil model preparation involved the fol­ lowing steps:

Table 1. Properties of Baskarp Sand No. 15 (after Ibsen & Bødker, 1994). Property

Unit

Value

D50 grain size D60/D10 Specific gravity, Gs Maximum void ratio, emax Minimum void ratio, emin

mm kN/m3 -

0.14 1.78 2.64 0.858 0.549

Figure 3. Model test setup: a) plan view, b) cross section A-A and c) cross-section B-B. Dimensions in mm.

319

I. Loosening of sand (15-20 min) by applying upward water flow from the bottom of the tank through the perforated tube. II. Pre-installation of pile to a soil depth, Zs = 200 mm. III. Vibro-compaction of sand. IV. Uniformity testing of sand model with mini­ ature cone. Pre-installation of the pile (step II) was carried out with the packer closed and with a small water flow through the inner flushing tube. The packer was kept closed during soil compaction to avoid sand particles filling the annulus between the casing and the middle tube. In step III the sand model was compacted by pushing a concrete vibrator vertically down to the geotextile layer before pulling it slowly up again. Figure 4 show the compaction grid/pattern with a center distance of 150 mm between each point. Compaction was carried out in two stages, starting with the “A” points followed by the “B” points. Figure 5 show a layout of the cone resistance testing positions. The cone had a diameter of 10 mm and apex angle of 60°. Penetration rate was about 5 mm/sec. Points B1, A2a, B2a and C2a where generally tested before each pile-drilling test to verify a consistent prep­ aration procedure and that the sand model was homo­ genous. After pile drilling, cone resistance tests were often also carried out in the points B3a, B4 and B5 to investigate the influence from drilling. All tests pre­ sented in this paper were carried out with a mean rela­ tive soil density (Dr) from 0.6 to 0.65. The density was calculated for each test based on the total dry mass of sand and the volume of sand in the tank. 2.4

Test program

A series of pile drilling tests were performed at the Norwegian Geotechnical Institute (NGI) in Oslo,

Figure 5. Layout for cone resistance testing of sand model. Black dots represent positions for testing and r the radial distance from the pile. Dimensions in mm.

Norway. An overview of the test program for this study is given in Table 2. Test A was carried out as a reference test by pushing the pile into the sand with­ out any rotation and flushing (i.e. displacement pile). The flushing flow rate (Q) varied for test B to E, while both penetration rate and pile rotation was kept con­ stant for all tests. The program included testing with air flushing as well. Despite several attempts with different starting depths and flushing pressure it was not possible to get transport of drill cuttings up through the pile. Due to low soil stresses and relative high air pres­ sures, air evacuated up along the outside of the pile, a behavior which is also frequently observed when drilling close to the soil surface in the field (Lande et al. 2020; Sandene et al. 2020). A starting depth of 40 cm and an air pressure of about 100 kPa was used in test F. All tests with water flushing followed the same procedure. First, the flushing started, and 5 seconds later the rotation and penetration was turned on at the same time. After the drilling was stopped, the packer was immediately closed to prevent sand

Table 2.

Figure 4. Grid for compaction of sand with concrete vibra­ tor. Dimensions in mm.

Test program.

Test

Flow rate l/min

Pressure kPa

Penetration rate mm/s

Rotation rpm

A B C D E F

1.5 2.0 3.0 5.0 -

-

2.5 2.5 2.5 2.5 2.5 1.5

0

20

20

20

20

20

320

15 60 100

particles flowing out from the pile casing. Subse­ quently, the cone resistance tests after the drilling were conducted, after which the pile was lifted out of the sand model. Then, the packer was opened to collect the sand particles in the pile casing. 3 RESULTS AND DISCUSSION The following section presents and discusses the obtained test results. Figure 6 show the load cell meas­ urements which provides a qualitative measure of the soil resistance against depth. The drilling was started at a soil depth, Zs of 200 mm (Figure 3b and 3c). Test A showed an immediate load increase to about 1400 N followed by an almost linear increase in penetration force with depth, resulting in a maximum value of about 4100 N at a drilling depth, Zd of about 260 mm. This equals a tip resistance in the same range as the cone resistance tests presented in the following. Small decrease in load was observed at about 160 mm and 220 mm drilling depths respectively. These differences from the linear trend could be explained by local inhomogeneity in the sand model after vibro­ compaction. Results from tests B to E indicate that when the water flushing flow rate (Q) is above a certain threshold the sand in front of the drill bit does not provide any resistance. This is likely explained by local fluidization of the sand in front of the pile tip like model test results of water jetting for piles reported by Tsinker (1988) & Shepley & Bolton (2014). For this reason, the load cell data show no load or negative load values for the tests C, D and E. This can be explained by the weight of the pile and rotation motor pulling on the load cell. For test B the flow rate (Q = 1.5 l/min) was likely too low to cause local fluidization, thus the soil remained some of its resistance and a maximum load of about 1300 N was observed. As the penetration load increased the friction between the pile and the soil reduced the rotation and the rotation almost stopped for the last 60 mm of the drilling.

Figure 6. Load cell measurements.

Figure 7 present measured change in pore pressure (Δu) against pile drilling depth (Zd) for all pore pres­ sure sensors (PP1 to PP6). Test A show a clear decrease in pore pressure as the pile was pushed into the sand. A maximum pressure reduction of about – 2.3 kPa in PP1 (Figure 7a) occurred rapidly after penetration started, before slowly increasing again during penetration, being about 0.5 kPa below the starting value at end of drilling. Similar trends were also observed in PP2, PP3 and PP5, however with less influence at greater distance from the pile. Only minor pressure reductions of about 0.2 kPa were measured in PP4 and PP6. The pressure reductions with Test A are likely explained by dilation effects in the sand sur­ rounding the pile tip and shaft as it was displaced by the penetrating pile (White & Bolton, 2004). All water flushing tests (Test B to E) resulted in excess pore pressures in the surrounding sand, how­ ever moderate compared to the applied input drilling fluid pressure up to 60 kPa in Test E. As expected the influence generally increased with higher flow rates. Test B (Q = 1.5 l/min) caused maximum pressure changes of about 0.5 kPa in PP1 and PP2 at 70 mm distance from the pile center. Test E (Q = 5 l/min) however caused pressure changes of about 1.7 and 2.8 kPa in PP1 and PP2 respectively. The influence grad­ ually decreases at greater radial distance from the pile. PP5 and PP6 at 210 mm distance from the pile center generally showed only minor changes during the tests. Measurements of vertical soil surface displacements (δv) against drilling depth are presented in Figure 8. Test A resulted in soil heave on all four LVDTs, as expected since the pile was pushed in. Test data show a maximum settlement of about 3 mm in LVDT1 (Figure 8a) and 1.8 mm in LVDT4 (Figure 8d). Results from Test B indicate some minor heave (0.1-0.2 mm) in all LVDTs except LVDT1 closest to the pile which settled about 0.3 mm. This behavior is most likely explained by soil displacements since the flushing was not able to fluidize and remove the sand in front of the drill bit. Test C and D had no impact on the LVDTs, while Test E gave about 0.5 mm settle­ ment in LVDT1. The cone resistance results provide indication of the soil area influenced by the pile drilling. Figure 9 shows cone resistance (qt) against penetration depth from the water level (Zw). Tests carried out in position B2a and B3a at about 90 mm distance from the pile (Figure 9a) show that an increase in flow rate reduced the cone resistance. Test E clearly stands out compared to the other water flushing tests, having an influence at pos­ ition with about 150-160 mm distance from the pile (Figure 9b). An interesting observation is that pushing the pile in without rotation and flushing (Test A) caused the lowest cone resistance in the sand with a significant influence at a distance of about 150-160 mm from the pile. This is explained by the pile penetration causing significant dilation (i.e. volumetric expan­ sion) in the soil. This dilative behavior agrees with results from triaxial tests on Baskarp sand No. 15

321

Figure 7. Change in pore pressure against drilling depth in a) PP1, b) PP2, c) PP3, d) PP4, e) PP5 and f) PP6.

Figure 8. Soil surface settlements against drilling depth for a) LVDT1, b) LVDT2, c) LVDT3 and d) LVDT4.

322

Figure 9. Cone resistance against depth for a) CPTs 90 mm from pile center, and b) 150-160 mm from pile center.

showing large dilation angles up to 18 degrees for low stress conditions (Ibsen et al. 2009). Figure 10 shows normalized flow rate (Qnorm) against normalized mass of drill cuttings (Mc,norm). Normalized flow rate is defined as:

where Q is the flushing flow rate in dm3/min, Apile is the cross-sectional area of the pile in dm2, and Vpen is the penetration rate in dm/min. The normalized mass of drill cuttings is defined as the ratio between the mass of drill cuttings (Mc) collected throughout the drilling and the theoretical mass of soil given by the installed pile volume (Mpile). This calculation ignores

potential drilling induced soil displacements and soil volume changes which is a simplification. A value lower than 1 indicates that the mass of drill cuttings is less than the theoretical one, meaning that the soil is likely replaced by the pile drilling. A value above 1 indicates that the mass of drill cuttings is higher than the theoretical mass, hence causing a potential soil volume loss. A value of 1 is defined as an “ideal” scenario. With Test A the pile was pushed into the sand with­ out no flushing, thus no drill cuttings were generated. Test B with a flow rate of 1.5 l/min resulted in Mc,norm = 0.85 indicating that the pile caused some soil dis­ placements like a driven closed-ended pile. This coin­ cides well with the observations of increased penetration resistance in the last 60 mm of drilling (Figure 6). Test C (Q = 2.0 l/min) seem to be close to an “ideal” flow rate for the given penetration rate and the model conditions with only about 7% excess drill cuttings compared to the installed pile volume (Mc,norm = 1.07). The results show a significant increase in drill cuttings transport with higher flow rates. Test E (Q = 5.0 l/min) caused a maximum of about 170% excess drill cuttings (Mc,norm = 2.70) indicating a major loss of soil volume (i.e. soil mass) around the pile. This likely explain the settlements observed in LVDT1 (Figure 8a). 4 CONCLUSIONS

Figure 10. Normalized flow rate against normalized drill cuttings transport.

This paper presents results from novel physical mod­ elling of pile drilling in saturated sand. The test set­ up made it possible to drill a miniature pile with sim­ ultaneous penetration, rotation and flushing with water through the pile. Installation effects from drill­ ing on the surrounding ground are identified and dis­ cussed based on monitoring data.

323

The results provide valuable knowledge and infor­ mation about the physical influence from flushing, and how it may affect the soil resistance, pore water pres­ sure, soil erosion and transport of drill cuttings. Tests with water flushing show that there is a clear relation between the flow rate and penetration rate, and the abil­ ity to achieve fluidization of the soil in front of the drill bit. If the flow rate is high enough fluidization reduces the penetration resistance similar to observations during pile jetting (Tsinker, 1988, Shepley & Bolton, 2014). Increasing flow rates caused larger excess pore pressures and increased the influence area in the sur­ rounding soil. Under this specific model conditions increased flow rate also generated more drill cuttings, with a maximum of 2.7 times the theoretical installed pile volume for the highest flow rate of Q = 5 l/min (Test E). Measurements of soil surface settlements did however only show small influences from the drilling, most likely due to the limited soil stress. By plotting the normalized flow rate against normalized drill cut­ tings transport it seems that a normalized flow rate from 10 to 15 is close to an “ideal” drilling i.e. Mc,norm equal or close to 1.0. The effect of more representative soil stresses is, however, an area that requires further research. Based on the results from this experimental mod­ elling, it is recommended to perform refined model tests under more representative stress conditions and pile drilling parameters. With the introduced frame­ work for the normalized drill cutting transport and flow rate the results could lead to practical recom­ mendations regarding the choice of drilling param­ eters. The introduced framework should also be investigated further through full-scale tests.

ACKNOWLEDGEMENTS The authors would like to acknowledge The Norwe­ gian Research Council and the eighteen partners in the research project REMEDY for funding of the model tests. The authors would also specifically acknowledge NGI’s staff with Axel Walta for the mechanical design of the model pile, Ole Petter Rotherud for help with data acquisition system and the excellent staff at NGIs workshop for help with practical issues related to the model test.

REFERENCES Ahlund, R. & Ögren, O. 2016. Pore pressures and settle­ ments generated from two different pile drilling methods. Master of Science thesis. Department of Civil and Architectural Engineering, Royal Institute of Tech­ nology, KTH, Stockholm. Asplind, M. 2017. Pore-water pressure and settlements gen­ erated from water driven DTH-drilling. A field study. Master of Science thesis. Department of Civil and Architectural Engineering, Royal Institute of Technol­ ogy, KTH, Stockholm.

Bredenberg, H., Jönsson, M., Isa, R., Larsson, M. & Larsson, E.L. 2014. Borrteknik för minimering av mark­ sättninger vid borrad grundläggning [Drilling technique for minimizing ground settlements from drilling of foun­ dation piles]. Bygg & Teknik 1/14. (In Swedish). Foglia, A. & Ibsen, L.B. 2014. Laboratory experiments of bucket foundations under cyclic loading. DCE Technical report, No. 177. Department of Civil Engineering, Aal­ borg University. Ibsen, L.B. & Bødker, L., 1994. Baskarp Sand No. 15. Data report 9301. Department of Civil Engineering, Aalborg University. Ibsen, L.B., Hanson, M., Hjort, T., & Thaarup, M. 2009. MC-Parameter Calibration of Baskarp Sand No. 15. DCE Technical Reports, No. 62. Department of Civil Engineering, Aalborg University. Karlsrud, K. & Andresen, L. 2008. Design and perform­ ance of deep excavations in soft clays. Proc. 6th Int. Conf. on Case Histories in Geotechnical Engineering. Arlington, Virginia 11-16th August 2008. Missouri Uni­ versity of Science and Technology. Paper No. 12. Konstantakos, D.C., Whittle, A.J., Regalado, C. & Scharner, B. 2004. Control of ground movements for a multi-level-anchored, diaphragm wall during excavation. Proc. 5th Int. Conf. on Case Histories in Geotechnical Eng. New York, NY, 13-17th April 2004. Missouri Uni­ versity of Science and Technology. Paper No. 5.68. Kullingsjø, A. 2007. Effects of deep excavations in soft clay on immediate surroundings – Analysis of the possibility to predict deformations and reactions against the retaining system. Doctoral thesis Chalmers University of Technol­ ogy, Göteborg, Sweden 2007. ISBN 978-91-7385-002-5. Lande, E.J, Karlsrud, K., Langford, J. & Nordal, S. 2020. Effects of drilling for tieback anchors on surrounding ground - results from field tests. Journal of Geotechnical and Geoenvironmental Engineering, 146(8): 05020007, ASCE. Doi:10.1061/(ASCE)GT.1943-5606.0002274. Langford, J., Karlsrud, K., Lande, E.J., Eknes. A.Ø & Engen. A. 2015. Causes of unexpectedly large settle­ ments due to deep excavations in clay. Proc. 16th Euro­ pean Conference on Soil Mechanics and Geotechnical Engineering. Edinburg 13-17th September 2015, ICE Publishing, London, pp. 1115–1120. Doi:10.1680/ ecsmge.60678.vol3.156. Mana, A.I. & Clough, G.W. 1981. Prediction of movements for braced cuts in clays. Journal of Geotechnical and Geoenvironmental Engineering, 107 (6), ASCE. pp. 759–777. Peck, R.B. 1969. Deep excavations and tunneling in soft ground. Proc. 7th Int. Conf. on Soil Mechanics and Foundation Engineering. Mexico City, 1969. State of Art Volume, pp. 225–290. Sandene, T., Ritter, S. & Lande, E.J. 2021. A case study on the effects of anchor drilling in soft, low sensitive clay and sandy, silty soils. Proc. 10th Int. Symposium on Geotech­ nical Aspects of Underground Construction in Soft Ground: Cambridge 28-30th June 2021. Rotterdam: Balkema. Shepley, P. & Bolton, M. D. (2014). Using water injection to remove pile base resistance during installation. Can­ adian Geotechnical Journal, 51(11), pp. 1273–1283. Tsinker, G.P. 1988. Pile Jetting. Journal of Geotechnical and Geoenvironmental Engineering, 114(3), pp. 326–334. White, D.J., & Bolton, M.D. 2004. Displacement and strain paths during plane-strain model pile installation in sand. Géotechnique, 54, No. 6, pp. 375–397.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Excavation of an artificial tunnel using compressed air N. Losacco Dipartimento di Ingegneria Civile, Ambientale, Edile, del Territorio e Chimica, Politecnico di Bari, Bari, Italy

M. Cafaro Salini Impregilo S.p.A, Milan, Italy

R. Marazzita NApoli CAncello Alta Velocità S.c.a.r.l., Rome, Italy

ABSTRACT: The High Speed/High Capacity (HS/HC) railway line between Naples and Bari is currently under construction in Southern Italy. The Napoli-Cancello stretch, at the beginning of the line, includes the Casalnuovo artificial tunnel, excavated through coarse-grained uncemented pyroclastic soil, mostly under the ground water level. The excavation will be carried out using a top-down approach, with the aid of compressed air to counterbalance the pore water pressure. This technique was chosen as an alternative to dewatering or execution of jet-grouting columns due to its reliability, flexibility and negligible impact on the pre-existing pore pressure regime. The design of the Casalnuovo tunnel excavation using compressed air is presented in this paper.

1 INTRODUCTION The new HS railway line project between Naples and Bari involves both infrastructural improvements (e.g. doubling of existing line, new double-track routes) and technological upgrades aimed at increas­ ing both the speed and the capacity with respect to the existing line, while at the same time improving the accessibility to the service in the crossed areas. The project is part of the TEN-T Corridor 5, “Scan­ dinavian-Mediterranean”, connecting Finland to Malta, and will provide an efficient link between the South-East of Italy and the existing HS railway net­ work in the rest of the country, hence facilitating the socioeconomic development of Southern Italy and contributing to transfer a large part of the traffic of people and goods from road to railway, with a remarkable effect in terms of reduction of CO2 emissions. The Casalnuovo artificial tunnel, approximately 2.3 km long, is part of the Napoli-Cancello stretch of the line, crossing the cities of Casoria, Casal­ nuovo and Afragola, at the outskirts of the city of Naples (Figure 1). It is a cut-and-cover tunnel, mainly excavated through coarse-grained pyroclas­ tic soils, to be constructed between reinforced con­ crete diaphragm walls using a top-down approach. The excavation does not interfere significantly with pre-existing structures, crossing an industrial area in the first part and subsequently running through

DOI: 10.1201/9780429321559-41

325

agricultural fields. Since the bottom of the excava­ tion is below the ground water table throughout most of the tunnel length, the contracting agency, Italferr S.p.A., required the excavation to be carried out in dry conditions so as to improve the continu­ ity of both the structure and the watertight layer. In order to prevent water inflow from the bottom during the excavation, the tender solution included the creation of a bottom plug made of jet-grouting columns. NACAV S.c.a.r.l., a joint venture between Salini Impregilo S.p.A. and Astaldi S.p.A., won the bid is-sued for the design of the Napoli-Cancello stretch by proposing an alternative solution, as requested by Italferr, consisting of performing the excavation under hyperbaric conditions to prevent water inflow, in place of the installation of jetgrouting columns. With respect to the tender design, the use of compressed air has some signifi­ cant advantages: – flexibility of the system, allowing an easy adapta­ tion to different ground water table conditions by changing the air pressure; – no risk of contamination of the ground water table; – elimination of the uncertainty connected to the execution of the jet-grouting treatment; – reduced work area required for the intervention; – increased overall safety.

following layers can be identified at the location of the examined cross section shown in the figure: 1. made ground layer (R): gravel and silty sand with brick inclusions, 3 m thick; 2. remoulded pyroclastic soil (DI): loose to medium-dense silty sand and sandy silt, with occasional peaty inclusions, 7 m thick; 3. recent pyroclastic deposits (PO): loose to medium-dense sand in silty and slightly clayey matrix with gravelly sub-layers, thickness larger than 15 m. Figure 1. Location of Napoli-Cancello stretch.

The excavation under hyperbaric condition will be carried out for a length of approximately 800 m, where the hydraulic head is significantly higher than the excavation bottom. The tunnel will be subdivided into compartments of variable length, between 20 m and 70 m, separated by airtight diaphragms. For each compartment, the air pressure required to lower the ground water table below the excavation bottom will be applied. Compressed air is commonly employed in bored tunnels, with both the traditional and the mechanised method (Semprich et al., 2003; Placzek, 2009). Two relatively recent examples of the application of com­ pressed air to the construction of artificial tunnels are the Audi Tunnel at Ingolstadt (Germany) completed in 2001, and the Allmend Tunnel in Lucerna (Switzer­ land), 2012.

Due to the coarse-grained nature of the encoun­ tered soils, the friction angle φ’ reported in Table 1 for each layer was deduced from the in-situ pene­ trometer tests. As an example, for the PO layer, Figure 3 shows the distribution of φ’ with depth obtained from Schmertmann (1978) for various cor­ relations of relative density DR with the SPT number NSPT. The Young’s modulus E’, instead, was calcu­ lated as 1/5 the small-strain modulus E0 inferred from the down-hole tests, considering an expected average shear strain level and a typical modulus decay curve for sandy soils. The results of the Lefranc tests shown in Figure 4 indicate that the permeability k of the pyroclastic cover (layers DI and PO) is in the range 10-6 – 10-3 m/s, with an average k = 10-5 m/s. The pore water pressure distribution is hydrostatic, with maximum level of the ground water table at 6.90 m below the ground surface with limited seasonal oscillations, as shown by piezometric measurements.

Table 1.

2 DESCRIPTION OF THE PROBLEM Four campaigns of geotechnical investigations were carried out at the site, spanning the period 1996­ 2008, including the execution of continuous coring bore-holes, installation of piezometers, CPTs and SPTs, down-hole tests and Lefranc tests. In few cases, laboratory tests, mainly direct shear and oed­ ometer tests, were also carried out. The analysis of the borehole logs allowed to reconstruct the geotech­ nical profile depicted in Figure 2, where the crosssection analysed in this paper is highlighted. he

Figure 2. Geotechnical profile of Casalnuovo tunnel.

Material properties for soil layers.

Layer

γ [kN/m3]

φ’ [°]

c’ [kPa]

E’ [MPa]

ν’ [-]

K0 [-]

R DI PO

18.0 16.0 16.0

30.0 31.5 34.0

0.0 2.5 5.0

40.0 60.0 136.0

0.2 0.2 0.2

0.50 0.48 0.44

Figure 3. Friction angle φ’ from SPT tests for PO layer.

326

2. construction of the diaphragm walls; 3. construction of the top slab; 4. backfilling above the roof slab up to the original ground surface level; 5. excavation below the roof slab with the applica­ tion of the required air pressure; 6. construction of the floor slab and of the lateral sheets; 7. deactivation of the air pressure. A watertight layer is installed between the soil and the slabs and between the finishing r.c. sheets and the diaphragm walls. The construction joints between the structural members have been studied in detail in order to enforce both water- and airtightness so as to minimize compressed air losses, hence opti­ mising the design and reducing the operational cost of the compressed air plant (Lunardi et al., 2019).

Figure 4. Permeability from all Lefranc tests along the stretch.

3 NUMERICAL MODEL The tunnel structure (Figure 5) is made of cast in situ reinforced concrete; the internal dimensions of the cross-section described in this paper are approximately 10.10 m x 7.00 m, with a 1.20 m thick roof slab and a bottom slab with variable thickness, ranging from 1.00 m at the sides to 1.60 m at the centre. The diaphragm walls are 17.50 m long and 1.00 m thick, with 0.35 m thick r.c. sheets casted immediately after the floor slab. The maximum excavation depth in the con­ sidered cross-section is 12.10 m, i.e. approximately 5.20 m below the maximum ground water table, hence an air pressure pair = 600 kPa in excess of the atmospheric pressure (assumed as a reference) is required to ensure that the excavation occurs with­ out any water inflow. The tunnel construction is carried out with a topdown approach, according to the following sequence: 1. unsupported excavation from the ground surface to the intrados of the roof slab (i.e. 3.40 m);

Figure 5. Casalnuovo tunnel typical cross section (dimen­ sions in m).

In this paper, a simplified simulation approach is proposed for the problem at hand, modelled using the Abaqus Finite Element software (Dassault Sys­ tèmes, 2014). In principle, the numerical simulation of the excavation under hyperbaric conditions would require running coupled hydro-mechanical computations with a two-phase fluid, in order to capture the flow of air into the soil and the conse­ quent desaturation. In partly saturated conditions, Abaqus allows to control suction as an independent variable but not directly the pressure of the gas phase in the soil, which is always assumed to be nil. Here, the analyses are carried out in an uncoupled fashion according to the conceptual scheme sketched in Figure 6, showing the distribu­ tion of pore water pressure acting on the dia­ phragms and the applied air pressure pair, assuming that, due to the large permeability of the coarsegrained soil involved in the excavation:

Figure 6. Conceptual application.

327

scheme

of

compressed

air

1. drained conditions hold at all times; 2. application of pair causes instantaneous desatur­ ation (i.e. Sr ≈ 0) down to the depth zw, below the design excavation bottom, where the initial hydrostatic pore water pressure pw equals pair. During the excavation, a fictitious pore pressure and external loads equal to the applied pair are applied above zw (model 1, see Figure 7) so as to ensure that both the effective stress in the soil, in the more gen­ eral sense of Bishop (1959) stress defined in Equation 1, and the total stress acting on the structural members are the same as if the injection of compressed air was explicitly accounted for. For a generic excavation step, the figure shows the applied distributed load equal to the compressed air pressure pair, plotted as arrows, and the prescribed hydraulic boundary condi­ tions. In Equation 1, σ is the total stress, ua and uw are respectively the air and water pressure in the pores and Sr is the saturation ratio.

With the proposed method, the soil between the dia­ phragms remains fully saturated. The analysis was repeated defining a peculiar water retention curve for the soil, similar to a step function, such that full desatur­ ation is achieved for very low values of suction, defined as s = -uw in Abaqus (model 2, see Figure 8). By pre­ scribing a very low suction s = 0.1 kPa to the soil above the design ground water table the Sr ≈ 0 condition is achieved, hence allowing to assess the effect of the unit weight of soil changing from the saturated (γ) to the dry value (γdry) due to desaturation, without the need to change fictitiously the material properties. In the figure, the prescribed distributed load equal to the compressed air pressure pair, shown as arrows, and the imposed hydraulic boundary conditions are displayed. The analyses are conducted in plane-strain condi­ tions. The FE mesh employed for the calculations is

Figure 7. Modelling strategy with saturated soil (model 1).

Figure 8. Modelling (model 2).

strategy with unsaturated soil

composed of eight-noded quadrilateral elements with reduced integration and with displacement degrees of freedom at each node. Elements with an additional d.o.f. for the pore water pressure in the corner nodes are employed to simulate the soil below the initial ground water table depth. The FE domain is 140 m wide and 80 m high; such a large mesh extent was chosen so as to minimise any boundary effects. A detail of the FE mesh in the vicinity of the tunnel is displayed in Figure 9. Tied

Figure 9. Detail of FE mesh in the proximity of the tunnel.

328

4 CONCLUSIONS

Figure 10. Comparison of bending moment and shear force in diaphragms for two modelling techinques.

contact is assumed between the tunnel structure and the surrounding soil; as shown in the figure, due refinement of the soil mesh, with side of the elem­ ents as small as 0.08 m, is enforced in the vicinity of the diaphragms, to allow shearing at the soilconcrete interface due to plastic strains developing in the first slice of soil elements. After the first calcu­ lation step, in which geostatic equilibrium is achieved, the analyses follow the phases of tunnel construction listed in section 2. A simple linear elastic-perfectly plastic material behaviour with a Mohr-Coulomb yield criterion has been assumed for the soil, whereas a linear elastic law has been employed for the reinforced concrete. The material properties adopted for the soil in the analyses are summarised in Table 1, while the fol­ lowing properties have been assumed for the con­ crete: γ = 25.0 kN/m3, E = 31.5 GPa, ν = 0.2. Figure 10 shows the profile of bending moment and shear stress in the diaphragms at the end of tunnel construction, after deactivation of the com­ pressed air pressure. The structural forces obtained with the two models are very similar, suggesting that, under the given assumptions, the desaturation of the soil does not have a significant influence on the mechanical response of the diaphragm. Hence, it is suggested to adopt the simpler approach defined in model 1.

The design of an artificial tunnel, to be constructed in hyperbaric conditions in order to avoid water inflow from the excavation bottom, has been described in the paper. As the excavation will be car­ ried out in coarse-grained pyroclastic soils, typically characterised by a low air-entry value, a simplified modelling strategy for this kind of problems has been proposed, assuming that the soil between the diaphragm walls is quickly desaturated by the appli­ cation of compressed air. In the proposed simplified approach, the soil is considered fully saturated and a fictitious pore water pressure is prescribed above the final ground water table in order to obtain the same effective stress in the soil and the same fluid pressure acting on the dia­ phragm walls as if the application of compressed air was explicitly modelled. A different, more sophisti­ cated approach, in which desaturation of the soil is prescribed below the excavation bottom, has also been attempted. The structural forces in the walls predicted with the two methods are consistent, con­ firming the validity of the proposed simplified approach. Further investigation, involving modelling of the two phases fluid flow, will be carried out in the future for further validation of the hypotheses assumed in this paper.

REFERENCES Bishop, A.W. 1959. The principle of effective stress. Teknisk ukeblad, 39: 859–863. Dassault Systèmes. 2014. Abaqus. Dassault Systèmes, Providence, RI, USA. Lunardi, G., Cassani, G., Bellocchio, A., Nardone, C., Cafaro, M., Ghivarello, G., Sorge, R. & Carriero, F. 2019. The design approach of cut & cover excavation in hyperbaric condition applied for Napoli/Cancello high speed railway. In D. Peila, G. Viggiani, & T. Celestino, eds. Tunnels and Underground Cities: Engineering and Innovation meet Archaeology, Architecture and Art; Proc. ITA-AITES World Tunnel Congress (WTC 2019), 3-9 May 2019, Naples. London: CRC Press. Placzek, D. 2009. The main principles of tunneling under compressed air. In M. Hamza, M. Shahien, & Y. ElMossallamy (eds.) The Academia and Practice of Geotechnical Engineering; Proc. 17th Int. Conf. on Soil Mechanics and Geotechnical Engineering (ICSMGE), Alexandria, 5-9 October 2009. IOS Press. Schmertmann, J.H. 1978. Guidelines for CPT performance and design. US Department of Transportation, Federal Highway Administration. Semprich, S., Scheid, Y. & Gattermann, J. 2003. Com­ pressed air tunnelling-determination of air requirement. In G. Beer (ed.), Numerical Simulation in Tunnelling: 249–301. Wien: Springer-Verlag

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

The use of adaptive smoothed finite-element limit analysis to seismic stability of tunnels H.C. Nguyen Department of Civil and Environmental Engineering, Imperial College London, Skempton Building, London, UK

ABSTRACT: A refinement strategy mesh is described for an upper bound procedure using smoothed finite element method and second-order cone programming. Central to adaptive mesh refinement strategy is the plastic dissipations gap of neighbouring elements which is directly derived from the upper bound solutions. The refinement scheme is then applied to assess how seismic conditions influence the stability of circular tunnels. In simulations, soil behaviour is assumed as a cohesionless frictional Mohr-Coulomb material in conjunction with the associated flow rule. Dependency of safety factors and failure mechanisms on the magnitude of seismic acceleration is intensively examined; results confrim that inclusion of psuedo-static seismic loading in computation simulations reduces the safety factor for the wightless soil. Moreover, the current performance proves its efficiency when producing better results of the safety factor than the existing standard finite element method and is able to reveal the slip-line failure mechanism as noted by Martin (2011).

u [ t ¼ , u \ t ¼ . Let u_ ¼ ½u; v T be plastic velocity or flow fields that belong to a space U of kinematically admissible velocity fields. According to the admissible displacement admissible displacement field u_ 2 U, such that:

1 INTRODUCTION Martin (2011) shows that use of adaptive-finiteelement analysis serves as as effective tool to ascertain the slip-line fields. After a few cycles of refinements, the adaptively refined mesh reveals the failure mechanism of tunnel that can be derived from the relevant slip-line fields as noted by Hill (1950). Noting that Martin (2011) only assesses the slip-line collapse mechanism associated with the static load, so this study extends the contribution of Martin (2011) by inclusion of seismic effects on tunnel stability (see Figure 1). Furthermore, the smoothing domain technique has been described and incorporated into limit analysis procedure, which aims to not only reduce the size of optimisation problem and also give better results of stability number when compared with the finite-element limit analysis. A series of computation performance have been run to provide plasticity solutions for either single or twin circular tunnels using this simple, but effective refinement scheme in both static and seismic conditions.

0 _ 5λþ Wext ðuÞ _ þ Wext _ Wint ð2Þ ðuÞ

ð1Þ

where λþ is the collapse load multiplier; W 0ext ðu_ Þ is the work of any additional loads f 0 ; g0 not subjected to the multiplier; Wext ðu_ Þ is the external work rate with a virtual plastic flow u and is expressed in the linear form as ð Wext ðuÞ ¼

ð f u_ dO þ T

O

t

gT u_ d

ð2Þ

_ is the internal plastic dissipation and Wint ð2Þ _ ¼ Ðof the two-dimensional domain O Wint ð2Þ _ _ O Dð2Þ dO in which the plastic dissipation Dð2Þ is defined by

2 UPPER BOUND ANALYSIS THEOREM Limit analysis bases upon the assumption that analysis is made on the perfect-plasticity material. Considering a rigid-perfectly plastic body of area O 2 R2 with boundary , which is subjected to body forces f and to surface tractions g on the free portion t of . The constrained boundary u is fixed and

_ ¼ max σ : 2≡σ _ ε : 2_ Dð2Þ ψðσÞ0

ð3Þ

with s represents the admissible stresses contained within the convex yield surface ψðσÞ and σε represents the stresses on the yield surface associated to

DOI: 10.1201/9780429321559-42

330

formulation is now defined by the following operation

where ruh are the compatible strains of the approximate fields, uh , and ¢k ðxÞ is a distribution (or smoothing) function that is positive and normalized to unity For simplicity, the smoothing function ¢k is taken as

Figure 1. Illustration of circular tunnel in seismic regions.

any strain rates 2_ through the plasticity condition. If , the collapse defining load multiplier λþ can be determined by the follow­ ing mathematical programming where Ak is the area of the smoothing domain

3 ESFEM In EsFEM, basing on the mesh of elements, we fur­ ther discretize the problem domain into smoothing domains based on edges of the elements such that and Oi \ Oj ¼ 0 for i ≠ j, in which Ned is the total number of edges of all elements in the entire problem domain. Moreover, EsFEM shape functions are identical to those in the FEM. However, instead of using compatible strains, the EsFEM uses strains smoothed over local smoothing domains. These local smoothing domains are constructed based on edges of elements as shown in Figure 2. A strain smoothing

j Ok , and is calculated by e k where Ne is the number of elements around the edge k (Nek ¼1 for the boundary edges and Nek ¼ 2 for interior edges) and Aje is the area of the jth element around the edge k. Since interior edges are formed by two neighboring elements, the smoothed strains in the smoothing domain Ok can be deter­ where dk1 mined by and dk2 are nodal displacement vectors of element e1 and element 1e2 , respectively, 1dk is the dis­ placement vector of the nodes associated with edge ~ kj ðj ¼ 1; 2Þ are the strain-displacement 1k, and B matrices defined by

with

Figure 2. Smoothing domain Ok connected to edge m of triangular elements.

is the smoothed version of shape func­ where tion derivative ; nα is the normal vector and rkj is boundaries of element j associated with edge k. It is worth noting that the EsFEM is different from the standard FEM by two key points: (1) FEM uses the compatible strain on the element, while EsFEM uses the smoothed strain on the smoothing domain; and (2) the assembly process of FEM is based on elem­ ents, while that of EsFEM is based on smoothing domain Ok .

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4 ESFEM DISCRETIZATION OF KINEMATIC FORMULATION In this study, the Morh - Coulomb failure criterion is used,

i.e.

. The plastic strains are assumed to obey the normality rule where the plastic multiplier μ is non-negative. Hence, the power of dissipation can be formulated as a function of strain rates for each domain Ok as formulation in Makrodi­ mopoulos and Martin (2007)

considering gap of plasticity dissipation between neighboring elements that is derived from upper bound limit analysis. The algorithm that requires plasticity deformation fields refines mesh in the regions where having high plasticity deformations is relatively straight forward. An “error indicator” is defined such that

Noting important that the finite element limit ana­ lysis provides sufficient information to calculate ηk and the large gap of dissipation energy between neighboring elements forms the boundary that slipline failure mechanisms will be formed.

where 5.2 Refinement strategy Calculation of the global plastic dissipation is rela­ tively straightforward by summing the plasticity deformations of all elements such that Introducing an approximation of the displacement and using the smoothed strains, the upper-bound limit analysis problem for plane strain can be formu­ lated as Noting that displacement fields from the upper bound solution render sufficient information that one can compare such energy of either each element or a group of elements with the total energy of assem­ bly of elements. Clearly, regions where the soil is collapsed are of the high magnitude of ηk . These elements are expected to refine using the criteria (see Dorfler, 1996) where the smoothed strains are used in Equation (4) and (4) instead of the compatible strains. 5 ADAPTIVE SCHEME In order to refine the mesh, there are two issues that decides the efficiency of refinement scheme. The first one is an appropriate error indicator which is herein based on the values of plastic dissipation of neighboring triangular elements obtained from finite limit analysis, while the second relates to refinement strategy. The newest node bisection which is adopted in this study would be shortly outlined in the follow­ ing section.

The magnitude of θ decides how many elements are refined in each iteration. In particular, smaller values of θ deduce a larger number of elements being refined in each adaptive cycle. One the region that are needed to refine are located, a simple refinement strat­ egy, so-called the newest node bisection, are adopted as illustrated as Figure 3. Details of this simple refine­ ment strategy can be referred to the prior contribu­ tions (Chen, 2008; Nguyen-Thoi et al. 2011). 6 RESULTS 6.1 Validation of refinement scheme

5.1 Error indicator based on values of plastic dissipation of triangular elements An error indicator that is conventionally used in solid mechanics is quantified by considering discrep­ ancies between numerical and recovery stresses. This principal is extended to soil mechanics by

332

Adaptive refinement scheme is validated by consid­ ering one of the most fundamental problem of soil mechanics, the failure mechanism of soil under foot­ ing strip. Figure 3 and Figure 4 show the initial and refinement meshes after five cycles of iterations. The

Figure 3. Illustrations of FEM and EsFEM using triangular mesh before and after refinements mesh.

new mesh reveals the slip-line collapse mechanism which can be derived from slip-line fields. After 8 iterations finite-element limit analysis gives values of collapse load, Nc ¼ 5:1693 while EsFEM limit analysis provides a better results of Nc ¼ 5:1493 (the exact solutions Nc ¼ z þ π ¼ 5:1416). Further ana­ lysis of this failure mechanism is beyond the scope of this study, but this analysis gives an additional confidence that this simple adaptive refinement serves as promising tool to analysis the plasticity solutions of circular tunnel. 6.2

Shallow circular tunnel

Let us consider a circular tunnel with diameter D and being rested at the depth H (see Figure 1). The stability number is widely used to assess the fol­ lowing terms: Ns ¼

ðσs - σt Þ cu

ð14Þ

where σs and σt are respective surcharged pres­ sure on the ground surface and the support pressure inside the tunnel. In limit analysis, for simplicity purpose to get the stability factor, the surcharged load and cohesion force are set to be 1 and the size of domain should be large enough for eliminating boundary effects on the plasticity solutions. Simul­ taneously, weight soil should be installed to be 1, for which the ratio γD=cu is easily calculated to assess the extend to which seismic effects on stability through the parameter αh. Smoothing techniques being not only used over the edge of element but also the node of triangle of elements (see Nguyen 2018) are adopted to examine the stability number. For comparison with the notable study (Davis et al. 1908; Sloan and Assadi, 1993, Martin, 2011) the case of weightless soil (γ ¼ 0) was analysed. Martine (2011) used OxLim gives 4:063 � N � 4:140. Use of NsFEM limit analysis gives N ¼ 4:1253 when performance on the mesh of 9016 NsFEM elements. This provides an additional confidence on the smoothing technique for computational analysis of

333

Figure 4. Initial mesh for limit analysis of footing strip using FEM and EsFEM.

Figure 5. Collapse of footing strip using FEM limit ana­ lysis and EsFEM limit analysis.

collapse load. Without adaptive refinement, compu­ tation of EsFEM limit analysis has been performed using the same mesh as NsFEM limit analysis. Results show that taking smoothing over nodes of element are converged faster than using smoothing domain over the edge of triangles (EsFEM limit ana­ lysis gives N ¼ 4:1253) Application this refinement mesh to seismic stability of circular tunnel (C=D ¼ 1 and γD=cu ¼ 0) is shown in Figure 6, where the initial mesh and the refinement mesh after four iterations are of 4511 and 11554 elements respectively. Because of limitation of power of

computer adaptive refinement scheme is terminated after a few cycles of refinements,so the slip-line fail­ ure mechanisms as noted by Martin (2011) are not totally emerged. Note importantly that care should be taken when using θ in Equation (13) to improve the solutions of callapse loads in each iteration of adaptive scheme. In addition, investigations into how the collapse load changes with the friction angle is examined and results are shown in Table 1. These dependen­ cies are also well illustrated in Figure 5 where an increase in friction angle coincides with a rise in

Figure 6. Initial and refinement meshes using the adaptive scheme (C=D ¼ 1, αh ¼ 0:2 and γD=cu ¼ 0): (a) the initial mesh; and (b) the mesh after 4 cycles of refinement.

334

Table 1.

Seismic stability of circular tunnel with H=D ¼ 1 using NsFEM limit analysis. αh

¢ ¢=5° ¢=10° ¢=20° ¢=30° ¢=35°

0 2.94 3.65 6.36 14.88 27.76

0.1 2.86 3.54 6.16 14.41 26.99

0.2 2.69 3.35 5.76 13.54 25.38

0.3 2.49 3.06 5.29 12.46 23.48

0.4 2.26 2.76 4.72 11.11 21.05

0.5 2.01 2.41 3.98 9.32 17.89

Figure 7. Dependency of Ns on ¢o for circular tunnel with H=D ¼ 1 and γD=c ¼ 0.

Figure 9. Changes in failure mechanisms of tunnel for

H=D ¼ 2; L=D ¼ 2; αh ¼ 0: (a) ¢¼ 10o ; (b) ¢¼ 20o ; and

(a) ¢¼ 40o .

Figure 8. Dependency of Ns on αh for circular tunnel with H=D ¼ 1 and γD=c ¼ 0.

the magnitude of stability number. Furthermore, dependency of displacement fields on levels of seis­ mic effect is assessed through changes in values of αh . Results reveal that stability of tunnel is highly sensitive to friction angle of soil as shown in Figure 8, which are in good agreement with results in the work of Sahoo Kumar (2012) who use FEM limit analysis to proceed the plasticity solutions for cir­ cular tunnel. Changes in plasticity distributions with ¢ are shown in Figure 9. As expected, collapse mechanisms tend to incline and expand as αh increases (see Figure 10). Clearly, inclusion of

335

Figure 10. Displacement fields and failure mechanisms of circular tunnels (C=D ¼ 1 and γ=cu ¼ 0): (a) αh ¼ 0; (b) αh ¼ 0:1; (c) αh ¼ 0:2; and (d) αh ¼ 0:3.

seismic effects on stability of tunnel made a changes in shape of the ground settlements.

REFERENCES

7 CONCLUSIONS This paper has shown combination of the finite elem­ ent analysis and refinement scheme for adaptive mesh refinement to capture the collapse regions of geotechnical stability problem. The smoothing tech­ niques has been implemented in the limit analysis in order to improve the accuracy of solution to the col­ lapse load. This simple adaptive scheme provides an effective tool to analysis the limit state of all manner of plane strain geotechnical problems. In addition, using node-based smoothing domains in limit ana­ lysis is likely to gain better results of the collapse load when compared with using the edged-based domains.

ACKNOWLEDGEMENTS This research is part of the TPS project. The author also are grateful to important supports provided by the Vied-Newton PhD scholarship and the Imperial College London Dixon PhD scholarship for his study in geotechnics section at Imperial College London.

CHEN, L. (2008). Short implementation of bisection in matlab. In Recent Advances In Computational Sciences: Selected Papers from the International Workshop on Computational Sciences and Its Education, pp. 318–332. World Scientific. Davis, E., M. Gunn, R. Mair, & H. Seneviratine (1980). The stability of shallow tunnels and underground open­ ings in cohesive material. Geotechnique 30(4), 397–416. Dörfler, W. (1996). A convergent adaptive algorithm for poisson’s equation. SIAM Journal on Numerical Ana­ lysis 33(3), 1106–1124. Hill, R. (1950). The mathematical theory of plasticity, clarendon. Oxford 613, 614. Liu, G.-R. & N. Trung (2016). Smoothed finite element methods. CRC press. Makrodimopoulos, A. & C. Martin (2007). Upper bound limit analysis using simplex strain elements and second-order cone programming. International journal for numerical and analytical methods in geomechanics 31(6), 835–865. Martin, C. (2011). The use of adaptive finite-element limit analysis to reveal slip-line fields. Géotechnique Letters 1(2), 23–29. Nguyen-Thoi, T., G. Liu, H. Nguyen-Xuan, & C. NguyenTran (2011). Adaptive analysis using the node-based smoothed finite element method (ns-fem). International Journal for Numerical Methods in Biomedical Engin­ eering 27(2), 198–218. Sahoo, J. P. & J. Kumar (2012). Seismic stability of a long unsupported circular tunnel. Computers and Geotech­ nics 44, 109–115.

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Upper bound analysis of seismic stability of tunnels using cell-based smoothed finite element H.C. Nguyen Department of Civil and Environmental Engineering, Imperial College London, Skempton Building, London, UK

ABSTRACT: Estimation of stability of tunnels is a long-standing subject of interest. In conjunction with analytical approaches and field observations, numerical approaches provide insight into the safety factor and its corresponding failure mechanism. This paper adopts an upper bound procedure associated with the cellbased smoothed finite element method and second-order cone programming to estimate the seismic stability of tunnels. In simulations, soil behaviour is assumed as the Mohr-Coulomb material and its plasticity deform­ ation increment is obeyed the associated flow rule. Variations in tunnel depths, distance between two tunnels were performed to investigate the dependency of safety factors and failure mechanisms on the tunnel geom­ etry. Moreover, changes in the magnitude of seismic acceleration were intensively created to examine the seis­ mic effect on tunnel stability.

1 INTRODUCTION Finite element limit analysis serves an effective means of estimations of collapse load and failure mechanisms as well. Accuracy of solutions to col­ lapse load lie in two themes, the forms of optimiza­ tion and discretization techniques. Optimization being casted as the second-order cone programming (SOCP) (Makrodimopoulos & Martin, 2006, 2007,2008; Le, 2018) proves its efficiency in solving the optimisation problems. Noting that variables used in the optimization problems depend the dis­ cretization technique; however, triangle elements are widely used in the limit analysis problem rather than using quadrilateral elements. This contribution adopts quadrilateral elements to discrete displace­ ment fields and the final form of limit analysis is also established as the SOCP problem. In addition, inclusion of seismic effects on tunnel stability are made, in this study, by varying the magnitudes of horizontal seismic acceleration, αh . A set of compu­ tational performances using the procedure to assess the stability of tunnel (see Figure 1) where the cohe­ sive force linearly increases with the depth as the form such that

where co and cz are the cohesive forces at the ground and the depth z, respectively. Intensive investigations into how variations in soil properties with the depth, the ratio, H=B and specific soil weight on the tunnel

stability and the collapsed mechanism are made. Concentration will be paid to the so-called stability number such that

where σs and σt are respective surcharged pressure on the ground surface and the support pressure inside the tunnel. Inclusion of seismic effects in simulations is made by introducing of additional work done proportional to horizontal seismic accel­ eration, αh . 2 GENERAL THEOREM OF UPPER BOUND ANALYSIS It is essential to know that limit analysis theory is founded on the hypothesis that behaviour of material is rigid-perfectly plastic, with increments of plastic­ ally admissible strain followed by the normality rule. To understand this approach, let us consider a rigidperfectly plastic body of area with boundary r, which is subjected to body forces f and to surface tractions g on the free portion rt of r. The con­ , strained boundary ru is fixed and . Let be plastic velocity or flow fields that belong to a space U of kinematically admissible velocity fields. According to the kine­ matic theorem, the structure will collapse if and only

DOI: 10.1201/9780429321559-43

337

Figure 1. Illustration of square tunnel in seismic regions.

load multiplier λþ and the corresponding failure mechanism.

if there exists a kinematically admissible displace­ ment field , such that:

3 USE OF CSFEM TO APPROXIMATE DISPLACEMENT FIELDS is where λþ is the collapse load multiplier; the work of any additional loads f 0 ; g0 not subjected to the multiplier; is the external work rate with a virtual plastic flow u_ and is expressed in the linear form as

Central to smooth finite element analysis is the com­ bination of standard finite element and strain smoothing scheme, which the smoothing domain is based on either the node, the edge or the nel of the element. Inclusion of smoothing techniques is advantageous in limit analysis which provides better values of collapse loads and faster convergences of solution than using standard finite element. Details of smoothing techniques can be refereed to the work of Liu & Nguyen (2016). If one consider a quadrilateral element which can be divided into n smaller domain inside the parent element, resulting

is the internal plastic dissipation of and the two-dimensional domain in which the plastic dissipation is defined by

in

. Smoothing strain of each cell,

is

a function of strain of parent element, 2h and a form of distribution function, ’ such that with σ represents the admissible stresses contained within the convex yield surface ψðσÞ and σε repre­ sents the stresses on the yield surface associated to any strain rates 2_ through the plasticity condition. If defining , the collapse load multiplier λþ can be determined by the follow­ ing mathematical programming

. It is noted that xC is a point in the domain Ωek . One can use different types of distribution function, ’ which must satisfy two conditions: (i) ; and (ii) c Noting that the compatible strain of the approxi­ mate fields, is approximated using uh . One can use the area of the smoothing domain Ωe to define the distribution function such that

Noting that the primary purpose of using limit analysis herein is to accommodate the collapse

If one applies the divergence theorem, then the smoothing strains can be written as

338

4 FORMULATION OF OPTIMISATION PROBLEMS IN THE FORM OF SECOND-ORDER CONE PROGRAMMING where rc is the boundary of smoothing domain as illus­ trated as Figure 2, and the normal vector nx such that

It is therefore that smoothing strain rates can be cal­ culated such that

The first step of limit analysis is described above, with the displacement field being constructed using CSFEM. This enables dissipation energy of whole domain to be formed using strain rates. Noting that the rate of dissipation is function of strain rates and soil properties. For simplicity, the increment of plas­ tic strains is calculated using the normality rule (i.e. , with the non-negative plastic multi­ plier, μ_ ). It is, in this study, assumed that soil behav­ ior is governed by the Morh - Coulomb failure criterion such that

d is the displacement vector of the nodes associated ~_ are the strain-displacement mat­ with the nel, and B rices defined by

in which c and ¢ are the respective cohesion force and internal friction angle of soil. The magnitude of dissipation energy for each domain i is identical to the form of the work of Makrodimopoulos and Martin (2007) such that

is introduced, Equation (13) can be If rewritten such that with

in which where is the Gauss point of boundary segment which has length lk and outward surface normal vector nk as illustrated in Figure 2. Introducing an approximation of the displacement and using the mean strains, the upper-bound limit analysis problem for plane strain can be formu­ lated as

Figure 2. Geometry definition of smoothing nel.

Noting that the last constraint in Equation (16) is expressed in the conic form, allowing a primal-dual interior-point algorithm specialized for SOCP to be employed to solve the optimisation problem.

339

5 LIMIT ANALYSIS OF STABILITY OF SQUARE TUNNEL Let us define the geometry of tunnel and soil proper­ ties before applying the cell-based smoothed finite element analysis to the tunnel stability and its failure mechanism. Considering a square tunnel that is of the width B and is buried at the depth H as previ­ ously illustrated in Figure 1. The cohesive force is varied as the form expressed in the Equation 1. Clearly many factors influence the magnitude of Ns ; however, this study is restricted to some of predom­ inant parameters Assessment of the case where there is an absence of seismic loads is firstly considered. Results reveal that the magnitude of Ns clearly increases as either ρ or increases for the case of weightless soil. This fea­ ture is reversed for the case that inclusion of soil weight is made in simulations. Interestingly, magnitude of Ns is either negative or positive, which depends upon the ratio of . The case, corresponds to the need for an internal compression loads that is required to stabilize the tunnel. For the constructing process of tunnel, for a specific soil properties, the

value of Ns provides a useful information as to calcula­ tions of the additionally supporting force that needs to apply to the inside face the tunnel. Evolutions of slipline failure pattern is shown in Figure 3 as the depth of tunnel increases. Figure 4 gives a view of variations in Ns with for the cases that soil weight is increased, while Figure 5 shows how Ns is dependent on variations in the cohesive force for three different values of . These values are in excellent agree­ ment with those obtained by Wilson et. al (2013) where significant higher number of elements used in their simulations. Clearly, Ns decreases as increases, meaning that σt exceeds the values of σs , leading to the negative values of Ns . In order to considering effects of friction angle on Ns , values of ¢ are varied in the modeling. This influence is revealed by considering both values of Ns and failure mechanisms which is illustrated in Figure 6. Expectation that Ns increases with

Figure 4. Variations in Ns with γB=co for H=B ¼ 1.

Figure 3. Changes in plasticity distribution with increasing H=B.

Figure 5. Variations in Ns with ρ for H=B ¼ 3.

340

Figure 6. Collapse loads and failure mechanisms when ¢ are varied: (a) ¢¼ 0o ; (b) ¢¼ 10o ; (c) ¢¼ 20o ; (b) ¢¼ 30o .

increasing ¢ is well captured, with a result of clear change in failure mechanisms where expansions of collapse fields are observed for the lower values of ¢. Inclusion of seismic effects on the stability makes a reduction in the magnitude of Ns . First, changes in Ns for a square tunnel are analyzed. Figure 7 shows that involvements of αh in the model results in

a drop in Ns . Second, visualization of changes in , αh was capplasticity distributions with both tured to provide an plausible explanation for the reduction in Ns with increasing αh . Clearly, inclinations of failure mechanisms are observed, fitting very well with limit-equilibrium theory. For the case of twin square tunnels, dependency of both stability number and the possibility that slip-line failure

Figure 7. Changes in failure mechanisms for the case both ¢ and αh are varied: (a) ¢¼ 0o ; γB=co ¼ 1; αh ¼ 0; (b) ¢¼ 0o ; γB=co ¼ 1; αh ¼ 0:1; (c) ¢¼ 20o ; γB=co ¼ 1; αh ¼ 0; and (d) ¢¼ 20o ; γB=co ¼ 1; αh ¼ 0:1.

341

Figure 8. Plasticity distributions of twin tunnels with H=B ¼ 2 as αh and L=B varied: (a) ¢¼ 10o ; γB=co ¼ 1; αh ¼ 0; L=B ¼ 1; (b) ¢¼ 10o ; γB=co ¼ 1; αh ¼ 0:1; L=B ¼ 1; (c) ¢¼ 20o ; γB=co ¼ 1; αh ¼ 0; L=B ¼ 2; and (d) ¢¼ 20o ; γB=co ¼ 1; αh ¼ 0:1; L=B ¼ 2.

mechanism are formed is clearly captured by this numerical simulations as shown in Figure 8. As expected, low values of deduce the smaller value of Ns , reflecting the influence of relative posi­ tions of two tunnels. Interestingly, failure mechan­ isms are altered as the distance between two tunnels rises. 6 CONCLUSIONS The smoothed finite element analysis using quadri­ lateral elements serves as a promising means of cal­ culations of the collapse loads and visualisations of failure mechanisms. To reveal the slip-line failure mechanisms, incorporation process of adaptive mesh refinement in the smoothed finite element limit ana­ lysis is required as noted in Martin (2011). Further­ more, in order to serve better the value of stability number, a quadrilateral element might be divided into smaller smoothing domains. Considerations of smoothing over either nodes or edges or elements might develop the plasticity solutions of tunnel stability.

ACKNOWLEDGEMENTS This research is part of the TPS project. The author is grateful to vital support provided by a Newton PhD scholarship and a Imperial College Dixon

scholarship for his study in geotechnics section at Imperial College London.

REFERENCES Le, C. V. (2017), ‘Estimation of bearing capacity factors of cohesive-frictional soil using the cell-based smoothed finite element method’, Computers and Geotechnics 83, 178–183. Liu, G.-R. & Trung, N. (2016), Smoothed finite element methods, CRC press. Makrodimopoulos, A. & Martin, C. (2006), `Lower bound limit analysis of cohesive-frictional materials using second-order cone programming’, International Journal for Numerical Methods in Engineering 66(4), 604–634. Makrodimopoulos, A. & Martin, C. (2007), ‘Upper bound limit analysis using simplex strain elements and second-order cone programming’, International journal for numerical and analytical methods in geomechanics 31(6), 835–865. Makrodimopoulos, A. & Martin, C. (2008), ‘Upper bound limit analysis using discontinuous quadratic displace­ ment fields’, Communications in Numerical Methods in Engineering 24(11), 911–927. Martin, C. (2011), ‘The use of adaptive finite-element limit analysis to reveal slip-line fields’, Géotechnique Letters 1(2), 23–29. Wilson, D. W., Abbo, A. J., Sloan, S. W. Lyamin, A. V. (2013), ‘Undrained stability of a square tunnel where the shear strength increases linearly with depth’, Com­ puters and Geotechnics 49, 314–325.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Investigation of the seismic performance of the complicated tunnel sections with non-uniform heights H. Nitta, S. Ito, T. Otsuka & S. Konishi Tokyo Metro Co., Ltd, Tokyo, Japan

K. Tsuno Railway Technical Research Institute, Tokyo, Japan

S. Tsuchiya Coms Engineering Corporation, Tokyo, Japan

K. Maekawa YOKOHAMA National University, Tokyo, Japan

ABSTRACT: Tokyo Metro Co. Ltd. operates 9 routes in the capital area of Japan; total length of tunnels is 166km. In Japan, subway tunnels were constructed by the cut & cover method and have horizontally oriented double cell box tunnels with rectangular cross sections. Because some tunnels had to be built under narrow roads in Tokyo, there exist some vertically oriented double cell box tunnels due to land acquisition problems. And there exist some parts with special cross sections changing gradually from the vertical or horizontal dir­ ection to the other direction, hereinafter called “tunnel sections with non-uniform heights”. The Earthquake Resistance Standard for Railway Structures in Japan (Railway Technical Research Institute. 2012 [3]) contains no explanation of the complicated structure such as the tunnel sections with non-uniform heights. To explain the earthquake resisting of the complicated structure, it is important to analyse behaviour of the ground and structure such as the complicated shape, and grasp the sectional force of the structure associated with the ground movement at the time of the earthquake. For this reason, the model ground and a model tunnel were made and shaking table tests using a shear soil vessel were carried out to simulate the ground movement situ­ ation and the tunnel behavior at the time of earthquake. And the reproduction analysis of the tests were carried out by FEM and to confirm validity of the analytical technique. In result, deformation behavior and sectional force of the tunnel sections with non-uniform heights associated with the ground movement at the time of the earthquake were made clear, and knowledge of the examination method of tunnel sections with non-uniform heights were obtained.

1 INSTRUCTIONS Subways are typically constructed below roadways, and when the open-cut method is used to construct the subway tunnels between stations, the most common configuration is a double-track box tunnel in a 1-layer-2-span structure. However, where sub­ ways run beneath narrow roadways, a 2-layer-1-span structure is required to ensure space for the doubletrack sections. Tunnels shown in Figure 1 also exist in the Tokyo Metro lines. In the transition zones between 1-layer-2-span structures and 2-layer­ 1-span structures, because the heights of the upper and lower slabs are different, tunnels in the parts are called “tunnel sections with non-uniform heights” in this research.

The Earthquake Resistance Standard for Railway Structures in Japan (Railway Technical Research Institute. 2012 [3]) shows descriptions of dynamic analysis methods and seismic deformation methods and for common box tunnels, namely the 1-layer­ 2span structure, and these methods are used in prac­ tice to evaluate the seismic performance of tunnels. However, for tunnel sections with non-uniform heights, evaluation method of their seismic perform­ ance has not been established yet. Therefore, for evaluating the seismic performance of the tunnels, it is important to analyses behaviour of the ground and structure with such a complex shape, and grasp the sectional force of the structure associated with the ground movement at the time of the earthquake. From the above, in this research dynamic shaking

DOI: 10.1201/9780429321559-44

343

tests on models of tunnel sections with non-uniform heights were conducted in an effort to fully under­ stand their behavior. In the shaking tests, the model ground and a model tunnel were made and Spectrum-II Level 2 seismic motion (“Spec-II L2 seismic motion”) as described in Seismic Resistance Stand­ ards was applied, and a full understanding of the dis­ tribution of bending moments and shear forces on the model tunnels were obtained. In addition, 3D finite element method (FEM) analyses were carried out to simulate the shaking tests in order to confirm the validity of the analysis method. We also com­ pared analytical results with test results in order to confirm the relevance of the methods of analysis, and gathered vital knowledge for establishing a method of investigating the seismic resistance of tunnel sections with non-uniform heights. 2 DYNAMIC SHAKING TESTS 2.1 Shaking test equipment, shear box and model tunnels Shaking test equipment (Railway Technical Research Institute. 2012 [3]) belonging to RTRI and a shear vessel were used to conduct shaking tests. The shear vessel is 3,000 mm wide, 1,100 mm deep, and 3,000 mm tall on the inside, and comprises 12 shear frames within the 2,400mm space from top to bottom. The shear frames are supported by a linear guide, and can be deformed according to the behavior of the ground. Acrylic sheets were used to create model tunnels of tunnel with non-uniform heights (Figure 2). The depth of model tunnel is about 1000mm. The elastic modulus of the acrylic was 3,090 MN/m2. The model tunnel is roughly 1/12 scale model of actual subway tunnels. In actual tunnel, member thickness differs according to the depth, but to simplify conditions of the test was applied, we assigned a uniform member thickness of 20 mm.

Figure 2. Dimensions of Model Tunnel Sections Tunnel and Shaking test equipment.

2.2 Model ground The basic ground material values (a dry sand founda­ tion composed of No. 6 Tohoku silica sand (D50 = 0.34 mm, Dmax = 16.9 kN/m3, Dmin = 13.8 kN/m3, Gs = 2.66) was used for the model ground. In addition, after placing the model tunnel in the shear vessel, soil was sprinkled and compacted ever 200 mm depth. The model tunnel was positioned such that there was 1,400 mm of space between the bottom edge of the model and the side wall of the shear vessel, and gave the model an earth covering of 610 mm. 2.3 Excitation method Figure 3 shows the input waveforms used for the shaking tests. The time-compressed input wave­ forms for Spec-II L2 seismic motion (maximum acceleration 955 Gal) were used in accordance with the law of similarity. In addition, the positive and negative excitation was applied, with left (with respect to model ground displacement) as the posi­ tive direction of loading and excitation. And the ground surface during excitation was measured and it confirmed that hardly any subsidence occurs. 2.4 Measurement items

Figure 1. Example of an Uneven Multi-Section Tunnel.

In order to fully understand the section force acting on each member of the model tunnel, displacement gauges were installed on each of the 45 shear frames

344

Figure 3. Input Seismic Waveform (Spec-II L2 Seismic Motion 955 Gal).

Figure 4. Strain Gauge installation Locations.

Figure 6. Diagrams of bending moment (Shaking Tests).

that comprise the shear vessel in an effort to fully understand the distribution of horizontal displace­ ment that occurs in the shear vessel in the depth dir­ ection. Figure 4 shows the strain gauges installation locations.

2.6

2.5

Result (Strain of model tunnel)

Figure 5 shows the hysteretic response of model tunnel strain during positive excitation under Spec-II L2 seis­ mic motion (maximum acceleration 955 Gal). The figure shows the strain of the outer surfaces of mem­ bers at G1, G20, G21 and G40 shown in Figure 4. The results show maximum strain at 3.4 seconds.

Figure 5. Model Tunnel Strain (Spec-II L2 Seismic Motion 955 Gal).

Result (Bending moment and shear forces)

Figure 6 shows the distribution of bending moments obtained from the strain gauges of model tunnel. These images show the variation in bending moments that occurred in response to excitation in a constant state after the construction of model ground at the timing of maximum strain at G20. Result of the tests revealed the tendency for nega­ tive bending moments to increase at the corners of the lower edge on the loading side and the corners of the upper edge on the opposite on side of the loading side. It was also able to be confirmed that the tendency for the bending moments of the center wall to increase. However, the bending moments increased at the upper and lower edges of the center wall under positive exci­ tation, and at the upper and lower edges of the center wall under negative excitation, which indicates distri­ bution that differs depending on the direction of excitation. Figure 7 shows the variation of shear forces that occurred in response to excitation in a constant state after the construction of model ground at 3.4 second. Result of the tests revealed that the tendency for shear forces to increase at the corners of the lower edge on the loading side and the corners of the upper edge on the opposite side of the loading side. In add­ ition, the shear forces increased at the middle part of

345

modulus are proportional to confined compression raised to the 0.3 and 0.5 powers, respectively. 3.2

Cases for analysis

Dynamic analysis was conducted on the uneven multi-section model. At first, a constant state was simulated after constructing the model ground for each of the model tunnels. Next, positive and nega­ tive excitation under Spec-II L2 seismic motion (maximum acceleration 955 Gal) was applied as shown in Figure 3. As for the shaking tests, excita­ tion in stages by sequentially increasing the acceler­ ation amplitude was used. 3.3

Figure 7. Diagrams of Shear Forces (Shaking Tests).

the center wall under positive excitation, and at the upper and lower parts of the center wall under nega­ tive excitation, which indicates distribution that dif­ fers depending on the direction of excitation in the same manner as the distribution of bending moments. 3 FEM SIMULATIONS 3.1

Methods of analysis

Simulation analysis through loading tests on model ground and model tunnels using 3D nonlinear ana­ lysis (Maekawa Koichi et al. 2003 [1]) for soil-RC structure systems was conducted with the intent to apply the analysis method to actual uneven multisection subway tunnels in future, and the relevance of the method of analysis was considered. This method of analysis makes it possible to trace crack­ ing and reinforcement bar yielding that result from opening and closing in multiple directions, which is nonlinear behavior peculiar to RC structures. A ground model that successively accounts for volumetric nonlinearity through nonlinear shear con­ fined compression dependence and dilatancy (Mohammad Reza Okhovat et al. 2009 [2]) was used. In this model, shear rigidity and the volumetric elastic

Analytical models

For the dynamic analysis, models based on the model tunnels and model ground were created and set the range of the models from +600 mm at the bottom of the lowest shear frame to +2,900 mm (the ground surface) in the vertical direction, and to 1,000 mm in the depth direction. The model tunnels and ground as solid elements were used, and rigid joint elements on the interface between the model tunnels and ground were set. In addition, pseudo-elements were installed on the sides between ground and side wall of shear vessel with the conditions of the shaking tests caused by the shear vessel. Also, joint elements on the inter­ face between the pseudo-elements and the ground elements were placed, and the shear rigidity from zero and restricting value was set to avoid the shear deformation of the ground. For configuring the pseudo-elements, the weight of the shear frames of the shear vessel was accounted and assigned rigid elastic moduli and extremely small shear rigidity. Furthermore, rigid truss elements to the nodal points at both edges of the shear vessel were joined. As for loads and boundary conditions, the bottom surface of the model was fixed and the vertical rollers were set on the sides. Side restrictions after simulating a constant state in consideration of the dead load were removed. Afterward, it was decided to input acceleration hysteretic waveforms. For the main dynamic parts, the waveforms at 0.002-second intervals between 0.5 seconds and 5.5 seconds was only input, and the Newmark-beta method (β = 0.36) was used to perform direct integration in consider­ ation of the hysteretic damping of materials and the numerical damping of high-frequency ranges. Figure 8 is an overview of the dynamic analysis element models. The models are divided into four parts perpendiculars to the model tunnel member axes, and into four equal parts in the transverse dir­ ection of the tunnels for the section member axes. 3.4

Input material values

For the model tunnels, the elastic modulus of 3,090 MN/m2 was assigned, and the models with elastic

346

elements were created with the unit volumetric weight and Poisson ratio were set to the nominal values of 1.181 g/cm3 and 0.43 cm3 (acrylic), respectively. For the model ground, relevant density to 70%, vis­ cosity to 0.0kN/m2, angle of internal friction to 40.4°, initial rigidity to 48.2 MN/m2 and unit volumetric weight to 15.7 kN/m3 were set. The basic skeleton curve for the shear stress-shear strain relationship was applied based on the results of the cyclic torsional shear tests conducted with con-fined compression of 20 kN/m2. For the hysteretic rules, Masing rules was used. Figure 8. Diagrams of the Analytical Models.

3.5

Result (strain of model tunnel)

Figure 9. Model Tunnel Strains (Analysis Results).

Figure 9 shows the hysteretic response of model tunnel strain and shear box displacement, respect­ ively, obtained from our analysis. The same method of compiling data as that for Figure 5 were used. Figure 9 shows that strain of the model tunnel increase at around 2.3 and 3.4 seconds for the uneven multi-section model, and the fact that the response values largely conform to the

Figure 10. Comparison with tests and analysis (bending moment).

Figure 11. Comparison with tests and analysis (share forces).

347

experimental values suggests that the results of the analysis simulated similar tendencies as the shaking tests. 3.6 Result (bending moments and shear forces) Figure 10 and 11 show the comparison with bending moments and shear forces of multi-section model obtained from our analysis and shaking tests. The analysis figures show the distribution of section forces caused by excitation after the simulation of a constant state at the timing during shaking tests of the maximum strain of elastic elements that corres­ pond to G20. The results of this analysis closely simulated the results of the shaking tests that model dynamic behavior during earthquakes. 3.7 Result (model deformation diagrams and strain contour diagrams) Figure 12 shows the distribution of section forces from the analysis and the diagram of model deform­ ation (with 10x magnification of deformations) at the same timing as in (b), as well as a ground shear

Figure 12. Deformation Diagrams and Strain Contour Dia­ grams (Dynamic Analysis Results).

strain diagram and a model tunnel maximum princi­ pal strain contour diagram. It was able to be confirmed that shear deformations of the center wall increased under great stress from the ground at the upper-right and lower-left parts of the model with respect to the center wall under

348

positive excitation in the dynamic analysis. This illus­ trates the need to appropriately consider the inter­ action of tunnel structures and the ground during earth-quakes. 4 CONCLUSIONS In this research, tests that model the dynamic action of earthquakes on models of tunnel sections with non­ uniform heights were conducted in an effort to fully understand their behavior in the transverse direction. Also, the numerical simulation based on dynamic 3D FEM analysis were carried out and effectiveness of the analysis was verified by means of the comparison with the result of analysis and dynamic shaking tests. We gained the following knowledge from this research: 1) It is cleared that negative bending moments have been caused by ground excitation at the corners of the lower edge of the loading side and the cor­ ners of the upper edge of the side opposite the loading side. And for tunnel sections with non­ uniform heights, the bending moments of the center walls increase, and the bending moments increase at the upper and lower edges of the center of the center wall under positive excita­ tion, and at the upper and lower edges of the upper and lower parts of the center wall under negative excitation, and that distribution differs de-pending on the direction of excitation. 2) It is cleared that shear force increasing has been caused by ground excitation at the corners of the lower edge of the loading side and the corners of the upper edge of the side opposite the loading

side. And for tunnel sections with non-uniform heights, the shear forces increase at the center of the center wall under positive excitation, and at the upper and lower parts of the center wall under negative excitation, and that distribution differs depending on the direction of excitation. 3) The results have shown that 3D static FEM ana­ lysis can roughly simulates results of static load­ ing test that difference of diagram of bending moment and shear force. 4) The results have shown that 3D dynamic FEM analysis largely simulates the dynamic behavior of model tunnels in the transverse direction in shaking tests that model dynamic actions during earthquakes. 5) The results have shown that 3D dynamic FEM analysis can clearly and visually show the impact of deformation modes and interaction of soilstructure systems during earthquakes, conditions that are difficult to fully understand through model tests alone..

REFERENCES [1] Maekawa Koichi et al. 2003. Nonlinear Mechanics of Reinforced Concrete. SPON Press:44–54. London: CRC Press [2] Mohammad Reza Okhovat et al. 2009. Nonlinear seis­ mic response and damage of reinforced concrete ducts in liquefiable soils. Journal of Advanced Concrete Tech­ nology. Vol.7, No.3: pp.439–454. Japan: JCI [3] Railway Technical Research Institute. 2012. Design Standards for Railway Structures and Commentary: Earthquake-Resistant Design. Japan:Maruzen

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Influence of the annulus grout on the soil-lining interaction for EBP tunneling M. Ochmański Faculty of Civil Engineering, Silesian University of Technology, Gliwice, Poland

G. Modoni Department of Civil and Mechanical Engineering, University of Cassino and Southern Lazio, Cassino, Italy

G. Spagnoli MBCC Group, Trostberg, Germany

ABSTRACT: Grouting of the annular space behind the tunnel lining is a delicate phase of the Earth Pressure Balance excavation process, being related with stress modification and deformation in the sur­ rounding soil. Normally operative parameters such as injection pressure, mechanical properties and hard­ ening rate of the grout are assigned with empirical rules relating them to the soil type, more than on mechanical schematization that analyses their effects. This paper examines the influence of the mechan­ ical characteristics of the annulus grout on the interaction between soil and lining activated by EPB tun­ neling. To this aim, a three-dimensional numerical model is created reproducing excavation, face pressurization, lining installation and tail void grouting with great accuracy. The computational model is validated against a tunnel built by the Metropolitan Rapid Transit Authority (MRTA) in Bangkok, selected thanks to the availability of experimental investigation and monitoring data. The hydromechanical response of the soil is simulated with a non-linear, irreversible, anisotropic hypoplastic model and the grout with a time dependent increase of stiffness and strength reproducing its hardening process. Final stiffness and hardening rate of the annulus grout are parametrically varied to investigate their role on the interaction between soil and lining and to identify a composition that minimizes the short and long-term effects on the surrounding environment. Results of the analysis are provided in terms of lining stresses, soil deformation and ground subsidence.

1 INTRODUCTION Mechanized tunnels are constantly increasing with respect to the conventional drilling and blast (D&B method) as it currently represents a cost-effective solution for tunnel constructions in case of long tunnels and well-known ground conditions (Nord 2006; Efron and Read 2012). After excavation with a rotating cutter-head, the cavity is supported by a segmental final lining continuously installed during the excavation process. As the excavation head is bigger than the external diameter of the lining, a gap is formed that must be filled (Figure 1). This is a delicate phase of the tunneling process as the compliance of the soil to the outer lining sur­ face may produce significant settlements at the ground level (Maidl et al. 1995; Peila et al. 2011). Together with stability during construction, back­ filling of the tail gap prevents the heave/flotation of the lining, takes early load in the build area, pre­ vents segmental misalignment and the rupturing of

gaskets, prevents/reduces water ingress to avoid secondary injections (costly). There are two basic types of annular grouts cur­ rently in use: thick mortar type grouts and highly mobile two-component grouts. Two-component grouts consist of an A and B component and are also referred to as a Bi-Component grout. The typical mix-design in a m3 system for a Bi-Component grout is very variable and changes from projects to projects but in general it consists of: 1. 2. 3. 4. 5.

Cement: 280-450 kg; Bentonite: 30-60 kg; Water: 730-860 kg; Retarder: 2-3 l; Accelerator: 60-80 kg (normally sodium silicate).

The A-Component is a stabilized grout containing 1 to 4 as shown above. The B-Component is a liquid accelerator that is added to the A-Component as it is being mixed into the annulus at the injection point. The performances required for the annulus grout

DOI: 10.1201/9780429321559-45

350

Figure 1. Sketch of the annulus grout through the grout channel.

consist in having a good flowability, good pumpabil­ ity over pumping distance, workability from 4 - 24 hours, no bleeding ( 0.4; 2. fR3k/fR1k > 0.5 According to the project requirement, a minimum value of fR1 and fR3 must be specified. We usually recommend the minimum perform­ ance class C35/45 3c according to fib Model Code 2010 for PSCL, meaning:

Figure 5. Three-point bending test on square panel with notch.

1. Characteristic compressive strength fck 35 MPa 2. Characteristic residual flexural tensile strength fR1k > 3.0 MPa 2.3.2 Alternative test to EN 14651: EFNARC [11] Three Point Bending Test on Square Panel with Notch A practical method to determine the tensile behav­ iour of SFRC for shotcrete applications is a threepoint bending test on square panels (Figure 5). This test combines the output of the EN14651 (Figure 4b) with the advantages of the EN14488-5 test. The same moulds can be used and due to the larger cracked section, the scatter is lower. This test method is promoted by EFNARC for the following main reasons: 1. The geometry and dimensions of the specimens as well as the spraying method, matches that of real structures. 2. The dimensions of the test specimen can be handed at a standard equipped laboratory (no excessive weights or sophisticated equipment needed).

Table 3.

3. The geometry is the same as in the plate test for Energy Absorption. 4. The plate can be sprayed on site. 5. No need to sawn a prism from a panel (no influ­ ence on the result). By analogy (Table 3) with EN 14651, this test defines the residual flexural strength (fR1, fR3) according to the updated international standard (fib Model Code 2010). The mechanical property obtained will serve as input for the dimensioning method. The slab specimens need to be prepared according to the regulations of EN14488-1. A mould with inner dimensions of 600 x 600 mm and an inner thickness of 100 mm has to be positioned within 20° of the vertical (unless another orientation has been specified) and sprayed with the same equip­ ment, operator, technique, layer thickness per pass and spraying distance as the actual work. Immediately after spraying, the thickness of the concrete specimens shall be trimmed to a 1000+5 mm. It is very important to make sure that the spraying side of the specimen is

Correlation table between CMOD, crack opening and deflection. EN14651

What Evaluation residual flex­ ural strength

3-point bending test on plates

CMOD mm

Crack Deflection opening mm mm

Crack opening mm

Deflection CMOD mm mm

0.50 1.50 2.50 3.50

0.45 1.36 2.27 3.18

0.41 1.23 2.05 2.86

0.63 1.89 3.16 4.42

0.41 1.23 2.05 2.86

389

0.46 1.36 2.30 3.22

perfectly flat; otherwise, problems can be caused during testing. The dimensions of the plates in a three-point bending [12-13] test on square panels are different than the dimensions of the beams in the EN14651 test. Because of this, the relation between the CMOD and the deflection is different as well. Three definitions need to be taken into account (see also Figure 6): 1. CMOD: crack mouth opening displacement: linear displacement measured at the bottom of the notch of the beam 2. Deflection: linear displacement, measured by a transducer, between the bottom of the notch and the horizontal line which connects the points located in the middle of the beam, above the supports 3. CO: Crack opening: linear displacement meas­ ured at the top of the notch of the beam

sprayed panel or EFNARC three-point bending test on square sprayed panel) have to be performed to assure the characteristics defined by the design. In addition, tests can be suggested to verify the fibre content or the fibre orientation. In order to check the compressive properties of the concrete, the same procedure adopted for ordin­ ary sprayed concrete should be followed. There are two methods to determine the tensile properties of the fibre reinforced concrete: 1. By performing a EN 14651 test on a section of a sprayed panel; 2. By performing an EFNARC three-point bending test on a square section of sprayed panel. The material should be classified according to fib Model Code 2010. To determine the characteristic values of the Fibre Reinforced Concrete residual strengths (fLk, fR1k and fR3k) it is suggested to perform at least 9 speci­ men tests according to EN14651 or EFNARC three point bending at 28 days of curing. The test results can be considered positive if:

This testing method has been investigated through an exhaustive RTT Program lead by the Ruhr Uni­ versity [14] for the CEN committee. There are some key points of conclusion: 1. Good Repeatability standard deviations and reproducibility standard deviations of EFNARCpanels in accordance with ISO 5725-2 2. Very low standard deviation between-lab Many tests are currently conducted to better assess the correlation with EN 14651 [12-13] 3 QUALITY CONTROL

1. The characteristic value of fR1k is higher than the values designed; 2. The ratio between fR3k and fR1k meets the design specifications; (If there are no specifica­ tions in the design, the strength ration of material may be higher.); 3. The fulfilment of the fib Model Code 2010 pre­ scription for substituting the traditional. 4. Reinforcement with fibre is verified (fR1k/fLk > 0.4 and fR3k/fR1k > 0.5). With:

A procedure for the control of Fibre Reinforced Con­ crete performance has to be defined in the design process. Usually a quality control procedure considers two steps: 1. Initial qualification of the material (trials testing) 2. Tests during the sprayed concrete lining produc­ tion (production testing)

In order to define the characteristic value from the tests results, the procedure suggested in Eurocode 2 [8] can be used. The average value mx and the coef­ ficient of variation Vx (Table 4 and Table 5) are defined:

Before starting the sprayed concrete lining, com­ pressive and bending tests (EN14651 cut from the

Figure 6. Definition of crack opening, CMOD and deflection.

390

Table 4.

Unknown Vx.

n

kn

3 4 5 6 8 9 10 12 15

3.37 2.63 2.34 2.18 2.01 1.96 1.92 1.87 1.82

Table 5. n

Table 6.

Known Vx.

Adopted mixes.

kn Mix name

3 4 5 6 8 9 10 12 15

1.89 1.83 1.80 1.77 1.74 1.73 1.72 1.71 1.70

MIX BEKAERT 1 MIX BEKAERT 2 MIX BEKAERT 3 MIX K&H 1

Concrete compres­ sive strength (MPa)

Fibre type

Fibre con­ tent (kg/m³)

55.3

4D 65/ 35BG 4D 65/ 35BG 4D 65/ 35BG 4D 65/ 35BG 4D 65/ 35BG 5D 65/ 60

25

55.3 86.6 46.0

MIX K&H 2 43.5 MIX K&H 3 42.3

The value of kn is defined according to ISO 12491, with the use of a Student’s distribution: with u0.05 fractile of the t- distribution for the probabil­ ity 0.05. Table 7.

40 25 25 40 40

EN14651 average results [MPa]. EN 14651 fL

fR1

fR2

fR3

fR4

2.7 3.5 4.4. 1.7 2.2 2.3

3.0 4.0 4.8 2.1 3.2 3.7

2.9 4.0 4.0 2.3 3.7 4.3

2.6 3.5 3.2 2.1 3.4 4.2

MPa Mix Bekaert 1 Mix Bekaert 2 Mix Bekaert 3 Mix K&H 1 Mix K&H 2 Mix K&H 3

4 COMPARISON OF THE RESULTS OBTAINED WITH THE TWO-TESTING SET-UP To compare the results obtained with EN 14651 and EFNARC tests, an experimental campaign was organized [15-16]. Tests were performed in two dif­ ferent laboratories: Bekaert Asia R&D Center and K&H GEOTECHNICAL LAB. In total six different mixes were prepared, mainly considering different fibre content, different concrete matrix, different type of fibres. The adopted mixes are summarized in Table 6. In the Bekaert Asia R&D Center tests six beams and sixpanels were cast for every mix; whereas in K&H GEOTECHNICAL LAB 12 beams and 12 panels were prepared for every mix. Table 7 and 8 show the average results in terms of fL, fR1, fR2, fR3, fR4 obtained with EN 14651 and EFNAR. Figure 7 shows the ratio between EN14651 and EFNARC average results for fL, fR1, fR2, fR3, fR4. Figure 8 shows the average value for all mixes of the results presented in Figure 7. It can be noted that the ratio is very close to 1.0 for all the quantities fL, fR1, fR2, fR3, fR4. This it means that the different between EN 14651 and EFNARC are limited.

Table 8.

5.1 5.3 7.6 3.3 3.4 2.9

EFNARC average results [MPa]. EFNARC fL

fR1

fR2

fR3

fR4

3.3 4.2 4.3 1.7 2.6 2.9

.3.7 4.8 4.6 2.1 3.7 4.5

3.5 4.6 3.8 2.1 3.7 4.5

3.0 3.9 3.0 1.9 3.5 4.2

MPa Mix Bekaert 1 Mix Bekaert 2 Mix Bekaert 3 Mix K&H 1 Mix K&H 2 Mix K&H 3

5.2 5.1 7.9 3.4 3.9 3.8

The same analysis is made considering only the three mixes having 25 kg/m3 of fibre. This is a typical content for sprayed concrete. In this case (Figure 9) the different between EN 14651 and EFNARC is very limited.

391

Table 9.

Standard deviation. EN 14651 LOP

CMOD 1 CMOD 2 CMOD 3 CMOD 4

MPa Mix Bekaert 1 Mix Bekaert 2 Mix Bekaert 3 Mix K&H 1 Mix K&H 2 Mix K&H 3

Figure 7. EN 14651/EFNARC average results.

0.14

0.54

0.62

0.63

0.50

0.24

0.65

0.94

0.93

0.73

0.34

0.77

1.00

0.80

0.61

0.91

0.56

0.71

0.76

0.73

0.3

0.7

1.0

1.1

1.0

0.9

0.7

1.1

1.3

1.3

EFNARC LOP

CMOD 1 CMOD 2 CMOD 3 CMOD 4

MPa Mix Bekaert 1 Mix Bekaert 2 Mix Bekaert 3 Mix K&H 1 Mix K&H 2 Mix K&H 3

Figure 8. Average value mixes.

Figure 9. EN14651/EFNARC average results for mixes with 25 kg/m³ of fibber.

ENFARC test in general give a characteriza­ tion of fibre reinforced concrete similar to EN 14651. EFNARC results in terms of fL, fR1, fR2, fR3, fR4 are slightly higher respect to the EN 14651: this is probably due to the higher

0.26

0.4

0.48

0.35

0.23

0.18

0.50

0.66

0.62

0.50

0.46

0.26

0.32

0.29

0.27

0.5

0.3

0.5

0.4

0.4

0.3

0.3

0.5

0.5

0.5

0.3

0.6

0.8

0.8

0.7

fracture area (almost 3 times higher in EFNARC respect to EN14651). The aforementioned statement is confirmed by the analysis of the standard deviation (Table 9) and the coefficient of variation of the results (Table 10). Table 4.3 and table 4.4 show the standard deviations and the coefficients of vari­ ation of fL, fR1, fR2, fR3, fR4 obtained with EN 14651 and EFNARC. Figure 10 shows the ratio between EN 14651 and EFNARC coefficient of variation results for fL, fR1, fR2, fR3, fR4. It can be noted as the coefficient of variation for fR1, fR2, fR3, fR4 are remarkably lower for EFNARC tests.

392

5 CONCLUSION

Table 10. Coefficient of variation. EN 14651 LOP

CMOD 1 CMOD 2 CMOD 3 CMOD 4

MPa Mix Bekaert 1 Mix Bekaert 2 Mix Bekaert 3 Mix K&H 1 Mix K&H 2 Mix K&H 3

0.03

0.20

0.20

0.22

0.20

0.05

0.19

0.23

0.23

0.21

0.05

0.18

0.21

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Figure 10. Ratio between EN 14651 and EFNARC coeffi­ cient of variation results.

Fibre reinforced sprayed concrete lining is a relevant material to be used for preliminary lining and for final lining. All the relevant standard concerning the prod­ uct, the testing, the design and quality control are today available to allow designers to specify the right performance. A good understanding of the material requires a complete information on the FRC material property. This is the reason why it is recommended to specify the Energy absorption and the residual strength in all cases. For struc­ tural use, mechanical performance of FRC must be verified according to the fib Model Code 2010 requirements and material properties determined based on three point bending test to limit the structural effect [17]. EFNARC test appears to be a suitable for charac­ terizing sprayed fibre reinforced concrete. The specimen adopted in EFNARC can be easily prepared with spay concrete since the same moulds for EN 14488-5 can be used. It has to remark that beams with the geometry proposed in EN 14651 are difficult to prepare with sprayed concrete and should not be representative of the actual material properties. The results obtained with EFNARC test can be used for the characterization of sprayed fibre reinforced concrete as demonstrated by the results of the laboratories tests. This is particularly evident for fiber content typically adopted in sprayed concrete. However further investigation has been conducted with different mix to consolidate this first conclu­ sion. Additional testing analyse will help to discuss this testing method. We recommend this test for all sprayed concrete application in order to check the minimum perform­ ance required by the Model Code 2010. Fibre reinforcement can substitute (also partially) conventional reinforcement at ultimate limit state if the following relationships are fulfilled: fR1k/fLk > 0.4; fR3k/fR1k > 0.5 A permanent sprayed concrete lining should be considered in the same way as any other per­ manent concrete structure. Hence, codes like Eurocode 2 [18] and ACI 318 [19], should be applied for the design and acceptance of the requirements for normal loading conditions in the long term. Quality and safety can be achieved using the rele­ vant product for the right use. The use of the finished material should be con­ sidered along with the test and performance criteria: 1. Post crack behaviour. 2. Match crack widths and deformation in the test to expectations in the project and durability requirements.

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REFERENCES [1] fib Model Code for Concrete Structures 2010 (2012), Final Complete Draft, fib bulletins 65 and 66, March 2012-ISBN 978-2-88394-105-2 and April 2012-ISBN 978-2-88394-106-9. [2] A new approach for Fibre reinforced Shotcrete under dynamic Load. Gagnon/Jolin Eight Inter­ national Symposium on Sprayed Concrete Norway June 208. [3] Gagnon, Antoine and Jolin, Marc (2017) ‘Specify­ ing and Testing Fibre Reinforced Shotcrete: Advances and Challenges’, in Shotcrete for Under­ ground Support XIII, Engineering Conference Inter­ national, p. 10. [4] EN 14487-1 (2005) ‘EN 14487-1 Sprayed concrete ­ Part 1: Definitions, specifications and conformity’. European Committee For Standardization, pp. 1–36. [5] ISO 13 270: Steel fibres for concrete — Definitions and specifications. (First edition 2013-01-15) I. S. Jacobs and C. P. Bean, “Fine particles, thin films and exchange anisotropy,” in Magnetism, vol. III, G. T. Rado and H. Suhl, Eds. New York: Aca­ demic, 1963, pp. 271–350. [6] Testing sprayed concrete – part 5: Determination of energy absorption capacity of fibre reinforced slab specimens”. EN144 Asquapro (2012). [7] ‘ASQUAPRO Technical booklet - Use of fibers in the reinforcement of sprayed concretes for tunnels temporary support’. Asquapro, pp. 1–51.

[8] Larive C. “Restrictive specifications for reinforced sprayed concrete for underground support”. World Tech Floor, 2013, Geneva, Asquapro. [9] Test method for metallic fibre concrete – measuring the flexural tensile strength (limit of proportionality (LOP), residual)”. EN14651: 2005, April 3rd, 2005. [10] RILEM TC 162-TDF (Oct 2003): Steel fibre concrete. [11] EFNARC. “Three point bending test on square panel with notch”. [12] EFNARC CREEP TEST PROCEDURE DESCRIP­ TION FOR SPRAYED CONCRETE AND TEST RESULTS WITH STEEL AND SYNTHETIC FIBRES – De Rivaz Spritzbeton-Tagung 2015, Alp­ bach Austria. [13] FIBRE SPRAYED CONCRETE RELEVANT TEST OF CHARACTERIZATION FOR DESIGN – De Rivaz Spritzbeton-Tagung 2012. [14] Ruhr University Bochum Faculty of Civil and Environmental Engineering Institute for Building Materials Short Report: RRT on Sprayed Concrete ­ WG 10- M.Sc. Sven Plückelmann. [15] Laval University test report 2019 Marc Jolin. [16] Roma University test report 2018 Alberto Meda. [17] Meda, Alberto and Rinaldi, Zila (2017) Technical report - Beam Tests on Fiber Reinforced Concrete according to EN14651 and ASTM C1609 Standard. [18] Eurocode 2. [19] ACI 318: Building Code Requirements for Reinforced Concrete.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Undrained seismic response of tunnels

E.A. Sandoval Universidad del Valle, Cali, Colombia

A. Bobet Purdue University, West Lafayette, USA

ABSTRACT: The seismic response of tunnels has been well-studied for structures placed in linear-elastic ground, but the results may be only applicable to tunnels placed in stiff soil or for moderate earthquakes. There is little information regarding the behavior of buried structures in soft soils, since they do not show a linear behavior, and for undrained conditions, where excess pore pressures accumulate during the earth­ quake. The paper presents the results of two-dimensional dynamic numerical analyses, under drained or undrained loading, to assess the seismic response of deep circular tunnels located far from the seismic source. It is assumed that the liner remains elastic and that the ground response can be approximated with a nonlinear elastoplastic constitutive model that includes excess pore pressures accumulation with cycles of loading. The liner deformations and load demand strongly depend on the flexibility ratio, F, which is a measure of the relative stiffness between the ground and the tunnel. Undrained conditions, compared with drained conditions, tend to reduce deformations for flexible liners and increase them for stiff tunnels early during the earthquake. With further cycles of loading, as the excess pore pressures increase, the differences in tunnel distortions between drained and undrained loading are reduced, i.e., they increase for flexible and decrease for stiff tunnels. Undrained loading, for stiff tunnels with F ≤ 2, produces larger thrust in the liner than drained loading. For more flexible tunnels, with F > 2, the behavior is the opposite: smaller axial forces for undrained loading. For tunnels with F > 2, the bending moments in the liner are not affected by the type of loading, drained or undrained, or by the magnitude of the excess pore pressures. For tunnels with F < 2, the bending moments increase from drained to undrained loading and with the magnitude of excess pore pressures.

1 INTRODUCTION Underground structures must be able to support static overburden loads as well as to accommodate additional deformations imposed by seismic motions. Soil-structure interaction and stress and dis­ placement transfer mechanisms from the ground to the structure, during a seismic event, have been explored extensively in the last few years. For most tunnels, perhaps excluding submerged tunnels, it seems that the most critical demand on the structure is produced by shear waves traveling perpendicular to the tunnel axis (Bobet, 2003; Wang, 1993; Merritt et al., 1985; Hendron and Fernández, 1983), which distort the tunnel’s cross section. Such distortions are called racking for a rectangular tunnel, and oval­ ing for a circular tunnel, as shown in Figure 1, and produce axial forces and bending moments to the structure. The seismic response of tunnels is a soil-structure interaction problem, where the tunnel modifies the deformation of the surrounding ground and thus

DOI: 10.1201/9780429321559-51

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demand and response depend on the relative stiffness between the ground and the tunnel support. Such interaction can be expressed by the flexibility ratio index, F, after Peck et al. (1972). The ratio F is a dimensionless parameter that quantifies the resist­ ance of tunnel and ground against distortion, under a state of pure shear, and so it is often used as a measure of the relative stiffness between the under­ ground structure and the surrounding ground. Previ­ ous studies have shown that this is the most important parameter controlling the seismic response of tunnels and that other parameters such as depth and shape have second-order effects (Bobet, 2003; Wang, 1993; Sandoval and Bobet, 2017, Sandoval and Bobet, 2020). While all this has been wellstudied for structures under drained conditions, there is little information regarding the behavior of buried structures under undrained conditions, i.e., when excess pore pressures generated are not dissipated. This paper provides results of 2D plane strain dynamic numerical analyses conducted using FLAC 7.0 (Itasca, 2011a) for deep circular tunnels

Figure 1. Racking and ovaling deformation of the tunnel cross section (after Owen and School, 1981).

subjected to cyclic shear. The liner is assumed to remain in its elastic regime without relative displace­ ment with the ground, i.e., with a tied interface. Non­ linear ground behavior under drained and undrained loading is considered. The effects of the flexibility ratio on the distortions of the cross section of the tunnel, and on the axial forces (thrust) and bending moments on the liner are investigated for both drained and undrained loading. For comparison pur­ poses, results of linear-elastic analysis for drained loading using one of the closed-form solutions (e.g. Bobet, 2010) are also included. 2 CYCLIC NONLINEAR ELASTOPLASTIC MODEL The constitutive model used in the paper is built based on the work by Jung (2009) and later by Kha­ sawneh et al., (2017). The model incorporates the well-known Masing’s rules, where the response of soil under cyclic loading follows a hysteretic behav­ ior (Hardin & Drnevich, 1972; Pyke, 1979; Ohsaki, 1980). The model is implemented in FLAC 7.0 (Itasca, 2011a) and is verified by comparing its pre­ dictions with results from simple shear laboratory tests at different scales, under drained and undrained loading. The model is rate-independent, as it is gen­ erally assumed for the seismic response of most geo­ materials and is defined within a small-deformation framework for incremental plasticity theory. The model includes four main characteristics: i) non­ linear stress-strain and hysteretic behavior during cyclic loading; ii) dependence of the very small strain shear modulus on confinement; iii) coupled shear-volumetric strains with excess pore pressures accumulation; and iv) a yield criterion and nonassociated plastic flow rule. The nonlinear behavior is simulated through a hyperbolic formulation. Equation 1 is used to cal­ culate the shear modulus as the soil deforms. In the equation, a factor n = 1 is used for the initial loading curve, a factor n = 2 reproduces the Masing’s formu­ lation, while n > 2 and n < 2 account for cyclic hard­ ening or softening, respectively, during unloading and reloading. The variables needed in Equation 1 are given in Equations 2 to 5.

where Go is the shear modulus at very small strains, r is the normalized shear strain, n is a scaling factor for the hysteresis loop, γoct is the octahedral shear strain, γoct is the octahedral shear strain, γoct,r is the reference octahedral shear strain, a and b are con­ stants that determine how the stress-strain relation deviates from the hyperbolic function, σ’m is the effective mean stress, eij is the deviatoric strain tensor, εij is the Lagrangian strain tensor, and ui, uj are the displacements along the xi, xj axes. The terms α and κ represent strength parameters and are discussed later. The dependence of the ground stiffness on the mean effective stress has been extensively discussed in the literature (e.g., Hardin & Drnevich, 1972; Hardin, 1978, Porovic & Jardine, 1994). Such behavior is considered in Equation 6, which is used to obtain the incremental small-strain shear modulus (dGo) as a function of the incremental effective mean stress (dσ’m).

The model provides a formulation that couples shear and volumetric plastic strains, needed to account for excess pore pressures accumulation during undrained loading. The formulation is based on Byrne (1991), after the work by Martin et. al. (1974). Equa­ tion 7 is used to obtain the plastic volumetric strains. The tendency to generate compressional volumetric strains during drained cyclic loading leads to positive excess pore pressures accumulation during undrained loading, because the soil cannot change volume.

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where is the incremental increase in plas­ tic volumetric strain per ½ cycle of loading, γoct is the octahedral shear strain, γoct-th is a threshold octahedral shear strain, εv is the accumulated volumetric strain, and c1, c2 are constants for the coupling between shear and volumetric strains that depend on the relative density or consistency of the ground. The model also includes an elastoplastic formula­ tion with a yield criterion and non-associated flow rule. Although well design tunnels should experience small to moderate strains, local yielding may occur due to the decrease in confinement caused by the excess pore pressures accumulation. The total strain increment is decomposed into its elastic and plastic components The stress incre­ ment (dσij) is computed from the generalized Hooke’s law using Equation 8. Yielding is defined with the Drucker-Prager (D-P) criterion, and the plastic strains are determined from a strain hardening law with a non-associated flow rule. Equations 9 and 10 show the yield function and the plastic potential, respectively. Equations 11 and 12 are used to deter­ mine the plastic strains, once yielding has occurred, invoking the consistency condition.

is

where Cijkl is the elastic modulus tensor, the incremental elastic strain tensor,

is the second invariant of the deviatoric stress tensor,

is the first invariant of the stress tensor, α is the D-P friction angle, κ is the D-P cohesion, and ψ is the interlocking component of strength, i.e., inter­ ference and dilation. The capabilities of the model are verified by com­ paring results of simulations with cyclic simple shear laboratory tests, at different scales, under drained and undrained loading. As an example, Figure 2 shows experimental results and the simula­ tions for an undrained cyclic simple shear test con­ ducted on Nevada sand by Chen (1995). The hysteresis loop and the normalized excess pore pres­ sures are compared in Figure 2. As seen in the fig­ ures, a good approximation is obtained with the constitutive model.

Figure 2. Comparison between simulations and results of laboratory undrained cyclic simple shear tests on Nevada sand, from Chen (1995).

3 TUNNEL DEFORMATIONS The effect of the flexibility ratio on the tunnel distor­ tions is investigated. Distortions for circular tunnels with 4.0 m in diameter and 0.40 m liner thickness are obtained for nonlinear ground under drained loading, and under undrained loading with different levels of excess pore pressures. More precisely, results for the first, fifth, fifteenth, and thirtieth peak of the cycles of loading are reported. At those peaks, the normalized excess pore pressures in the free field (i.e. at the loca­ tion of the tunnel, but without the tunnel structure), Δu/p’o, are 0.02, 0.21, 0.39 and 0.53, respectively. Δu is the excess pore pressures in the free field and p’o is the initial effective confinement stress. For comparison purposes, results from the closed-form solution for drained linear-elastic ground provided by Bobet (2010) are included. The mesh has dimensions 300 m x 100 m and the tunnel is placed at 75 m depth. The depth of the model is selected following the recom­ mendations from Itasca (Itasca, 2011b) to avoid the effect of waves reflection from the free surface. Reflec­ tions from the bottom and sides are minimized by pla­ cing quiet and free-field boundaries. These are absorbing boundaries built up in FLAC, which use independent dashpots in the normal and shear direc­ tions to avoid or decrease energy radiation (Itasca, 2011b; Lysmer & Kuhlemeyer, 1969). A dynamic input velocity with amplitude equal to 0.1 m/s and frequency of 2.5 Hz is imposed at the bottom of the model. The amplitude corresponds to a strong earthquake with magnitude VI in the Modified Mercalli Intensity Scale, according to the classification provided by Wald et al. (1999). The frequency is repre­ sentative of far-field motions, whose predominant fre­ quencies range between 0.1 and 10 Hz, according to Dowding (1985). For the ground, a shear modulus (G) and a Poisson’s ratio (ν) equal to 80 MPa and 0.25 are assumed, respectively. For the liner, a value of 0.15 is used for the Poisson’s ratio (νs). The initial flexibility

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ratios range from 0.10 to 15, following the definition provided by Peck et al. (1972), given in Equation 13. Due to the stiffness degradation during the dynamic loading, the flexibility ratios are reduced, from 0.07 to 0.09, and from 9.0 to 14.1, respectively. The reduction in values is variable because those are related to the ground stiffness, which changes between drained and undrained loading, and between the different cycles in undrained loading.

where Em is the Young’s modulus of the medium, Is is the moment of inertia of the liner per unit length, and R is the radius of the tunnel. Figure 3 shows the results of the analyses in terms of normalized distortions of the tunnel versus flexibility ratio, for both drained and undrained loading. The inset in the figure contains results for stiff tunnels, i.e. for F ≤ 1. The normalization is done with respect to the distortions that would occur in the free field (i.e. with­ out the tunnel). As seen in the figure, for the nonlinear ground under drained or undrained loading, the nor­ malized distortions increase with the increase of the flexibility ratio, as has been previously reported from linear-elastic analysis (e.g., Bobet, 2010; Wang, 1993). It can also be seen in Figure 3 that larger normalized distortions are observed for drained loading using the nonlinear ground model (open circles) than for the linear-elastic model (dashed line). This result is expected, as the nonlinear ground deforms more and so there is a larger deformation demand on the tunnel. It is also seen in Figure 3 that, early during the earth­ quake, the undrained condition (filled markers) tends

Figure 3. Effect of flexibility ratio on distortions of circular tunnels under drained or undrained loading.

to reduce deformations for flexible structures and increase them for stiffer structures, compared to drained loading. However, when pore pressures in the ground increase with cycles of loading, the normalized tunnel distortions increase for flexible liners and decrease for stiff liners. For large magnitude of the excess pore pressures (see for example the inverted tri­ angle that represents the 30th peak), the normalized dis­ tortions for flexible tunnels can be even larger than those from drained loading, while for stiff tunnels, they are smaller. The threshold flexibility ratio from where the normalized distortions change from increasing to decreasing with cycles of loading is at F = 2, approxi­ mately. Interestingly, this is the same threshold found using linear-elastic analyses (Bobet, 2010). The behavior discussed above can be explained as the combined effect of the stiffness degradation around the tunnel and in the free field. The free field distor­ tions always increase with the excess pore pressures due to the stiffness degradation triggered by the reduc­ tion of the mean effective stress (σ’m = p’). The abso­ lute tunnel distortions also increase with the increase of the excess pore pressures; however, the magnitude depends on the flexibility ratio. For flexible tunnels, there is an important increase in the absolute distortions of the tunnel, which is larger than the increase in the distortions in the free field. For stiff tunnels, on the other hand, the increase in the distortions is almost negligible. As a result, the normalized distortions increase for flexible tunnels and decrease for stiff tun­ nels. As mentioned, the threshold flexibility ratio for the change in behavior is for F = 2. 4 LOADING ON THE LINER Figure 4 shows axial (thrust) forces of the liner for the same cases discussed in the previous section, i.e., for drained loading and for undrained loading with nor­ malized excess pore pressures (Δuff/p’o) between 0.02 and 0.53. As suggested by Einstein and Schwartz (1979), the axial forces are normalized by the product of the input shear stress and the radius of the tunnel (τR). Similar to what was discussed for the distortions in the previous section, the flexibility ratios decrease with the number of cycles, i.e., with the level of excess pore pressures in the free field. As a result, for each ini­ tial flexibility ratio, the far-right marker corresponds to the drained loading and the far-left corresponds to the larger cycle considered for that flexibility ratio. Figure 4 shows that the drained linear-elastic analysis gives very similar values for the normalized axial forces than assuming a nonlinear ground, if the tunnels are flexible, with initial F ≥ 2. However, for stiff tunnels, with F < 2, larger thrusts are obtained with the nonlinear ground. As discussed in the previous section, early during the earthquake, the undrained loading produces much larger thrusts than the drained loading. Neverthe­ less, different to that seen with the distortions, the mag­ nitude of the excess pore pressures, i.e., the duration of the earthquake, does not introduce significant changes

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undrained) and on the magnitude of the excess pore pressures, i.e., on the number of cycles of loading, during undrained loading. However, for stiff tunnels with F < 2, the differences in the bending moments between drained and undrained loading increase as the flexibility ratio decreases, with larger magnitudes for undrained loading. Nevertheless, the magnitude of the excess pore pressures, i.e., the duration of the earth­ quake, has little effect on the bending moments. 5 CONCLUSIONS

Figure 4. Effect of flexibility ratio on axial (thrusts) forces of circular tunnels under drained or undrained loading.

to the axial forces (the maximum difference for the data shown is 6%). Figure 5 shows the bending moments normalized by the product of the input shear stress and the square of the radius of the tunnel (τR2), again, following the suggestion by Einstein and Schwartz (1979). The figure shows that, for flexible tunnels (F ≥ 2), the bend­ ing moments decrease with the flexibility ratio, and are independent of the drainage condition (drained or

Figure 5. Effect of flexibility ratio on bending moments of circular tunnels under drained or undrained loading.

The paper evaluates the drained and undrained seismic response of deep circular tunnels. Numerical simula­ tions are done using the commercial code FLAC 7.0, where it is assumed that plane strain conditions apply to any cross section perpendicular to the tunnel axis. A new elastoplastic hyperbolic constitutive model is developed, implemented and verified in FLAC. The model considers the stiffness degradation of the ground with deformation, the dependency of the small-strain shear modulus on confinement, the coupling between shear and volumetric strains, and yield with plastic deformations. The effects of the relative stiffness between the tunnel and the ground on the distortions of the tunnel cross section, and on the loading on the liner are inves­ tigated, for drained and undrained loading. The normal­ ized tunnel distortions increase when the flexibility ratio (F) between the tunnel and the ground increases. Larger distortions are obtained for nonlinear than for linear-elastic ground, with differences increasing as the tunnel becomes more flexible, i.e., when F increases. Early during the earthquake, undrained loading condi­ tions tend to reduce deformations of flexible structures and increase them for stiffer structures, compared to drained loading. However, when pore pressures in the ground accumulate with cycles of loading, distortions of flexible liners increase and of stiff liners decrease, thus the differences with those from drained loading are reduced. For larger magnitudes of the excess pore pressures, the normalized distortions for flexible tun­ nels can be larger, or smaller for stiff liners, compared to those under drained loading. As previously found using linear-elastic ground, the threshold flexibility ratio from where the normalized distortions change from increasing to decreasing with cycles of loading is at F ≈ 2. Larger axial forces and bending moments are obtained for stiff tunnels (F ≤ 2), for both drained and undrained loading. The demand decreases as the tunnel becomes more flexible, i.e., as the flexibility ratio increases. The undrained loading produces larger axial forces than the drained loading for stiff tunnels (F ≤ 2). In more flexible tunnels (F > 2), the behavior is the opposite. i.e., smaller thrusts are obtained for undrained loading. The magnitude of the excess pore pressures in the free field (or the duration of the earthquake), does not produce significant changes in the axial forces, irre­ spective of the flexibility ratio. The drainage condition,

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drained or undrained, or the magnitude of the excess pore pressures in the free field during undrained load­ ing do not affect the bending moments in flexible tun­ nels (F > 2). For stiffer tunnels (F ≤ 2), the bending moments are larger for undrained than for drained loading. The results highlight the importance of considering nonlinear ground and excess pore pressures accumula­ tion in the ground (if applicable) when the seismic response of tunnels is investigated. The flexibility ratio, which changes with the ground deformation, is the most important parameter controlling the seismic response. For stiff structures (F < 2), important axial forces and bending moments are produced, with larger magnitudes for undrained loading. However, the tunnel distortions are small. For flexible liners (F ≥ 2), the loading on the liner decreases, but the distortions become important. For very flexible tunnels, with F > 10, the structural behavior is mainly controlled by the tunnel deformations, given that the axial forces and bending moments are very small. Such distortions are drastically affected by the type of loading (drained or undrained) or by the magnitude of the excess pore pressures during undrained loading.

ACKNOWLEDGEMENTS The financial support provided by Colombia-Purdue Institute for Advanced Scientific Research (CPI), Col­ ciencias (Colombia), Colfuturo (Colombia) Universi­ dad del Valle (Colombia) and Purdue University (United States) are highly appreciated.

REFERENCES Bobet, A. 2003. Effect of pore water pressure on tunnel support during static and seismic loading. Tunnelling and Underground Space Technology 18(4): 377–393. http://dx.doi.org/10.1016/S0886-7798(03) 00008-7. Bobet, A. 2010. Drained and undrained response of deep tunnels subjected to far-field shear loading. Tunnelling and Underground Space Technology 25 (1): 21–31. http://dx.doi:10.1016/j.tust.2009.08.001. Byrne, P. M. 1991. A cyclic shear-volume coupling and pore pressure model for sand, Proc. Second Inter­ national Conference on Recent Advances in Geotech­ nical Earthquake Engineering and Soil Dynamics, 47–55, St. Louis, Missouri, USA, 11–15 March 1991. Chen, Y. -R. 1995. Behavior of a fine sand in triaxial, tor­ sional and rotational shear test, PhD thesis, Univ. of California, Davis, CA, USA. Dowding, C. 1985. Earthquake response of caverns: empir­ ical correlations and numerical modeling, Proc. 1985 Rapid Excavation and Tunneling Conference, 1: 71–83, New York, New York, USA, 16–20 June 1985. Einstein, H. H. & Schwartz, C. Simplified analysis for tunnel supports, Journal of the Geotechnical Engineer­ ing Division, 105 (GT4): 499–518. Hardin, B. O. 1978. The nature of stress-strain behavior for soils, In: Specialty Conference on Earthquake Engineer­ ing and Soil Dynamics ASCE, 3–90 Pasadena, Califor­ nia, USA, 19–21 June 1978.

Hardin, B. O. & Drnevich, V. P. 1972. Shear modulus and damping in soils: design equations and curves. Journal of the Soil Mechanics and Foundations Division 98 (7): 667–692. Hendron Jr., A. J. & Fernández, G. 1983. Dynamic and static design considerations for underground chambers, In: Seismic Design of Embankments and Caverns, ASCE Symposium, 157–197, Philadelphia, Pennsylvania, USA, 16–20 May 1983. Itasca. 2011a. FLAC Fast Lagrangian Analysis of Con­ tinua, Ver. 8.0. Minneapolis, MN, USA: Itasca Consult­ ing Group. Itasca. 2011b. FLAC Fast Lagrangian Analysis of Con­ tinua, Ver. 8.0, Dynamic Analysis (Fifth ed.). Minneap­ olis, MN, USA: Itasca Consulting Group. Jung, C. M. 2009. Seismic loading on earth retaining struc­ tures. PhD thesis, Purdue University, West Lafayette, Indiana, USA. Khasawneh, Y., Bobet, A., & Frosch, R. 2017. A simple soil model for low frequency cyclic loading. Computers and Geotechnics 84: 225–237. http://dx.doi:10.1016/j. compgeo.2016.12.003. Lysmer, J. & Kuhlemeyer, R. L. 1969. Finite dynamic model for infinite media. Journal of the Engineering Mechanics Division 95 (4): 859–878. Martin, G. R., Finn, W. L., & Seed, H. B. 1974. Fundamen­ tals of liquefaction under cyclic loading. Technical report, The University of British Columbia, Soil Mech­ anics Series No. 23. Masing, G. 1926. Eigenspannungen und verfestigung beim messing, In: Proceedings, Second International Con­ gress of Applied Mechanics, 332–335, Zürich, Switzer­ land, 12–17 September 1926. Merritt, J. L., Monsees, J. E., & Hendron Jr, A. J. 1985. Seismic design of underground structures. In Proc. 1985 Rapid Excavation and Tunneling Conference, 1: 104–131, New York, New York, USA, 16–20 June 1985. Ohsaki, Y. 1980. Some notes on Masing’s law and non­ linear response of soil deposits. Journal of Faculty of Eng., The University of Tokyo Ser B 105 (4): 513–536. Peck, R. B., Hendron, A. J., & Mohraz, B. 1972. State of the art of soft-ground tunneling, Proc. 1972 Rapid Exca­ vation and Tunneling Conference, 1: 259–286, Chicago, Illinois, USA, 5–7 June 1972. Porovic, E. & Jardine, R. 1994. Some observations on the static and dynamic shear stiffness of Ham River sand. In: Proceedings of the International Symposium on Pre­ failure Deformation Characteristics of Geomaterials, 25–30, Sapporo, Japan, 12–14 September 1994. Pyke, R. 1979. Nonlinear soil models for irregular cyclic loadings. Journal of the Geotechnical Engineering Div­ ision 105(GT6): 715–726. Sandoval, E. & Bobet, A. 2017. Effect of frequency and flexibility ratio on the seismic response of deep tunnels. Underground Space 2 (2): 125–133. Sandoval, E. & Bobet, A. 2020. Effect of input frequency on the seismic response of deep circular tunnels. Soil Dynamics and Earthquake Engineering 139 (2020): 106421. doi:10.1016/j.soildyn.2020.106421. Wald, D. J., Quitoriano, V., Heaton, T.H. & Kanamori, H. 1999. Relationships between peak ground acceleration, peak ground velocity, and modified Mercalli intensity in California, Earthquake Spectra, 15 (3): 557–564. doi:10.1193/1.1586058. Wang, J. N. 1993. Seismic design of tunnels. A simple state-of-the-art approach. Technical report, Parsons Brinckerhoff Monograph 7.

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Preliminary evidences on the influence of grains micro-structural features on the TBM tools wear D. Sebastiani & S. Miliziano Department of Structural and Geotechnical Engineering, University of Rome “La Sapienza”, Rome, Italy

G. Guida Department of Civil and Evironmental Engineering, Politecnico di Milano, Milan, Italy

F. Casini Department of Civil and Informatic Engineering, Università degli Studi di Roma “Tor Vergata”, Rome, Italy

ABSTRACT: The wear of the excavation tools during the mechanized tunnelling in coarse grained soils is an extremely widespread phenomenon that should be taken into account to properly develop an excavation project. In fact, the excessive tools wear reduces the penetration speed, increases the energy consumption and requires stops necessary to replace the worn tools. Thus, to optimize the process, the tools consumption rate should be necessarily predicted in advance. Different works have focused the attention on the relationship between the tools wear and the quartz content of the soil to be excavated. In this research, a step forward is made focusing the attention on the role played by the morphology of the grains in the wearing process. The paper includes some experimental results of abrasion tests carried out on different sand samples. The results gathered lead to evaluate the influence of grain morphology of coarse soils on the wear of TBM tools. The experimental data reported are part of a wide research project aiming to link the mechanical behaviour of coarse soils with the grain morphology in different engineering applications.

1 INTRODUCTION In the last century the diffusion of tunnel boring machines (TBM) has led to the development of an ever-increasing number of tunnels built in extremely difficult contexts, linked to the strong interaction with the historical heritage of large metropolitan areas (Pirone et al., 2019; Rampello et al., 2012) or to the hydraulic and geotechnical conditions of the underground context (Losacco & Viggiani 2019; Boldini et al., 2018; Russo et al., 2012). Among the various risks to which a TBM is sub­ jected during the tunnel excavation, the wear of the excavation tools and the metal carpentry of the cutter-head must certainly be mentioned. It is caused by the prolonged rubbing, particularly rele­ vant in the presence of coarse soils or rocks and can lead to the slowing down of excavation activities, the need to replace excavation tools more fre­ quently than planned or unexpected stops. These eventualities entail delays in the excavation activ­ ities, additional costs and risks for workers linked to the intervention in the excavation chamber through the TBM hyperbaric chamber. The ability to predict and control the phenomenon of wear

during the design phase and the possibility of plan measures to mitigate this risk are therefore funda­ mental for the optimization and safety of tunnelling projects. During the digging operation, the wear recorded on the TBM cutter-head is a phenomenon widely studied in literature (e.g. Nilsen et al., 2006; Ghar­ anbagh et al., 2011). Barzeghari et al. (2015) stud­ ied the main parameters affecting the abrasion phenomena, while Jakobsen (2014) predicted the excavation tools life based on a large set of experi­ mental data. Finally, the design of the excavation tools in term of shape and materials is reported in Küpferle (2017). From the experimental point of view, different authors have used the abrasion tests to quantify the TBM tool wears on soil and rock samples (Käsling & Thuro, 2010; Gharanbagh et al., 2011; Jakobsen et al., 2013; Sebastiani et al., 2016). Based on the literature data, the relation between the quartz content and the TBM tools is widely investigated. While, there is a lack of understanding on the role played by the morphology of coarsegrained soils. The latter is due to two reasons: it is considered a less relevant factor; it is missing

DOI: 10.1201/9780429321559-52

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a simple and effective tool able to proper character­ ize the shape of coarse materials. In this work, the method proposed in Guida et al., (2019b) is applied to characterize the morphology of grains and a first attempt is made to relate the grain shape to the TBM tools wear. This study is part of a wide experimental research activity developed to point out the relation between the morphology of grains, the physical and the mechanical properties of coarse soils. To this aim, several samples of three sands (see Table 1) were tested at the Geotechnical Laboratory of Sapienza University of Rome. The following quantities has been determined: the grain size distribution, the min­ eralogical composition; the shape, angularity and surface roughness of the grains. In the following is reported a description of the experimental apparatuses used and of the tests per­ formed. Then, the attention is draw to the fractal analysis of particle contour and to a recent method developed to characterize the grain shape at multiscale based on it. Finally, the abrasiveness measured is related to the grain shape descriptors obtained with the fractal method, emphasizing the role played by the three scale of morphology on the observed behaviour. 1.1

Soil samples

Two of the sands selected where retrieved from sites where tunnelling projects were running, specifically Rome sample from Rome Metro C project and Milan sample from Milan M5 project. While the third sand, Colleferro is a natural sand retrieved from a site in the Lazio region (see also Guida et al., 2019a). They are silica sands with different mineralogical compos­ ition: Colleferro is composed by quartz (48%), calcite (15%), albite (15%), microcline (13%), and others (9%); Milano has the greater quartz content (72%) plus albite (16%), orthoclase (6%) and others (6%), the mineralogical composition of Rome is missing. Uniform samples of the three sands, with grain size ranging between 1.0-1.5 mm (US sieve size No 14-18, respectively), were tested both in the natural state (N) and in an artificially grinded state (G)

obtained in laboratory through a mechanical crush­ ing process performed by throwing grains of size greater than 1.5 mm and collecting and sieving frag­ ments with grain size ranging between 1.0-1.5 mm. In Table 1 are listed the soil samples tested together with the intrinsic properties measured, such as the minimum and maximum void ratio, emin and emax, according to ASTM-D4254 (2006). The values obtained within this experimental work well agreed with those measured by Guida et al. (2019a) on natural grains of the same sands. Further, the values of emin and emax for grinded materials are higher than those of natural sands (Table 1). This is probably due to the increasing shape irregularity of the grinded grains (Miura et al., 1997; Cho et al., 2006; Simoni & Houlsby, 2006; Yang & Luo, 2015; Altuhafi et al., 2016; Suh et al., 2017; Guida et al., 2019a). 1.2

Analysis of grains morphology

The grain morphology has been quantified with the method developed by Guida et al. (2019b), based on the fractal analysis of particle contour obtained by two-dimensional images of grains. The fractal method is based on the measurement of the contour length using set of sticks as unit of different sizes. Figure 1 reports a typical results of the fractal ana­ lysis of three objects with different shape –a circle; a Toyoura sand grain; the fractal of Koch– in terms of contour length p, as a function of stick unit b, both normalized by the equivalent diameter D (diameter of the equivalent sphere). The circle is characterized by the smallest con­ tour length for a given area. Its contour achieves an asymptote of p/D~3.14 already for stick lengths of b/D~0.4, while the Toyoura grain and the Koch snowflakes contours continue to increase decreas­ ing the unit length in accordance to the definition of fractal outline (Turcotte, 1986). In particular, the natural sand grains are usually characterized by a bi-linear trend with slope m and μ.

Table 1. Intrinsic properties of the natural (N) and grinded (G) sand.

Sand Colleferro Milano Roma

N G N G N G

Gs

emin

emax

[-]

[-]

[-]

2.72

0.68 0.79 0.60 0.67 0.72 0.82

1.05 1.21 1.13 1.21 1.14 1.26

2.70 2.73

Figure 1. Fractal analysis of particle contour.

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The fractal of Koch is characterized by a linear trend of slope m=μ=0.216, representing its fractal dimen­ sion. The contour length increases with the complex­ ity of the grains examined and with the decreasing of the stick length. The interpretation of the fractal curves reported in Figure 1 allows to define three multiscale morpho­ logical descriptors of contours (after Guida et al., 2019b): 1) M (macro-scale parameter) is defined as the ratio Δcircle/Δ, where Δ is the increase of contour per­ imeter between b/D=1 to b/D=bm/D, and Δcircle is that related to the circular shape equal to ~1.14. It is a descriptor of circularity and assumes values between 0 and 1, where 1 represents the perfect circle. 2) m (meso-scale parameter) is the slope of the linear trend corresponding to the larger stick length. It is a descriptor of angularity and it assumes values ranging between 0 and 1, where 0 indicates a perfect roundness of corners (e.g. the circle). 3) μ(micro-scale parameter) is identified by the second linear slope of the plot. It describes the roughness, it is bounded between 0 and 1, where 0 indicates a perfect smoothness of the grain texture (e.g. the circle). The scale bm/D is the boundary value between the two linear trends of the grains contour (see Figure 1). It distinguishes the scale length of angularity to those of roughness and it represents the characteristic scale of asperities. 1.3

Abrasion test apparatus

The tribometer is used to perform the wear tests (see Figure 2 and Figure 3). It consists of a metal plate, where the sample is positioned, rotating around a vertical axis. At a fixed distance of 1 mm from the metal plate are placed two prismatic shaped tools of different materials made ad hoc for this study (the material used for the described tests is Plexiglas).

Figure 2. Simplified scheme of the tribometer.

Figure 3. Image of the tribometer.

To proper replicate the contact between the excavation tools and the soil is installed a spring, with known stiffness, on the vertical axis between the fixed frame and the tool. With the aim of investigate the effect of different contact pres­ sures, the tests were performed using two springs of different stiffness K1 = 6.57 kN/m and K2 = 80.4 kN/m. The test consists in placing the tool in contact with the soil sample in rotation for a time of 10 min. Then, by weight difference, the tool wear is expressed as weight loss divided by the contact area (mg/cm2). 2 RESULTS AND DISCUSSION 2.1

Morphology characterization

Natural grains and grinded grains of diameters ran­ ging between 1.0 and 1.5 mm were analysed to appreciate the change in morphology induced by a mechanical crushing. The number of grains ana­ lysed are fifty natural and fifty grinded, selected from the samples tested. The grains images are acquired with an optic microscope at a magnification of 50x, then they are processed with a Matlab Code developed to this aim (Guida et al., 2019b). Figure 4 reports selected results of twelve grains’ contour of each sand, six natural and six grinded. There are qualitative differences in shape between natural and grinded contours: grinded grains appear more elongated and angular than natural grains. Table 2 reports the morphological descriptors quantified with the fractal method. After grinding, it is possible to note:

403

evidence is also in accordance with experimental data presented on the same materials in Guida et al. (2019a), which support the hypothesis that an assembly of more regular, circular and smooth particles may arrange with a denser con­ figuration compared with irregular ones Cho et al.(2006). 2.2

Figure 4. Example of contour extracted from each set of grains analyzed.

Table 2. Results of fractal analysis of particle contour on N and G grains of sand.

Colleferro Milano Roma

N G N G N G

M [-]

m [-]

μ [-]

bm/D [-]

0.79±0.15 0.77±0.12 0.81±0.09 0.71±0.10 0.80±0.10 0.69±0.12

0.16±0.06 0.19±0.09 0.15±0.05 0.14±0.08 0.17±0.08 0.16±0.08

0.02±0.01 0.02±0.01 0.02±0.00 0.04±0.01 0.02±0.01 0.02±0.00

0.25 0.25 0.24 0.18 0.25 0.21

i) a slight reduction of morphology circularity (M tends to decrease according to the less circu­ lar overall shape of grinded grains); ii) the grains angularity of Colleferro m tends to slightly increase, while for the other sands it does not show substantially changes; iii) μ have a less evident trend, probably because of the too small scale referring to.

Abrasion test results

The abrasion tests were performed on the sam­ ples reported in Tables 1 using two springs of stiffness K1 and K2. The average values of the consumption are summarised in Table 3. Grinded sand samples exhibit a greater value of abrasion compared to the natural specimens. Further, a strong correlation exists between the stiffness of the springs (at contact between tool and soil sample, see Figure 2) and the measured tools wear. This effect can be ascribed to the increas­ ing normal contact forces developed between the tool and the sand grains. This observation agreed with different literature experimental results (Bar­ zegari et al., 2015) and with the evidences recorded during the in situ tunnels excavated with TBM (Amoun et al., 2017). The authors show an increase of tool wear with the contact pressure between soil and tools. As expected, the stiffer systems (K2) apply a higher load on the sand (higher sand-plate contact stresses), resulting in a bigger amount of abrasion >1.8 mg/cm2 in all the tests performed. Moreover, an increasing tool wear is measured for the artificially grinded samples compared to naturalspecimen. These are preliminary experimental evi­ dences supporting the hypothesis that also the morphology of grains plays a role in the abrasiveness of a sand together with mineralogy. Despite Milano has a higher quartz mineralogy content (meaning of greater hardness of the con­ stituent grain material) than Colleferro sand, it results less abrasive (see Table 3). The less abra­ siveness is in this case determined by the more regular shape of the Milano grains respect to Colleferro ones. Table 3. sand.

The characteristic size of asperities bm/D is for all sands around 0.2. This indicates that the angularities are related to a scale length range between 1-0.2 D, while roughness to a scale μ and thus the grains profile has angular corner but smoother texture. The change of morphology features is related to the higher void ratios of the grinded sam­ ples (see Table 1). Figure 5 shows the evolution of the void ratio limits emax and emin versus the parameters M (a) and m (b). The data shown a good correlation: they decrease with M and increase with m. This

Results of abrasion tests on N and G grains of Abrasion [mg/cm2] K1

Colleferro Milano Roma

404

N G N G N G

1.192 1.347 1.061 1.204 1.357 1.439

K2 2.235 2.500 1.864 1.918 2.622 2.939

Figure 5. Correlation between morphological descriptors of sand grains and the minimum and maximum void ratios.

2.3

Abrasiveness versus grain morphology

Figure 5 reports the abrasiveness as a function of the morphological descriptors M (a) and m (b). The interpretation yields the following findings: – the sand abrasion clearly increases as M reduces, because the increasing rubbing of the contacts, accordingly to the toll wear; – Roma and Milano show an increase in abrasion with a decreasing in m, while Colleferro exhibits an increase of abrasion with increasing m. – The increase of abrasiveness after grinding for Milano and Roma sands could be related to the rele­ vant decreases of M combined with a slight decrease of m. While Colleferro sand exhibits a less evident reduction of M after grinding and an increase of m. This different behaviour observed can be also related to the different mineralogy of the grains playing a rule in the change of morphological descriptors. From these preliminary results emerges a direct relation between abrasion and M, while further

investigation must be done to identify the relation with m. Further, for Colleferro sand, the effect of the appreciable increase of m is realistically the main reason of the abrasion increment, being very small the reduction of M. The two morphologic descriptors play an opposite role in respect to the abrasiveness increment: inversely related to M and directly related to m. As first attempt, to summarize the relation between the abrasiveness and the morphological descriptors, the macro (M) and meso (m) scale parameters are combined into a single descriptor of regularity Rm, defined as their arithmetic average:

� ¼ 1 - m is the complementary of where m angularity m. The descriptor of regularity Rm is dimensionless and assumes values ranging between 0 (when M = 0 and m = 1) and 1 (when M = 1

Figure 6. Correlation between abrasion and morphology parameters M and m and Rm.

405

and m = 0). Its definition take inspiration by the regularity parameter introduces by Cho et al. (2006). The authors define the regularity as the arithmetic mean between sphericity (macro-scale) and round­ ness (meso-scale). Small abrasiveness is expected for Rm values close to 1, while the highest abrasion is expected for value of 0. Figure 6 shows the correlation between sand abrasion and regularity of the grains morphology, Rm. Natural grains, character­ ized by higher values of Rm (0.82-0.83) > 0.8 are systematically less abrasive than grinded grains, characterised by more irregular morphology Rm (0.77-0.79) < 0.8. It is interesting to note how the grains of Milano sand are definitely the less abrasive, but also the less angular (lower values of m) and in particular their natural grains the most circular and regular (greater values of M and Rm). 3 CONCLUSIONS The paper investigates the factors influencing the abrasiveness of sand in relation to the wear of the excavation tools during the mechanized tunnelling. Three different sands were tested and the results compared. The tools wear is strongly related to the frame stiffness regardless the nature of the sand. Stiffer is the system, greater the contact forces between grains and metallic plate, and consequently higher the tool consumption. A greater thrust force of the TBM advancement may lead to a stiffer system and to a more abrasion of the excavation tools. From this preliminary study, interesting correl­ ation have been found between the tools wear and the grain morphology. The latter has been quantified with the fractal analysis of particle contour, through three multiscale morphological descriptors describ­ ing the overall form, the angularity and the rough­ ness of the particles. The effect of morphology of sand grains has been investigated comparing the samples composed with natural grains, with the artificially grinded specimens with the same grain size distribution. Moving from natural to grinded samples, the abrasion increases as well as the irregularity of grains. The comparisons let emerged correlations between the abrasiveness of sands and the grains morphology: in particular, it tends to increase with the decrease of grains circularity M, and on the other hand it tends, less clearly, to increase with the increasing angularity of grains (descriptor m). Finally, experimental data showed that the abrasiveness is well correlated with a single par­ ameter descriptor of grains regularity, Rm, which combine sphericity (macro-scale, M) and round­ ness (meso-scale, m). As Rm increases, the regu­ larity of grains increases and the abrasiveness decreases.

REFERENCES Altuhafi, F. N., Coop, M. R. & Georgiannou, V. N. 2016. Effect of particle shape on the mechanical behaviour of natural sands. J. Geotech. Geoenviron. Engng 142(12), 04016071. Amoun, S., Sharifzadeh, M., Shahriar, K., Rostami, J. and Azali, S.T., 2017. Evaluation of tool wear in EPB tun­ neling of Tehran Metro, Line 7 Expansion. Tunnelling and Underground Space Technology, 61, pp.233–246. Barzegari G., Uromeihy A. and Zhao J. 2015. Parametric study of soil abrasivity for predicting wear issue in TBM tunneling projects. Tunnelling and Underground Space Technology, 48, pp.43–57. Boldini, D., Losacco, S., Bertolin, S., Amorosi, A. 2018. Finite element modelling of tunnelling-induced dis­ placements on framed structures Tunn Undergr Sp Tech 80, 222–231. Cho, G. C., Dodds, J. & Santamarina, J. C. 2006. Particle shape effects on packing density, stiffness, and strength: natural and crushed sands. J. Geotech. Geoenviron. Engng.,althuaf 132(5), 591–602. Gharanbagh E.A., Rostami J., Palomino A.M. 2011. New soil abrasion testing method for soft ground tunnelling applications. Tunnelling and Underground Space Tech­ nology 26(5) 604–613. Guida, G., Sebastiani, D., Casini, F., & Miliziano, S. 2019a. Grain morphology and strength-dilatancy of sands. Géotechnique Letters, 1–19. Guida, G., Viggiani, G.M.B. & Casini, F. 2019b. Multiscale morphological descriptors from the fractal analysis of particle contour. Acta Geotechnica, 15(5), 106–108. Jakobsen P.A., Bruland A., Dahl F. 2013 Review and assessment of the NTNU/SINTEF Soil Abrasion Test (SAT™) for determination of abrasiveness of soil and soft ground. Tunnelling and Underground Space Tech­ nology 37, 107–114. Jakobsen P.A., Estimation of soft ground tool life in TBM tunnelling. Norwegian University of Science and Tech­ nology, Trondheim. 2014. Käsling H. & Thuro K. 2010. Determining abrasivity of rock and soil in the laboratory. Engineering Geology, Technische Universität München. Küpferle J., Röttger A. and Theisen W. 2017. Excavation tool concepts for TBMs–Understanding the materialdependent response to abrasive wear. Tunnelling and Underground Space Technology, 68, pp.22–31. Losacco, S., Viggiani, G.M.B. 2019. Class A prediction of mechanised tunnelling in Rome. Tunn Undergr Sp Tech 87, 160–173. Miura, K., Maeda, K., Furukawa, M. & Toki, S. 1997. Physical characteristics of sands with different primary properties. Soils Found. 37(3), 53–64. Nilsen B., Dahl F., Raleight P., Holzhauser J. 2006. Abra­ sivity of soils in TBM tunnelling. Tunnels & Tunnelling International, March. 36–38. Pirone, M., Carriero, F., Sorge, R., Sebastiani, D., Miliziano, S., Foti, V. 2019. The management of the soil conditioning process for the excavation of the Rome metro C line. Tunnels and Underground Cities: Engin­ eering and Innovation meet Archaeology, Architecture and Art- Proceedings of the WTC 2019 ITA-AITES World Tunnel Congress. Rampello S., Callisto L., Viggiani G.M.B. (2012). Evaluat­ ing the effects of tunnelling on historical buildings: the example of a new subway in Rome. Geomechanics and Tunnelling, 5(3), 275–299.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Front-face pressure drop during the standstill phase for EPB mechanized tunnelling in coarse-grained soils D. Sebastiani & S. Miliziano Department of Structural and Geotechnical Engineering, University of Rome “La Sapienza”, Rome, Italy

A. Bezuijen Department of Civil engineering, Gent University, Gent, Belgium/Deltares, Delft, The Netherlands

ABSTRACT: Mechanized tunnelling with EPB-TBMs bases its success on the ability to guarantee a uniform and constant front-face pressure during the excavation, using the excavated soil properly treated with chemical additives. This front-face pressure, measured with pressure sensors inside the working cham­ ber, is kept constant during the excavation phase by balancing the amount of incoming soil, the foam injected, and the amount of soil extracted through the screw conveyor in the working chamber. During the standstill phase, required to install the tunnel lining elements, particularly in the coarse-grained soils, the pressure inside the chamber tends to decrease until the next resumption of the excavation. This effect, mainly linked to the permeability of the soil outside the TBM, tends to increase settlements induced on the surface and, there­ fore, should be minimized. In this paper, a very simple calculation model proposed by Bezuijen and Gerheim Souza Dias in 2017 for tunnelling in saturated sandy soils, is applied to an Italian tunnelling project. A comparison of model predictions and data recorded real time on the site during the excavation is illustrated and discussed. The predictive ability of the calculation model proposed for saturated sand seems to be useful also for TBM applications in gravelly soil under the phreatic surface, highlighting the weight of the main factors affecting the specific boundary value problem; moreover, it can be a useful tool to address the problem and to individuate technical countermeasures finalized to minimize the phenomenon and its negative effects.

1 INTRODUCTION Tunnel Boring Machines (TBM) and Earth Pressure Balance (EPB) technology is currently often used for the mechanized excavation of tunnels particularly in urban areas. Among several advantages, including the excavation performances and the safety condi­ tions, an important role is played by the ability to apply and control a support pressure to the excava­ tion front-face. The pressure is transferred to the front-face through the excavated soil, suitably injected with chemicals, quite often foam. The pres­ sure is real time regulated modifying the amount of soil entering into the working chamber, through the control of the thrust force and cutterhead rotation speed, and the amount of conditioned soil extracted by the screw conveyor, through the control of its rotation speed. The tunnel construction provides a sequence of excavation phases followed by standstill phases, in which the lining is installed. During that standstill phase pressure reduction is measured regularly, espe­ cially in coarse-grained soils and already in the very first minutes after the end of the excavation phase.

This phenomenon is particularly critical in urban areas where the interaction with pre-existing buildings can lead to damages if the earth pressure in the working chamber reduces excessively. Thus, the ability to pre­ dict the reduction of pressure applied to the front-face during the standstill phase is very important to quantify the risk and to study and put in place remediation and mitigation actions. To predict the evolution of pressure in the excava­ tion chamber during the standstill phase, Bezuijen & Dias (2017) proposed a rather simple straightforward analytical model based on a simple seepage scheme for sandy soils. In the present work, the model is applied to the Italian case of tunnelling project of the Milan Metro line M4, where the coarse soil is rich of gravel, with the aim of validating it also in a very coarse gravelly soil. After a brief description of the model and the Pro­ ject of Milan Metro line M4, the paper presents a comparison between the measured pressure evolution in the standstill phase and the model predictions. A discussion of the results is proposed together with an interpretation of the strong and weak points of the model.

DOI: 10.1201/9780429321559-53

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2 BACKGROUND 2.1 Coarse-grained soil conditioning process It is commonly established that the conditioning of coarse-grained soils has the purpose of modifying its physical and mechanical characteristics to make it suit­ able for excavation and, in particular, to increase poros­ ity, decrease permeability, increase compressibility (Bezuijen, 2002), to reduce the internal friction angle and abrasiveness (Sebastiani et al. 2016). A proper conditioning is necessary to bring the coarse-grained soil over the maximum void ratio, emax. Under these conditions, the grains of sand and gravel are no longer all in contact with each other, the overall behavior can be associated with that of a fluid, the effective stress state is zero, the shear stresses are reduced to low values, the pressure sensors inside the excavation chamber measure values that correspond with the verti­ cal stress, the pore pressures of water and gas are prac­ tically equal. Naturally, this is the limit condition of a perfect soil conditioning in which all the coarsegrained soil granules are homogenized a single mixture. These conditions are easily reachable through the injection of water and foam for sandy soils, but if the percentage of gravel begins to be relevant, large differences begin to appear between the characteris­ tics of the conditioned ground in the upper part and in the lower part of the excavation chamber. 2.2 The pressure applied to the front-face during EPB-TBM tunnel excavation Having very clear the direct correlation between pressure applied to the front-face during the tunnel excavation and induced surface settlement, from the very first application of EPB-TBM in Europe, increasing interest has grown around the possibility to control the pressure in the working chamber. Several researchers have developed calculation methods to determine the value of the pressure required to limit surface settlements while only few attempts were performed to predict the trend of the pressure into the working chamber during the stand­ still phase of the excavation process, the evolution of the features of the conditioned soil in this phase and, consequently, the evolution of the pressure applied to the front-face. From the studies of Anagnostou & Kovári (1996) developed simultaneously with the first EPB-TBM applications in Europe in Turin and Valencia, several relevant studies were performed taking advantage of the development of increasingly accurate numerical models and the remarkable development of the ability to acquire and process field data (Jancsecz et al. 1999). This enormous development has always been con­ trasted with a number of elements of complexity intrinsically inherent in the interaction between the excavation of a new tunnel, the soil and the existing structures on the surface. Without claiming to be

exhaustive, we can mention a) the heterogeneity and unpredictability of the soil to be excavated, b) the variability of the surrounding hydraulic conditions, c) the variability of the boundary conditions, d) the three-dimensionality of the problem, e) the depend­ ence of the soil response on highly variable excava­ tion operating parameters. 2.3 The decrease in pressure in the working chamber during the standstill phase In recent years, the use of TBM-EPB has also been widespread in soils like gravel, resulting frequently in form of decrease of pressures in the excavation chamber. This forces the workers to delicate emer­ gency operations necessary to maintain the correct pressure in the excavation chamber. These emer­ gency operations are generally based on the injection of foam or bentonite in the excavation chamber. In this perspective, the development of simple models able to predict the pressure evolution during stand­ still takes on considerable interest as an element cap­ able of joining on the one hand the more operational aspects linked to the TBM excavation parameters and on the other complex numerical models. The reduction of pressures in the excavation chamber is certainly a complex phenomenon as result of the superposition of different effects. Some of the main phenomena that occur during the stand­ still are: i) the collapse of foam bubbles; ii) the seep­ age of fluid outward the excavation chamber. The duration of normal standstill phase is usually quite short (20-30 min), thus the bubbles collapse can be neglected and a model taking in account only seepage of the liquid towards the outside of the TBM, can be developed. For the standstill phase, the subject of this study, a first attempt to model the evolution of pressure in the excavation chamber in saturated sandy soils, was made by Bezuijen & Dias (2017), considering several previ­ ous studies developed by Bezuijen (2002 and 2005) and Bezuijen et al. (2006) and based on the experi­ mental tests developed by Bezuijen et al. (1999). The model is based on observations that the soil inside the working chamber is homogeneously com­ posed by soil, water and air (foam bubbles). Just out­ side the front of the TBM, the saturated soil experiences the penetration of foam into the soil which is, in case of coarse-grained soils, determined by the groundwater flow and not by the properties of the injected foam. Several studies, performed among the others by Bezuijen et al. (1999) Yu et al. (2017) and Wu et al. (2018), demonstrated the tendency of the foam to stay into the working chamber due to the relatively large diameter of the bubbles. On the opposite, the water can flow away from the working chamber with velocity depending very much by soil permeability. Considering the schematic section shown in Figure 1 in the working chamber, in which the starting pressure is the value actually measured at the end of

409

contribution of each of these in point A x m in front of the face is found by integration over both θ and r. With the aforementioned assumption, the evolu­ tion of the pressure in the working chamber is described by Eq. 1 (Bezuijen & Dias, 2017):

where: – p is the pressure in the working chamber, – pw is the pore pressure far from the TBM, – p0 is the pressure in the working chamber at the end of the excavation phase, – R is the radius of the TBM, ­ – k is the permeability of the saturated soil, – pavg is the average between the initial pressure in the mixing chamber and the pore water pres­ sure far from the tunnel, – c is the products of pa x Vair, where pa is the pressure of injection of foam (considering an average pressure at the axis of the tunnel) and Vair is the volume of injected air through the foam injection system.

Figure 1. Sketch TBM with foam and pore pressures in the

TBM and the soil. Adapted from Bezuijen & Dias (2017).

drilling and tends to decrease during standstill until it reaches the equilibrium with the pore pressure at a distance from the TBM (Δp=0), at the same elevation. The consequence of the seepage of water from the excavation chamber is the tendency of the foam to expand to compensate for the loss of volume. The pressure drop in the excavation chamber can be easily calculated assuming that the air into the bub­ bles behave as perfect gas, following the Boyle’s law. The model, as anticipated, is based on the hypoth­ esis that the chamber is connected to the soil and that only the interstitial water can seep away under the hydraulic gradient; the flow is considered spherical symmetric to the half sphere in front of the TBM. Other basic hypotheses are the permeability of the conditioned soil is lower than that of the soil outside the TBM and the one-dimensional filtration of the fluid part only and the consequent variation of the volume of the foam bubbles inside the working cham­ ber to compensate the lost volume. The flow into the soil and pressure distribution in the soil at the tunnel axis is calculated according with the scheme in Figure 2, from Bezuijen (2002), where the tunnel face is composed out of a number of point sources and the

3 MILAN METRO LINE M4 PROJECT 3.1

The Milan metro line M4 project will connect the his­ toric center of Milan with the eastern part (Linate Air­ port) and the western part (Lorenteggio and Stazione San Cristoforo), integrating the urban transport network with a new line. Its construction includes 21 new sta­ tions and about 15 km of line tunnels having 6.36 m of excavation diameter for external sections, and an exca­ vation diameter 9.15 m for the central stretch. The con­ struction of the tunnels is performed using EPB-TBMs. In the stretch here studied, the excavation diameter is 6.36 m and the segments width is 1.4 m. 3.2

Figure 2. One-dimensional flow hypothesis scheme from Bezuijen (2002).

General overview

Geology

From the geological point of view, the Milan subsoil is characterized by a homogeneous layer of alluvial soil, belonging to the unit called “Fluvio glaciale Wurm”, which extends from the first few meters below the ground level to about 70 m. It consists of very coarse soils composed predominantly by gravel, more than 50%, and sand (Figure 3). In the stretch analyzed in this study, the tunnel is under water table. The pore pressure regime in the soil is hydrostatic and the piezometric surface is located 6.5 m above the axis of the tunnel. From the execution of a large number of Lefranc permeability in situ tests, soil permeability values at the TBM excavation depth vary along the path in the range of 10-5 ÷10-4 m/s. In the excavated sections

410

Table 1. Average values of the main conditioning param­ eters during the excavation phase (rings n. 2005, 2006, 2007).

Ring number

FER (-)

FIR (%)

Vair (m3)

2005 2006 2007

6.2 6.3 7.7

52.7 54.5 83.9

18.94 20.07 32.81

Table 2 reports the average values of the pressures in the excavation chamber, at different elevations, at the end of the excavation phase. p0high is the measured pressure in the higher part of the excavation chamber, p0mid, in the middle part, and p0low in the lower part. Small differences were recorded between the sensors on the right and those on the left of the excavation chamber. The reported values are average values of the measurements recorded on each pair of sensors, placed at the same elevation. As showed in Figure 4, the elevation of the couple of sensors placed in the upper part of the chamber is 2.25 m above the tunnel axis while, in the lower part, they are placed 2.25 m below. For the selected rings, Figure 5 shows the evolution of the measured pressure at the three elevations (phigh, pmid, plow,). At the end of each excavation phase, a marked reduction in chamber pressure values was

Figure 3. Grain size distribution curves of coarse-grained soil from Milan.

analyzed in the following, the results of the Lefranc tests provided permeability values in the range between 1·10-4 and 4·10-4 m/s. 3.3 Earth pressure evolution into the excavation chamber The excavation was carried out without significant problems maintaining an average advancement speed of 40 mm/min. The average employed values of thrust force (10 MN) and torque (3.2 MNm) were well below the TBM limits. For the management of the soil conditioning pro­ cess, preliminary laboratory test results, information from the previous excavation of the Milan M5 line and data acquired during the excavation start-up phase were considered. Average values of the Foam Expansion Ratio (FER) (Eq. 2) between 8 and 9 and values of the Foam Injection Ratio (FIR) (Eq. 3) of 60% were maintained.

where Vfoam is the volume of injected foam, Vliquid is the volume of liquid (water and surfactants at 2.0%) used to generate the foam and Vsoil is the volume of the excavated soil. Three consecutive excavation steps were selected for which all the measurements of the excavation parameters and pressures in the excavation chamber are available and have regular and reliable trends. Table 1 presents the average values of the soil condi­ tioning parameters together with the total volume of injected air into the soil, Vair, from the 7 injection lines, during the excavation phases carried out to install the rings n. 2005, 2006 and 2007.

Table 2. Pressure values at the end of excavation phase.

Ring number

p0high (kPa)

p0mid (kPa)

p0low (kPa)

2005 2006 2007

105.6 112.0 96.53

152.4 165.1 155.39

173.8 209.7 206.6

Figure 4. Scheme of the working chamber sensor placement.

411

Table 3. Percentage of residual pressure within the work­ ing chamber referred to the pressure at the beginning of standstill phase, 1200 s after the end of the excavation phase.

Ring number

phigh/p0high (%)

pmid/p0mid (%)

plow/p0low (%)

2005 2006 2007

49.2 56.0 68.9

77.5 81.8 83.8

83.2 86.3 85.8

kPa for the center and lower sensors, 40 kPa for the upper part of the excavation chamber. Taking into account the position of the sensors, the initial pressure values and those at the end of the assembly are compatible with a weight of the unit of volume approximately equal to 18 kN/m3, slightly decreasing with the height and value of K=1 (liquid status of the conditioned material).

Figure 5. Pressure decrease during the standstill phase; ring numbers 2005, 2006 and 2007.

systematically measured already in the first few minutes. Irrespective of the position of sensors, the pres­ sures tend gradually to decrease with a decreasing gradient. This trend was always more marked in the upper part of the excavation chamber. After 1200 - 1500 s from the beginning of stand­ still phases (the elapsed time necessary for the instal­ lation of the precast concrete lining segments), the pressure reduction was about 50% in the upper part of excavation chamber and only about 20% else­ where (Figure 6 and Table 3). Despite the differences in percentage reduction, the pressure chance during the standstill phase is quite similar irrespective of the sensor position: 30

4 PREDICTION VS MEASUREMENTS AND DISCUSSION The calculation model described in par. 2.2 was applied to the case of the Milan M4 Metro line and in particular 3 sections selected. Monitoring data recorded in the central part of the excavation cham­ ber and model results are presented in Figure 7. Real soil conditioning parameters were considered for the calculation and, being available, for the perme­ ability k, for a first evaluation a value of 1·10-4 m/s was used resulting from Lefranc tests performed just in the analyzed section. Since the result of the model strongly depends on the value of the permeability con­ sidered, and this value strongly varied along the path of the tunnel and very difficult to be estimated on site with a high level of precision, a parametric analysis was carried out to verify how the trends of the pres­ sure in the excavation chamber vary with the variation of the value of k within the range of values measured on site (between 1·10-4 and 4·10-4 m/s). The comparison showed the good ability of the model to reproduce the trend of the pressures recorded. The model also highlights the strong dependence of pressure evolution to the value of coefficient of permeability. Increasing the value of k by a factor of 4, after 1200 the pressure reduction changes passes from about 25-30 kPa to 70-80 kPa. 5 CONCLUSIONS AND FUTURE DEVELOPMENTS

Figure 6. Decrease of pressure measured during the stand­ still phase; ring numbers 2005 and 2006.

The model proposed by Bezuijen & Dias (2017) for sandy soils seems to reproduce with good accuracy the decreasing trend of the average pressures in the excavation chamber over time for the tunnel excava­ tion in sandy gravels performed in the stretch of

412

a part of foam, and therefore of air, comes out of the excavation chamber, which is unlikely in the case of fine sands, but more likely instead for the gravels. Furthermore, another important aspect that must be considered is the possibility that part of the foam bubbles collapse during the excavation phase (Sebas­ tiani et al. 2019) due to the rotation and during the standstill phase due to natural phenomena of reduc­ tion of the thickness of the foam walls producing a leakage of air or an accumulation of air volumes in the upper part of the excavation chamber. These aspects, which are difficult to reproduce with models like the one used in this study without greatly increasing the level of complexity of the system, can explain the small differences between the model predictions and the experimental data. In spite of the shortcomings mentioned, the model, as it is now is a useful tool to estimate the pressure decrease that can be expected in the mixing chamber of an EBP during ring building. This can be used to evaluate the face stability at the reduced pressures.

REFERENCES

Figure 7. Comparison between measurements and model results for pressure values (middle pressure gauges) in the working chamber of different values of permeability considered.

Milano underground studied in this paper. It should also be underlined that the model does not take into account the possibility that, together with water,

Anagnostou, G. and Kovári, K., 1996. Face stability condi­ tions with earth-pressure-balanced shields. Tunnelling and underground space technology, 11(2), pp.165–173. Bezuijen, A., Schaminee, P.E.L. and Kleinjan, J.A., 1999. Additive testing for earth pressure balance shields. In Twelfth European Conference on Soil Mechanics and Geotechnical Engineering (Proceedings). The Nether­ lands Society of Soil Mechanics and Geotechnical Engin­ eering; Ministry of Transport, Public Works and Water Management; AP van den Berg Machinefabriek; Fugro NV; GeoDelft; Holland Railconsult (No. Volume 3). Bezuijen, A., 2005. The influence of soil permeability on the properties of a foam mixture in a TBM. In Tunnelling. A Decade of Progress. GeoDelft 1995–2005 (pp. 50–57). CRC Press. Bezuijen, A., 2002. The influence of permeability on the properties of a foam mixture in a TBM. 4th Int. In Symp. On Geotechnical Aspects of Underground Con­ struction in Soft Ground-IS Toulouse. Bezuijen, A. and Schaminee, P.E., 2006. Simulation of the EPB-shield TBM in model tests with foam as additive. In Tunnelling A Decade of Progress GeoDelft 1995–2005 (pp. 157–163). Taylor & Francis. Bezuijen, A., Talmon, A.M., Joustra, J.F.W. and Grote, B., 2006. Pressure gradients and muck properties at the face of an EPB. Proceeding of Geotechnical aspects of underground construction in soft ground, pp.195–201. Bezuijen A. & Gerheim Souza Dias T. 2017. EPB, chamber pressure dissipation during standstill. In Proceeding IV International Conference on Computational methods in tunneling and subsurface engineering (Vol. 1, pp. 225–231). Jancsecz S., Krause R., Langmaack L., 1999. “Advantages of soil conditioning inshield tunnelling: Experiences of LRTS Izmir”. Proc. of the Conf. Challenges for the21st century, Allen, et al. (eds), Balkema, Rotterdam, pp 865–875. Yu, H., Mooney, M.A. and Bezuijen, A., 2017, November. A simplified chamber pressure model for EPB TBM tunneling in granular soil. In Geotechnical Aspects of

413

Underground Construction in Soft Ground: Proceedings of the 9th International Symposium on Geotechnical Aspects of Underground Construction in Soft Grounds (IS-São Paulo 2017), April 4–6, 2017, São Paulo, Brazil (p. 127). CRC Press. Sebastiani, D., Passeri, D., Belardi, G. and Miliziano, S., 2016. Experimental study of coarse soil properties influ­ encing soil abrasivity. Procedia Engineering, 158, pp.9–14.

Sebastiani, D., Vilardi, G., Bavasso, I., Di Palma, L. and Miliziano, S., 2019. Classification of foam and foaming products for EPB mechanized tunnelling based on half-life time. Tunnelling and Underground Space Tech­ nology, 92, p.103044. Wu, Y., Mooney, M.A. and Cha, M., 2018. An experimen­ tal examination of foam stability under pressure for EPB TBM tunneling. Tunnelling and Underground Space Technology, 77, pp.80–93.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

A miniature EPB TBM for use in a geotechnical centrifuge

C.J. Shepheard, A.S.N. Alagha, G.M.B. Viggiani & S.K. Haigh Department of Engineering, University of Cambridge, UK

ABSTRACT: Tunnelling is becoming an increasingly frequent solution to the need to provide services and infrastructure in growing urban environments. Engineers must be able to predict the effects of the construction of new tunnels on surrounding structures both at surface and at depth. The interaction between new tunnels and existing structures has been investigated extensively in the past by physical and numerical modelling. However, physical models have been limited by the methods used to simulate tunnel construction, often ignor­ ing the complex three-dimensional effects of tunnelling processes. This paper describes the design and devel­ opment of a novel miniature tunnel boring machine to be used in the Cambridge geotechnical centrifuge in soft ground. When driven in-flight, the model tunnel boring machine will permit to simulate the main pro­ cesses occurring around the shield, thus allowing real-time study of the development of ground movements and their effect on nearby structures.

1 INTRODUCTION Increasing urban populations and rapidly growing cities require the provision of more supportive infra­ structure. Whilst tunnelling has long been the solu­ tion to this problem, the progressively more congested underground urban environment is caus­ ing tunnels to be constructed very close to founda­ tions and other tunnels. Prior to construction, tunnelling induced ground movements should be predicted and their effects on the structures and ser­ vices assessed, to make sure that tunnelling does not cause unacceptable damage to surrounding and over­ lying structures. In soft ground, the most common form of Tunnel Boring Machine (TBM) today is the Earth Pressure Balance (EPB) (Mair 2008). This is a closed shield, providing support to the face of the excavated tunnel, thus limiting ground movement and volume loss. The main features of an EPB shield are illustrated in Figure 1. The steel shield, housing the working chamber and the drive motor, terminates with a rotating cutterhead that excavates the soil, loosening it with knives, disk cutters and scrapers, and extracts it through openings into the excavation chamber. To facilitate advancement of the shield and avoid jamming, the diameter of the cutterhead is somewhat larger than the diameter of the shield and the shield itself is slightly tapered or telescopic, with a decreasing diameter from face to tail. The thrust from the cutterhead and the pressure exerted at the face by the excavated soil contribute to support the tunnel face.

A screw conveyor, or cochlea, extracts the muck from the excavation chamber. The rate of earth dis­ charge through the screw conveyor and the rate of advance of the TBM control the pressure at the face. If needed, for better control of pressure and muck discharge, the soil in the excavation chamber can be conditioned with the addition of foams and chemical additives. A bulkhead, or pressure wall, divides the excava­ tion chamber from the rear of the shield, where a segment erector, or gripper, mounts the precast segments of the permanent lining, and a set of hydraulic jacks or thrust cylinders push the shield forward reacting against the already built lining. Unavoidably, although significantly reduced if compared to traditional tunnelling, some ground movements are still caused by TBM operation. The total volume of soil that, due to deformations induced by tunnel excavation, flows through the the­ oretical excavation boundary, is called ground loss or volume loss and is normally expressed as a percentage of the nominal tunnel volume. The main sources of short term ground loss in shield tun­ nelling are due to face extrusion, overcut, and lining deformation, although usually the latter can be neg­ lected with respect to the other two because of the high stiffness of the lining. Face extrusion is due to the stress relief induced by the removal of soil at the tunnel face and can be con­ trolled to some extent by adjusting and maintaining an adequate value of the face support pressure throughout excavation. The overcut component of volume loss includes several contributions connected to cutterhead

DOI: 10.1201/9780429321559-54

415

does provide useful data. However, the cost of these studies makes them fairly rare, and pure greenfield conditions cannot usually be established. 2.2

Figure 1. Main features of an EPB TBM (railsystem.net, 2014).

overcut, shield taper and tail void. This is the physical gap existing at the tail of the shield because of tail piece thickness and clearance for the erection of seg­ mental lining (Figure 2) and, unless promptly grouted, maybe the main source of volume loss. The movements of the ground surrounding the shield, manifest at surface as a settlement trough extending directly above the tunnel and ahead of the tunnel face. The displacement field induced by tun­ nelling is inherently three-dimensional and timedependent, as it evolves as tunnel face advances and, for fine grained soils, as excess pore water pressures induced by excavation dissipate. 2 MONITORING THE TUNNELLING PROCESS A number of previous studies have sought to study and/or model this phenomenon, however results so far have been limited in a number of ways. 2.1

Field studies

Evidently, field monitoring of the ground and foun­ dations near the construction of a new tunnel (e.g., Bakker et al. 1999, Selemetas 2005, Fargnoli et al. 2013) will be able to capture the full process and

Figure 2. Shield geometry and sources of overcut (after Loganathan, 2011).

Centrifuge modelling

Centrifuge modelling can be used to investigate the effects of tunnel construction on particular ground conditions. The benefit of such studies are that they are repeatable and can be highly instrumented with minimal soil disturbance. The complexity of centri­ fuge testing and the demands of increased gravity levels means that modelling the tunnelling process is generally highly simplified by employing a volume loss system to produce the resultant ground move­ ments (e.g. Grant & Taylor 2000, Jacobsz et al. 2004, Marshall et al. 2012). In the most common version of this this method, a thin rubber membrane is wrapped around a mandrel and inflated with incompressible fluid; during flight, the fluid is slowly extracted, thus redu­ cing the diameter of the ‘tunnel’ and producing volume loss. Usually performed in plane strain con­ ditions, this method does not account for the threedimensional effects of tunnelling. Gue et al. (2017) sought to rectify this by designing a staged volume loss system, where fluid was progressively extracted from compartments to replicate the advance of the TBM. While modelling volume loss does well replicate the end result effect of tunnelling – ground move­ ment and settlement – it eliminates the complex effects of the tunnelling process such as rotation of the cutterhead, forward advance of the TBM and for­ mation of tail void. 2.3

Reduced scale models

Some studies have attempted to address the threedimensionality of the tunnelling process through the construction of reduced-scale TBMs at 1g. The bene­ fit of such projects is the extensive monitoring of tunnel advance in the longitudinal direction under controlled conditions, although the limitations of 1g testing with regards to accurate soil stresses should be noted. Due to the size of such set-ups and the associated cost of construction and testing, few have been constructed (Bel et al. 2015). Indeed, in a thorough review of the published literature, only three individual reduced-scale TBM models for 1g were identified: (1) Xu et al. 2011, (2) Bel et al. 2013, (3) Fang et al. 2015, see Table 1. These reduced scale TBM models excavate soil through a rotating cutterhead powered by a motor, and are advanced by a drive system at the rear. Although these models do replicate more accurately the tunnelling process than the simple volume-sink centrifuge models, they still do not include all the major aspects of ground movement occurring around

416

Table 1.

Comparison of 1g reduced scale models. Diameter

Model

Material

m

Overcut

Tail void

(1) (2) (3)

silty clay sand sand

0.40 0.55 0.50

no no yes

no no yes

the shield (Mair & Taylor 1997). Indeed, the volume loss associated to cutterhead overcut and shield tapering and the existence of a gap between the shield tail and the lining are not included in two of the three models. 2.3.1 Reduced scale centrifuge model Only one study so far (Nomoto et al. 1999) has sought to create a reduced scale centrifuge model for use in a geotechnical centrifuge. The TBM model created for this study used a two-stage process to replicate tunnel construction. First, the shield would be driven into dry sand to excavate the tunnel, and then an outer tube would be withdrawn to produce a tail void. The model had a diameter of 0.1 m and was tested at 25 g in dry sand, modelling a prototype tunnel of 2.5 m diameter. Even this model does not manage to capture most of the tunnelling processes except for the creation of the tail void at realistic soil stresses, and is limited by the inability to model the continuous tunnelling process. The creation of the tail void in the reverse direction to the advance of the shield is likely to have affected the surrounding sand in a manner not consistent with actual tunnelling processes. 3 DEVELOPMENT OF MINI-TBM To address the errors and limitations in the studies so far and discussed above, a new reduced scale model TBM (mini-TBM) for use in the Cambridge

geotechnical centrifuge has been developed. It will be used in a series of centrifuge tests in greenfield conditions and then to study tunnel-pile interactions. As this research is interested primarily in the tunnel construction process, the long-term ground movements due to consolidation are not included in the design. Also, at this point, accurate modelling of lining stiffness has not been considered, because, as mentioned above, ground losses due to lining deflec­ tions are generally much smaller than those due to stress relief at the face and overcut. In order to reproduce the main sources of ground loss, the focus of the design of this model TBM is to produce the correct tunnel face conditions and overcut, including shield tapering and tail void. The com­ pleted model is intended to be a ‘kit of parts’ whereby individual features may be swapped to account for different ground conditions and or detailed geometries (cutterhead overcut and shield tapering). In particular, cutterheads with lower and higher opening ratios will be used for sands and clays, respectively. 3.1

Mini-TBM geometry

To reproduce accurately the sources of volume loss, the geometry of the mini-TBM is critical. The full design of the final model is given in Figure 3. Three main components of the model replicate the prin­ ciple features of EPB shield tunnelling: a cutterhead with an overcut, behind which sits a conical collar as the shield, and then a ‘lining tube’ with a reduced diameter from the back of the collar to produce the tail void. 3.1.1 Lining tube In initial planning for the design of the mini-TBM, the diameter of the tunnel to be studied was a limiting factor. A size must be chosen that would model a tunnel large enough to require an EPB TBM, yet small enough that it could be modelled on the Turner Beam Centrifuge without the results being affected by boundaries of the soil container. In addition, due to the motors necessary to operate the

Figure 3. Final design of mini-TBM.

417

mini-TBM, a sensible scale factor N had to be defined for which motors with the required power were available. N = 50 was chosen to account for this limitation and the dimensions of other previous tunnel models on this centrifuge inspired the early specification for diameter of the model as 80 mm (or 4 m at prototype scale). The actual lining tube for the mini-TBM has a diameter of 76.2 mm and 3 mm thick walls. This tube was chosen as the relatively thick walls allowed to machine down the diameter at one end for attachment of the shield sleeve. It is clear that this represents an extremely stiff lining at prototype scale. 3.1.2 Screw conveyor and hosting tube To transfer the excavated soil from the working chamber a screw conveyor is used. While in a field scale TBM the screw conveyor is usually at an angle and takes in soil from the bottom of the working chamber, this is not an integral part of the TBM design. Therefore, a horizontal screw was chosen, due to limited space. A horizontal screw was also employed by Nomoto et al. (1999), Xu et al. (2011) and Fang et al. (2015) in their model TBMs. A 40 mm diameter screw auger is hosted within a close fitting brass tube to contain the excavated soil. The auger is centred by a bearing at the rear and attachment to the centre of the cutterhead at the front. Three brass rings support the brass tube con­ taining the auger to keep it centred within the lining tube (Figure 3) and are soldered to the tube to ensure they remain in place. Six small holes are drilled in each ring (Figure 4) to allow for the wires of instru­ mentation or tubing for soil conditioning to pass inside the mini-TBM.

lining allows for it to be swapped out for other geometries in the future. Using the relationship between shield length, L and tunnel diameter, DL proposed by the Inter­ national Tunnelling Association (2000), an L/DL ratio of about 1.5 was adopted. At model scale this corresponds to a shield length of 114 mm. For ease, this was rounded down to 100 mm for the final collar design giving a ratio of 1.3. The tail void was created in this model by creat­ ing a step down in radius between back of the collar and the lining tube, simulating tail piece thickness and clearance for the erection of segmental lining. From a review of published details of TBM geom­ etries, it was found that the usual difference in diam­ eter between the shield and the lining is of the order of 6-7% of the shield diameter. In this case a radial tail void of 2.5 mm (outer diameter at shield rear: 81.2 mm) creates a 6.6% change in diameter. TBM shields are usually telescopic, with a circumference that decreases in two to three steps away from the cutterhead. This is to ensure ease of movement and control through the soil in combin­ ation with the overcut and reduce thrust force on the TBM (Ji et al. 2008). A review of tapering values quoted in the literature suggests usually a gradient of around 0.1-1% from behind the overcut to the tunnel lining (Ji et al. 2008, Liu et al. 2014, Schivre 2015). For the mini-TBM a 0.5% gradient was chosen, with the radius of the collar increasing from 81.2 mm at the tail to 82.2 mm at the junction with the cutterhead over the 100 mm length (Figure 5).

3.1.3 Shield collar The conical collar was designed to create the tail void at the rear of a shield TBM. The collar slots onto one end of the lining tube which was turned down radially by 1.5 mm to hold the collar in place. Having the shield as a separate component to the

3.1.4 Cutterhead In a TBM, the cutterhead face provides support to the ground ahead of the shield, while also allowing excavated ground into the working chamber. There­ fore, the critical parameter for the cutterhead in this model was the opening ratio (ratio of the open area in and the total cutterhead area) and hatch design. Different cutterheads were designed for use in sands and clays. As clays have undrained shear strength, they can provide some self-support and a higher opening ratio can be used (60-80%), while sands require more face support and thus a lower

Figure 4. Rear of mini-TBM showing screw auger, hosting tube and brass ring inside lining tube.

Figure 5. Finished brash conical shield component: (a) side view, (b) front view.

418

opening ratio (30-35%) (Berthoz et al. 2018). In this project, the same hatch pattern was used for both clay and sand but with opening ratios of 40% and 30% respectively (Figure 6 a and b, respectively). The cutterhead must also create the overcut, which is the small difference in diameter between the cutterhead and the nearest point of the shield to allow clear­ ance for the passage of the shield (Dowden & Cass 1991). The magnitude of the radial overcut varies somewhat in the literature. With the overcut described as the percentage difference between the cutterhead radius and the closest point of the shield, the literature suggests it is between 0.2% and 1.6% (Bezuijen & Talmon 2008, Ramoni & Anagnostou 2011, Schivre 2015, Hasanpour et al. 2017, Cording 2018). For the mini-TBM an overcut ratio of about 1% was chosen, corresponding to a diameter of 83 mm. However, manufacturing error produced cutterheads of 85 mm diameter (thus reducing slightly the open­ ing ratios), as such the final mini-TBM design has a 1.4 mm radial overcut corresponding to 70 mm at prototype scale. This is on the large side but within the range of values quoted in the literature. In order to ensure the cutterhead remains centred with respect to the body of the mini-TBM, a 6 mm deep channel has been milled into the reverse with an outer diameter of 82.2 mm and inner diameter of 68.2 mm. This is lined with PTFE to ensure smooth rotation. The cutterhead and auger are designed to rotate as one to simplify the model, with the auger attached to the cutterhead with a screw. At this time, the cutterheads are not equipped with cutting tools. For the sand cutterhead, it is

Figure 8. Conical working chamber component.

Figure 9. (a) working chamber, (b) face of sand mini-TBM.

believed these will not be a necessary addition, as preliminary tests showed sand flowing well through the openings. For the clay cutterhead, simple scrap­ ing tools will need to be attached to guide clay into the working chamber. 3.1.5 Working chamber A working chamber component was manufactured from brass to guide the excavated soil from the face into the auger to be conveyed away. As the diameter of the outside of the hatches is larger than that of the brass conveyor tube, this component has a conical interior design. To aid in centring the auger and cut­ terhead, the chamber is solid outside of the recessed chamber to support the conveyor tube at the head of the mini-TBM (Figure 8) Figures 9 (a) and (b) show photographs of the working chamber component and a front view of the assembled face for the sand cutterhead, respectively. Finally, Figure 10 shows the fully assembled miniTBM. 4 SUMMARY AND CONCLUSIONS

Figure 6. Design details for clay and sand cutterheads.

Previous studies of tunnelling effects on nearby ground have yielded some useful information and highlighted the issues that should be considered when constructing new tunnels. However, these studies to date have all been limited in some way

Figure 10. Top view of full mini-TBM.

Figure 7. Finished sand cutterhead.

419

and the results do not accurately represent the real mechanisms at work. In this paper, the development and design of a new mini-TBM have been presented. The model TBM can replicate the major physical process and connected source of volume loss occurring around the shield during flight in a geotechnical centrifuge, hence permitting accurate modelling of the entire tunnelling process.

REFERENCES Bakker, K.J., De Boer, F., Admiraal, J.B.M., and Van Jaarsveld, E.P. 1999. Monitoring pilot projects using bored tunnelling: The Second Heinenoord Tunnel and the Botlek Rail Tunnel. Tunnelling and Underground Space Technology, 14(2): 121–129. doi:10.1016/S0886­ 7798(99)00025-5. Bel, J., Branque, D., and Wong, H. 2013. NeTTUN Deliverable 9.2: Review of the state-of-the-art on laboratory studies on the impact of tunnelling and experimental program. Bel, J., Branque, D., Wong, H., Viggiani, G., and Losacco, N. 2015. Experimental study on a 1g reduced scale model of TBM: impact of tunnelling on piled structures. Geotechnical Engineering for Infrastructure and Development: XVI European Conference on Soil Mechanics and Geotechnical Engineering,: 413–418. doi:10.1680/ecsmge.60678. Berthoz, N., Branque, D., Wong, H., and Subrin, D. 2018. TBM soft ground interaction: Experimental study on a 1 g reduced-scale EPBS model. Tunnelling and Under­ ground Space Technology, 72: 189–209. Pergamon. doi:10.1016/j.tust.2017.11.022. Bezuijen, A., and Talmon, A.M. 2008. Processes around a TBM. Cording, E.J. 2018. Muir Wood Lecture 2018: Monitoring and Controlling Ground Behavior at the Source Recent Applications to Pressurized Tunneling. Dowden, P.B., and Cass, D.T. 1991. Shielded TBMs: matching the machine to the job. In Rapid Excavation and Tunnelling Conference. Seattle. pp. 787–805. Fang, Y., Yang, Z., Cui, G., and He, C. 2015. Prediction of surface settlement process based on model shield tunnel driving test. Arabian Journal of Geosciences, 8(10): 7787–7796. doi:10.1007/s12517-015-1800-0. Fargnoli, V., Boldini, D., and Amorosi, A. 2013. TBM tunnelling-induced settlements in coarse-grained soils: The case of the new Milan underground line 5. Tunnelling and Underground Space Technology, 38: 336–347. Elsevier Ltd. doi:10.1016/j.tust.2013.07.015. Grant, R.J., and Taylor, R.N. 2000. Tunnelling-induced ground movements in clay. Proceedings of the Institu­ tion of Civil Engineers, Geotechnical Engineering, 143 (1): 43–55. doi:10.1680/geng.2000.143.1.43. Gue, C.Y., Wilcock, M.J., Alhaddad, M.M., Elshafie, M.Z. E.B., Soga, K., and Mair, R.J. 2017. Tunnelling close

beneath an existing tunnel in clay-perpendicular undercrossing. Geotechnique, 67(9): 795–807. doi:10.1680/jgeot.SiP17.P.117. Hasanpour, R., Schmitt, J., Ozcelik, Y., and Rostami, J. 2017. Examining the effect of adverse geological condi­ tions on jamming of a single shielded TBM in Uluabat tunnel using numerical modeling. Journal of Rock Mechanics and Geotechnical Engineering. doi:10.1016/ j.jrmge.2017.05.010. Hose Supplies,. 2014. Herrenknecht Tunnel Boring Machine (TBM) animation. Jacobsz, S.W., Standing, J.R., Mair, R.J., Hagiwara, T., and Sugiyama, T. 2004. Centrifuge Modelling of Tunnelling Near Driven Piles. Soils and Foundations, 44(1): 49–56. doi:10.3208/sandf.44.49. Ji, Q.Q., Huang, Z.H., and Peng, X.L. 2008. Analysis on influence of conicity of extra-large diameter mixed shield machine on surface settlement. In The Shang­ haiYangtze River Tunnel: Theory, Design and Construc­ tion. Edited by R. Huang. Taylor & Francis, Shanghai, China. pp. 237–274. Liu, C., Zhang, Z., and Regueiro, R.A. 2014. Pile and pile group response to tunnelling using a large diameter slurry shield - Case study in Shanghai. Computers and Geotechnics, 59(October): 21–43. Elsevier Ltd. doi:10.1016/j.compgeo.2014.03.006. Mair, R.J. 2008. Tunnelling and geotechnics: new horizons. Géotechnique, 58(9): 695–736. doi:10.1680/ geot.2008.58.9.695. Mair, R.J., and Taylor, R.N. 1997. “Theme Lecture: Bored tunnelling in the urban environment,” Plenary Session 4. In 14th International conference on soil mechanics and foundation engineering. pp. 2353–2385. Marshall, A.M., Farrell, R., Klar, A., and Mair, R. 2012. Tunnels in sands: the effect of size, depth and volume loss on greenfield displacements. Géotechnique, 62(5): 385–399. doi:10.1680/geot.10.P.047. Nomoto, T., Imamura, S., Hagiwara, T., Kusakabe, O., and Fujii, N. 1999. Shield tunnel construction in centrifuge. Journal of Geotechnical and Geoenvironmental Engin­ eering, 4(125): 289–300. doi:10.1061/(ASCE)1090­ 0241(1999)125:4(289). Ramoni, M., and Anagnostou, G. 2011. The Interaction Between Shield, Ground and Tunnel Support in TBM Tunnelling Through Squeezing Ground. Rock Mechan­ ics Engineering, 44(1): 37–61. doi:10.1007/s00603-010­ 0103-8. Schivre, M. 2015. Frejus highway tunnel. Selemetas, D. 2005. The response of full-scale piles and piled structures to tunnelling. (December). WG No. 14 Mechanised Tunnelling - International Tunnelling Association. 2000. Recommendations and Guide­ lines for Tunnel Boring Machines (TBMs). In ITA Report. Xu, Q., Zhu, H., Ding, W., and Ge, X. 2011. Laboratory model tests and field investigations of EPB shield machine tunnelling in soft ground in Shanghai. Tunnelling and Underground Space Technology, 26(1): 1–14. doi:10.1016/j.tust.2010.09.005.

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Urban tunnelling in glacial soils: Tunnel de Champel, Geneva W. Steiner, T. Witschi & A. Ferrari B+S AG, Bern, Switzerland

ABSTRACT: The 1622 m long Tunnel de Champel in Geneva, Switzerland forms part of the rail-link CEVA (Cornavin, Eaux-Vives, Annemasse) to France. The plateau of Champel is formed by glacial deposits of interglacial and postglacial deposits from the Arve and Rhone glaciers and rivers, with 10 to 30 m overburden. The sections with tunnelling were relatively short therefore conventional tunnelling was selected with a pipe-roof umbrella and face stabilization with fibre-glass anchors. The behaviour of the advancing tunnel was modelled with a three-dimensional Finite Element model. The excavation was monitored in the tunnel, geodetically, inverse extensometers at the face and horizontal inclin­ ometers in the roof umbrella, and at the ground surface with settlement and geodetic monitoring. The build­ ings on the surface were statically analysed and their risk to settlement evaluated. The comprehensive site investigation, detailed analyses and meticulous construction procedures proved decisive for a safe construction.

1 DESCRIPTION OF THE PROJECT In December 2019 the 16 km long rail link CEVA in Geneva entered in service after a century long history. CEVA is the acronym for Cornavin, the Swiss Main railway station, Eaux-Vives the former French terminal station in Switzerland and Annemasse in Savoie, France. The 1.6 km long Tunnel de Champel is one of the key elements of this cross-city railway line. (Figure 1). The line and its main features have been described in detail by Witschi et al. (2013). Here the focus will be on geotechnical and construction issues of the mined sections of 502 m and 926 m on both sides of the 194 m long station in the centre and the experience gained during the construction of these tunnels. The access to the western portal was initially by a temporary bridge over the Arve river from the flat western bank to the steep eastern riverbank. 2 GROUND CONDITIONS 2.1

Geologic conditions

Geologic conditions are governed by quaternary gla­ cial deposits of various ages of the Rhône and Arve glaciers and rivers that have their confluence in the region of Geneva. The longitudinal section is shown in Figure 1. The deposited sediments are rather complex, and the geologic service has developed a classification system for the glacial deposits. The

major water table lies several meters below the invert of the tunnel and is controlled by the Arve and Rhone rivers as well as Lake Geneva (Lac Léman). 2.2

Site investigations

The underground was explored with 14 cored borings from 24 to 50 m depth, to at least 10 m below the invert of the tunnel. Standard penetration tests (SPT) were executed at 3 m intervals. In the zones with smaller overburden 8 super heavy dynamic penetrometer tests (DPSH) were carried out to determine the thickness of the soft retreat layer. Particular attention was given the presence of sus­ pended water tables in the section of the tunnel, as these water pockets might flow out during tunnel advance and lead to loss of ground ahead of the tunnel face and large surface settlements. Humid layers with a thickness of several decimetres to one­ meter thickness were detected approximately every two to three meters. In the old gravel layer (Cailloutis morainiques) eleven pressure-meter tests and four in the Moraine were carried to determine the deformation properties of the layers in the tunnel section. 2.3

Geotechnical properties

The geotechnical design properties were derived from the results of the site investigation, field and laboratory tests. For tunnel design three major differ­ ent layers (Table 1) were defined.

DOI: 10.1201/9780429321559-55

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Figure 1. Longitudinal section with geologic conditions and buildings above the tunnel.

Table 1.

Geotechnical parameters used in design.

Parameter

Old gravel Cailloutis Symbol Unit morainique

Classification USCS

Unit Weight

γsat

Friction

—

GM, GP

Glacial Moraine retreat GM-GC CL,ML

SM

SM­ SC

23.2

22.7

21.5

φ‘k

kN/ m3 °

34

34

27

ψ‘

kPa

10

6

0

c‘k

MPa 0….100

5

5

ME

MPa 100…300

80… 150

10

ME‘

MPa 300…1ʹ500

240… 750

30

2.3.2 Moraine This moraine was deposited during the last ice-age (Würm) and consists of gravel, subordinate sand with high content of silt and clay fraction and is grey to brownish coloured. In the eastern section the lower part lacks the gravel and cobbles. In this layer high SPT blow counts with NSPT > 50 were determined; the moraine is very dense. Four pressure-meter tests had also been carried out. The layer is very impervious due to the high content of fines and the high density. Sus­ pended water tables are present within this moraine.

angle Angle of Dilatancy Cohesion Intercept Initial

modulus Reloading

modulus

2.3.3 Glacial retreat and transition zone The moraine is covered by transition layers that in part are glacial till and silty clay and clay layers deposited in a frozen stage by the retreating glacier. These layers are unconsolidated and compressible and located several meters (Figure 1) above the exca­ vated tunnel section, except near the eastern portal. 3 GENERAL TUNNEL DESIGN AND TENDER

2.3.1 Old gravel (Cailloutis morainiques) This layer of interglacial gravel forms the base of the plateau and the lower part of the tunnel always stays in this layer (Figure 1), overall it is rather homoge­ neous and mainly consists of slightly silty gravel with cobbles to 200 mm and is very dense, with number of blow counts NSPT > 50. Above the water table there are cemented layers, sand and open gravel layers pre­ sent. Core drilling produced rock flour above the water table, thus the grain-size distributions may be misleading. Geotechnical classification is mainly GM, GW and GP. The layer has a large permeability, K >1….2.5 10-2 m/s, as determined by pumping tests below the groundwater table.

The Plateau de Champel has been built over during the last century with buildings up to eight stories high of different construction periods and types. The tunnel had to be driven below these existing build­ ings through soil, with 10 to 30 m overburden. The main access to the tunnel was on the eastern bank of the Arve river with a 50 m high steep slope (falaise). Access had to be gained first with a bridge over the river and the construction of the portal cut. The station in the middle of the tunnel subdivides the tunnel in two rather short 509 and 928 m long sections. In addition, an enlarged section was neces­ sary at the western entrance to the station of Cham­ pel underneath existing buildings.

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As for mechanical tunnelling with a shielded TBM a rather long installation and start area is needed, a TBM solution was not considered economical. 3.1

Urban constraints

The plan view of the western section (Figure 2) shows that the tunnel lies beneath numerous 8-story apartment buildings of different construc­ tion periods and types. Some with concrete struc­ tures. At the eastern side of the station, a building from the first part of the 20th century with masonry walls and wooden floor beams had to be under crossed. The eastern section (Figure 3 3) had to cross to a smaller extent below existing buildings; however, some crossings were diagonally, which had the potential of causing asymmetric deform­ ations to the buildings. At the eastern end, with the poorer ground conditions, the alignment passes between two rows of apartment buildings. Directly above the tunnel axis were one story garage buildings. 3.2

Access to the western portal

The main site installations were on the west bank of the Arve River (Figure 2). The portal is on the steep east bank (Figure 4). For gaining access to the portal area a temporary bridge had to be constructed south of the future railway alignment. The slope had to be secured before the portal cut was excavated in stages with continuous support with soil nails, shotcrete, micro-piles and tiebacks from above the tunnel. This part of the tunnel was a major geotechnical project.

Figure 4. Portal below the cliff and bridge over Arve River with temporary bridge and auxiliary piers for sliding-in the permanent railway bridge.

3.3

Tunnel design

Geometric conditions already practically excluded the use of a shielded TBM. Ground conditions would require positive, either slurry or earthpressure balance face support. The presence of very pervious layers might lead to loss of slurry and face support. For an Earth Pressure Balance shield the cemented layer would form problem­ atic obstacles. Considering these factors, the use of a TBM had been excluded for the tunnel de Champel. Settlements at the surface had to be limited to a few centimetres to prevent damage to overlying buildings; this meant that a settlement trough with a loss of ground of less than 0.4% had to be achieved. A construction method with continu­ ous roof, lateral and invert support (Figure 5) was selected. 3.3.1 Cross section and construction sequence The cross section with conical pipe umbrella and face support gives an interior concrete liner of vari­ able thickness (Figure 5). The minimal liner thick­ ness is 0.4 m and reaches 1.45 m (Figure 5) at the end of a stage (Figure 6).

Figure 2. Alignment of western section between western portal and Champel station showing built-up area on surface.

Figure 3. Alignment of tunnel between Champel station and Eaux-Vives station with built-up zones.

Figure 5. Cross Section of tunnel with variable thickness liner.

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Figure 6. Longitudinal section through regular tunnelling method selected at Tunnel de Champel.

3.3.2 Roof umbrella The roof umbrella consists of 51 to 77, 15 m long, pipes, with 140 mm diameter and 10 mm wall thickness, drilled at an angle of 5 to 7° to the tunnel wall; drilling took 25 min per boring. The void between pipe and drill-hole were grouted through the four valves per meter of pipe with a pressure of 3 to 5 bar. Grout take was about 350 l per pipe and lasted about 5 minutes. The following umbrella was drilled after 10 meters of successive excavation phases of initially 1 m length, later increased to 1.25 m. 3.3.3 Face support Face support was with 20 m long Fiberglass bolts, 38 to 47 bolts per excavation stage, with a structural capacity of 550 kN, grouted into the borehole, with a minimal overlap of 10 m. Grout take of the cement grout, with a W/C ratio = 0.8, was 500 l. Time for drilling was 35 minutes and grouting took 10 minutes. After each phase the face was supported by a 7 cm layer of fibre reinforced shotcrete. 3.3.4 Drainage borings For dewatering saturated sand pockets during each phase 3 sub horizontal 15 m long drainage boring with a maximal inclination of 5° were planned with a PVC filter pipe. None proved necessary.

steel foot to the lattice girders. The placement of the invert with shotcrete had to be achieved no later than 3.75 m behind the face. 3.3.7 Enlarged section The western end of Champel station was located dir­ ectly beneath residential buildings and therefore could not be built with the top down method of the station. The station had to be constructed as tunnel over a length of 30 m with a cross section of 160 m2. The maximum span is 17.2 m, the height 12.5m and the overburden to the foundation of the buildings is 10 m and 14 m to the ground surface. Initially the construction of the double layer pipe umbrella was planned from within the station towards the tunnel drive from the west portal. The construction and scheduling constraints within the station led to the search for another solution with driving of the enlarged tunnel from west. The pipe roof had to be inclined at a larger angle to the tunnel. Consequently, the excavated length had to be reduced to 6 m, which resulted with the 12 m long pipes in a double layer (Figure 7) pipe umbrella. The double layer pipe roof umbrella was stiffer and compensated for the larger span. The last pipe umbrella was fixed into the slurry trench wall of the station box in order to limit settlements. Settlements of a few millimetres only were measured.

3.3.5 Excavation and support sequence The tunnel was excavated initially with 1 m long stages (rounds), which could be increased to 1.25 m with the experience gained during tunnelling. With each round a four-rod lattice girder with dimen­ sions adjusted to the actual cross-section of the tunnel was placed, which was followed by steel-fibre reinforced shotcrete layers placed in stages of 7 to 10 cm thickness up to the final thickness of 400 mm. 3.3.6 Invert support The invert is supported with a wide-flange steel beam HEB 200 that connects with a prefabricated

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3.3.8 Eastern portal in station At the eastern side of the station the tunnel excava­ tion started again and passed underneath an 8-storey building, dating from the 1910’s. For 30 m the tunnel was constructed with the double roof umbrella and 6 m stages in order to limit the settlements. 3.3.9 Tender The tender design was based on the above described description with focus on details of execution of the grouting the pipes of the roof umbrella and the bolts of the face support. If these voids would not be

Figure 7. Tunnelling through the enlarged section with double roof umbrella and shortened length of phases.

grouted fully, a loss of ground of 1% would result with associated settlements. 4 DESIGN ANALYSES The project started in 2004 and the effect of sur­ face settlements was identified as the major risk for damaging the existing structures. Other important risks were face instabilities or worst collapse to the surface. The first initial designs started with estimating surface settlements with the empirical solutions. The practical available numerical methods in 2004 were plain strain FE elements, the three-dimensional effects were esti­ mated with correction factors. The empirical methods (Mair et al. 1996, O’Reilly et al. 1982) were judged more reliable. The results of these initial analyses showed that a loss of ground less than 0.4% must be achieved. When the design entered execution phase three-dimensional finite element programs became available and analyses with PLAXIS 3D were carried out. 4.1

to complex and on the other data the buildings were mostly not sufficiently documented. The initial liner from shotcrete was modelled linear elastic and considers the development of the stiffness with time (Möller 2006) based on an advance rate of 1m/day. The contact between liner und soil was mod­ elled with interface elements. The piles of the roof umbrella (Figure 8) were modelled as “embedded piles”. The modelling of the face support with 30 to 50 fibre-glass bolt per excavation phase would substan­ tially increase the computational effort, therefore the face support was modelled by an equivalent distributed load acting on the face (Figure 9) along the axis of the tunnel. The necessary anchors in the face to avoid face instabilities were determined beforehand with limit equilibrium models (Anagnostou 1999). Tunnel advance was modelled (Ruse 2004) as round length of 1 m with corresponding phases of placement of liner thickness.

Three-dimensional finite element analysis

With the FE-code PLAXIS-3D (Brinkgreve et al. 2012) the effect of tunnelling along the entire should be simulated as realistically as possible considering all the support elements (Figure 8) and excavation and construction phases. The soil was modelled with the material model: “Hardening soil with small-strain stiffness” (HS-small). The strength of the soil is considered with a friction angle ɸ’, an angle of dilatancy ψ and the cohesion intercept c’ (Table 1). The stiffnesses of the soil layers were described by the initial loading and the reloading modulus. The existing surface buildings or structures were modelled as area loads (Figure 8) on the level of the foundation. The detailed modelling was on one hand

425

4.2

Parametric studies

The precision and the variability on the geotechnical properties (Table 1) and the influence of their variation was studied in parametric studies. The effect of the soil stiffness and variable face support is shown in Figure 9. Different combinations of soil stiffness were used to simulate the effect on settlements (Figure 10) pre­ dicting longitudinal and transversal settlement troughs. 5 MONITORING AND OBSERVATIONS 5.1

Monitoring of the tunnel advance

Monitoring of the tunnel advance is the key element for a successful completion with the required min­ imal effect on the surrounding construction.

Figure 8. 3D Finite Element Model used for modelling face support and settlement.

Figure 9. Deformation at face as function of soil stiffness and equivalent support pressure of face bolts.

Convergence measurements were carried out during every excavation phase. The maximum convergence measured convergence was 4 mm, which confirmed the stiffness of ground and liner. The vertical deformations of the roof umbrella were monitored (Figure 11) with sub-horizontal inclinometers installed in the roof. The measured and computed deformations compare favourably, mostly around 10 mm, whereas the shape deviates slightly. The deformations ahead of the face were moni­ tored with reverse head extensometer (RH) and gave values of 3 to 10 mm thus confirming the computed values (Figure 10) and the high ground stiffness. The monitoring during tunnel advance was the key elem­ ent to control the deformations at the surface and adapt the construction procedures in a time.

5.2

Surface settlements

In the zone where the tunnel was in the dense layer of the old gravel or the moraine, the measured settle­ ments varied between 0 to 10 mm. In Figure 12, settlements along the tunnel axis and across are com­ pared to the predictions that were substantially higher. In the eastern area where overburden was smaller (6 m) and consisted of glacial retreat over 200 m length, as predicted, larger settlements of 60 mm were observed. The settlement through was 10 m wide and the structures were one story garages. The garages needed repairs. In this area during drill­ ing of the roof umbrella some heave was observed, as the drilling for the placement of the pipes dis­ placed the ground.

426

Figure 10. Computed settlements for different stiffness of the soil (unloading and reloading modulus).

5.3

Figure 11. Measurements with inclinometers in pipe roof.

Verification of soil-structural interaction

The overlying buildings had foundations of different age and types and had to be studied and their suscep­ tibility to settlement damage was assessed by an independent mandate. Each building was assigned a susceptibility class and assigned limit values of total and differential settlements. These limiting values were compared to the predicted deformations and assigned risk classes. Those buildings that were assigned to an unacceptable risk class were subjected to an in-depth structural analysis, e.g. with finite element analyses. And to more reliably determine the risk classes and propose damage mitigating measures, if necessary. In case of Tunnel de Champel all examined build­ ings could be assigned a lower risk class and the deformation limits be adjusted accordingly. In total

Figure 12. Settlements measured at the ground surface compared to computed values with 3D Finite Elements.

427

the damage to existing structures by tunnelling was minimal.

placement the invert strut was stipulated in add­ ition to the maximal ring closure distance.

6 ADAPTATION OF TUNNELLING

7 CONCLUSIONS

The planned support procedure was essentially kept over the entire construction time. The full-face exca­ vation procedure with a face of more than 100 m2 was an extremely large cross section for soil. The full-face construction avoided different intermediate phase of stress changes, which would accumulate more deformations and lead to more settlements and damage to structures above the tunnel. The work safety in the tunnel should not be compromised. There were, however, different changes made during the tunnel driving based on the experience or in one case due to an accident at the face and subsequent work stoppage.

The construction of the Tunnel de Champel in Geneva through glacial soils proved challenging. The full-face excavation with support with a pipe roof umbrella and fibre reinforced anchor in the face proved successful. The proper grouting of the voids around the pipes and the ground are an important issue and were part of the success of the method. The detailed investigation of the underground with cored borings and in-situ tests:

6.1

Double roof umbrella

In the enlarged zone and the eastern portal zone at Champel station the roof umbrella was changed to a double pipe roof as described in the section on design. 6.2

Roof support in the zone with soft retreat

In the eastern zone of the tunnel close to EauxVives station, it was foreseen to complement the pipe roof umbrella with jet-grouting columns to close the gaps between the pipes. Tests were only possible, once the tunnel was under construction. The tests showed that grout escaped to the sur­ face and grout columns could not be formed to the required diameter. Instead also a double pipe roof umbrella was constructed. 6.3

• Standard Penetration Tests: SPT • Dynamic penetrometer tests: DPSH • Pressure-meter tests for determining the soil stiff­ ness (deformation properties) • Permeability tests • Piezometers

Local instabilities at the face and sidewalls

In late 2014 an accident close to the face with materials falling out led to a work stoppage and gave rise to intensive discussions regarding the tunnelling concept. Finally, it was decided to implement at the side walls more closely spaced pipes that were in the tender only foreseen spor­ adically. Before the tender documents these sup­ port measures had been foreseen at the same distance as in the crown. The lower side walls received a support placed ahead of the excavation, which made the con­ struction area safer. As a benefit the round length could be increased from 1.0 to 1.25 m, as sug­ gested by the contractor, after an additional static verification. This had the additional benefit that the rate of advance was increased with increased safety. As a further measure to increase safety near the face and reduce the risk, i.e. the time of stay near the face, a minimal distance of

were key elements for characterizing the subsoil conditions and formed the basis for the design with the use of advanced numerical methods, i.e., threedimensional finite element programs. The prediction of the deformation of the support elements (Pipe roof) and the ground (face of the tunnel) allowed to verify during tunnel advance the performance of the tunnel. In case of unsatisfactory performance, the construction procedures could have been modified. The monitored surface settlements were less than the computed ones. In many sections where the tunnel lies in the over-consolidated, inter­ locked stiff old gravel (Cailloutis morainique) or the moraine, measurements showed no surface settlement. The characteristics of these soils are difficult to model in numerical analyses. In a few sections, settlements were measurable, the example in Figure 12 shows computed settle­ ments of 25 mm vs. 10 mm measured.

ACKNOWLEDGMENT The Tunnel has been built by CEVA an organization of the Swiss Federal Railways SBB-CFF and the Canton of Geneva. Design, General and Site supervi­ sion was carried out by GECA the association of: • • • •

Stucky SA, Lausanne B+S AG, Bern EDMS SA, Petit Lancy GLM Gysi Leoni Mader AG, Zürich

B+S AG was responsible for the design of the tunnel. Tunnel contractor was Marti Construction SA, Geneva and Marti Tunnel AG, Moosseedorf.

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The authors thank the owner CEVA for giving permission to publish this paper.

REFERENCES Anagnostou, G. 1999. Standsicherheit im Ortsbrustbereich beim Vortrieb von oberflächennahen Tunneln. Sympo­ sium „Städtischer Tunnelbau: Bautechnik und funktio­ nelle Ausschreibung“, Zürich. Brinkgreve, R.B.J., Engin, E. & Swolfs, W.M. 2012. PLAXIS 3D 2012. Plaxis bv, Delft. Mair, R.J. Taylor, R.N. & Burland, J.B. 1996. Predic­ tion of ground movements and assessment of risk of building damage due to bored tunnelling. Proceed­ ings International Symposium on Geotechnical

Aspects of Underground Construction in Soft Ground, London, 713–718. Möller, S,C. 2006. Tunnel induced settlements and struc­ tural forces in linings. PhD thesis, Institut für Geotech­ nik, Universität Stuttgart. O’Reilly, M.P. & New, B.M. 1982. Settlements above Tun­ nels in the United Kingdom – their magnitude and pre­ diction. Tunnelling 82, Institution of Mining and Metallurgy, London, 173–181. Ruse, N.M. 2004. Räumliche Betrachtung der Standsicher­ heit der Ortsbrust beim Tunnelvortrieb. PhD thesis, Institut für Geotechnik, Universität Stuttgart. Witschi, T, Steiner, W.& Ferrari, A, 2013. Urban, geo­ technical and construction challenges for the realiza­ tion of the CEVA Tunnel de Champel in Geneva, Proc. World Tunnel Congress 2013, Geneva, 861–868.

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Simplified stress-strain models applied to data from triaxial and pressuremeter tests on London Clay P.J. Vardanega University of Bristol, Bristol, UK

M.D. Bolton & S.K. Haigh University of Cambridge, Cambridge, UK

R.W. Whittle Cambridge Insitu Ltd, Cambridge, UK

A. Klar Technion – Israel Institute of Technology, Haifa, Israel

M.G. Williamson Mott MacDonald Group, Singapore

ABSTRACT: The Mobilisable Strength Design (MSD) philosophy has been used in various applications related to underground construction, e.g. for analysis of deep foundation and retaining wall performance. MSD requires simple models for the stress-strain behaviour of soils. The use of a mobilisation factor on undrained strength to limit soil mobilisation was introduced in BS8002 in 1994. To assist with MSD calcula­ tions, the mobilisation strain framework (MSF) has been developed to allow geotechnical engineers to account for the non-linear behaviour of fine-grained soils in routine geotechnical design. In this paper, triaxial and pressuremeter test data from the London Clay deposit are analysed, using the MSF, to study the effects of anisotropy on both the mobilisation strains and non-linearity exponent. The implications for design of under­ ground constructions are also discussed.

1 INTRODUCTION 1.1

Mobilisable strength design

This paper presents a summary of some recent advances in simplified soil models that can be used in Mobilisable Strength Design (MSD). Simplified stress-strain models are compared with pressuremeter and triaxial data from the London Clay deposit. The effects of past stress history and shear mode are discussed in this paper. Considerable effort has been undertaken in recent years to characterise the stress-strain behaviour of fine-grained soils in an attempt to improve soil-structure interaction calcula­ tions. Characteristic values for soil parameters are required when using Eurocode 7 (CEN, 2004) for the purposes of design. Arguably, most geotechnical engineers have little problem assigning characteristic values for strength parameters e.g. undrained shear strength (cu) or friction angle (ϕ'peak, ϕ'cv, ϕ'res). Recently the idea of using characteristic stress strain curves to complete simplified soil deformation

calculations has entered some design codes and man­ uals (BSI 2015a, 2015b). The importance of stressstrain (particularly small strain non-linearity) for geotechnical design is well explained in Atkinson (2000). However, it can be argued that the moderate strain region (defined in Vardanega & Bolton (2011a) to be 0.2 ≤ τmob/cu ≤ 0.8 where τmob is the shear stress sufficiently smaller than the undrained shear strength cu) is relevant to many geotechnical constructions and therefore studying the effects of the stress-strain behaviour confined to this region is worthy of continued study. 1.2

London Clay deposit

Considerable tunnelling work has been carried out in London Clay over many years (e.g. Gourvenec et al., 2005 and Mair, 2008). Advanced mechanical testing on London Clay has been performed by many researchers (e.g. Ward et al. 1959, 1965, Parry 1960, Bishop et al. 1965, Skempton et al. 1969, Atkinson 1975, Yimsiri 2001, Gasparre 2005, Nishimura 2005

DOI: 10.1201/9780429321559-56

430

Hight et al. 2007 and Gasparre et al. 2007). The data from this work has often been used to calibrate numerical models for use in the London clay deposit (e.g. Simpson et al. 1979, Simpson 1992 and Ellison et al. 2012). Vardanega & Bolton (2011b) analysed triaxial compression test data from Gasparre (2005), Yimsiri (2001) and Jardine et al. (1984) using the mobilisation strain framework (MSF) set out below.

Vardanega and Bolton (2011a) showed that the following power-law expression captures the shear stress-strain behaviour of intact clays and silts in only three parameters (Eq. 3):

1.3

where, γM=2 is the engineering shear strain required to mobilise 50% of the strength, b is a curve fitting parameter and M is a mobilisation factor (see BSI 1994). Equation 3 was successfully used along with a concentric cylinder model (e.g. Randolph & Wroth 1978, Fleming et al. 2009) to develop simple expressions for the settlement of bored piles in London Clay (Vardanega et al. 2012a, Williamson 2014, Kolodiy et al. 2015, Vardanega 2015, Williamson et al. 2017, Crispin et al. 2019 and Voyagaki et al. 2019). Klar and Klein (2014) presented an alternative exponential function (Eq. 4):

Study aims

This paper has the following aims: (i) study the effects of past stress history (indicated by sample depth) on the MSF parameters of London Clay, taken as a case study; (ii) determine if transformation models (e.g. Phoon & Kulhawy 1999a, 1999b) can be determined linking the MSF parameters with depth for the London Clay deposit and (iii) compare the values of the MSF parameters for three test types: triaxial compression (CIUC), triaxial exten­ sion (CIUE) and pressuremeter (PMT). 2 SOIL STRESS-STRAIN MODELLING 2.1

Empirical models

Hollomon (1945) proposed a simple equation of the form shown as Equation 1 to model the strainhardening behaviour of metals:

which was used to develop an MSD-inspired for­ mulation for volume loss due to tunnelling. 2.2

where, σ = stress, ε = strain, K and n are curve fit­ ting parameters with n representing the ductility of the material. Jardine et al. (1984) and Jardine et al. (1986) pro­ posed the six-parameter model (Eq. 2) to describe the undrained shear behaviour of some North Sea clays:

where, Eu is the strain-dependent undrained Young’s modulus, εa is the axial strain in a triaxial test (taken as 2/3 the engineering shear strain, γ) and A′, B′, C′, α and Γ are curve-fitting parameters. For geotechnical work the ε50 (strain to half the undrained shear strength) was recognised by Matlock (1970) as a useful normalising parameter for soil strains and proposed a power-law model although the non-linearity exponent was assigned a fixed value (Zhang & Anderson, 2017). The use of the ε50 (axial strain to 50% strength mobilisation) was also dis­ cussed in the review of Wroth et al. (1984).

Influence of past stress history

Undrained shear strength of fine-grained materials is strongly affected by overconsolidation ratio (OCR) (e.g. Henkel 1956, Mayne 1985, Mayne 1988, Mayne & Stewart 1988, Mayne 2001 and Mayne et al. 2009), strain rate (e.g. Kulhawy & Mayne 1990) and shear mode (Chen & Kulhawy, 1990 and Mayne et al. 2009). Vardanega and Bolton (2011a) presented the ana­ lysis of a geo-database of 115 stress-strain curves on 19 clays and silts. An average b of about 0.6 was determined and a standard deviation (SD) of 0.15 computed. The multiple linear regression analysis yielded the following expression (coefficient of determination, R2 = 0.44, number of data-points used to generate the correlation, n = 97):

where, IP = plasticity index, patm = atmospheric pressure, p'0 = confining stress and C = a regression constant (approximately 0.011). Equation 5 was used by Vardanega & Bolton (2011a) to propose that:

431

(also reviewed in Beesley & Vardanega 2020b: for further details on RFG/TXCU-278 see also the thesis of Beesley 2019): Given that many of the sources of data analysed in Vardanega and Bolton (2011a) did not report the OCR of the materials, a suite of triaxial tests on reconstituted kaolin were conducted. Vardanega et al. (2012b) pre­ sented test data for a kaolin and showed that γM=2 cor­ related strongly with increase in OCR (R2 = 0.815, n = 18) (Eq. 7) while the b value was correlated with OCR (R2 = 0.591, n = 18) (Eq. 8) albeit with a lower R2 value.

[R2 = 0.71, n = 50, standard error, SE = 0.0031, probability that no correlation exists, p < 0.001]

[R2 = 0.46, n = 25, SE = 0.0099, p < 0.001] The next section investigates the effect of sample depth in the London Clay deposit on the MSF parameters. Vardanega and Bolton (2011b) analysed triaxial compression data from samples from the London Clay deposit and reported that a function of γM=2 decreasing with depth (d) (R2 = 0.46, n = 17) could be valid (Eq. 9):

2.3

Influence of K0 and shear mode

Vardanega & Bolton (2011a) showed how Equation 3 could be ‘shifted’ to better account for the ‘K0­ effect’. This technique was subsequently used by Li & Bolton (2014) to analyse retaining walls in sand. Later Vardanega (2012), Vardanega & Bolton (2016) (while discussing Casey et al. (2016) and Casey (2016) presented the following modification of Equation 3 for K0 data:

where, τ0 = initial shear stress and γref,K0 = shear strain to mobilise B = 0.5. Beesley & Vardanega (2020a) assembled and ana­ lysed a database (named RFG/TXCU-278) of recon­ stituted fine-grained soils to study in part the effect of shear mode on the γM=2, γref,K0 (referred to as γ50 with the shear mode denoted as a subscript for the rest of this paper) and the b values from the mobil­ isation strain framework. Beesley & Vardanega (2020a) found the following correlations (Eq. 11 and Eq. 12) for the γ50 parameter as related to a comparison of compression and extension testing

3 ANALYSIS 3.1

Triaxial tests

Vardanega & Bolton (2011b) analysed a collection of CIUC tests (n=17) on London clay samples sourced from the literature. In the previous work three London Clay sites were represented: Canon’s Park (Jardine et al. 1984), Kennington Park (Yim­ siri 2001) and Heathrow Terminal 5 (Gasparre 2005). In Mayne’s discussion to Vardanega et al. (2012b) the relevant cu value needs to be used in the normal­ isation e.g. if an extension test is performed then cu (extension) is required (Vardanega et al. 2013). Given the importance of examining the effect of shear mode (Vardanega et al. 2013, Beesley 2019 and Beesley & Vardanega 2020a, 2020b) comparison with triaxial extension test data is warranted. Figure 1 and Table 1 shows the analysis of exten­ sion test data (n=18) from Hight et al. (2007), Gas­ parre (2005) and Nishimura (2005) on London Clay samples from Heathrow Terminal 5 (see also Klar & Klein 2014 where the summary of this analysis was first reported). Figure 1 shows the data on plotted log-log axes and the straight-line fits obtained (for the data generally in the range of 0.2τmob/cu to 0.8τmob/cu). Table 1 shows the ε50 values derived in two ways: from the fit of the function (scaling off the curve) and from the derived log-log relation (the method that was used in Vardanega and Bolton, 2011a, 2011b) – for these tests the results are essen­ tially the same. For the subsequent analysis of the triaxial data in this paper the γ50 and b values from the fitted log-log curve are used. Figure 2a shows that the γ50 CIUC values range from 0.0043 to 0.0118 with an average value of 0.0075. The computed SD value for this dataset (n=17) is 0.00023 and the COV = 31%. As also reported in Vardanega & Bolton (2011b) a correlation with depth (decreasing OCR) (cf. the

432

This trend follows that observed for reconstituted soils: γ50 increases with increasing OCR (Vardanega et al. 2012b and Beesley & Vardanega 2020a). It should be noted that the R2 for an exponential fit to the data from Figure 2a does result in a slightly higher R2 value of 0.46 (cf. Eq. 9 and Vardanega & Bolton 2011b): a power fit is shown in this paper. Figure 2b shows that the b CIUC values range from 0.38 to 0.83 with an average value of 0.57. The com­ puted SD value for this dataset (n=17) is 0.12 and the COV = 21%. As also investigated in Beesley & Vardanega (2020b) a correlation with depth (decreas­ ing OCR) (cf. Butcher & Lord 1993) is found (Eq. 14).

Figure 1. Normalised Extension Tests performed on high quality samples of London Clay from Heathrow Terminal 5: data from Hight et al. (2007), Gasparre (2005) and Nishi­ mura (2005).

Table 1. Analysis of extension test data from Hight et al. (2007), Gasparre (2005) and Nishimura (2005). Fit of function

Log-log relation

Test*

d (m)

b

ε50

b

ε50

B2c 11.35m B2b 19.96m A3 37.4m A2 48.67m B2c 10.6m B2a 23m B2a 27.9m 7gUE 25.4gUE 23gUE TE1 TE2 TE3 TE4 TE6 TE7 TE8 TE9

11.35 19.96 37.4 48.67 10.6 23 27.9 7.1 25.4 23.2 20.9 16.1 27.9 8.2 13.6 10.6 29.1 28.2

0.31 0.37 0.47 0.52 0.36 0.43 0.57 0.21 0.48 0.26 0.40 0.42 0.63 0.44 0.59 0.45 0.50 0.51

0.00042 0.00110 0.00259 0.00224 0.00189 0.00186 0.00102 0.00100 0.00363 0.00028 0.00062 0.00036 0.00248 0.00419 0.00348 0.00162 0.00064 0.00397

0.31 0.36 0.46 0.48 0.39 0.47 0.56 0.21 0.53 0.26 0.40 0.46 0.62 0.42 0.61 0.43 0.50 0.51

0.00042 0.00110 0.00262 0.00225 0.00188 0.00186 0.00103 0.00097 0.00380 0.00028 0.00062 0.00036 0.00249 0.00428 0.00350 0.00155 0.00064 0.00398

Max. Average Min. SD ** COV (%) ***

48.7 21.6 7.1 10.9 50

0.63 0.44 0.21 0.11 25

0.00419 0.00186 0.00028 0.00130 70

0.62 0.44 0.21 0.11 25

0.00428 0.00187 0.00028 0.00133 71

Interestingly this trend is the reverse of Equation 8 (Vardanega et al. 2012b). Figure 3a shows that the γ50 CIUE values range from 0.0004 to 0.0064 with an average value of 0.0028. The computed SD value for this dataset (n=18) is 0.00020 and the COV = 71%. No correlation with depth was observed. Equation 11 suggests that on average for reconstituted soils γ50 CIUC is about 1.3 times that of γ50 CIUE. However, for this natural clay dataset, the difference is approximately a factor of 2.7. Figure 3b shows that the b CIUE values range from 0.21 to 0.62 with an average value of 0.44. The SD value for this dataset (n=18) is 0.11 and the COV = 25%. A correlation with depth (decreasing OCR) (cf. Butcher & Lord 1993) is found (Eq. 15).

As for Equation 14, this is the reverse of the trend shown as Equation 8 (Vardanega et al. 2012b). It is also observed that the average b CIUE value is lower than the average b CIUC value for the London Clay data analysed here – as similar trend is shown for the RFG/TXCU-278 database (see Beesley 2019 and Beesley & Vardanega 2020b). 3.2

* rows 1-7 (data from Hight et al. 2007); 8-10 (data from Gasparre 2005) and 11-18 (data from Nishimura 2005); ** standard deviation; *** coefficient of variation expressed as a percentage.

data from Gault Clay reported in Butcher & Lord 1993) is found (Eq. 13).

Pressuremeter tests

Pressuremeter tests in London clay have been reported by various researchers (e.g. Wood & Wroth 1977, Marsland & Randolph 1977, and Bolton & Whittle 1999). Figure 4 shows the analogous MSF parameters derived from pressuremeter test information from the 1992 Crossrail boreholes (see Shuttle & Jefferies 1996 who also made use of some of this dataset). Figure 4a shows that the γ50 PMT values range from 0.0017 to 0.0057 with an average value of

433

Figure 2. (a) γ50 CIUC versus depth for London Clay; (b) b CIUC versus depth for London Clay (data from Jardine et al. 1984, Yimsiri 2001, Gasparre 2005) (database originally analysed in Vardanega & Bolton 2011b).

Figure 3. (a) γ50 CIUE versus depth for London Clay; (b) b CIUE versus depth for London Clay (data from Hight et al. 2007, Gasparre 2005, Nishimura 2005).

Figure 4. (a) γ50 PMT versus depth for London Clay; (b) b 1992 (data from Cambridge Insitu Ltd)).

PMT

434

versus depth for London Clay (Crossrail boreholes from

Table 2.

MSF analysis of Crossrail pressuremeter tests at six locations in London Clay: variation of γ50PMT . Site statistics

Site

Max.

Average

Min.

SD

COV (%)

n

Davies Street (119.45mTD) Farrington (115.4mTD) Paddington (1) (126.4mTD) Paddington (2) (126.4mTD) Royal Oak (121.7mTD) Falconberg Court (Tottenham Court Rd) (124.45mTD) ALL

0.00498 0.00341 0.00537 0.00498 0.00565 0.00486 0.00565

0.00355 0.00257 0.00352 0.00362 0.00387 0.00340 0.00347

0.00216 0.00210 0.00169 0.00225 0.00230 0.00201 0.00169

0.00086 0.00043 0.00093 0.00083 0.00076 0.00082 0.00087

24 17 26 23 20 24 25

17 12 21 10 19 17 96

Table 3.

MSF analysis of Crossrail pressuremeter tests at six locations in London Clay: variation of bPMT . Site statistics

Site

Max.

Average

Min.

SD

COV (%)

n

Davies Street (119.45mTD) Farrington (115.4mTD) Paddington (1) (126.4mTD) Paddington (2) (126.4mTD) Royal Oak (121.7mTD) Falconberg Court (Tottenham Court Rd) (124.45mTD) ALL

0.74 0.68 0.75 0.66 0.71 0.70 0.75

0.65 0.59 0.67 0.62 0.63 0.63 0.64

0.56 0.53 0.56 0.58 0.52 0.49 0.49

0.048 0.051 0.056 0.023 0.058 0.054 0.055

7 9 8 4 9 9 9

17 12 21 10 19 17 96

0.0035. This average value lies between the aver­ age value for the γ50 CIUE and γ50 CIUC values (and is closer to the γ50 CIUE value). The computed SD value for this dataset (n=96) is 0.00087 and the COV = 25%. No obvious correlation with depth is observed. Figure 4b shows that the b PMT value range from 0.49 to 0.75 with an average value of 0.64. This average value lies above the average value for both the b CIUE and b CIUC data (and is closer to the b CIUC value). The computed SD value for this dataset (n=96) is 0.06 and the COV = 9%. No obvious correlation with depth is observed. It is interesting that the COV values of the MSF param­ eters for the pressuremeter data (Figure 4) are lower than those for the compression (Figure 2) and extension test data (Figure 3). Tables 2 and 3 show the statistical variation of γ50 PMT and b PMT for each of the six London clay locations. The COV values for each PMT location are generally lower than for the triaxial locations, especially for the b values. (The two Canon’s Park tests are not considered in the following discussion as the sample size is too small to compute sensible COV values.) The Kennington Park tests γ50 CIUC values have a similar COV to some of the pressuremeter sites of around 26% while the Heathrow Ter­ minal 5 tests γ50 CIUC have a COV of around 31% with a COV of 71% for the γ50 CIUE values. This analysis is interesting as we may postulate that either the pressuremeter causes less disturbance

(testing and sampling) than the triaxial testing sampling regime or the pressuremeter sites more homogeneous than Heathrow Terminal 5 Kennington Park especially with respect to compliance (ductility).

and are and soil

4 SUMMARY AND CONCLUSIONS This paper has reviewed the use of the MSF for both triaxial and pressuremeter test data for data from various sites in the London Clay deposit. The fol­ lowing conclusions are made: (a) The MSF can describe both triaxial and pres­ suremeter test data from a natural clay deposit (in this case London Clay); (b) The γ50 CIUC values are on average three times higher than the γ50 CIUE values for the triaxial data set analysed in this paper (this difference is much greater that observed for the reconstituted test data from the RFG/TXCU-278 database (Beesley 2019 and Beesley & Vardanega 2020a)); (c) Correlations with depth were found for the γ50 CIUC, b CIUC and b CIUE data (although the com­ puted R2 values are relatively low) but not for the γ50 CIUE, γ50 PMT and b PMT data; (d) The COV values of the γ50 and b values for the pressuremeter sites are generally lower than those for the triaxial sites (especially for b). This could be due either to disturbance effects

435

or be an indication of site heterogeneity. In any event, the variation in the MSF values from state-of-the-art triaxial tests on high quality cores is potentially a matter of concern. If such cores can become irremediably damaged as they are recovered, extruded, trimmed and tested, the greater stiffness of pressuremeter tests may be more relevant to the design of underground facilities. Anisotropy would, how­ ever, remain a concern. A future comparison of all three test types on the same site at the same depth(s) would be useful to advance the MSF.

ACKNOWLEDGEMENTS The first author thanks Dr M. Beesley for her helpful comments and work on revising the MSF and Dr A. Gasparre for providing her triaxial test data for analysis. Data Availability Statement: This research has not generated new experimental data.

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European Conference on Soil Mechanics and Geotech­ nical Engineering. ICE Publishing, London, UK, vol. 7, 3777–3782. Vardanega, P.J. & Bolton, M.D. 2011a. Strength mobiliza­ tion in clays and silts. Canadian Geotechnical Journal 48(10): 1485–1503, Corrigendum, 49 (5):631. Vardanega, P.J. & Bolton, M.D. 2011b. Predicting shear strength mobilization of London clay. In (Anagnostopoulos, A. et al. eds.) Proceedings of the 15th European Conference on Soil Mechanics and Geotechnical Engineering vol. 1. 487–492. Vardanega, P.J. & Bolton, M.D. 2016. Discussion of “Undrained Young’s Modulus of Fine-Grained Soils” by B. Casey, J.T. Germaine, N. O. Abdulhadim, N. S. Kontopoulos and C. A. Jones. Journal of Geotech­ nical and Geoenvironmental Engineering 142 (10): [07016023]. Vardanega, P.J., Williamson, M. & Bolton, M.D. 2012a. Bored pile design in stiff clay II: mechanisms and uncertainty. Proceedings of the Institution of Civil Engineers – Geotechnical Engineering 165(4): 233–246, Corrigendum, 166 (5):518. Vardanega, P.J., Lau, B.H., Lam, S.Y., Haigh, S.K., Madabhushi, S.P.G. & Bolton, M.D. 2012b. Laboratory measurement of strength mobilization in kaolin: link to stress history. Géotechnique Letters 2(1): 9–15. Vardanega, P.J., Lau, B.H., Lam, S.Y., Haigh, S.K., Madabhushi, S.P.G. & Bolton, M.D., Mayne, P.W. 2013. Discussion: Laboratory measurement of strength mobil­ isation in kaolin: link to stress history. Géotechnique Letters 3(1): 16–17. Voyagaki, E., Crispin, J., Gilder, C., Nowak, P., O’Riordan, N., Patel, D. & Vardanega, P.J. 2019. Ana­ lytical Approaches to Predict Pile Settlement in London Clay. In: El-Naggar H. et al. (eds) Sustainability Issues for the Deep Foundations. GeoMEast 2018. Springer, Cham, Switzerland, 162–180. Ward, W.H., Marsland, A. & Samuels, S.G. 1965. Proper­ ties of the London Clay at the Ashford Common Shaft: In-situ and Undrained Strength Tests. Géotechnique 15 (4): 321–344. Ward, W.H., Samuels, S.G. & Butler, M.E. 1959. Further studies of the properties of London Clay. Géotechnique 9(2): 33–58. Williamson, M.G. 2014. Tunnelling Effects on Bored Piles in Clay. Ph.D. thesis, University of Cambridge, Cam­ bridge, UK. Williamson, M.G., Elshafie, M.E.Z.B., Mair, R.J. & Devriendt, M.D. 2017. Open-face tunnelling effects on non-displacement piles in clay – part 1: centrifuge mod­ elling techniques. Géotechnique 67(11),983–1000. Wood, D.M. & Wroth, C.P. 1977. Some laboratory experi­ ments related to the results of pressuremeter tests. Géo­ technique 27(2): 181–201. Wroth, C.P., Randolph, M.F., Houlsby, G.T. & Fahey, M. 1984. A Review of the Engineering Properties of Soils with Particular Reference to the Shear Modulus. O.U.E. L. Report No. 1523/84 (Soil Mechanics Report No. SM049/84), Department of Engineering Science, Uni­ versity of Oxford, Oxford, UK. Yimsiri, S. 2001. Pre-deformation characteristics of soils: anisotropy and soil fabric. Ph.D. thesis, University of Cambridge, Cambridge, UK. Zhang, Y. & Anderson, K. 2017. Scaling of lateral pile p-y response in clay from laboratory stress-strain curves. Marine Structures 53: 124–135.

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Propped cantilever wall stability design with the ‘What You Design Is What You Get’ method – background and development C.K.S. Yuen Transport for New South Wales, Sydney, Australia

ABSTRACT: Stability is a crucial consideration in the design of embedded cantilever walls. There are many methods available for this design. However, the choice for a suitable method is not easy as the mean­ ing of stability differs significantly between methods. Communication of the calculated wall stability could pose false security between designers, asset owners and operators. Designers often regard active earth pres­ sures as loads and passive earth pressures as resistances, but is this correct? Following a rational examin­ ation a postulation is made on the nature of earth pressures on embedded structures. This leads to the formulation of a new model based on limit equilibrium to calculate wall stability. The formulation over­ comes the issues with the existing methods and is shown to be consistent for soil spectrum maintaining a constant margin for a given stability design. The proposed method relates the Factor of Safety as a function of the restoring moment capacity at critical equilibrium and What You Design Is What You Get (WYDIWYG). For example if the required factor of safety is two then the design ensures the restoring cap­ acity of the wall is two times that at critical equilibrium. It provides a new platform for objective assess­ ment of wall stability. The approach is reliable, economical and suitable for conforming design to Australian Standard AS 5100 Bridge design.

1 INTRODUCTION

limit equilibrium and that is the focus for discus­ sion in this paper.

With the advent of urbanisation in Sydney, Australia, there is a significant increase in demand for basement excavation to accommodate such facilities as car park­ ing and other underground infrastructures. Embedded retaining walls are a most popular support type for such excavations. For these walls to be used satisfac­ torily, they need to be constructed into the ground to mobilise sufficient lateral resistance for stability. In general these walls are installed to an economical depth for stability and functional performance, which includes strength and deformation. An optimum bal­ ance between these considerations is hence needed. The lateral earth pressure distribution on the wall is a fundamental consideration for either stability or functional performance. The consideration of earth pressure distribution on these aspects is quite differ­ ent though. For functional performance such as wall deformation, bending moment and shear load designs, the walls are at stable equilibrium, the pressure distribution due to soil-structure inter­ action is an appropriate choice. There are ample lit­ erature discussion on such pressure distributions and factors affecting them e.g. Potts & Fourie (1984, 85), and readers should refer to them separ­ ately. For stability the pressure distribution is at

DOI: 10.1201/9780429321559-57

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2 ISSUES WITH THE CURRENT STABILITY DESIGN Stability assessment has been a contentious issue for many decades. Earth pressure distribution on the wall is idealised to facilitate calculation of loads for stabil­ ity design. Most engineers are familiar with the free and fixed earth support models, shown in Figure 1, for propped and free cantilever walls. Due to practical difficulty to locate the pivot point in the fixed earth model free earth support becomes a favourable simpler model to use where the earth pressure on the retained side is treated as load and that on the excavation side as resistance. The CP2 (1951) is a classical example. Variations of the CP2 model have formed the basis for most common methods such as the BSC Piling Hand­ book (1988) (BS), Strength Factor (SF) and Burland et al (1981) methods (BP). Figure 2 shows a schematic distribution of the earth pressures used by these methods where the shaded parts are the earth pressure modification represented by the indi­ vidual model.

Figure 1. Idealised earth pressure distribution for stability design of propped and free cantilever walls.

Figure 2. Schematic pressure distribution used by the common methods. Table 1.

Comparison of existing design methods. Case1:Effective Stress Case2:Total Stress FoS FoS

Model CP2 BS BP SF3 WYDIWYG5 1 2 3 4 5

Calc4 1.63 2.66 1.72 1.26 1.84

Acceptable Calc4 1

1.5-2.0 2.0 1.5-2.01 1.2-1.52 -

1.36 2.01 1.19 1.15 1.55

Acceptable 2.0 2.0 2.0 1.4-2.02 ­

the acceptable factor varies with strength of the materials and nature of the works these values vary with design standards the acceptable factor varies with method of analysis FoS computed using commercial package WALLAP, except the WYDIWYG values FoS calculated using the proposed method

It is apparent from the results that for a problem with the same margin of stability a wide range of FoS values are calculated, between 1.26 and 2.66 for the effective case, and between 1.15 and 2.01 for the total stress case. It is not possible to assess the quantum of stability offered in a design by the sheer calculated value, nor is it possible to compare between design cases to gauge adequacy of a design without also referencing a particular design method. The results of the proposed method are also included in the table for comparison purpose. In Australia stability of soil supporting structures for major infrastructure developments is designed using the Australian Standard AS 5100 Bridge design (2017). The ultimate geotechnical capacity is multiplied by a geotechnical reduction factor to ensure the product is larger than the design action effect. The geotechnical reduction factors, which are similar to the reciprocal of FoS, are defined in the Standard. The equivalent FoSs thus have a minimum value for design conformance. Those methods that consistently calculate FoS values lower than these minimum values thus cannot be considered for con­ formance reasons. Apart from the above, some of the methods are also known to calculate unrealistic results (Burland et al 1981, Simpson & Powrie 2001, Yuen 2019). This is particularly for some total stress conditions where increase in wall embedment does not bring a corresponding increase in the calculated FoS. For these reasons, there is still a need to clarify between the design communities and stakeholders on what margin of stability is being used in designs, both objectively and consistently. This could also help re­ establishing an optimal stability margin for modern day developments. 3 IS EARTH PRESSURE A LOAD OR A RESISTANCE

Table 1 compares their calculated factor of safety (FoS) for two cases. The first case is for a 5m deep supported excavation in a uniform soil with an internal friction angle of 35 de grees propped at the top of the wall with a toe embedment of 2.28m. A 10kPa sur­ charge is considered at the crest level of the wall and no groundwater is considered. The second case is similar to the first one with 5m deep excavation, but in a uniform clay with an undrained cohesion of 40kPa and a toe embedment of 1.49m below the exca­ vation level. A minimum fluid pressure of 5kPa is also considered in the total stress case.

Factor of Safety for stability assessment is expressed as a ratio of available resistance to design load. Traditionally earth pressure is treated either as a load or as a resistance depending on the relative movements between the soil and the struc­ ture. If the structure is moving away from the soil then the active pressure developed is considered as a load and if it is moving against the soil then the passive pressure is a resistance. This concept has been used on L-shape retaining walls and other gravity retaining structures. For embedded canti­ lever walls the situation is similar but only at limit equilibrium condition. For acceptable stability the propped cantilever wall needs to be installed beyond the critical equilib­ rium depth. Some noticeable changes in the earth pressures could be detected as the embedment is

439

As the wall embedment increases beyond the crit­ ical depth, the wall stability improves from a critical state to a stable state. The increase in stability is derived from the extra embedment beyond the crit­ ical depth. Logic follows that the earth pressure on the retained side of the extra embedment should not be treated as a disturbing load. It serves as a component to provide a reserve capacity for stabil­ ity such that the wall can sustain additional load without falling over. However, the pressure is acting in the disturbing direction. To mitigate this one has to also consider the earth pressure on the excavation side. As the passive pressure acts coherently with the active pressure, it is reasonable to consider that the net of these pressures is responsible for the add­ itional capacity. 4 FORMULATION OF THE ‘WHAT YOU DESIGN IS WHAT YOU GET’ WYDIWYG METHOD

Figure 3. Lateral earth distributions on the retained side (a and b) and on the excavation side (c & d) (After Potts & Fourie, 1984).

increased. Pott & Fourie (1984) presented the lateral earth pressure distributions of a 20m sheetpile wall for a range of excavation depths. The pressure distri­ butions are partly reproduced in Figure 3. Of interest is the excavation at 13.26m. Although the example had not calculated the corresponding crit­ ical embedment depth for such excavation depth one could expect it to be shallower than 20m, and the shapes of the pressure distributions on the wall would be similar to that at 15.26m, line 4 Figure 3, which was at critical equilibrium for the wall described in the paper and the lateral earth pressures were at limiting states on both sides of the wall. The exception is for a local area near the prop. For the 13.26m excavation with wall toe at 20m, line 3 Figure 3, the wall is stable with FoS equals to two. The pressure distribution on the retained side is close to that at critical equilibrium roughly above the critical equilibrium depth. More sig­ nificant pressure increases over the active limit occur below such depth. A similar observation could be made on the excavated side except the earth pressure decreases approximately below the critical equilibrium depth.

The above discussion of lateral pressure distribu­ tions on a propped cantilever wall is for a stable condition. For stability the pressure distribution is idealised to be at their limiting conditions. As such one would appreciate that the distributions above the critical depth would also be at limiting conditions. Here the critical depth is the depth at which the wall is at critical equilibrium. Section 3 demonstrates that the earth pressures reach their limiting states at this depth, i.e. active limit on the retained side and passive limit on the excavation side, and the earth pressure on the retained side to this critical depth is the disturb­ ing earth pressure on the wall. These observations constitute the first two assumptions below. Inevit­ ably earth pressures continue beyond the critical depth, but as discussed, such pressure on the retained side is not considered contributing to wall disturbance. It acts coherently with the pas­ sive pressure as a group to provide the observed stability, and the net effect of them contributes to the improved wall stability. This forms the 3rd assumption. The maximum reserve capacity that the ground can provide for stability occurs when the earth pressures on the extra embedment beyond the critical depth are at their respective pressure limits. On the basis of this, the assumptions for establish­ ment of the WYDIWYG model, shown in Figure 4 for stability assessment, are summarized as: • The disturbing load on the active side of the wall extends below the excavation level but only to the depth at which the wall is at critical equilibrium. • The earth pressure on the passive side remains at passive limit above the critical depth. • The embedded wall beyond the critical depth pro­ vides a ‘reserve’ capacity to stability and such

440

Using the conventional overturning factor of safety definition,

Since Mrestoring = Mpcr + Mr, and Mdisturbing = Macr = Mpcr The factor of safety can be re-written as:

For a general case in layered materials it may be shown that Figure 4. The WYDIWYG model.

capacity is derived from the ‘net’ limiting earth pressures from either side of the wall.

where the summation is of all layers of materials; i layers to the critical depth and j layers to the toe of the wall; the apostrophe indicates the effective moment capacity Due to constraint of this paper, example calcu­ lation is not included but it suffices to say that calculation of the embedment depth to a given FoS may be accomplished in a two-step process. The critical embedment depth is first calculated and from there the critical restoring capacity is determined. Then this is followed by proportion­ ing the additional wall embedment to give the required reserve capacity using equations (3) and (4) or (5).

Where: Kaa is the active earth pressure on the retained side of the wall Kpp is the passive earth pressure on the excavated side At the design embedment the wall is at stable equilibrium, the restoring moment capacity from the excavation side must be larger than that from the retained side. Hence take moment about the support, the reserve moment capacity, Mr, is:

where: Mp is the total moment on the excavation side Ma is the total moment on the retained side Expanding Equation (1) as follows and re-group:

5 DISCUSSION ON THE CHARACTERISTICS OF THE MODEL

where: Mpcr = the moment above the critical depth on the excavated side Mpp = the moment below the critical depth on the excavated side Macr = the moment above the critical depth on the retained side Maa = the moment below the critical depth on the retained side At critical depth the wall is in equilibrium with Mpcr = Macr, therefore Equation (2) may be simpli­ fied as follows:

It may be apparent from the simple derivation above that the FoS on geotechnical stability is made a function of the restoring moment capacity at critical equilibrium. In another words the stabil­ ity calculation is using the critical equilibrium con­ dition at the design excavation depth as a unique measurement reference. Given this, one can dir­ ectly associate the margin of stability included in the design and what you designed is what you get. For example if the design factor of safety is two then the design stability is two times what is avail­ able at critical equilibrium. A stronger foundation material offers higher lateral resistance and hence requires a shallower embedment and vice versa. Where soil layering occurs, the model can take into account of the changes in resistance and adjust the embedment corresponding to the

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Figure 5. Parametric studies - effective stress.

Figure 6. Parametric studies – total stress.

required restoring capacity. One may notice that the design margin is maintained constant across the material strength spectrum, whether it is for total or effective stress analysis. This is different from some of the methods where the factors need to be adjusted in response to change of soil condi­ tions. The parametric studies of the examples used for Table 1 are shown in Figures 5 and 6. It may not be a fair comparison of the methods as the FoSs considered are somewhat different between methods and this echoes on the issue of choices for appropriate FoS values for design discussed earlier in the paper. The plots serve to confirm a unique characteristic of the proposed method that it is consistent under different soil conditions and over a range of material strengths. 6 DISCUSSION AND SUMMARY The continued urban development in Sydney trig­ gers increased demand for excavation supports for infrastructure development. Some of the existing stability assessment methods are fraught with

problems and may not be applicable to the Austra­ lian Standard design requirements. A review of the existing design methodologies indicates that there is a need to differentiate between earth pressures and loads and this has led to the formulation of the WYDIWYG method which overcomes the issues associated with stability designs using the existing models. The role of the extra wall embedment beyond the critical depth has not previously been analysed. The fact that it is solely responsible for stability suggests that the earth pressures over this part of the wall are neither only a load nor only a resistance. This is cru­ cial in the development of the model. At stable con­ dition the change in pressures from their limiting states indicates that the increased embedment brings about a reserve in restoring stability, i.e. the wall is able to take on additional load without failure. The wall reaches the restoring capacity when these pres­ sures are returned to their limits. Since the limiting pressures are acting coherently it is reasonable to consider that their net pressure on this part of the wall is responsible for this reserve capacity. The model also introduces a unique way to calculate the FoS by measuring it against the critical condition at the design excavation depth. It may be shown that the formulated stability design is suitable for con­ forming design to the Australian Standard AS 5100 Bridge design. It is worthwhile to indicate that the aforementioned pressure distribution approach is used for stability calculations and the wall strength also needs to be checked to withstand for the ideal­ ized pressure conditions. For other wall load and deformation calculations, e.g. bending moments and shear forces, the conventional soil structure inter­ action earth pressure distributions should also be considered for more realistic results. For the WYDI­ WYG method, this process is usually undertaken after the final wall embedment is determined. Or it may be completed simultaneously if the method is codified within a suitable calculation program. Limited sensitivity studies on the proposed method are undertaken. The results demonstrate that the new formulation calculates consistently throughout the soil strength spectrum. The stability margin is constant against a given design requirement, whether the materials are medium dense or dense sand, firm or very stiff clay. The formulation is free of numerical problems. The calculation is stable, independent of the stress state of the soil parameters used. For asset owners and operators the proposed cal­ culation provides improved certainty and clarity to stability designs balancing between economics of construction and margin of stability in design. Using the WYDIWYG method, a FoS of two would ensure the restoring capacity is two times the capacity avail­ able at limit equilibrium. The measure of stability is hence tangible and comprehensible. Consistent assessment of stability and meaningful comparison of stability performance may be made.

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ACKNOWLEDGEMENT The author would like to thank Filomena Manalo for assistance on word processing, Transport for New South Wales for the permission to publish this paper. The opinions expressed in this paper are those of the author and do not necessarily represent those of Transport for New South Wales.

REFERENCES Australian Standard AS 5100:2017. Bridge design. Stand­ ards Australia. BSC Piling Handbook 1988, 6th Edition. British Steel Corporation. Burland, J.B., Potts, D.M. & Walsh, N.M. 1981. The over­ all stability of free and propped embedded cantilever retaining walls, Ground Engineering, July 1981

Civil Engineering Code of Practice No.2 1951. Earth Retaining Structures. Institution of Structural Engineers Potts, D.M. & Fourie, A.B. 1984. The behaviour of a propped retaining wall: results of a numerical experiment. Geotechnique 34(3), 383–404. Potts, D.M. & Fourie, A.B. 1985. The effect of wall stiff­ ness on the behaviour of a propped retaining wall. Geotechnique 35(3), 347–352. Simpson, B. & Powrie, W. 2001. Embedded retaining walls: theory, practice and understanding. Perspective Lecture for 15th International Conference on Soil Mech­ anics and Geotechnical Engineering, Istanbul WALLAP, Geotechnical design software – Earth retaining structures by Geosolve. Yuen, C. K. S. 2019. What you design is what you get – An economical way to the stability design of propped canti­ lever walls. 13th Australia New Zealand Conference on Geomechanics, Perth.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Multi-objective optimisation design for composite tunnel linings using non-dominated sorting genetic algorithm W. Zhai Department of Geotechnical Engineering, Tongji University, Shanghai, China Department of Civil Engineering, University of Birmingham, Birmingham, UK

D. Chapman & A. Faramarzi Department of Civil Engineering, University of Birmingham, Birmingham, UK

H. Huang & D. Zhang Department of Geotechnical Engineering, Tongji University, Shanghai, China

ABSTRACT: A multi-objective optimisation design method for composite tunnel linings is proposed in this paper. The geometry and material parameters of the inner and outer components of tunnel linings are treated as design variables, while the lining safety and cost value are considered as design objectives. Such a multiobjective optimisation problem is solved by integrating a computational model into the non-dominated sorting genetic algorithm. The solutions are obtained as a set of optimised design points, named the Pareto front, and all the designs from this set provide a trade-off relationship between the design objectives. In addition, the concept of a knee point is employed to provide the most recommended design which can be located on the Pareto front using the bending angle approach. Subsequently, a detailed case study is presented to demonstrate the application and advantages of the proposed method. It is shown that the proposed optimisation approach can help to achieve simultaneously a higher safety level and better cost efficiency for the design of composite tunnel linings.

1 INTRODUCTION Over the past few decades, a huge number of tunnels have been constructed using tunnel boring machines (TBMs) all around the world. Assembled by precast reinforced concrete (RC) segments, the linings of these shield-driven tunnels can be quickly con­ structed and often work as a permanent supporting system. However, due to complex geological condi­ tions and wide range possible external loads, a variety of defects have been found in many operat­ ing tunnels. To solve this problem, composite tunnel lining systems are often adopted in many projects to achieve a higher support capacity and also to provide improved water proofing. For examples, linings composed of an outer segmental lining and an inner cast-in-situ concrete layer have been applied as the main type in recent years by Austrian Railways ÖBB (Strappler et al., 2012). Linings composed of an inner steel lining and an outer back-fill concrete lining are widely used for water distribution pressure tunnels (Hachem & Schleiss, 2011, Pachoud & Schleiss, 2015). Retrofitting damaged segmental tunnel linings by constructing a secondary steel or

RC lining within the existing tunnel to form a new composite lining system has also been applied in many tunnel repair projects (Chang et al., 2001, Kir­ iyama et al., 2005, Van Empel et al., 2006, Shao et al., 2016). Although the composite lining has been widely used in industry, there is still room for improvement on design aspects. A current design procedure for composite tunnel linings has been proposed in the design guidelines published by ITA (2000). Accord­ ing to this guideline, first, the member forces of lining components are calculated using the bedded frame model or elastic equation method. And then, a given design is considered to be acceptable if it can pass the check of section safety. One of the main disadvantages in this design approach is that the given design to be checked is predetermined by the designer according to the engineering characteristics and professional experience. This might work well for the case of single layer lining with few design variables. However, as the number of design vari­ ables increases in the case of composite linings, there would potentially be other designs satisfying the safety check but with lower costs.

DOI: 10.1201/9780429321559-58

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Multi-objective optimisation tools provide an alter­ native approach to solve such problems. There has been an increase use of this approach in engineering fields (Marler & Arora, 2004, Tang et al., 2019). Unlike a single-objective optimisation problem, which has only one optimal solution, a multi-objective opti­ misation problem yields a set of optimal solutions, namely, the Pareto front. This set of Pareto solutions provides a trade-off relationship between conflicting design objectives, among which none of them is better than any other ones in terms of all the objectives. In some cases, if the design variables are fewer and objective functions are less nonlinear, some con­ ventional optimisation methods or even an exhaustion method can be applied. These conventional optimisa­ tion methods achieve solutions mainly based on algo­ rithm that convert multi-objective problems to singleobjective problems by emphasising just one particular objective at one time. However, once a larger number of design variables are involved and the objective functions become more nonlinear, these methods sometimes introduce unacceptable computation times and don’t perform well with respect to robustness (Deb et al., 2002). Over recent decades, genetic algo­ rithms (GAs) have shown strong capability in solving such multi-objective optimisation problems efficiently and precisely. GA optimisation methods don’t require gradient information and are effective regardless of the characteristics of the objective functions (Marler & Arora, 2004). There are various GA-based multiobjective methods, among which the non-dominated sorting genetic algorithm (NSGA-2) proposed by Deb et al. (2002) is widely adopted for many engin­ eering problems for it efficiency and robustness (VoDuy et al., 2017). In this paper, an analytical model to analyse com­ posite tunnel linings is proposed based on the theory of elasticity, where the lining response can be obtained using different design parameters as inputs. Subse­ quently, the proposed analytical model is integrated with the NSGA-2 to form a coupled optimisation tool, and hence obtain a Pareto front. Finally, the proposed optimization design method is illustrated by a case study, and its advantages are shown by comparison between the Pareto designs and other feasible designs. 2 ANALYTICAL MODEL FOR COMPOSITE TUNNEL LINING DESIGN Assuming plane strain condition and linear elastic materials, an analytical model is proposed to com­ pute internal forces within the composite tunnel lining system. A schematic of the model is illustrated in Figure 1, the ground rock or soil is treated as an infinite plane with a circular hole, while the RC lin­ ings are treated as two thick-walled cylinders. The lining dimensions are indicated by inner section radius R3, inner lining thickness til and outer lining

Figure 1. An analytical model for composite tunnel lining.

thickness tol. Using the theory of elasticity, force and displacement of the inner and outer linings can be evaluated with given boundary and compatibility conditions. A general stress function for plane stress and strain problems in the polar coordinate was given by Timoshenko et al. (1970) in the form of the Airy function as Equation (1).

Where r and q are the coordinates, the constants an, bn, cn, etc. are unknown parameters to be solved by satisfying the equilibrium equations, compatibility con­ ditions and boundary conditions for a certain problem. By using this Airy function, Bobet (2001) and Park (2004) presented elastic solutions for the ground response with a circular tunnel excavation. The stress components of the ground in Figure 1 are given by Equation (2):

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where sr g, sqg are stress in the ground along radial and circumferential direction, is the shear stress. From Equations (2), the corresponding strain components can be derived using Hooke’s law in the plane strain condition. Subsequently, the dis­ placement components can be obtained by inte­ grating the strain-displacement relationship as shown in Equations (3).

where ur g and uqg are the displacement of ground in radial and circumferential directions, Eg is the elastic modulus of ground, vg is the Poisson’s ratio of the ground, gg is the unit weight of the ground, k is the coefficient of lateral earth pressure, h is the distance from the tunnel centre to the ground sur­ face, and a0g, c1g, d1g, etc. are the unknown con­ stants to be solved. Based on the elastic solutions for a thickwalled cylinder (Timoshenko et al., 1970), stress and displacement components within outer lining are given by Equation (4) and (5) respectively:

where Eol and vol are the equivalent elastic modulus and Poisson’s ratio of the outer lining, and a0°l, b0°l, c1°l, d1°l etc. are unknown constants to be solved. Here, the equivalent elastic modulus of the RC lining is calculated according to the concrete grade (Col) and steel reinforcement ratio (rol). Similarly, the stress and displacement components for the inner lining can be expressed by changing the superscript or subscript in Equations (4) and (5). In this paper, the interface between the ground and the outer lining and the interface between the outer and inner lining are assumed to be frictionless and in con­ tact. Thus, following boundary conditions are applied: At interface r=R1:

At interface r=R2:

At interface r=R3: Thus, all the constants in Equation (2)-(5) can be solved and the stress and displacement fields are obtained. In the optimisation design process, the axial force and moment of the lining sections are the main consideration with respect to the safety level of tunnel in terms of lining bearing capacity. At any section with angleθ, the axial force and section moment in the inner lining can be calculated by Equations (6), while those in the outer lining can be calculated by Equations (7).

Þ

446

According to the design code for RC structures tunnel lining (2008), the bearing capacity [N] of eccentric compression RC lining sections is deter­ mined by Equation (8). Subsequently, the safety factor of each lining can be estimated by ratio of [N] to N, and safety factor of the composite lining system is adopted using Equation (9). Figure 2. Formula for the multi-objective optimisation design for composite tunnel linings.

where [N] is the capacity of the RC beam given cross section parameters, fc and fy’ are compressive strengthen of concrete and steel bar, As and As’ are the steel section area in the compression and tension area, h, b, x, as are dimension parameters. The cost value is evaluated by equation (10) according to the volume and price of different materials in the lining construction.

where Vc and Vs are the total volumes of concrete and steel, Cc and Cs are unit price of concrete and steel determined from project conditions. 3 MULTI-OBJECTIVE OPTIMIZATION DESIGN METHOD The design process for the composite tunnel linings, presented in Section 2, can be formulated as a multiobjective optimisation formula as shown in Figure 2. It contains six different design variables and two design objectives. Initially, a parent population (P0) is generated, which is actually a matrix composed of a variable matrix VP with size of pop×n and an objective matrix OP with a size of pop×m. VP is generated ran­ domly within the domains of each design variables, and then OP can be obtained using the proposed computation model.

Subsequently, an offspring population Q0 is gen­ erated based on the obtained parent matrix P0, which is also composed of a variable matrix VQ with a size of pop×n and an objective matrix OQ with a size of pop×m. VQ is created from VP using a crossover and mutation operator (Deb, 2001), and then the corres­ ponding objective OQ is obtained by the proposed computation model. Thus, a combined population R0=P0∪Q0 contain­ ing 2×pop row vectors is formed, each vector repre­ sents a design point. These vectors are sorted by performing a non-dominated check on every design point (Deb et al., 2002). All non-dominated design points are marked as rank one, denoted as F1 in Figure 3. Then, non-dominated design points relative to the rest of design point within R0 except F1 are marked as rank two, denoted as F2. This process is repeated until all design points are ranked. After sorting, the population size is reduced to pop by just delivering the design points with the higher ranks into the next parent population P2. The diversity of non-dominated solutions is guaranteed during this procedure. Considering Fl is the last nondominated set after which no other sets could be accommodated in P2, all the design points within Fl are sorted in descending order according to crowd­ ing-distance comparison principle (Deb et al., 2002), and only those falling in the range of P2 are kept. Finally, the new parent population obtained is delivered into the next round of evolution. This loop continues until i=gen. All the design points within the final population Pi are desired Pareto solutions and can be illustrated in the objective space as a Pareto front. 4 CASE STUDY In this section, the proposed multi-objective opti­ misation design method for composite lining is illus­ trated by applying it to an actual tunnel project. This tunnel is from Shaxi to Luotian, which is a section of the Water Allocation Project in Pearl River Delta, Guangdong, China. It is a 2.4km long shield-driven tunnel with a double-layer composite lining system constructed in a strong or fully weathered rock mass.

447

Figure 3. Illustration of the non-dominated sorting genetic algorithm (modified from Deb et al. (2002)).

The outer is a precast RC segmental lining and the inner section is a cast-in-situ RC linings. Based on the engineering geological investigation and initial planning, some fundamental parameters for the tunnel lining design are listed in Table 1. The concrete grades, steel reinforcement ratio and lining thickness of both the inner and outer linings are considered as design variables to be optimised. All these variables are constrained by predefined design domains in terms of the lower and upper bounds, as shown in Table 2. As shown in Figure 4, the current populations are represented by green dots, while five thousand feas­ ible designs are randomly generated and presented as light grey dots for illustration purpose. At the initial stage, gen=0, the initial populations are widely spread over the objective space. By

Table 1. Design parameters for the surrounding rock and tunnel profile of the Shaxi-Luotian tunnel project.

Surrounding rock

Tunnel

Parameter

Symbol Value Unit

Elastic modulus Poisson ratio Unit weight Coefficient of earth pressure Depth Inner water pressure TBM excavation radius

Eg νg γg

2.5 0.3 20

GPa kN/m3

k

0.5

-

h

28

m

pw

0.2

MPa

R1

4.25

m

Table 2. linings.

Design variables for tunnel inner and outer Parameter

Concrete grades Cast-in-situ RC Steel ratio inner lining Thickness Concrete grades Precast RC Steel ratio outer lining Thickness

Symbol Domain Unit Cil ρ il t il Col ρol t ol

10~35 0.5~4 30~70 40~60 0.5~4 30~50

­ % cm % cm

generation gen=3, the populations are starting to be optimised and moving towards a boundary, which is the Pareto front. By generation gen=10, the popula­ tions have gathered close to the Pareto front, but are quite crowded. From generation gen=20 to 100, it can be observed that the populations are getting more and more dispersed along the Pareto front. By generation gen=100, the new generated front are smooth and well-dispersed along this boundary, and this can therefore be adopted as the Pareto front in this case. The Pareto front obtained is shown in Figure 5. All the points on this Pareto front are opti­ mised solutions, since none of them is superior to others with respect to all the objectives. They provide a trade-off relationship for the designers. A final design can be easily determined from the Pareto front, with a given preference in terms of desired cost or safety level. Although the absolute best design doesn’t exist in such optimised results, there is usually a knee point

448

Figure 5. Pareto front and comparison of different design pints.

Figure 4. Evolutionary process at different generations: (a) gen=0; (b) gen=3; (c) gen=10; (d) gen=20; (e) gen=50; (f) gen=100.

presented on the Pareto front. The knee point is defined as the most preferred solution within the obtained Pareto front, since it requires an unfavor­ able sacrifice in one objective to achieve a little improvement in another objective (Deb & Gupta, 2011). Therefore, if no specific preference is offered, the recommended design can be provided with help of the concept of the knee point. The bending angle approach proposed by Deb & Gupta (2011) is used in this paper to locate the knee point on the obtained Pareto front. As shown in Figure 5, the knee point is noted as a .red triangle, which is actually located at the pos­ ition reflecting a change in deflection on the Pareto front. In addition, a feasible design point within the

Table 3.

design domain and two other design points from the Pareto front are highlighted for comparison pur­ poses, and their specific values are listed in Table 3. The feasible design point has a safety factor of 3.50 and a cost of 3.22×104 RMB. Comparing this feas­ ible design with the Pareto design-1, the safety factor is increased by 52.3% to 5.33 with almost the same cost volume. Comparing the feasible design with the Pareto design-2, the cost is reduced by 54.3% to 1.47×104 RMB with almost the same level of safety. Therefore, it is obvious that the knee point presents a good compromise considering both the lining cost and safety. In addition, it is noted that although the case shown in this article is a tunnel with a segmental outer lining, the proposed multi-objective optimisa­ tion design methodology could be extended for com­ posite tunnel linings constructed using sprayed concrete. This could be achieved by substituting the computational model in this paper by another com­ putational model (either a numerical or an alternative analytical solution, as sprayed concrete lined tunnel tend to be non-circle, otherwise the analytical model presented in this paper could be used), with different design variables appropriate for sprayed concrete linings.

Comparison between different design points.

Design Points

t il (mm)

Cil (C)

ρ il (‰)

t il (mm)

Cil (C)

ρ il (‰)

SF

Cost (RMB)

A feasible design point Pareto design point-1 Pareto design point-2 Knee point

50 70 51 70

20 30 30 30

5 5 5 5

50 5 30 50

60 57 40 40

25 15.5 5.5 7

3.50 5.33 3.51 4.97

3.22 3.25 1.47 2.56

449

CONCLUSIONS A multi-objective optimisation design method for com­ posite tunnel linings using the NSGA-2 approach is proposed in this paper. First, the analytical computa­ tional model is derived for the calculation of composite lining performance, based on which the tunnel safety can be evaluated. Subsequently, the bi-objective opti­ misation formula for composite lining design is formed with both safety level and cost considered as design objectives. The Pareto front can be obtained using the NSGA-2 approach and the proposed computational model. Finally, a case study is presented for a better understanding of the method and to demonstrate the proposed optimisation design method for composite tunnel linings. The following conclusions can be drawn: (1) The multi-objective optimisation design model for composite tunnel linings is formulated, with consideration of the cost and safety level as the design objectives. Such problem can be effi­ ciently solved by integrating the NSGA-2 approach and the analytical computational model. (2) The Pareto front presents all the optimised solu­ tions, which provides a trade-off relationship between the two design objectives. An appropri­ ate design can be obtained with a given prefer­ ence in either cost or safety level. In addition, the recommended design on the Pareto front can be located with the help of concept of the knee point. (3) A comparison between the Pareto designs and a ‘feasible’ design shows the benefits of using proposed optimisation design method, as a higher level of lining safety can be achieved with a better cost efficiency

ACKNOWLEDGEMENT The authors gratefully acknowledge the support of Chinese National Science Foundation Committee

Program (No. 51538009), and the Engineering and Physical Sciences Research Council via the Self Repairing Cities grant (EP/N010523).

REFERENCES Bobet, A. 2001. Analytical solutions for shallow tunnels in saturated ground. Journal of Engineering MechanicsAsce, 127, 1258–1266. Chang, C.T., Wang, M.J., Chang, C.T. & Sun, C.W. 2001. Repair of displaced shield tunnel of the Taipei rapid transit system. Tunnelling and Underground Space Technology, 16, 167–173. Deb, K. 2001. Multi-objective optimization using evolution­ ary algorithms, John Wiley & Sons. Deb, K. & Gupta, S. 2011. Understanding knee points in bicriteria problems and their implications as preferred solution principles. Engineering Optimization, 43, 1175–1204. Deb, K., Pratap, A., Agarwal, S. & Meyarivan, T. 2002. A fast and elitist multiobjective genetic algorithm: NSGA-II. IEEE transactions on evolutionary computa­ tion, 6, 182–197. Hachem, F. E. & Schleiss, A. J. 2011. A review of wave celerity in frictionless and axisymmetrical steel-lined pressure tunnels. Journal of Fluids and Structures, 27, 311–328. ITA, I. T. A. 2000. Guidelines for the design of shield tunnel lining. Tunnelling and Underground Space Tech­ nology, 15, 303–331. Kiriyama, K., Kakizaki, M., Takabayashi, T., Hirosawa, N., Takeuchi, T., Hajohta, H., Yano, Y. & Imafuku, K. 2005. Structure and construction examples of tunnel reinforce­ ment method using thin steel panels. Nippon Steel Tech­ nical Report, 92, 45–50. Marler, R. T. & Arora, J. S. 2004. Survey of multi-objective optimization methods for engineering. Structural and Multidisciplinary Optimization, 26, 369–395. Pachoud, A. J. & Schleiss, A. J. 2015. Stresses and Dis­ placements in Steel-Lined Pressure Tunnels and Shafts in Anisotropic Rock Under Quasi-Static Internal Water Pressure. Rock Mechanics and Rock Engineering, 49, 1263–1287. Park, K. H. 2004. Elastic Solution for Tunneling-Induced Ground Movements in Clays. International Journal of Geomechanics, 4, 310–318.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Support pressure transfer at a slurry supported tunnel face due to time dependent decrease of soil permeability C. Zhao School of Civil Engineering, Sun Yat-sen University, China

Z. Zizka Metroprojekt Praha a.s., Prague, Czech Republic

B. Schoesser & M. Thewes Department of Civil and Environmental Engineering, Institute for Tunnelling and Construction Management, Ruhr University Bochum, Germany

A.A. Lavasan Department of Civil and Environmental Engineering, Chair of Soil Mechanics, Foundation Engineering, and Environmental Geotechnics, Ruhr University Bochum, Germany

ABSTRACT: Within mechanized tunnelling, the excavation chamber of a slurry pressure balanced (SPB) shield is filled with the bentonite suspension as support medium. A compressed air cushion is applied to gen­ erate a hydrostatic fluid pressure, which corresponds to the support pressure required for the stability of the tunnel face due to earth pressure and groundwater pressure. In reference to the support pressure, the bentonite suspension penetrates into the grain structure and transfers the support pressure onto the soil. In the case of bentonite suspensions, the hydrostatic support pressure is transferred to the grain skeleton time-dependent if a zone is formed at the surface or to a certain penetration depth in the near-surface area of the soil whose permeability is smaller than that of the soil itself. In this zone, the pressure difference between the support pressure and the pressure of the soil to be stabilised is converted into an effective stress acting on the grain structure. The formation of a zone of low water permeability in the soil depends on the penetration behaviour of the support medium. This paper presents the experimental results of advanced column tests, where the decrease of the soil permeability due to the penetration of bentonite suspension is measured over time for various combin­ ations of soil and bentonite suspensions. Boundary conditions of the soil are described as particle size dis­ tribution, bulk density and initial permeability according to Darcy’s law, bentonite suspension is rheologically characterized in terms of yield point and viscosity. In order to examine the impact of the permeability evolutions in front of the slurry shield TBM, numerical investigation is carried out. The finite element numerical model is calibrated and validated in accordance with physical and rheological observa­ tions along with laboratory measurements from the advanced columns test under controlled hydromechanical boundary conditions. After verification of the numerical 2D model, it is extended to a full scaled mechanized tunnelling 3D problem that to analyse the hydraulic flows and effective stress vari­ ations in the soil in front of the TBM and their influence on the face pressure transfer process from the TBM to the soil skeleton.

1 FACE SUPPORT OF SLURRY PRESSURE BALANCED (SPB) SHIELDS Among the mechanized tunneling machines, slurry pressure balance (SPB) shields are known as a reliable excavation technology in non-cohesive soils under groundwater table and unstable tunnel face conditions. Utilizing pressurized bentonite sus­ pensions (slurries) as support medium in the

excavation chamber, the main advantage of a SPB shield is the exact control of the support pressure. In such conditions, the tunnel face is actively supported preventing surface deformations and other ground movements. The SPB shield is a system for driving with full face excavation and liquid-supported working face. The ground is loosened by the cutting wheel rotating in the excavation chamber, which is completely filled

DOI: 10.1201/9780429321559-59

451

with liquid (Figure 1). The soil material is mixed with the liquid and pumped off through a delivery line. At the working face, the bentonite suspension interacts with the soil. Via the excess pressure of bentonite sus­ pension in the excavation chamber, above the hydro­ static level, the bentonite suspension penetrates into the surrounding soil. During the penetration process, reac­ tion forces are generated in the soil: Penetration of the suspension into the soil creates a flow force in the pore space. By filtering out bentonite particles at the entrance surface to the grain skeleton, a membrane with a planar force transmission is built up. The yield point of the bentonite suspension generates shear stres­ ses at the grain surfaces. The pore water pressure increases due to the reduced permeability of the soil. The sum of the proportions of the reaction forces cor­ responds to the applied supporting pressure. To achieve a stabilised tunnel face, two key condi­ tions must be fulfilled. First, a sufficiently large slurry pressure in the excavation chamber of the shield has to be applied with reference to the acting boundary condi­ tions. The required pressure in the excavation chamber can be determined by various methods, (e.g. Anagnos­ tou & Kovari (1994), Jancsecz & Steiner (1994)) or following the recommendation of the German Tunnelling Committee for the calculation of face support pres­ sure (Zizka & Thewes 2016). Second, the slurry excess pressure in terms of the difference between the slurry pressure and the groundwater pressure must be trans­ ferred at the soil skeleton to counteract the earth pres­ sure. In practice, the German standard DIN DIN4126 (2013)is often used to predict the support pressure transfer at the soil skeleton. As summarized by Zizka, Schoesser, Thewes, & Schanz (2018), Morgenstern & Amir-Tahmasseb (1965) conducted one of the first attempts to analyse the interaction between bentonite slurry and soil for the purposes of open trench stabilization. The authors pointed out that at the time of publishing their paper, several mechanisms such as hydrostatic pressure, arch­ ing of the soil and electro-osmotic forces were dis­ cussed as the main mechanisms responsible for the slurry support of non-cohesive open trenches. Weiss (1967) performed laboratory experiments dealing with

Figure 1. Principle of Slurry Pressure Balanced (SPB)

Shield.

(source: Herrenknecht AG)

stabilization of non-cohesive soils by bentonite slurry. In his experiments, he visualized the penetration behaviour of slurry into the pores of soil and suggested that the stabilisation behaviour depends on the equiva­ lent pore diameter of soil and the yield point of the slurry. Weiss (1967) summarizes the stability condition of a trench in non-cohesive soils as three complemen­ tary conditions: • yield point of slurry is required to achieve the equilibrium of forces on a single soil grain • slurry pressure must exceed the groundwater pressure • slurry excess pressure must counterbalance the earth pressure Further research on the stability of slurry-sta­ bilized trenches was conducted by MuellerKirchenbauer (1972). The author performed both, experiments and theoretical analysis. He stated that penetration of slurry in the pores of soil influences the slurry pressure transfer on the soil skeleton. The penetration behaviour of slurry is determined by the grain size distribu­ tion of soil and the stagnation gradient of slurry. He distinguishes two cases of slurry-soil inter­ action (Figure 2). The first is the formation of a filter cake at the soil’s surface (Type I) and the second is slurry penetration inside soils pores without any aggregation of slurry particles at the surface (Type II). In Type II, the shear resistance of the pore channel wall is activated. The shear resistance is in equilibrium with the excess hydraulic head of the slurry in the trench at the end of the penetration process. The pres­ sure transfer Type II consist of both, penetration zone and the filter cake (Zizka, Schoesser, Thewes, & Schanz 2018). For tunnelling in coarse and permeable soils, the pressure transfer according to type II is the applic­ able principle (Figure 2). Here, the bentonite suspen­ sion penetrates into the ground. The movement is induced by the acting pressure and leads to the inter­ action between the suspension and the surface of the soil particles. Depending on the flow velocity, a certain shear rate acts within the suspension. At the contact areas, shear stresses of the magnitude of the yield point of the suspension are transferred to the soil particles. When the penetration depth has become so large that the integral of the transferred shear stresses is in equilibrium with the difference between the suspension pressure and the prevailing earth pressure, the penetration process stagnates. For the design of a bentonite suspension, the rheo­ logical parameters need to be adapted to the boundar­ ies of the in-situ soil. Here, the relevant geological characteristics of non-cohesive soils cover the descrip­ tion of the pore space in terms of porosity, permeabil­ ity, compactness and effective or characteristic grain size d10 gained from the particle size distribution.

452

Figure 2. Different types of pressure transfer in diaphragm wall stabilization according to Mueller-Kirchenbauer (1972) and Zizka et al. (2018).

2 EXPERIMENTAL INVESTIGATIONS IN SOIL COLUMN Figure 3. Experimental set-up of soil column including levels of measurement [Zizka (2019)].

The set-up can be designated as column test and con­ sists of a slurry cylinder, a soil cylinder (internal diameter 40cm) and a reservoir with free surface for the discharged fluid from the soil cylinder. The dis­ charge reservoir is located on a scale. The scale is connected to a computer for continuous data logging during the experiment. One pore pressure sensor (PWD) continuously monitors pressures in the slurry cylinder and 7 other sensors are placed in the soil cylinder. Two total stress sensors are located at two measurements levels in the middle of the soil cylinder (Figure 3). Data from both sensor types are transferred to the computer. The data are logged every 0.25s. 2.1

Another requirement for obtaining reprodu­ cible results of experiments involving bentonite slurry is the monitoring of its physical and rheo­ logical properties. Before every experiment, the slurry was tested according to DIN4127 (2014) and API13B (2014). The used methods for determination of rheological properties were ball harp according to von Soos (static yield point), Marsh funnel (Marsh time) and direct-reading viscometer (apparent viscosity, plastic viscosity and Bingham yield point). The resulting average values of slurry characteristics are outlined in Table 2.

Test programme

The experimental program was designed with the aim of investigating the deep slurry penetration. Simultaneously, the slurry had to stagnate before the second measuring level in the experimental device. Another requirement is that relatively high injection pressures representing the slurry excess pressures are used. This is due to easier tracking of soil reaction during slurry penetration. Finally, one slurry concentration with concentration of 60 kg/ m3 ), one soil fraction (grain size 1–2mm) and level of injection pressure (40 kPa pressure drop over the sample) were chosen based on preliminary tests. The chosen injection pressure corresponds to a realistic range of slurry excess pressure at a real slurry shield. Three experimental runs were conducted to assure the reproducibility of the results. 2.2

Materials

A coarse, uniformly graded sand of fraction 1-2mm with relatively high characteristic grain size (d10 ) was chosen for the investigation basis (Table 1). The installation and compaction of the soil in the cylinder was controlled to obtain highly similar porosities among all investigated samples.

Table 1 .

Properties of soils used in experimental studies.

Soil fraction [mm]

1.00 - 2.00

Density [g/m3 ] d10 [mm] Porosity [-] Compaction ration [-] Permeability kf [m/s]

1.57/(1.59) 1.15 0.41 0.63/(0.72) (5-11)*10-3

Table 2 . studies.

Properties of suspensions used in experimental

Suspension

60 [kg/m3 ]

Density [g/m3 ] Yield point (static) [Pa] pH value [-] Marsh time (1l) [s] AP Viscosity [mPa*s] PL Viscosity [mPa*s] Bingham YP [Pa]

1.031-1.036 58-59 9.4-9.9 49-52 19.1 2.9 15.5

453

3 EXPERIMENTAL RESULTS AND DISCUSSION The time-dependent slurry penetration depth was evaluated first (see Figure 4). The original intention was to assess the timedependent slurry penetration depth in the soil sample visually. However, this turned out to be difficult due to very shallow penetration depths. Therefore, only the final penetration depth was determined directly visually during disassembling of the set-up and sub­ sequently, the time-dependent penetration was obtained based on scaling of the final penetration by the time-dependent volume discharge of the fluid from the experiment measured by the scale. Figure 5. Time dependent slurry penetration – comparative values according to Krause (1987) and Zizka (2019).

where lðtÞ is the time dependent slurry penetration depth (m), lmax is the final penetration depth (m), VðtÞ is the volume of discharged fluid at time t (m3 ) and Vmax is the volume of discharged fluid at the end of experiment (m3 ). If the slurry penetration depth were to only be defined by the volume of discharged water and the porosity of compacted soil, an inaccuracy will be included. This inaccuracy was mentioned by Krause (1987). He stated that using this approach, the fil­ trated water, i.e., the amount of slurry from which the slurry particles were filtered out, would unrealis­ tically increase the slurry penetration depth. Thus, using the combined approach of visual determination and outflow scaling, as described above, improves the accuracy of the results (Figure 5). Further, pore pressure distribution at various timespans from the beginning of penetration will be evaluated. This is shown for timespan t = 3s, t = 10s in Figure 6 and for timespan t = 40s and t = 60s in

Figure 4. Time dependent slurry penetration for different injection pressure - absolute values [Zizka (2019)].

Figure 6. Distribution of pore pressure in soil sample at different stages of slurry penetration depending on different injection pressure (timespan 3-10sec) [Zizka (2019)].

Figure 7. The distributions are scaled by the injec­ tion pressure in general. The only deviation to the scaling can be noticed for the timespan t=3s for com­ binations IV and VI, when the distributions are com­ parable. The encountered deviation from the trend cannot be reasonably explained. The regular scaling is observable also for the later stages of the slurry penetration in in Figure 7. For all combinations, approximate linear pore pressure distribution inside the slurry-penetrated zone could be measured. To compare experimental results with FE modelling, time-dependent permeability coefficient has to be evaluated. For this purpose, Darcys0 law was util­ ized. Theoretical background for this application of Darcy’s is discussed by Zizka (2019). The coefficient of permeability kf can be directly determined timedependently from experiments as an instantaneous

454

Figure 8. Comparison of the penentration depth [m] in ref­ erence to the pressure drop [bar] resulting from experi­ ments and theoretical calculations based on German Standard DIN4126 (2013).

Figure 7. Distribution of pore pressure in soil sample at different stages of slurry penetration depending on different injection pressure (timespan 40-60sec) [Zizka (2019)].

relationship between pressure gradient and discharge (Equation 1) within very small time-steps. The por­ osity of soil is assumed to be constant in Equation (1). Hence, the entire change of hydraulic resistance of soil during interaction with slurry is expressed by the permeability coefficient. The measured time dependent pressure difference is obtained by the pore pressure sensors form the experiment (PWD). Macroscopic flow path in the set-up is adopted to be 40cm. Moreover, the outflow from the set-up is dir­ ectly measured by the scale. The obtained timedependent permeability is shown in Section 4. The German standard DIN4126 (2013) offers a formula to predict the penetration depth of a bentonite suspension in a specific soil (Eq. 2). Here, the yield point of slurry, characteristic grain size of soil and the resulting slurry stagnation gradi­ ent for the penetration model are related. It is also possible to calculate slurry penetration while know­ ing the slurry stagnation gradient. In DIN4126 (2013) the empirical fitting factor (a) in Eq. (2) is set between 2 and 3.5. If the value a = 2 is chosen, the evaluation of the referenced experiments by Kilchert & Karstedt (1984) considers standard deviation of the results. If a = 3.5 is taken, it corresponds to the evaluation of referenced experimental results consid­ ering average values.

yield point of the slurry (Pa), Δs is the slurry excess pressure (kPa) and lmax;calc is the slurry penetration depth (m). Figure 8 shows the comparison of the calculated penetration depth according to Eq. (2) and the experimental results of the penetration depth in the soil column. Note that the difference between pres­ sure drops over the soil sample and injection pres­ sure is approx. 0.1 bar. As was expected, higher injection pressure caused a deeper slurry penetration. Additionally, the point showing the zero slurry pene­ tration for the zero pressure was added to the dia­ gram based on logical derivation. The diagram is showing the relationship between the slurry penetra­ tion depth and the pressure drop over the soil sample. It turned out that the slurry penetration depth is approx. linearly dependent on the pressure drop. The linear relationship in Figure 8 is indicating that the methodology for pressure steered repenetration as described in section 2.1 can be used for this slurry-soil combination without significant deviations from realistic re-penetration at the tunnel face with soil cutting. Moreover, Figure 8 shows that the interaction between slurry and soil in this com­ bination does not have to be assessed on the micro­ level. The linear dependence between pressure drop and penetration logically means that the retention of slurry particles in the pore space does not cause sig­ nificant change in the pores space geometry (bulk flow of slurry). This can be confirmed also by Figure 8 showing negligible dependency of the comparative slurry penetration depth development on the injec­ tion pressure. 4 2D NUMERICAL SIMULATION OF COLUMN TEST

where fso is the slurry stagnation gradient (kN/m2 / m), a is the empirical factor from experiments 2–3,5 (–), d10 is the characteristic grain size of soil (10 per­ cent passage in sieve analysis) (mm), τf ;s is the static

In order to simulate the aforementioned column test, 2D finite element (FE) analysis is conducted in this section via the commercial FE-code Plaxis. The

455

model geometry is shown in Figure 9. The bottom of model is fully fixed in both horizontal and vertical directions. For the left and right boundaries, only ver­ tical displacements are allowed while the vertical dis­ placement is restricted at the top surface of the model. For the hydraulic boundary conditions, only the top surface allows excess pore pressure dissipa­ tion. The water level is assumed at the model surface. According to the experimental description, the soil domain can be divided into two parts, lower penetration zone and upper non-penetration zone. In the present work, the penetration zone is modeled with time-dependent permeability, and the timedependent permeability is defined based on the

measured permeability as presented in Figure 10. It is worth mentioning that permeability of penetration zone is calculated based on the Darcy’s law and flow is assumed normal to the bedding planes (Sridharan & Prakash 2002):

where keq is the equivalent permeability of entire domain, n is the number of soil layer, ki and Li are permeability and thickness for each layer, respectively. Hardening Soil (HS) model is applied in the pre­ sent work to describe the soil behavior. The HS model parameters are presented in Table 3, which is derived based on the triaxial and oedometer tests conducted in Schoesser (2004). For more details on HS model, the reader is referred to Schanz, Vermeer, & Bonnier (1999). In order to model the face pressure at the beginning of the experiment, a dummy elastic layer with very high stiffness is adopted in the FE-model, water head is increased by 3.6 m in this layer to model the con­ stant 36 kPa excess pressure at the bottom the soil

Figure 10. Measurement permeability and assumed perme­ ability in FE-model.

Table 3 . Input parameters of the used soil constitutive model - the hardening soil model.

Figure 9. Geometry of the 2D axisymmetric model.

Parameter

Description

Value

Unit

’0 ψ0 c0 Eref 50 Eref oed Eref ur pref m l

Friction angle Dilatancy angle Cohesion Secant stiffness Tangent stiffness Elastic stiffness Reference stress Exponent power Elastic Poisson’s ratio

42.18 14.47 0 50 50 150 100 0.5 0.2

[� ] [� ] [kPa] [MPa] [MPa] [MPa] [kPa] [-] [-]

456

Figure 11. Time distribution.

dependent

excess

pore

pressure

Figure 12. Excess pore pressure at different measurement points.

domain. The soil has initial permeability of 2.12E­ 2 m/s. After that, the permeability of penetration zone decreases based on Figure 10, while the permeability of upper non-penetration zone keeps constant. Figure 11 shows the distribution of excess pore pressure with time. As seen, when the pressure applied at t = 0s, the generated excess pore pressure linearly distributed along the column. Due to the decrease of permeability with time at the lower part of the device, the excess pore pressures gradually accumulate in this penetration zone. Figure 12 presents the numerical cal­ culated excess pore pressure at different measurement points compared to the real measurements. It can also be observed that the pressure distribution tends to be stable after 10s which is consistent with the excess pore pressure distribution shown in Figure 11 . Further­ more, the numerically predicted pore pressure distribu­ tion agree well with the real measurements at 40s and 60s. It should be noted that there is a large discrepancy between the predicted and measured values at t = 10s. This might be attributed to the measurement inaccur­ acy at the beginning of the experiment.

model. The model geometry is shown in Figure 13. The tunnel has a diameter of D=10 m, and the overbur­ den depth is 10 m. The mechanical boundary condi­ tions at the bottom and outer boundaries of the model are defined by restricting deformations in the normal directions, whereas the in-plane displacements are allowed. There is no mechanical fixity on the model’s top surface. Only the model surface allows excess pore pressure dissipation. The applied soil constitutive model and corresponding material properties are identi­ cal as that in the 2D model. The tunnel properties can be found in Zhao, Hölter, König, & Lavasan (2019). It is assumed that tunnel face has reached a certain position and it starts the next round soil excavation. A penetration zone with 20 cm thickness is assumed in front of tunnel face based on the column test results, and its time-dependent perme­ ability is assumed identical as that in the 2D model. In the initial step of numerical simulation, the applied face pressure is defined as the summation of lateral earth pressure and steady water pressure, which means there is no movement at the tunnel face. Then the face pressure increases 36 kPa imme­ diately along the tunnel face to induce slurry penetra­ tion. After certain time period (t =0s, 1s, 2s, 5s, 10s, 40s, 60s), the applied face pressure decreases to 60% of the initial value. Consolidation analysis is con­ ducted for another 60s to evaluate the maximum horizontal displacement due to reduction of face pressure. It is found that when t =0s, which means penetration has not yet occurred, 0.96 m maximum horizontal displacement is observed at tunnel face after consolidation analysis. After that, lower value of maximum horizontal displacements are obtained at tunnel face due to decrease of permeability in the penetration zone as shown in Figure 14. This proves that the permeability reduction due to slurry penetra­ tion can improve the tunnel face stability, and signifi­ cant improvements in the mechanical behavior of the excavation face. As see, the maximum horizontal

5 3D NUMERICAL ANALYSIS OF TUNNEL FACE STABILITY In order to provide guideline for the real-scale tunnel­ ing problem, influence of soil permeability on the tunnel face stability is investigated in a 3D tunnel

Figure 13. Geometry of the 3D model.

457

REFERENCES

Figure 14. Maximum horizontal displacements due to 40% reduction of face pressure, uh;max;t0 =0.96 m.

displacement is reduced to a large extent (i.e. 80%) after about 40s of applying face pressure. 6 CONCLUSIONS The support of the tunnel face within slurry pressure balanced (SPB) shields depends mainly on the ability of the bentonite suspension to form a zone of reduced permeability in non-cohesive soil. This paper presents the experimental and numerical results of the decrease of the soil permeability in reference to the penetration depth of the bentonite suspension over time. The numerical investigations carried out to assess the impact of considering permeability evolutions in front of the slurry shield TBM on the soil deform­ ations due to water transport in front of the TBM. The 2D simulation verified the experimental results and observations while the 3D model analysed the hydraulic flows and effective stress variations in the soil in front of the TBM. Based on the results obtained from 3D numerical simulations, it was found that the face pressure transfer process from the TBM to the soil skeleton would be significantly improve when the slurry penetrates into the cohesionless soil. Numerical simulation of this process indicated that a 3D simulation of the excavation process along with the permeability update due to slurry penetration can realistically address the coupled system behavior.

ACKNOWLEDGEMENT The content of this paper results from the fruitful col­ laboration of the subprojects A5 Adaptive Constitutive Modeling of Soil with Special Consideration of Destructuration and A6 Local Transient Face Support within Hydro-Shields, both part of the Collaborative Research Centre – SFB 837 at Ruhr-University Bochum founded by DFG (Deutsche Forschungsge­ meinschaft). The authors wish to express their grati­ tude to Joerg Sahlmen (CEO of SFB 837) and Samy Tong (TLB) for their continuous support.

Anagnostou, G. & K. Kovari (1994). The face stability of slurry-shield-driven tunnels. Tunnelling and Under­ ground Space Technology 9(2), 165–174. API13B (2014). Recommended Practice for Field Testing Water-Based Drilling Fluids. American Petroleum Institute. DIN4126 (2013). Nachweis der Standsicherheit von Schlitzwaen­ den. Deutsches Institut fuer Normung DIN e.V. DIN4127 (2014). Erd- und Grundbau – Schlitzwandtone fuer stuetzende Fluessigkeiten, Anforderungen, Pruef­ verfahren, Lieferung, Gueteueberwachung. Deutsches Institut fuer Normung DIN e.V. Jancsecz, S. & W. Steiner (1994). Face support for a large mix-shield in heterogeneous ground conditions. In in Papers presented at the 7th International Symposium Tunnelling 94, pp. 1–19. Springer US. Kilchert, M. & J. Karstedt (1984). Band 2 Standsicherheits­ berechnung von Schlitzwaenden nach DIN 4126. Wies­ baden Berlin, Bauverlag GmbH (Beuth-Kommentare). Krause, T. (1987). Schildvortrieb mit fluessigkeitsgestuetzer Ortsbrust. Ph. D. thesis, Technische Universitaet Braun­ schweig, Germany. Morgenstern, N. & I. Amir-Tahmasseb (1965). The stability of a slurry trench in cohesionless soils. Geotechnique 15 (4), 387–395. Mueller-Kirchenbauer, H. (1972). Stability of slurry trenches. In Proceedings of 5th European Conference on Soil Mechanics and Foundation Engineering, Tokyo, pp. 543–553. Balkema. Schanz, T., P. Vermeer, & P. Bonnier (1999). The hardening soil model: Formulation and verification. In Proceedings of 1st International PLAXIS Symposium on Beyond 2000 in Computational Geotechnics, pp. 281–296. Balkema. Schoesser, B. (2004). Untersuchungen zur Entwicklung und Uebertragung von Tangentialspannungen am Umfang von Vortriebsrohren im nichtbindigen Lockergestein. Ph. D. thesis, Ruhr-University Bochum, Germany. Sridharan, A. & K. Prakash (2002). Permeability of two-layer soils. Journal of Geotechnical Testing 25(4), 443–448. Weiss, F. (1967). Die Standfestigkeit fluessigkeitsgestuetz­ ter Erdwaende. Muenchen: Verlag Ernst u. Sohn Bauingenieurpraxis 70. Zhao, C., R. Hölter, M. König, & A. Lavasan (2019). A hybrid model for estimation of ground movements due to mechanized tunnel excavation. Computer-Aided Civil and Infrastructure Engineering 34(7), 586–601. Zizka, Z. (2019). Stability of slurry supported tunnel face considering the transient support mechanism during excavation in non-cohesive soil. Ph. D. thesis, RuhrUniversity Bochum, Germany. Zizka, Z., B. Schoesser, M. Thewes, & T. Schanz (2018). Slurry shield tunneling: new methodology for simplified prediction of increased pore pressures resulting from slurry infiltration at the tunnel face under cyclic excava­ tion processes. International Journal of Civil Engineer­ ing: Special Issue Current trends and challenges in subsurface engineering 15(4), 387–398. Zizka, Z. & M. Thewes (2016). Recommendations for face support pressure calculations for shield tunnelling in soft ground. Technical report, Deutscher Ausschuss für unterirdisches Bauen (DAUB) German Tunnelling Committee (ITA-AITES), Cologne, Germany.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

SEM deformation prediction and observation by 3D numerical analysis H. Zheng, M. Mooney & M. Gutierrez Colorado School of Mines, Golden, Colorado, USA

C. Bragard Traylor Bros., Inc., Evansville, Indiana, USA

ABSTRACT: This paper summarizes a computational modeling effort carried out during real-time construc­ tion of a cavern to predict ground movements and surface settlement. The Regional Connector Transit Corri­ dor (RCTC) tunnel project in Los Angeles required the construction of a 90-m-long, 11-m-wide and 18­ m-high crossover cavern. The cavern was constructed by sequential excavation method (SEM) at relatively shallow depth. The SEM cavern was excavated after twin 6.7 m diameter tunnels were excavated via earth pressure balance shield machine through the cavern profile. A three-drift seven-stage excavation configuration was designed and implemented to control ground movements within allowable limits. A 3D numerical model was developed (FLAC3D) to simulate the SEM construction process. The model parameters were calibrated during the initial excavation of the left drift. Thereafter, the model predictions of ground movements (conver­ gence and settlement) provided good agreement with field measurements, providing confidence to the con­ struction team as it proceeded.

1 INTRODUCTION Rapid urbanization opens up huge demands on infrastructure systems. We are seeing increased use of the sequential excavation method (SEM), also commonly known as the New Austrian Tunneling Method (NATM) in urban tunneling (Romero 2002). This method was started 200 years ago and was originally conceived by Rabcewicz (Von Rab­ cewicz 1964). The basic concept of SEM is to favorably utilize the ground-support interaction for achieving efficiency and economy with a controllable influence on the environment. How­ ever, instead of mobilizing a high degree of ground self-support for rock tunneling (Høien et al, 2019, Lunardi 2008) in relatively rural areas where deformation control was of secondary significance, urban SEM tunneling with shallow depth, weak ground and presence of sensitive structures requires more rigorous control and risk management. Com­ putational modeling is a powerful tool in the per­ formance-based engineering analysis and design of tunnel structures. However, the application of numerical simulation on SEM tunneling still faces many challenges. A complex construction process, uncertainty in ground properties and various types of support combinations all demand more robust consideration and simulation approaches. The present paper demonstrates the application of 3D computation modeling in predicting behavior of

an SEM cavern constructed in downtown Los Angeles. The SEM cavern was excavated after twin 6.7 m diameter tunnels were constructed via earth pressure balance shield machine through the cavern profile. A three-drift seven-stage excavation config­ uration was designed and implemented to control ground movements within allowable limits. The field measurements were interpreted to characterize the displacement behavior caused by SEM tunneling. Model input parameters were calibrated by field measurements through initial excavation process and adaptive model was used for the following prediction.

2 PROJECT OVERVIEW 2.1 Regional Connector Transit Corridor (RCTC) tunnel The Regional Connector Transit Corridor is a 3-km-long light-rail line that will connect the existing Gold line to the existing Blue and Expo lines to improve mobility within Downtown Los Angeles and throughout the Greater Los Angeles area. The entire RCTC project contains 1.6 km of twin bored tunnels, 1.1 km of cut-and-cover tun­ nels, three cut-and-cover stations, 0.3 km of atgrade alignment and a crossover cavern (Herranz et al. 2016).

DOI: 10.1201/9780429321559-60

459

the formation is generally poorly cemented and weak to very weak, according to unconfined com­ pressive strength (UCS) tests results (3.4 to 5.5 MPa) (Ulusay 2014), and is perceived to behave similarly as stiff to hard clay. Therefore, the for­ mation is generally considered as a continuum for the purpose of analysis. The actual ground conditions encountered during construction have confirmed that the SEM cavern has been exca­ vated in fresh Fernando Formation. The exposed ground exhibits good stand-up time and no facestability issue has been encountered. The geotechnical properties presented in Table 1 are obtained based on all available field investi­ gation results and laboratory testing data from the corresponding boreholes within SEM cavern area plus boreholes comprising similar stratigraphy along the tunnel alignment. The effective strength parameters (cohesion and friction angle) are selected from triaxial test results and direct shear test results. Based on piezometric measurements, a perched groundwater table exists around 4m below ground surface and is expected to fluctuate seasonally about ±1:5m. In addition, the regional water level is

The case study presented in this paper focuses on the construction works of a mined crossover cavern, as one of the most challenging and risky parts of the RCTC project. The location and the tunnel geometry are shown in Figure 1. The mined crossover cavern is 90-m-long, 11-m-wide and 18-m-high, located adjacent to the east end of the 2nd St/Broadway Sta­ tion, under 2nd Street between Spring and Main streets. The cavern was constructed by sequential Excavation Method (SEM) with two-side wall threedrifts configuration and is followed by the twin bored tunnels excavation. 2.2

Geological and geotechnical conditions

The project alignment traverses the southeastern end of the Elysian Park Hills and the ancient Los Angeles River floodplain. The SEM cavern sec­ tion will encounter artificial fill and geologic units that range in age from Miocene to Holo­ cene. From oldest to youngest in geologic age are the Pliocene-age sedimentary strata of the Fernando Formation and Pleistocene-age allu­ vium. Along the cavern alignment, artificial fill soils (Af), about 2 to 3 m thick, are underlain by about 3.3 to 4.6 m thick of coarse-grained allu­ vial deposits (Qal2). The coarse-grained alluvium is underlain by Fernando Formation bedrock, which comprises an upper layer of moderately to highly weathered material with a slightly wea­ thered to fresh layer underneath. The entire cavern is excavated through more competent slightly weathered to fresh Fernando Formation, as shown in Figure 1. Fernando Formation is comprised predominantly of massive siltstone, with some interbeds of sandstone and conglomer­ ate (Lamar 1970). Two boreholes were drilled within the cavern area for the preliminary geo­ logical survey. And with adjacent exploratory borings in similar ground condition, it shows that

Table 1.

Summary of average geotechnical parameters. c0 ’0 (kPa) (º)

E (MPa) K0

ν

Soil type

γ (kN/m3)

Artificial Fill (Af) Alluvium (Qal2) Weathered Fernando Formation (Tf1) Fresh Fernando Formation (Tf2)

18.9

0

28

28

0.5

0.35

18.5 18.9

10 75

30 25

57 156

0.5 0.6

0.35 0.4

19.3

180

26

200

0.65 0.4

Figure 1. Alignment and monitoring layout of the SEM cavern.

460

present within the Fernando Formation. Due to the low permeability of Fernando Formation (around 2 x 10-9 m=s), the groundwater inflow into the cavern during excavation is unlikely to happen except while encountering discontinuities or large varying levels of rock mass fracturing. During the excavation, the exposed Fernando Formation encoun­ tered with a massive composition and with little presence of predominant discontinuity sets and there has been no groundwater infiltration into the tunnel Bored tunnels and buildings Twin-bored tunnels were constructed prior to the excavation of the SEM cavern. These tunnels were excavated by an earth pressure balance machine (EPBM) with an excavated diameter of approximately 6.7m, installing 533mm thick, 5.74m inner diameter precast concrete tunnel lining (PCTL) with double gaskets. Three existing buildings are located in the proximity of the crossover cavern area. In add­ ition, the cavern is overlain by an LA County storm drain- The Central District Storm Drain. The location of the three buildings is shown in Figure 1 and a summary of the structures is listed in Table 2.

Table 2.

Building

Building information. Number of stories

LAPD Head- 10 quarters Metropolitan 1 Building Higgins 10 Building

Equivalent building surcharge (kPa)

Height (m)

Foundation depth (m)

48.7

6.8

52.6

6.1

6.1

35.1

38.4

8.2 5.5

112.8

2.3

Construction sequences

The cavern construction follows the completion of the excavation for the 2nd/Broadway Station and twinbored tunnel. The excavation of the cavern commenced at the end of May 2018. The tunnel was driven by the SEM, with a three-drift seven-stage scheme. A detailed cavern excavation sequence is shown in Figure 2. The left and right drifts are divided into top heading and bench and the center drift includes an invert section. Before the start of SEM excavation, the bored tunnel was partially filled with low-strength cementitious material to provide cushion for demolished segmental lining and maintain stability during sequential excava­ tion. To protect the crown excavation line of the cavern from localized ground instability, the first 18m of the cavern was excavated under a 53-unit pipe umbrella system installed from the station side. Two rows of 18m-long, 8.6-mm-thickness, 127-mm-outer diameter canopy pipes cover an arc area of 90 degrees, symmet­ rically distributed from tunnel centerline. The excavation started from the left drift, followed by the right drift, each divided into top heading (LTH/ RTH) and bench (LBEN/RBEN). Then the center drift excavation was separated into two main stages, the center top heading (CTH) excavation was followed by the center bench (CBEN) and invert (CINV) advancing. The temporary sidewalls were demolished at a certain distance from the ring closure section. In order to miti­ gate successive disturbance due to multiple drifts exca­ vation, trailing distances were remained between each drift during the construction process. The average trail­ ing distances between left and right drift, right and center drift are 21 m and 35 m respectively. The cavern initial support comprises a 305-mm­ thick initial perimeter shotcrete lining except the center invert applying cast-in-place concrete as shown in Figure 3 (b). The final lining will be cast-in-place concrete with a minimum thickness of 457 mm. The initial support system generally consists of steel-fiber reinforced shotcrete with localized wire mesh and

Figure 2. Excavation sequence for the SEM cavern.

461

Figure 3. Initial support system.

lattice girders spaced at 1 m as shown in Figure 3 (a). The initial lining was designed for minimum 34.5MPa 28-day compressive strength. In order to monitor the progressive strength of initial lining, sam­ ples were taken on-site for several key locations and 1, 3, 7 and 28 days strength were obtained through the compression strength tests (Figure 4). The advance length for side drifts excavation was initially started with 1 m and modified to 1.5 m after 70 m excavation of left drift to accommodate PCTL demolition of bored tunnels. Even though the extension of advance length leads to an increased unsupported length per round, a more effective installation was achieved, avoiding unfavorable local over-excavation, also reduce the total cycle time (~7 hours reduction for two PCTL length compared with previous steps) with­ out causing any additional ground deformations. The advance lengths for center top heading and center bench/invert are 1 m and 2 m, respectively. 2.4

Instrumentation

SEM follows the principle of the observational method, which requires a comprehensive monitoring program. An extensive instrumentation system has been implemented for monitoring ground deformation

response, for the influence of existing structures, and to measure the performance of initial shorcrete lining. Instrumentation locations and types are shown in Figure 1. There are 11 monitoring cross-sections (MCS) set for surface displacement and 13 MCS for cavern initial support convergence. Other instrumen­ tation types include arrays of multipoint borehole extensometers, piezometers, tiltmeters and structural monitoring points. The surface monitoring was car­ ried out by using automated total stations (AMTS). The recorded data were transferred in real-time via a web-portal-based instrumentation data management system. They were stored in a database consulted remotely by the project team (contractor, designer, project owner). During the construction process, the field measurements were carefully interpreted and analyzed to consider necessary adapting the construc­ tion sequence and adopting toolbox items for the deformation control. 3 NUMERICAL MODELING The numerical investigation was conducted and updated during the entire process of construction. The numerical calculations aimed at simulating SEM tunneling and predicting the deformation behavior for future construction stages were carried out employing the finite difference (FD) code FLAC3D. The use of 3D modeling was able to capture threedimensional stress-strain feature for multiple excava­ tion faces without assumptions of stress release ratio before the support activation. 3.1

The geometry of the 3D model is shown in Figure 5. A discretization with a total number of 300,000 poly­ hedral-shaped zones has been generated. The cavern region was subdivided into refined zones in order to account for high-stress gradients developed in that region. The top surface of the model that represented the ground surface was free to move. The bottom boundary was fixed, and four vertical boundaries were constrained in normal direction movement. 3.2

Figure 4. Shotcrete strength test results.

Model geometry

Constitutive model

An elastic perfectly plastic model based on the Mohr-Coulomb (MC) failure criterion was adopted to represent the layered ground. Although the mech­ anical behavior of the ground could be more pre­ cisely simulated by advanced constitutive model, due to limited geotechnical tests available, more uncer­ tainty would be introduced to the model as a result of more parameters required by complex constitutive models. Considering the ground condition, plastic zone is limited around the opening and MC model should have a satisfying performance simulating the undrained behavior. The geotechnical parameters assigned in the model are shown in Table 1. It

462

Table 3. Parameter

Properties for structural elements. Storm Drain Cavern shotcrete tunnel PCTL initial support

Elastic modu- 21.5 lus (GPa) 23.5 Unit weight (KN/m3)

Poisson’s ratio 0.2

Figure 5. Finite Difference (FD) model mesh and geometry.

should be noted that the initial selection of Young’s modulus mainly depends on unload-reload modulus obtained from the pressuremeter tests, due to the fact that the primary response of the ground in the vicin­ ity of the opening is unload-reload process during the SEM construction. Then the value of Young’s modu­ lus was tuned through back-analyzing the monitoring data obtained from the initial left drift construction. Due to the low permeability of Fernando Forma­ tion (in the order of 10-9 m/s), the SEM excavation was simulated under undrained condition. The undrained (short-term) response was simulated in FLAC3D using wet approach (Itasca 2011), in which a coupled system is analyzed and drained geotech­ nical parameters were used, fluid modulus was given a realistic value (2 GPa), and pore pressure was gen­ erated as a result of mechanical deformations. The structural elements were modeled with linear elastic behavior. Concrete structural elements for the storm drain tunnel and PCTL were modeled using three-node shell elements attached to the ground zones via links representing the ground-structure interface. The initial lining of the cavern was modeled by the liner elements, considering the aging effect of shotcrete strength as shown in Figure 6. The lattice girder and

Figure 6. Age-dependent shotcrete lining and umbrella canopy pre-support system.

31.7 23.5

3.4 to 28.9 (Time

dependent)

23.5

0.2

0.2

wired mesh were not modeled explicitly. The main mechanical properties of structural elements are listed in Table 3. The elastic modulus of concrete is calcu­ lated based on the concrete compressive strength (Standard 2011). The properties of the shotcrete liner were updated based on the test results along the con­ struction process as shown in Figure 4. The umbrella canopy used for pre-support cavern crown was modeled by pile structural elements, and they were considered as the straight segments with uniform bi-symmetrical cross-section between two nodes as shown in Figure 6. 3.3

Existing structures

Building structures were not modeled explicitly, instead, they were considered by excavation to the foundation level and application of the corres­ ponding building surcharge pressure (Table 2). The storm drain tunnel was simulated by a single step excavation and installation of its concrete lining. Since the bored twin tunnels were constructed before the SEM cavern excavation, the disturbed ground stress and strain state are needed to con­ sider as a prior condition. The bored tunnels were modeled by a multi-stage step by step pro­ cedure (Mooney et al, 2016) as shown in Figure 7: (1) At each excavation stage, the mod­ eled excavation advances by deactivating soil elements of one ring advance (1.5 m); (2) The support pressure at tunnel face was modeled by a 263 kPa pressure at the springline level with 8.8 kN/m3 vertical gradient, which are selected based on TBM monitoring data during the con­ struction in the cavern section; (3) The TBMrecorded annulus pressure is applied to the radial boundary along the 10.5 m length of the TBM shield. Instead of modeling the TBM shield expli­ citly, it was represented by a pressure boundary prior to the lining installation; (4) The segmental lining represented by shell elements was installed at 10.5 m behind the tunnel face, with a hardening grout annulus modeled by solid elements inserted between the PCTL and ground material. This tunnel excavation model sequence was continued until the last ring installed at the end of the model.

463

a round length of 1m, the center bench and invert were modeled with 2 m round length. The lagged distances between each drift were set to average values in the real case. 4 NUMERICAL RESULTS COMPARED WITH MEASUREMENTS In order to characterize the displacement behavior and validate the adaptive 3D simulation, the measure­ ments of an instrumented section, including ground surface and in-tunnel convergence, are compared with the results obtained from numerical analysis. Figure 7. FlAC3D model of EPBM tunneling.

3.4

Construction sequence modeling

The simulation of the construction sequence was kept conformity with the actual scenario described in Section 2.4. Since the flashcrete layer was sprayed immediately right after the excavation, the 1st layer support liner was assumed wished-in-place in the model (activated after ground material removal). Two inches flashcrete applied on tunnel face was not considered here. Deformable links were utilized to connect liner elements installed at different stages. The excavation of the side drifts was simulated in 60 steps with an excavation length of 1m. In each step, the bench was excavated and supported with a delay of 6.7 m following the top heading. The excavation of center top heading was simulated in 60 steps with

4.1

Surface settlement

Figure 8 shows a comparison between numerical model results and measurements for progressive ground surface deformation at the monitoring section SEM 00+50 (15 m away from the start section). The surface settlements caused by TBM excavation are less than 1.2 mm calculated from the numerical model. In the field, the instrumentation devices did not record any noticeable deformations induced by bored tunnel construction. Thus, here we only dis­ cuss the relative deformations caused by SEM construction. The ground surface movements above cavern area were monitored throughout the entire construction process. Also shown in this figure are model results obtained from corresponding monitoring points, plot­ ted against the excavation faces distance to the moni­ toring section. The surface settlement started to develop when left drift began to excavate, which is

Figure 8. Comparison of computed and measured surface settlement at SEM 00+50.

464

at 15 m away from the monitoring section. On arrival of the left top heading at the monitoring section, sur­ face settlement reached 10% of its final value, which can be noted as pre-settlement due to left drift exca­ vation. After left drift passed, right drift started, the settlement was continuously growing with an increas­ ing gradient resulted from a superposed disturbance form two side drifts. Upon the arrival of right drift, it reached 38% of the total settlement. After the right drift passed 20 m away from the monitoring section, the surface settlement behaved a trend to be stable. Then it quickly started a new settlement trend due to the center top heading excavation. On arrival of the center top heading at the monitoring section, 61% of total surface settlement has been developed. Further advancement of the center top heading gradually increased surface settlement and ceased its effect when the center top heading drove approximately 15 m beyond the monitoring section. The pre­ settlement due to center bench and invert excavation was initiated 14 m ahead of monitoring points. The remaining 33% of total surface settlement happened during the final stage of excavating center bench and invert. Advancing 28 m after the ring closure of the primary support, ground surface settlements grad­ ually ceased development. As a comparison in Figure 8, it indicates that the numerical model results generally show a good agreement with field measurements. After calibrat­ ing Young’s modulus and shotcrete strength at the initial stage of construction, the model very well predicts the value of surface settlements as well as the trends during different drifts advancing. Except that it underestimated the part caused by right drift excavation and overestimated the final

stage settlement a little bit. The model results indi­ cate that the right-side monitoring point gained more surface settlement than the left-side. How­ ever, there is not so much difference from in situ measurements. Since the model has the assump­ tion that surface buildings were modeled by exca­ vation to the foundation level, the right-side monitoring point is closer to the vertical boundary of represented building. Whereas the interaction between ground and building structure is more complicated in reality, which will affect the verti­ cal propagation of deformation. 4.2

Cavern initial support

Figure 9 illustrates the measurements and predictions of initial lining convergence at left drift along the three drifts’ excavation process. The results show clearly the effects of excavation on other drifts on relaxing the deformation shape of the left initial lining. Since the pre-deformation and loss displacement (Lunardi 2008) cannot be measured before the instal­ lation of the monitoring instrument, the measured convergence is only part of the total radial displace­ ment. In most cases, displacement calculated by numerical simulation exceeds the measured conver­ gence (Li et al. 2016). Herein, the liner deformation from numerical model was recorded after the stage of installing initial support, in order to have a better comparison with the measured data. Although part of the loss displacement resulted from the time lag for in-tunnel survey is still unable to capture, gener­ ally, the patterns of measured tunnel convergence are very well described by the FD model.

Figure 9. Comparison of left drift convergence at SEM 00+50.

465

5 CONCLUSION

REFERENCES

This paper presents a case study of an SEM cavern construction in urban soft ground settings. A numerical analysis was carried out in real-time construction to predict ground movements and surface settlement. In this case the overall surface settlements reached 18-20 mm. Surface deform­ ations caused by the excavation of side drifts, center top heading and center invert are 50%, 17% and 33%, respectively. It has demonstrated the predictive capabilities of 3D FD analysis con­ ducted for the excavation and support of the SEM cavern. By calibrating the elastic modulus and construction parameters, the FD model with relatively simple Mohr-Coulomb elastic-perfectly plastic constitutive model not only reproduced the recorded deformations but also well predicted the further step response both in magnitude and in shape. Further refinements of the computational model are needed to understand how uncertainties in the stratigraphy and construction variables affect the prediction performance of the SEM tunneling.

Herranz, C., Penrice, D., Lianides, J., and Horvath, Z., 2016. SEM Crossover Cavern in Downtown L.A. Geotechnical and Structural Engineering Congress, 2043–2053. Høien, A.H., Nilsen, B., and Olsson, R., 2019. Main aspects of deformation and rock support in Norwegian road tunnels. Tunnelling and Underground Space Tech­ nology, 86 (January), 262–278. Itasca, F., 2011. V5. 0, fast lagrangian analysis of continua in 3 dimensions, user’s guide. Itasca Consulting Group, Minneapolis, Minnesota, 308–318. Lamar, D., 1970. Geology of the Elysian Park-Repetto Hills Area, Los Angeles County, California. Li, P., Zhao, Y., and Zhou, X., 2016. Displacement character­ istics of high-speed railway tunnel construction in loess ground by using multi-step excavation method. Tunnelling and Underground Space Technology, 51, 41–55. Lunardi, P., 2008. Design and construction of tunnels: Ana­ lysis of Controlled Deformations in Rock and Soils (ADECO-RS). Springer Science & Business Media. Mooney, M.A., Grasmick, J., Kenneally, B., and Fang, Y., 2016. The role of slurry TBM parameters on ground deformation: Field results and computational modelling. Tunnelling and Underground Space Technology, 57, 257–264. Von Rabcewicz, L., 1964. The new austrian tunnelling method. Water Power, 65. Romero, V., 2002. NATM/SHOTCRETE FOCUS-NATM in soft ground–A contradiction in terms? Views on NATM and its application to soft-ground tunnelling dis­ pelling some misconceptions about this sometimes con­ troversial. World Tunnelling, 15 (7), 338–344. Standard, A.A.C.I., 2011. Building Code Requirements for Structural Concrete (ACI 318-11). In: American Con­ crete Institute. Ulusay, R., 2014. The ISRM suggested methods for rock characterization, testing and monitoring: 2007-2014. Switzerland: Springer.

ACKNOWLEDGMENT This research was supported by the Univer­ sity Transportation Center for Underground Transportation Infrastructure (UTC-UTI) at the Colorado School of Mines under Grant No. 69A3551747118 from the U.S. Department of Transportation (DOT). The support from Traylor Bros., Skanska, Mott Macdonald and LA Metro is greatly appreciated.

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Face stability of slurry shield-driven tunnel in an aquifer

T. Xu & W.H. Zhou State Key Laboratory of Internet of Things for Smart City & Department of Civil and Environmental Engineering, University of Macau, Macao, China

A. Bezuijen Department of Civil Engineering, Ghent University, Ghent, Belgium Deltares, Delft, The Netherlands

ABSTRACT: In this article the face stability of slurry shield-driven tunnel in an aquifer is investigated. The influences of slurry infiltration at the face and soil layering are taken into account. The results show that due to the infiltration an additional margin for the support pressure is required. At larger slurry infiltration distance or larger cover depth to tunnel diameter ratio (C/D), a higher margin is required. It also appears that soil layer­ ing affects the hydraulic gradient that stabilise the face of cohesionless soil. TBM excavation in a semiconfined aquifer (compared to an unconfined aquifer) leads to higher excess pore water pressures in the soil around the TBM and a reduced infiltration of slurry into the soil in front of the TBM. The excess pore water pressures in a semi-confined aquifer are present over a larger area than in an unconfined aquifer, increasing the risk of a blow-out during excavation.

1 INTRODUCTION The tunnel-boring machine (TBM) tunnelling tech­ nique has been developed to construct tunnels that require strict settlement control, e.g. in urban areas with a large amount of buildings, historic areas etc., or where the soil is very soft (e.g. when crossing a river or estuary). When TBM tunnelling is per­ formed below the water table, especially in cohe­ sionless soil with high water pressure, effective support at the face plays a key role in the face sta­ bility. Practice proved that, in such a condition, sup­ port suspension without additive(s) is insufficient to stabilise the face (Bezuijen, 1996; Hölscher, 2008), and may lead face collapse. If the face pressure is higher than the pore pressure in the soil, there will be no flow into the excavation chamber. However, the face will not be stable, because the hydraulic gradient is too low (Van Rhee & Bezuijen, 1992). The face stability and limitation of the groundwater flow have to be achieved by pressurising the ben­ tonite slurry for slurry TBM or foam for earth­ pressure-balance shield (EPB shield) at the face. It should also be noted that too large slurry or foam pressure may lead to face blow-out. Therefore, carefully determining the support pressure is important. Various methods in analysing face stabil­ ity and determining the minimal and maximum

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support pressures can be found in literature, such as limit equilibrium analysis (Horn, 1961; Jancsecz & Steiner, 1994; Qarmout et al. 2019), numerical modelling (Zhang et al. 2015; Liu et al. 2019), physical modeling (Chambon et al. 1991; Ahmed & Iskander, 2012; Chen et al. 2013). Just little atten­ tion was given to the infiltration of support suspen­ sion in existing literature (e.g. Bezuijen et al. 2001; Broere, 2001; Broere & van Tol, 2001; Broere, 2015). Due to the pressure difference between the exca­ vation chamber of the TBM and the ground, support suspension will infiltrate into the soil and thus there will be a flow in front of the face. In such a situation, part of support pressure applied through the support suspension at the tunnel face will transfer into excess pore water pressure in the soil. The induced lowering of effective support pressure should be avoided (Broere, 2015). However, this is hard to be achieved. This study therefore aims to investigate the effect­ iveness of the support, global stability of the face under infiltration of slurry into the soil ahead of the tunnel face and soil layering and seepage and their influences on the macro- and micro-stability of the face. The study focuses on slurry shields, but the findings are also valid for earth-pressure-balance shields (EPB shields).

2 MACRO-STABILITY OF THE FACE 2.1

Basic framework

The basic framework for analysing the face stability considering the effect of infiltration is based on the wedge model that first proposed by Horn (1961), and then followed by many others (e.g. Anagnostou & Kovari; 1994; Jancsecz & Steiner, 1994; Broere, 2001; Broere & van Tol, 2001; Perazzelli et al. 2014; Broere, 2015). Figure 1 shows a sketch of the basic framework of the model of face stability. The sup­ port pressure E(ϑ) is derived from the equilibrium of forces that act at the wedge as shown in Figure 2 (Broere, 2015):

where G is the weight of the soil wedge (kN), Pv is the vertical load from the soil prism (kN), ϑ is the sliding angle between the wedge and the horizontal plane (°), ϕ is the friction angle (°), c2 is the cohesion (kPa) of soil in the tunnel face area, T is the shear resistance force on the vertical triangular plane of the wedge (kN), and D is the tunnel diameter (m). The maximum support pressure can be deter­ mined by derivation of E(ϑ) to ϑ:

where ϑcr is the critical sliding angle (°). The forces acting on the wedge can be obtained as follows: Weight of the wedge:

with γ2,av being the average unit weight of the soil in the tunnel face area (kN/m3). Vertical load from the soil prism:

Figure 1. Wedge stability model (Broere, 2015).

Figure 2. Equilibrium of forces that act at the wedge (modified after Zizka & Thewes, 2016).

468

with σv, crown the total vertical stress in tunnel crown (kPa).

a critical velocity that the advancement rate is equal to the infiltration velocity. If the advance rate is slower than the infiltration velocity, the assumption that a filter cake is formed during drilling is valid. If the advance­ ment rate of the TBM is higher than the infiltration vel­ ocity, the cutting wheel of the TBM will cut off the all the slurry infiltrated soil and some non-infiltrated soil and carry it into the excavation chamber. In both situ­ ations, soil-slurry-mixture rather than ‘clean slurry’ will be present in the gap between the face and the excavation chamber. Hence, depending on the density of the slurry, there will be either no filter cake or a little filter cake formation at the face but a continuous infiltration (Xu & Bezuijen, 2019). For low infiltration rates the pressure in the soil close to the TBM is equal to face pressures. For higher infiltration rates there is still excess pore water pressure in the soil, but it is now determined by the TBM:

with γ1,av the average unit weight of the soil in the tunnel crown (kN/m3).

where A is the cross-sectional area of the silo (m2); U the circumference length of the silo (m); K1 the coefficient of the lateral earth pressure in the area of the soil prism (-). Resistance force on the vertical triangular plane of the wedge is:

, K0 is the with coefficient of lateral earth pressure at rest (-) and Ka is the coefficient of Rankin’s active earth pres­ sure (-). 2.2

Effective support ratio

As pointed by Bezuijen et al. (2016) and Xu & Bezui­ jen (2019), during excavating, the advancement rate of TBM partially determines the face stability. There is

with n the porosity of the soil (-); vp the pore fluid velocity (m/s) and D the tunnel diameter (m). As a result, part of support pressure will be trans­ ferred into excess pore water pressure (Bezuijen et al. 2001; Broere, 2001; Xu & Bezuijen, 2018). The net force acts at the face to support the wedge therefore is less, see Figure 3. Bezuijen et al. (2001) argued that the support is less effective in the situ­ ation with excess pore water pressure. The excess pore pressure will also create a vertical gradient over the block CDEFKLMN (see Figure 3) resulting in a reduction of the force from this block on the

Figure 3. Pressure distribution over infiltration zone and excess pore water pressure (modified after Bezuijen, 2001).

469

triangle. COB (2000) and Broere (2001) showed a significant increase in the minimum allowable tunnel face pressure to achieve a stable front, for the cases of the Second Heinenoord Tunnel and Botlet Rail Tunnel. Zizka & Thewes (2016) proposed a concept for the global stability of the wedge, namely the factor (F) defined as a ratio of area of the infiltrated zone and the wedge:

with

As the pressure is linear along the depth, we have:

where Se, Sw are the area of infiltrated zone within the wedge and the area of the total infiltrated zone (m2); emax,crown, emax,invert, the maximum infiltration distance in crown and invert (m). The maximum infiltration distance (emax) during mud spurt can be determined by the infiltration tests (Xu & Bezuijen, 2019a). Here, a more general expression is given. As shown in Figure 4, the two lines, the inclined sliding line (L1) and the front infiltration line (L2) are:

The areas that infiltrated zone within the wedge (Sw) and the total infiltrated zone (Se) are:

2.3

The joint point of L1 and L2 is (x0, y0):

Figure 4. Influence of infiltrated zone in the global stability.

Case study

To investigate the effects of the infiltration distance, soil strength, tunnel diameter and cover depth on the effective support ratio, a case study of a tunnel bored in a homogeneous soil is presented: tunnel diameter D = 10 m, cover depth C = 20 m, slurry pressure Δp = 50 kPa, water table depth h = 3 m, effective specific weight γ’ = 11 kN/m3, dry specific weight γd = 17 kN/m3, saturated specific weight γs = 21 kN/ m3, cohesion c = 0 kPa, friction angle ϕ = 36°. Equa­ tions (8) to (13b) based on Figure 4 are used to cal­ culate the effective support ratios. This is valid during standstill, but not during drilling. Figure 5 shows that infiltration distance has sig­ nificant influence in the effective support that acts at the tunnel face. For small infiltration distances only (< 0.5 m), the effective support ratio is high (> 80 %). For large infiltration distance (> 6.0 m), the effective support ratio (< 20%) can be neglected. The results were based on the low viscosity of sup­ port suspension. Support suspension of higher vis­ cosity can lead to a shorter infiltration distance and hence a higher effective support ratio. Also for a low soil permeability the effective support ratio is limited, but now because of excess pore water pres­ sure. Furthermore, the internal friction angle of the soil in front of the face has limited influence in the effective support ratio because the induced change in sliding angle caused by is small, see Figure 6. Cover depth and tunnel diameter may influence the required support pressure too. For coarse sands

470

Figure 5. Effective support ratio with infiltration distance for various friction angles.

Figure 8. Effective support ratio with infiltration distance for various C/D with the constant D.

shown in Figure 7, a larger value of C/D results in smaller effective support ratio for a given certain infiltration distance. Figure 8 shows that for a certain tunnel diameter, cover depth has little influence on the effective support ratio. 3 SOIL LAYERING AND MICRO STABILITY Soil layering may affect the pressures at the tunnel face and thus the stability of the tunnel face. Two common types of soil layering are shown in Figure 9. For an unconfined aquifer, a condition of homoge­ neous soil may be assumed. The piezometric head at the face can be approximated by (Bezuijen, 2001):

Figure 6. Sliding angle against internal friction angle of the soil in front of the tunnel face.

where ϕ the piezometric head (m) at a distance x (m) in front of the tunnel face, ϕ0 is the piezometric head at the tunnel face (m), and D (m) the tunnel diameter, assuming a piezometric head of zero far from the tunnel in the pore water. Derivation of Equation (14) to x leads to:

At the tunnel face x = 0, we have:

Figure 7. Effective support ratio with infiltration distance for various C/D with the constant C – coarse sand.

For a semi-confined aquifer, the piezometric head at the face can be calculated with (Bezuijen & Xu, 2018):

471

Figure 9. Sketch groundwater flow caused by shield tunnelling. (After COB, 1998, drawings by A. Bezuijen jr, used with permission).

where Q is the discharge at the face (m3/s), k the permeability of the soil (m/s), H the height of aquifer (m), r the distance from the centre of the tunnel face (m), λ the leakage length (m). K0 is the modified Bessel function of zero kind, see Appendix A. Equation (17) is valid when the height of the aquifer is more or less equal to the tunnel diameter. Otherwise, the calculation by Bezuijen & Xu (2018) has to be followed. An approximate gradient at the tunnel face x = 0 was given by Bezuijen & Xu (2018):

To stabilise a vertical face of cohesionless sand, a constant drag force with a gradient of i ≥ 2 has to be maintained (van Rhee & Bezuijen, 1992). Equations (16) and (18) show that a larger tunnel diameter leads a lower gradient. For a tunnel with diameter of 5 m or smaller in an unconfined aquifer, a support pressure of 50 kPa (piezometric head of 5 m) at the face will result in i ≥ 2. This meets the minimum required gradi­ ent. However, in case of a tunnel with diameter of 10 m or larger, a gradient i ≤ 1 will be achieved for a support pressure of 50 kPa in an unconfined aquifer. Therefore, support with pure water fails to stabilise the face. In a semiconfined aquifer, the condition will be more com­ plicated. The gradient depends on the support pressure, the tunnel diameter, the height of

472

aquifer and the leakage length. For a 10 m diameter tunnel in a 30 m high aquifer, a support pressure of 50 kPa (piezometric head of 5 m) a gradient i < 2 will be achieved at the face. This gradient is unable to achieve a stable body of cohesionless grains. For a 5 m diameter tunnel, the gradient will be larger than 2 and thus a stable tunnel face can be achieved. Apart from the micro-stability consequence, soil layering has some other influences in the sta­ bility: The excess pore water pressures and thus the infiltration in front of the TBM can be differ­ ent for various soil layering. Bezuijen & Xu (2018) have shown that TBM excavation in a semi-confined aquifer leads to higher excess pore water pressures in the soil around the TBM during excavation. Furthermore, the high pore water pressures decrease the stability of the tunnel face and therefore an increased face pres­ sure will be necessary compared to the uncon­ fined situation. Bezuijen & Xu (2018) showed that for a semi-confined aquifer, the excess pore water pressures caused by drilling are present over a larger area than in an unconfined aquifer. This may increase the risk of a blowout (Aime et al. 2004). Furthermore, the infiltration velocity of the slurry in the soil in front of the TBM will be slower during both excavation and the stand­ still. Both the larger extent of the excess pore water pressures and the lower infiltration velocity of the slurry will decrease the stability of the tunnel face (Bezuijen & Xu, 2018). 4 CONCLUSIONS Based on the analyses of macro- and micro- stability of the face of slurry shield-driven tunnel in an aquifer, the following conclusions may be allowed to be drawn.

The larger infiltration distance or the larger ratio of cover depth to the tunnel diameter (C/D) at a certain cover depth, the higher additional margin for the support pressure is required. Soil layering influences the hydraulic gradient at the face and thus the face stability. In an unconfined aquifer, a larger tunnel leads to a lower hydraulic gradient at the face. In a semiconfined aquifer, a larger tunnel also leads to a lower gradient but the condition will be more complicated. It also depends on the height of aquifer and the leakage length etc. Generally, for a tunnel with the same diameter and support pressure, the gradient at the face in an unconfined aquifer will be higher than that in a semiconfined aquifer. Furthermore, TBM excavating in a semi-confined aquifer leads to higher excess pore water pressures in the soil around the TBM and a reduced infiltration of slurry into the soil in front of the TBM during both excavation and the standstill; The excess pore water pressures are present over a larger area, increasing the risk of a blow-out during excavation.

ACKNOWLEDGEMENT The research is funded supported by Science and Technology Development Fund, Macao Special Administrative Region of China (File numbers: FDCT/0035/2019/A1 and FDCT/193/2017/A3).

REFERENCES Aime, R., Aristaghes, P., Autuori, P., Minec, S. 2004. 15 m diameter tunneling under Netherlands polders. In: ITA Proceedings of Underground Space for Sustainable Urban Development, Singapore, Elsevier. Ahmed, M. & Iskander, M. 2012. Evaluation of tunnel face stability by transparent soil models. Tunnelling and Underground Space Technology, 27(1), 101–110. Anagnostou, G. & Kovári, K. 1994. The face stability of slurry-shield-driven tunnels. Tunnelling and Under­ ground Space Technology, 9(2): 165–174. Bezuijen, A., 1996. Inventarisatie Boorprojecten met “Loop­ zand”. Technical Report 15, Boren Tunnels en Leidingen. Bezuijen A., Pruiksma J.P., Meerten H.H. van, 2001. Pore pressures in front of tunnel, measurements, calculations and consequences for stability of tunnel face. Proc. Int. Symp. on Modern Tunneling Science and Techn. Kyoto. Bezuijen, A., Steeneken, S. P., Ruigrok, J. A. T., 2016. Monitoring and analysing pressures around a TBM, In: The 13th International Conference Underground Con­ struction, Prague. Bezuijen, A. & Xu, T. 2018. Excess pore water pressures in front of a tunnel face when drilling in a semiconfined aquifer. ITA-AITES World Tunnel Congress, Dubai, 2–10.

Broere, W. 2001. Tunnel Face Stability & New CPT Appli­ cations. Ph.D. thesis, Delft University of Technology, Delft, The Netherlands. Broere, W. 2015. On the face support of microtunnelling TBMs. Tunnelling and Underground Space Technology 46, 12–17. Broere, W., van Tol, A. 2001. Time-dependant infiltration and groundwater flow in a face stability analysis. In: Adachi, T., Tateyama, K., Kimura, M. (Eds.), Modern Tunneling Science and Technology. Balkema, 629–634. Chambon, P., Corté, J., Garnier, J., König, D. 1991. Face stability of shallow tunnels in granular soils. In: Ko, H., McLean, F. (Eds.), Centrifuge ’91. Balkema, Rotterdam, 99–105. Chen, R. P., Li, J., Kong, L. G., Tang, L. J. 2013. Experi­ mental study on face instability of shield tunnel in sand Tunnelling and Underground Space Technology 33: 12–21. COB (Centre for Underground Construction). (2000. Second Heinenoord tunnel evaluation report, COB report K100-06. Gouda, The Netherlands: COB. Horn, N. 1961. Horizontaler Erddruck auf senkrechte Abschlussflächen von Tunnelröhren. In: Landeskonfer­ enz der Ungarischen Tiefbauindustrie, 7–16. Hölscher, P. 2008. Procesverbetering Aanleg Ondergrondse Infrastructuur –Microtunnelling. Technical Report TC161, Centrum Ondergronds Bouwen. G. Jancsecz, S. & Steiner, W. 1994. Face support for a large mix-shield in heterogenous ground conditions. In: Tun­ neling ’94, Institution of Mining and Metallurgy, London, 531–550. Liu, X. Y., Wang, F. M., Fang, H. Y., Yuan, D. J. 2019. Dual-failure-mechanism model for face stability ana­ lysis of shield tunneling in sands. Tunnelling and Under­ ground Space Technology 85: 196–208. Perazzelli, P., Leone, T., Anagnostou, G. 2014. Tunnel face stability under seepage flow conditions. Tunnelling and Underground Space Technology 43: 459–469. Qarmout, M., König, D., Gussmann, P., Thewes, M., Schanz, T. 2019. Tunnel face stability analysis using Kinematical Element Method. Tunnelling and Under­ ground Space Technology 85, 354–367. van Rhee, C. & Bezuijen, A. 1992. Influence of seepage on stability of sandy slope. ASCE Journal of Geotechnical Engineering 8: 1236–1240. Xu, T. & Bezuijen, A. 2018. Analytical methods in predict­ ing excess pore water pressure in front of slurry shield in saturated sandy ground. Tunnelling and Underground Space Technology 73: 203–211. Xu, T. & Bezuijen, A. 2019. Bentonite slurry infiltration in sand: filter cake formation under various conditions. Géotechnique, 69(12), 1095–1106. Zizka, Z. & Thewes, M. 2016. Recommendations for face support pressure calculations for shield tunnelling in soft ground. Cologne, Germany: Deutscher Ausschuss fuer unterirdisches Bauene. V. German Tunnelling Com­ mittee (ITA-AITES). Zhang, C., Han, K., Zhang, D. 2015. Face stability analysis of shallow circular tunnels in cohesive–frictional soils. Tunnelling and Underground Space Technology, 50, 345–357.

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APPENDIX

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Numerical analysis of Double-O-Tube shield tunneling in Shanghai D. Zhou & L. Zdravković Department of Civil and Environmental Engineering, Imperial College London, London, UK

ABSTRACT: Double-O-Tube (DOT) shield tunnelling is a technology developed to enable a more effi­ cient use of the underground space and to shorten the time of tunnel construction. Its primary application in the past two decades has been in Japan and China, involving about twenty engineering cases, six of which are on the Shanghai metro system. This paper presents advanced finite element analysis of one of the DOT tunnelling cases on the Shanghai metro, focusing in particular on the development of a realistic ground model based on a systematic characterisation of the available laboratory and field experimental evidence.

emphasis is placed on the characterisation of the soil behaviour and initial ground conditions, based on the experimental evidence collected from the literature.

1 INTRODUCTION The first appearance of a double circular shield tunnel dates to 1987, when it was patented in Japan. Since then the technology, known as the Double-O-Tube (DOT) shield tunnelling, has been applied primarily in Shanghai, Taiwan and Japan. It employs an earth pressure balance machine (EPBM) with multiple cutters which rotate in opposite directions to advance the con­ current boring of two tubes (Figure 1). This tunnel design achieves a more efficient use of the underground space, in particular in congested urban environments, which was the principal reason for its introduction on the Shanghai Metro system in 2002 (He et al. 2008). Six DOT tunnel sections have been constructed on the Shanghai Metro since then, in the ground conditions com­ prising generally soft silty clay and clayey silt layers (SGEAEB, 2002), and high ground water levels. The latter, together with the high perme­ ability of the soil, has contributed significantly to further complexity of DOT tunnelling in Shang­ hai in terms of the permeability conditions of the tunnel lining. Due to the high building and popu­ lation density in this urban area, accurate and robust assessment and prediction of the DOTinduced ground movements, lining forces and effects on the neighbouring infrastructure are essential. This paper presents advanced numerical ana­ lyses of DOT tunnel construction in soft ground conditions of Shanghai, based on the geometry of the metro Line 6. The analyses are hydromechanically coupled and performed using the Imperial College Finite Element Program ICFEP (Potts & Zdravković 1999, 2001). A particular

2 GROUND CONDITIONS 2.1

Ground profile

Shanghai is located in south-east China, at the estuary of the Yangtze River, and is therefore underlain by soft deltaic deposits. The geotech­ nical profile of this area is characterised by the high ground water table and thick soft clay layers. Tunnels are reasonably shallow, with typ­ ical depths of tunnelling at 10 – 20 m below the ground surface (Shen et al., 2014), where the soil type is mainly muddy clay (5h layer in Table 1), as defined in SGEAEB 2002. The ground profile in the numerical model is inter­ preted with 7 principal layers defined in Table 1, with a maximum depth of 35 m. The character­ isation of the soil mechanical behaviour for this study relied on limited data found in the litera­ ture and on the number of interpretations of both laboratory and field experiments reported in the literature. Table 1 summarises the values of bulk unit weight, γsat , and drained strength parameters c0 (effective cohesion) and ¢0 (angle of shearing resistance), presented in Sun (2007). 2.2

Ground water

The ground water table (GWT) varies from 0.5 m to 1.5 m beneath the ground surface (Luo et al. 2015). The numerical model adopts GWT at 0.5 m depth,

DOI: 10.1201/9780429321559-62

475

Figure 1. EPBM for DOT tunnelling (photo: geor.co.jp).

Table 1.

Shanghai ground profile (Sun, 2007).

Layers

Soil properties Depth c0 (kPa) ¢0 (°) (m) γsat (kN/m3)

Backfill Silty clay Sandy silt Muddy silty clay Muddy clay Silty clay – 1 Silty clay – 2

0~2 2~6 6~7 7~10 10~20 20~25 25~35

17.45 17.8 18.2 17.2 17.1 17.54 18.10

10 18 3 13 14 16 10

25 19.3 35 16 16 14 29

Figure 2. Interpreted permeability profile of Shanghai deposits.

for a conservative estimate of effective stresses in the soil. The pore water pressure distribution is hydrostatic both below and above the GWT, imply­ ing a small suction (5 kPa) at the ground surface, and the soil is considered fully saturated. The perme­ ability is found to be anisotropic and relatively high (Figure 2), the latter being a consequence of the high silt content (Zhang & Huang 2014). The coefficient of permeability in the vertical direction, kv , is gener­ ally higher than that in the horizontal direction, kh . 3 NUMERICAL MODEL 3.1

Soil modelling and initial ground conditions

A critical state-based Modified Cam Clay (MCC) model (Roscoe & Burland 1968) is considered appropriate for simulating the behaviour of soft clays. An extended MCC model (Potts & Zdrav­ ković 1999) has been adopted here to represent each of the layers identified in Table 1. Although this is an effective stress model, it is possible to specify an undrained strength profile for a clay soil from the model input parameters, as given in Equation 1:

; where σ0vi is the initial vertical effective stress in the ground, calculated from the bulk unit weights given in Table 1 and the adopted pore water pressure pro­ file; gðθÞ is the inclination of the critical state line; θ is the Lode’s angle; OCR is the overconsolidation ratio; is the earth pressure coefficient at rest for a normally consolidated (NC) state; K OC 0 is the earth pressure coefficient at rest for an overconsolidated (OC) state; λ is the gradient of the normal compres­ sion line and κ is the gradient of the swelling lines in the v - ln p0 plane, v being the specific volume and p0 the mean effective stress. The extended MCC model adopts a MohrCoulomb hexagon for the shape of the yield surface in the deviatoric plane (instead of the original circu­ lar shape), implying a constant value of ¢0 irrespect­ ive of the direction of loading (i.e. irrespective of the Lode’s angle, θ). The compression, λ, and swelling, κ, parameters for each layer were interpreted with a help of data from one-dimensional (1D, oed­ ometer) compression tests, as no isotropic compres­ sion tests were available. In this respect, 1D compression and swelling data were available for the

476

Table 2.

Compression parameters. Soil properties

Layers

Depth (m) λ

κ

l

Backfill Silty clay Sandy silt Muddy silty clay Muddy clay Silty clay – 1 Silty clay – 2

0~2 2~6 6~7 7~10 10~20 20~25 25~35

0.018 0.018 0.003 0.026 0.045 0.045 0.010

0.33 0.32 0.32 0.32 0.33 0.26 0.29

0.115 0.115 0.0216 0.171 0.291 0.205 0.065

Muddy clay, from which λ ¼ 0:291 and κ ¼ 0:045 were interpreted (Shen et al. 2008, see Table 2), resulting in the ratio λ=κ ¼ 6:47. For all other layers only 1D compression data was available, from which the λ values were interpreted, while the κ was then calculated by adopting the same λ=κ ratio as for the Muddy clay above (summarised in Table 2, together with the Poisson’s ratio, l, values). The undrained shear strength, Su , was inter­ preted from triaxial compression tests and cone penetration tests (CPT), after Lee et al. 1999, and presented in Figure 3 with a dashed grey line. Given the scatter in the data (in the ori­ ginal reference), it was considered that the

Figure 4. Profile of earth pressure coefficient at rest for the Shanghai ground conditions.

indicated strength increases above 5 m and below 22 m depth are unrealistic. The numerical model adopted the profile shown with a solid black line, which has a constant gradient of Su increase with depth and a smaller Su at the ground surface. This profile is consistent with Equation 1 and with a profile of the earth pres­ sure coefficient at rest, K0 , derived in Figure 4. The final extension of the MCC model involves coupling with a small strain stiffness overlay, for a more accurate simulation of ground movements. In this respect, the Imperial College Generalised Small Strain Stiffness model (ICG3S, Taborda & Zdravkovic 2012) was used to describe the dependency of the soil tangent shear stiffness, Gtan , on both stress and strain levels, as given by Equation 2:

Figure 3. Profile of undrained shear strength for the Shang­ hai ground conditions.

The model is calibrated against the available small strain stiffness measurements (Zhang et al. 2017) and shown in Figure 5, with parameters b ¼ 0:9, nG ¼ 1, Ed;r ¼ 0, a ¼ 0:121% and

477

Table 3.

Mechanical parameters of C50 concrete.

Symbol Description

C50 (MPa)

fck

32.5

fcm Ecm fcd ftk ftd

Characteristic cube compressive strength Mean compressive strength Elastic modulus Design compressive strength Characteristic tensile strength Design tensile strength

50.0 3.45E4 23.0 2.65 1.90

Figure 5. Normalised shear stiffness.

RG;min ¼ 0:1. The generalised deviatoric strain is defined as 3.2

Modelling of DOT lining

A single DOT ring consists of 11 segments (1.0 m wide in the out of plane direction), as shown in Figure 6, including 8 normal segments, one column segment, one small seagull segment

and one large seagull segment. The segments are pre-fabricated and made of C50 reinforced con­ crete, according to the Chinese code GB50010­ 2010 (2010), with the mechanical properties of the concrete summarised in Table 3. During their installation, the segments are connected with M27 bolts, in both the transverse and longitudinal directions. In the first instance, for the analyses presented here, the DOT lining was modelled as an elastic material, adopting the Young’s modulus of con­ crete E ¼ 34:5 GPa and the Poisson’s ratio l ¼ 0:2. 3.3

Figure 6. Cross-section of a DOT lining

Finite element analysis

Being a linear structure, the geometry of the prob­ lem was discretised in plane strain conditions, adopting an average depth of the tunnel axis of 13.7 m, as reported in Sun (2007) for the Shanghai Metro Line 6. The finite element mesh is shown in Figure 7, indicating the position of the tunnel axis and the GWT, as well as the boundaries between layers. The soil is discretised with 8-noded iso­ parametric quadrilateral elements. All soil layers are assumed to consolidate, with respective

Figure 7. Finite element mesh.

478

magnitudes of permeability as shown in Figure 2, and elements therefore have a pore pressure degree of freedom at corner nodes. The segments of the tunnel lining and joints between the seg­ ments are discretised with 3-noded beam elements. The base analyses presented in this paper assume an elastic and continuous ring, hence the joints, although existing in the mesh, have the same prop­ erties as the lining segments. The vertical boundaries of the mesh are pre­ vented from moving horizontally, while a zero force is prescribed in the vertical direction. The bottom boundary is prevented from moving in both the vertical and horizontal directions, while the top boundary is free to move and therefore has both vertical and horizontal forces at surface nodes prescribed as zero. In terms of hydraulic boundary conditions, zero change in pore water pressure is prescribed at vertical boundaries, while zero flow is prescribed at the bottom boundary (i.e. impermeable). The simulated construction sequence of the DOT tunnel is based on the realistic time-scale of the Shanghai Metro Line 6 construction and is summar­ ized in Table 4. After initialising the stresses in the ground, the soil elements within the tunnel boundary are excavated (removed from the mesh) in 50 incre­ ments over the period of 12 hours, matching the actual tunnelling speed of around 7-8 m/day. During the excavation, a normal stress of 2.5 kPa per incre­ ment is applied around the excavated boundary, simulating the support of the EPBM. A total of 125 kPa normal pressure is applied with this process, until the end of excavation. The lining construction is then carried out within a single analysis increment and subsequently the applied normal pressure is reduced to zero. Consolidation is allowed to continue into the long term.

Table 4.

An important aspect of analysis is the hydraulic boundary condition applied around the tunnel bound­ ary. As the soil is more permeable compared to, for example, stiff London clay (k ¼ 10-9 - 10-11 m/s; Chandler et al. 1990), no short-term suction is gener­ ated in the soil after the excavation. The lining instal­ lation includes an installation of a waterproof membrane, however this does not fully prevent the seepage of water across the tunnel boundary. Accord­ ing to Technical code for waterproofing of under­ ground works (GB/T 50108-2008), the maximum allowable water infiltration ratio is 0.1 l/(m2×d). The analyses have therefore considered the scenarios of the tunnel boundary after excavation being imperme­ able, fully permeable and partly permeable, adopting the above flow rate (base analysis). 4 RESULTS AND DISCUSSION The predicted ground movements and lining forces are presented and compared, where possible, against the available measurements on Line 6. In general, the engineering experience from the Shanghai Metro construction has shown that the ground movements tend to slow down within about 100 days postconstruction, after which the settlement rate is con­ sidered to have diminished (Ao & Zhang., 2011). Therefore, the available field measurements are pre­ sented only at 100 days post-construction. Figure 8 shows the measured and predicted (from the base analysis) surface settlement profiles. The field data are taken at the positions of two lining rings, R130 and R100 (Wei et al., 2011), both of which are slightly deeper than the average depth of the tunnel axis at 13.7 m, which was adopted in the analyses. The average maximum settlement is about 30 mm, the deeper ring moving slightly less than the

Excavation and construction sequences of the base analysis. Time

Boundary conditions on the tunnel boundary

Inc

Stage

ΔT

Total time Stress boundary

Hydraulic boundary

0 1~50 51 52 53 54~59 60~63 64 65 66~70 71~80 81~85

Initialising ground stresses Full excavation of DOT Lining construction Short term consolidation Short term consolidation Short term consolidation Short term consolidation Short term consolidation Short term consolidation Short term consolidation Short term consolidation Long term consolidation

0 10h 1h 1d 4d 3d 2d 10d 8d 72d 900d 77yrs

/ / / 1d 5d 8d 10d 20d 28d 100d 1000d 80yrs

/ Impermeable Impermeable Impermeable Impermeable Impermeable Impermeable Impermeable Impermeable Infiltration rate at 0.1 l/(m2×d) Infiltration rate at 0.1 l/(m2×d) Fully permeable

/ +2.5kPa/inc. (125kPa in total) Stress stays (125kPa) Stress stays (125kPa) Stress stays (125kPa) -25kPa/inc. (-125kPa in total) / / / / / /

479

Table 5.

Ground loss (%) at different depths. Stage

Depth (m)

End of excavation

100 days

1000 days

0(surface) 2 6 7 10

0.74 0.74 0.74 0.74 0.74

0.97 0.79 0.79 0.79 0.79

1.27 0.82 0.81 0.80 0.80

Figure 8. Measured and predicted (base analysis) surface settlement profiles.

shallower. The predicted maximum settlement of the model is 26 mm, which, being the shallowest depth, agrees with the measured range. The model also shows a much smaller settlement (around 16 mm) at the end of excavation, as well as only a slight increase of the maximum settlement (by about 10%) from 100 days to 1000 days post-construction. This too is in agreement with the above reported engin­ eering experience of tunnel construction in Shanghai that most of the ground movement happens within the first 100 days after tunnel construction. Further interpretation of the results of the base analysis is shown in Figure 9, which plots ground settlements at the end of excavation along the layer boundaries above the tunnel (see Figure 7 for loca­ tions of those). As expected, the settlement profiles become narrower with depth, with the maximum settlement also increasing. The calculated ground loss, expressed in terms of the volume of the soil within each of the settlement profiles, is summar­ ised in Table 5. The values at the end of excavation are practically the same, indicating that the applied excavation rates, which were simulated in the ana­ lysis, render the excavation process practically undrained. The structural forces in the lining, bending moment and axial (hoop) force, were measured in the long term, after about three years postconstruction and plotted in Figure 10 for half of the lining (Hu & Zhang, 2008). The numerical predic­ tion from the base analysis is also taken in the long

Figure 9. Settlement profiles at different depths below ground surface (tunnel axis depth 13.7 m; base analysis).

Figure 10. Measured and predicted lining forces in the long term.

term. It indicates a reasonably good agreement with the measured bending moment, in terms of both the magnitude and the distribution around the lining per­ imeter. The predicted axial force, while being com­ pressive, as measured, is smaller in magnitude. This needs further investigation as the initial assumption adopted in this study was to simplify the lining into a continuous elastic ring, without taking account of its segmental nature. Finally, a brief assessment is considered here of the effect that the prescribed hydraulic boundary condition at the tunnel boundary may have on the predicted ground movements. Apart from the base analysis, where this boundary is partially permeable as explained above, two more analyses were per­ formed, prescribing in one an impermeable bound­ ary (i.e. zero flow of water across the boundary) and in the other a fully permeable boundary (i.e. zero pore water pressure at the boundary). With respect to Table 4, the different boundary condition is applied from increment 51 to increment 80. During the first 50 increments of excavation the tunnel boundary in all three analyses is impermeable. The resulting surface settlement profiles are com­ pared in Figure 11, at 100 days after the end of exca­ vation. As expected, the maximum settlement calculated from the analysis with an impermeable lining is the smallest, but in this case comparable to the settlement resulting from adopting a partially permeable lining. The shapes of the settlement

480

Figure 11. Predicted surface settlements for different hydraulic conditions at the tunnel boundary.

troughs are also comparable. However, if the lining is fully permeable, the maximum settlement and the shape of the settlement trough depend on the size of the modelled domain. The settlement trough at end of excavation is the same for all three analyses, as the tunnel boundary is considered impermeable. 5 CONCLUSIONS This paper presents a numerical model, developed in the finite element code ICFEP (Potts & Zdravković, 1999; 2001), for the analyses of double-o-tube tun­ nelling technology as applied in the construction of the Shanghai Metro Line 6. Although the tunnel lining was simplified in this initial study and the available soil data was scarce, the established ground model, in which the soil behaviour is simu­ lated with an extended Modified Cam Clay model, shows significant level of accuracy in predicting the measured ground movements and structural forces. Further refinements, in particular in the modelling of the segmental nature of the concrete lining, will follow in subsequent studies.

ACKNOWLEDGEMENTS The work presented here is part of the PhD project of the first author at Imperial College London, who is financially sponsored by the China Scholarship Council. Their supported is greatly appreciated.

REFERENCES Ao, R. & Zhang, Y. 2011. Analysis of consolidation settle­ ments caused by shield tunnelling. Rock and Soil Mech­ anics, 32(7): 2157–2161. (in Chinese)

Chandler, R., Leroueil, S. and Trenter, N. 1990. Measure­ ments of the permeability of London Clay using a self-boring permeameter. Géotechnique, 40(1): 113–124. GB/T 50108. 2008. Technical code for waterproofing of underground works, Beijing: Standardization Adminis­ tration of China. He, C., Teng, L. and Yan, J. 2008. The double-o-tube shield tunnel in Shanghai soil. In: Geotechnical Aspects of Underground Construction in Soft Ground: Proceedings of the 6th International Symposium. Shanghai: 313–318. Hu, X. & Zhang, Z. 2008. Internal forces of DOT shield-driven tunnel lining. Chinese Journal of Geotech­ nical Engineering, 30(2): 172–180. (in Chinese) Lee, K., Ji, H., Shen, C., Liu, J. and Bai, T. 1999. Ground Response to the Construction of Shanghai Metro Tunnel-Line 2. Soils and Foundations, 39(3): 113–134. Luo, C., Shen, S., Han, J., Ye, G. and Horpibulsuk, S. 2015. Hydrogeochemical environment of aquifer groundwater in Shanghai and potential hazards to under­ ground infrastructures. Natural Hazards, 78(1): 753–774. Potts, D.M. & Zdravkovic, L. 2001. Finite element analysis in geotechnical engineering: application, London: Thomas Telford. Roscoe, K.H. & Burland, J. 1968. On the generalised stress-strain behaviour of ‘wet’ clay. J Eng Plast: 553–609. Shanghai Geological Environmental Atlas Editorial Board (SGEAEB). 2002. Shanghai Geological Environmental Atlas (SGEA). Beijing: Geology Press. Shen, S., Horpibulsuk, S., Liao, S. and Peng, F. 2008. Ana­ lysis of the behaviour of DOT tunnel lining caused by rolling correction operation. Tunnelling and Under­ ground Space Technology, 24(1): 84–90 Shen, S., Wu, H., Cui, Y. and Yin, Z. 2014. Long-term settlement behaviour of metro tunnels in the soft deposits of Shanghai. Tunnelling and Underground Space Technology, 40): 309–323. Sun, T.L. 2007. Research on the ground movement induced by the disturbance of multi-circular shield construction and its control technology PhD Thesis. Tongji Univer­ sity, Shanghai. (in Chinese) Taborda, D.M.G. & Zdravkovic, L. 2012. Application of a Hill-Climbing Technique to the Formulation of a New Cyclic Nonlinear Elastic Constitutive Model. Computers and Geotechnics, 43: 80–91. Wei, G., Zhu, K. and Chen, W. 2011. Ground settlement induced by double-O-tube shield tunneling under differ­ ent construction conditions. Chinese Journal of Geotechnical Engineering, 33(3): 477–482. (in Chinese) Zhang, Z. and Huang, M. 2014. Geotechnical influence on existing subway tunnels induced by multiline tunneling in Shanghai soft soil. Computers and Geotechnics, 56: 121–132. Zhang, J., Wang, W., Xu, Z., Li, Q. 2017. Laboratory test of small-strain characteristics of typical Shanghai cohe­ sive soils. Rock and Soil Mechanics, 38(12): 1001–1008. (in Chinese)

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Physical modelling of transient processes at the slurry supported tunnel face during shield excavation Z. Zizka METROPROJEKT Praha a.s., Prague, Czech Republic

B. Schoesser & M. Thewes Ruhr-Universität Bochum, Bochum, Germany

ABSTRACT: Stability of the tunnel face is a key design aspect for slurry shield tunnelling. Two require­ ments have to be fulfilled during excavation in order to achieve a stable tunnel face. The first requirement is a sufficient slurry pressure in the excavation chamber of the shield to balance the pore water pressure and the earth pressure. Moreover, the fraction of the slurry pressure, which exceeds the pore pressure, has to be applied as effective stress onto grains of the soil, whereas the efficient transfer of effective stress can be for­ mulated as the second requirement. The pressure transfer theory currently being employed in the practice expects a filter cake formation on the tunnel face or a slurry penetrated zone through which the pressure trans­ fer is assured. This assessment approach was developed originally for diaphragm walls and therefore the tran­ sient tunnel excavation process is neglected. The effects of the transient slurry penetration on the support pressure transfer under cyclic soil excavation at the tunnel face are investigated in this paper. Case A and Case B of the interaction at the tunnel face are introduced based on local comparison between slurry penetra­ tion and tool cutting depth. Case A stands for higher cutting depth than slurry penetration depth, while Case B represents shallower cutting depth than slurry penetration depth. Particular focus is given to the Case B of interaction in this paper. The interaction at the tunnel face are investigated in advanced 1-g experiments. The aim of the investigation is to characterize the pressure transfer and resulting tunnel face support efficiency for various combinations of slurry penetration and excavation scales. The obtained efficiencies are compared to contemporary used design methods in practice.

1 INTRODUCTION At the time of introduction of the slurry shield, the theories to describe the face support were taken from diaphragm wall technology, in which the bentonite slurry supports the open trench. The pressure transfer in diaphragm wall technology is summarized in DIN 4126 (2013) and by Kilchert & Karstedt (1984). The theory was implemented in tunnel face stability cal­ culations by Jancsecz & Steiner, 1994 and Anagnos­ tou & Kovari, 1994. The pressure transfer theories were updated by Broere & van Tol (2000), Bezuijen et al. (2001) and recently also by Xu (2019). This paper aims to physically model the cutting process at the tunnel face during excavation with a slurry shield that is important for the assessment of the transfer of slurry excess pressure into effective stress in soil. The paper focus on the Case B of inter­ action. In Case B, the cutting depth of a single tool is shallower than the depth of slurry penetrated soil, when observed locally (Figure 1). Case B results in a partial removal of the pressure transfer mechanism

with every pass of the cutting tool. Due to the only partial removal of the pressure transfer mechanism, the occurring change in the reactions is comparably less abrupt than in Case A. Thus, lower grade of het­ erogeneity of the pressure transfer mechanism can be observed within a cutting track and consequently on the entire tunnel face. It can be derived that the pres­ sure transfer will take place by pressure drop over the partially or fully formed pressure transfer mechanism. It can be here expected that the entire slurry excess pressure will be transferred, as the equilibrium condi­ tion requires. Considering the efficiency of the pres­ sure transfer, it is equally important as in Case A that the transfer occurs within the soil wedge as mentioned also in the theory by Anagnostou & Kovari (1994). The fundamental question in Case B of pressure transfer is the ratio of pressure transferred inside and outside of slurry penetrated zone. Theories used in practice (DIN 4126, 2013) assume that the transfer occurs entirely within the slurry penetrated zone and the gradient of slurry pressure (slurry stagnation gra­ dient) within this zone is believed to be constant.

DOI: 10.1201/9780429321559-63

482

Figure 1. Different interactions with the pressure transfer mechanism: Case A – left, Case B – right (Zizka et al., 2017).

It is expected for Case B of interaction that slurry re-penetration is the governing process at the tunnel face. Re-penetration is the process in which slurry penetrates the pore space of soil that already contains slurry fluid (including particles) from the previous excavation cycle. Due to the already present slurry particles in the soil skeleton, it is expected that repenetration lead to a shallower depth compared to primary slurry penetration. The first insight into transient slurry penetration processes at the tunnel face was suggested by Ana­ gnostou & Kovari (1994). They proposed that the infiltration of slurry takes place simultaneously with the removal of ground during excavation leading to a “quasi-steady state” after several excavation cycles. The quasi steady-state requires that global average slurry penetration velocity becomes equal to advance rate of the machine. This is also assumed by Bezuijen et al. (2016). According to Anagnostou & Kovari (1994), the quasi-steady state is reached by one of two possible processes. As a first option, when the average slurry flow velocity is initially lower than the advance rate, the penetration distance will be gradually reduced until the equilibrium is reached (=quasi-steady state). As second option in contrast, when the average infiltration velocity at some particular time is higher than the advance rate, the penetration distance will increase over the course

of time and, consequently, the average infiltration velocity will decrease towards equilibrium. For testing of slurry penetration behavior at the tunnel face, two new developed set-ups are introduced in this paper. Consequently, the results are presented and conclusions in terms of slurry penetration distance during excavation and slurry stagnation gradient during excavation with Case B are formulated. 2 CHARACTERIZATION IN THE COLUMN TEST 2.1

Set-up

The set-up can be designated as column test and con­ sists of a slurry cylinder, a soil cylinder (internal diameter 40 cm) and a reservoir with free surface for the discharged fluid from the soil cylinder. The dis­ charge reservoir is located on a scale. The scale is connected to a computer for continuous data logging during the experiment. One pore water pressure sensor (PWD) continuously monitors pressures in the slurry cylinder and 7 other PWD sensors are placed in the soil cylinder. Two total stress sensors are located at two measurements levels in the middle of the soil cylinder (Figure 2). Data from both sensor types are transferred to the computer. The data are logged every 0.25 s. In the developed set-up, two stiff plastic grids stabilize the soil sample, which is com­ pacted to a prescribed ratio (porosity). The effective surcharge stress induced by the grids on the soil sample is adjustable by screws and can be controlled. A further aim of the grids during slurry penetration is to ensure uniform distribution of slurry before enter­ ing the soil sample at the bottom and uniform outflow of displaced water at the top. The boundary between the bottom grid and the cylinder circumference is sealed. The chosen layout of sensors enables to track

Figure 2. A) Column test for investigation of re-penetration (Zizka, 2019), b) RUB tunnelling device.(Kuepferle et al., 2018), PWD=pore pressure sensor, WZ=total stress sensor c) Grain distribution curves of soils used in the investigation.

483

the reaction of soil during slurry penetration inside and outside of the slurry penetrated zone. The location of the sensors makes the tracking possible for only a certain number of combinations of slurry concentration, injection pressure and soil, due to particular target penetration depth. 2.2

Methodology

The details concerning homogeneous soil sample preparation can be found in Zizka (2019).The greatest challenge of the physical modelling of the repenetration is to achieve the same conditions that are present at the tunnel face during excavation. Either the depth of the primary penetration or the timespan at which the re-penetration starts must be chosen for the experiment. Moreover, in order to keep the pro­ cess realistic, the timespan for the duration of repenetration must be determined. It was concluded in the first section that the slurry in Case B reaches between the tool passages a particular penetration depth, which corresponds exactly to the cutting depth of the excavation tool at single passing (ptool = lr – see also Figure 3). This was designated on the global

level as a requirement for the equilibrium state during excavation (quasi-steady state) because of the move­ ment of the slurry penetration front that follows the movement of the shield. If the slurry would achieve a smaller penetration depth, no penetration of the sus­ pension would occur after some cutting cycles. If, in contrast, the depth of penetration during the repenetration would be greater in comparison to the cut­ ting depth, an infinitely high penetration depth of the slurry would occur after some cutting cycles. Hence, the slurry re-penetration depth needs to correspond to cutting depth of a single tool per passing and needs to be achieved during the timespan between two subse­ quent tool passes. These variables can be obtained from the evaluation of machine data from the excava­ tion. The determination of the primary penetration depth (lp), before the re-penetration starts, needs to be conducted parametrically. The re-penetration will be started at different timespans, from the beginning of the primary penetration until the previously formu­ lated spatial and time condition for equilibrium state during excavation are fulfilled (ptool = lr). Due to the absence of a cutting mechanism in the developed column test, the re-penetration has to be simulated in pressure-controlled steps. The principle of this method requires the increase of the slurry pressure at the start of the re-penetration (s2) in a way that the pore pressure in the soil at the adopted tool cutting depth (ptool) becomes exactly the same as the slurry pressure during the previous primary penetration (s1) in the chamber (Figure 2). The time­ span for starting the re-penetration has to be para­ metrically determined here, as described above, in order to fulfil the boundary conditions for a realistic re-penetration depth. The described approach for pressure-controlled repenetration is valid for certain for certain boundary conditions. In general, it is assumed that the concentra­ tion of slurry particles up to the end of primary pene­ tration stage is evenly distributed inside the slurry penetrated zone. This distribution is expected to remain during re-penetration, so the slurry flows inside the soil pores as a “bulk fluid”. As realistic approach, the boundary conditions are valid for coarse soils with corresponding large pore space. Hence, the investiga­ tion of the primary slurry penetration has to provide evidence, if the fulfilment of the requirements can be expected for the investigated soils. The main benefit of the methodology is the realistic time-scheduling of the experiment in comparison to the conditions at the tunnel face that avoids any unwanted slurry solidifica­ tion in the soil skeleton (Zizka, 2019). 3 CHARACTERIZATION IN RUB TUNNELLING DEVICE

Figure 3. Principle of the pressure-controlled repenetration experiment in the column test – movement of injection pressure p1 further in soil, the moving distance corresponds to the tool excavation depth (Zizka, 2019).

3.1

Set-up

To overcome the limitations and simplifications of the column test described in the previous section,

484

the RUB tunnelling device (Küpferle et al., 2016) was developed. It enables to investigate the transi­ ent slurry penetration process under cyclic excavation. A simplified cutting wheel is fixed at the end of a shaft (Figure 2-b-2 and 2-b-3) inside the soil cylinder and is propelled by the lathe. Cylindrical steel pins represent simplified cutting tools placed at the star-shaped cutting wheel. A single pin is always located within one cutting track. It is expected that the pins disturb the pressure transfer mechanism (slurry penetrated zone) in a similar fashion as scrapers. The fixing of the shaft to the lathe enables it to move forward and rotate the cut­ ting wheel while simultaneously creating the desired helix-movement of the cutting tools. The length of used pins depended on the experimental run 1 – 1.6 cm long. Thus, it has to be assured that the slurry penetration is deeper than 1 – 1.6 cm at the start of the excavation in order obtain Case B of the interaction. 3.2

Methodology

First of all, the soil cylinder, which will be later exca­ vated, is filled with water-saturated and compacted soil. The free area between lid and tunnel face repre­ sents a slurry chamber and is therefore filled with slurry (Figure 2-b). The slurry chamber and the center of the cutting wheel are fed by two separate slurry pipes. After closing, sealing and fixing the soil cylin­ der on the lathe, a defined pressure is applied to the entire cylinder using the slurry feed pipes. The pres­ sure is monitored by pore water pressure sensors in both the soil cylinder and slurry cylinders (PWD). After a defined period of time, the drainage port is opened. The slurry starts to penetrate into the soil and the support mechanisms begin to build up. The forma­ tion of the pressure transfer mechanism can be veri­ fied by the measured constant pore pressure in the slurry chamber and decreasing pore pressure in the soil outside of the slurry penetrated zone. Consequently, when the pressure transfer mechan­ ism is entirely built-up, the machine is started, and the cutting process begins. As pointed out before, a desired penetration rate (PR) and revolutions per minute of the cutting wheel (RPM) can be adjusted within a limited range. During excavation, the outflow from the soil cylinder and pore water pressures are measured and logged every 0.25 s (Zizka, 2019). Note that the drainage remains open during excavation. 4 MATERIALS AND TESTING PROGRAMME The materials used in the tests referenced in this paper are characterized in Table 1 (soils) and Figure 2.c and Table 2 (bentonite slurry).

Table 1.

Soils used in the experiments.

Soil fraction (no.) [mm] 3

Density [g/cm ] d10 [mm] Porosity [-] Water permeability coefficient [m/s]

Table 2.

(1)

(2)

1.00-2.00

0.063-4.00

1.58 1.15 0.40 (5-11)*10-3

1.63 0.07 0.39 6*10-4

Properties of used bentonite slurries. B1

Type Bentonite concentration

6% (60 kg in 1 m3)

Density [g/cm3] Yield point (ball harp) [Pa] pH [-] Marsh time 1000, 1500 [s]

1.037 58 9.4 55/105

4.1

Tests with column device

In the column device, only the poor graded coarse grained soil (1) was tested using the methodology of pressure controlled re-penetration. The pressure con­ trolled re-penetration would not be feasible for well graded Soil (2) (as turned out form the primary pene­ tration test). The testing pressure 0.4 bar was con­ sidered. First excavation scale assumed tool passing in each 60s and in the second the tool passes every 100s. In both scales, the tool cuts 40 mm of the slurry penetrated zone. Bentonite slurry from Table 2 was employed. Next to slurry re-penetration, primary penetration was also investigated for the outlined combinations. Each combination was reproduced three-times. 4.2

Tests with RUB tunneling device

Both soil (1) and soil (2) were tested with RUB tun­ neling device. The testing pressures were 0.2 bar and 0.4 bar for soil (1). For soil (2) the pressures were increased to 0.2 bar and 0.8 bar. Due to the size of the device, it was not possible to induce a realistic tool cutting depth per pass (per rota­ tion). On the other hand, the used lathe did not allow for RPMs realistic for shield machines. Hence, follow­ ing parameters characterizing the cutting process were adopted: • RPM = 71, PR = 0.1 mm/rev, AR = 7.1 mm/min It resulted in a pass of a cutting tool through a particular point at the tunnel face each 0.85 s and a tool penetration per passing 0.1 mm.

485

5 RESULTS, INTERPRETATION AND COMPARISON WITH OTHE AUTHORS The conditions at the tunnel face during excavation and Case B of the interaction are discussed. The slurry primary penetration test shown that the timedependent reaction of soil could be observed. The development of time-dependent primary slurry penetration depth for soil (1) is shown in Figure 4. It was found using the column device, that excess pore pressure had linear distribution inside the slurry pene­ trated zone during primary penetration of soil (1) (Figure 5). Therefore, it was designated Case B-1. In contrast, the distribution of excess pore pressure in the slurry penetrated zone was non-linear for soil (2) during the primary penetration in the column test. It is shown in Figure 7. Therefore, it was designated as Case B-2. The investigation of the re-penetration in the column tests consequently showed fundamentally dif­ ferent behavior in comparison to primary penetration.

Figure 4. Slurry penetration depth during excavation in soil fraction 1-2 mm for two excavation scales compared with primary penetration depth (Zizka, 2019).

Figure 5. Excess spore pressure distribution during pri­ mary penetration and re-penetration in advanced column test for soil (1) (Zizka, 2019).

Eq. 1 shows the definition of the slurry stagnation gradient based on DIN 4126, which governs the pressure transfer process in Case B. Based on the stagnation gradient, pressure transfer efficiency can be determined. The procedure is utilized later in this section.

where fs0 = Stagnation gradient of slurry [kN/m3], a = Empirical factor from the experiments 2 or 3.5 [-], d10 = Characteristic grain size of soil (10 % pas­ sage in sieve analysis) [mm] τf,s = Static yield point of slurry [Pa], Δs = Slurry excess pressure [kPa], lmax,calc [m]=maximal slurry penetration depth. 5.1

Slurry penetration depth

There is a presumption that the slurry penetration during excavation depends on the distribution of pore pressure in the slurry penetrated zone (stagna­ tion gradient of slurry) during primary penetration. For the soil fraction 1- 2 mm (1), the column test demonstrated, that the distribution of pore pressure was almost perfectly linear regardless of slurry injec­ tion pressure (Zizka, 2019). Thus, it could be expected that the slurry penetration depth during excavation will be reduced in comparison to primary penetration. Figure 4 shows the experimentally determined primary slurry penetration depth within the column test is shown for soil fraction 1-2 mm (1) – dotted line. In Figure 4, the fluctuations of slurry penetra­ tion depths for two excavation scales are also shown (orange and blue line). The two particular excavation scales are reaching a different total slurry penetration depth during an excavation cycle. For the excavation scale 100s adopting a tool passing every 100 s and the cutting depth 40 mm, the maximum reached penetration depth equals to the maximal primary slurry penetration depth. The reduced slurry penetra­ tion depth was obtained for the faster excavation scale adopting a tool passing within every 60 s. Based on the results, the theoretical prediction by Anagnostou & Kovari (1994) could be confirmed that the equilibrium point can be reached by slurry penetrating in average faster during primary slurry penetration than a shield excavation only due to repenetration. The decrease of average slurry penetra­ tion velocity due to re-penetration can be easily derived from Figure 4. Broere & van Tol (2000) also expect a reduced slurry penetration depth during excavation. Their calculation approach would also deliver different slurry penetration depths for soil fraction 1-2 mm (1) for the two investigated excavation scales. The aver­ age penetration depth within a cycle in front of the tunnel face would be 0.12 m for the scale with tool

486

passing every 60 s and 0.14 m for the tool passing every 100 s. Hence, the approach reflects the influ­ ence of the excavation scale properly in comparison to the experimental results. But it predicts slightly lower slurry penetration depth. It has to be kept in mind that the approach by Broere & van Tol (2000) does not consider eventual differences in the cutting depth of tools per passing. Moreover, the approach expects that the slurry excess pressure, which did not drop over the slurry penetrated zone transforms in groundwater flow. This expectation could not be confirmed for soil fraction 1-2 mm (1) due to the increased stagnation gradient during penetration pro­ cess (see next section). Bezuijen et al. (2001) and Bezuijen et al. (2016) do not focus on the prediction of slurry penetration depth during excavation in detail. Bezuijen et al. (2016) assume that the slurry penetration front moves together with the slurry shield. Accordingly, the slurry should adopt the velocity of movement from the shield. This assumption corresponds to the equilibrium state (=quasi-steady state) mentioned in the introduction. The column experiments for the investigation of the re-penetration were conducted while physically simulating a passing of a single cutting tool. Thus, no interaction between cutting tracks was considered in the column experiments. It could be expected that the interaction would cause a sort of smearing effect, but the overall mechanics of moving of slurry pene­ tration front with the shield will be preserved. The argument can be supported by the statement that zones immediately after cutting and zones with longer times since cutting do not differentiate dra­ matically in terms of their hydraulic properties (Zizka, 2019). The expectation stated in the intro­ duction for reduction of slurry penetration distance during excavation could be confirmed. The previously discussed reduction of the penetra­ tion depth during excavation determined in column test could not be confirmed by observations in the RUB tunnelling device for soil (1). The slurry penetra­ tion depth cannot be determined directly inside the device and had to be predicted based on pore pressure distribution. The result is explainable due to excavation scale adopted in the RUB tunnelling device. Moreover, from the global pore pressure distribution, a local slurry penetration depth cannot be determined with a high accuracy. The slurry penetration depth in soil fraction (2) 0.063-4 mm is not discussed here, because it could only be evaluated based on pore pressure distribution within the RUB tunnelling device. Hence, it is assessed within the next section together with the stagnation gradient. 5.2

expressed by inclination of an imaginary line placed between excess slurry pressure in chamber at the machine and a point with zero excess pressure in the ground. It is assumed that the slurry penetrates up to the point with zero excess pressure. As could be expected from the development of the penetration depth (Figure 4), not only the depth during repenetration is changing in soil 1-2 mm, but also the stagnation gradient is changing. To determine the stag­ nation gradients more properly, the measured excess pore pressure has to be evaluated. In Figure 5, the pore pressures are shown as scaled percentages of the pore pressure at the injection point to improve the comparison. It turns out that the pore pressure distributions for soil 1-2 mm are almost linear. Thus, using the linear regression based on least square method, the stagna­ tion gradient for the particular experimental combin­ ation and adopted time can be determined. The average obtained pore pressure distribution (Figure 5) and stagnation gradients (Figure 6) are shown. In Figure 6, the measure stagnation gradients are shown. Additionally, a stagnation gradient calculated using Eq. 1 acc. to DIN 4126 (2013) is also shown here. The calculated gradient distinguishes an assump­ tion for the empirical factor “a” in Eq. 1. It can be seen that the calculation adopting a=2 is significantly under­ estimating the measured gradients, while the formula from DIN 4126 (2013) with a=3.5 is approximating the final stagnation gradients very well. Generally, it can be seen that the stagnation gradients obtained for the re-penetration are slightly higher than for the pri­ mary penetration. This could be expected due to lower penetration depth as discussed in the previous section. Anagnostou & Kovári (1994) suggested an approach to calculate the stagnation gradient while considering the influences of excavation. The stagna­ tion gradient during excavation was calculated for two excavation scales from the final gradient

Stagnation gradient of slurry

The stagnation gradient of slurry is defined according to Eq. 1 as slurry excess (pore) pressure drop divided by slurry penetration distance. Thus, the gradient is

Figure 6. Obtained slurry stagnation gradient in soil (1) (Zizka, 2019).

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measured in combination IV at t=60s (primary pene­ tration). The comparison of the calculated gradients with measured during primary penetration and repenetration is shown in Figure 6. The comparison shows that the theory by Anagnostou & Kovári (1994) shows only a very limited increase in the stag­ nation gradient due to excavation, which is lower in comparison to corresponding settings of the repenetration experiments. In contrast, Broere & van Tol (2001) consider the equal stagnation gradient (fs0) both for static case and excavation. In their theory, the gradient calculation adopts the approach included in DIN 4126 (2013) while considering a=3.5. Thereby, the theory by Broere & van Tol (2000) do not reflect the obtained experimental results. A different situation appears when the stagnation gradient during excavation is evaluated for soil frac­ tion (2) 0.063-4 mm. Note that the results were obtained from investigation by RUB tunneling device. Excess pore pressure during excavation in RUB tunnelling device are shown Figure 7. The soil fraction (2) 0.063-4 mm was character­ ized with non-linear pore pressure distribution inside the slurry penetrated zone during primary slurry penetration (Zizka, 2019). Using the same evaluation methodology as for soil fraction (1) 1-2 mm based on linear regression, the stagnation gradient was determined. The stagnation gradient is shown together with pore pressure distribution in Figure 8. It can be seen that DIN 4126 (2013) delivered comparably higher stagnation gradient than meas­ ured in the column test for soil (2) under static con­ dition at 120 s since the experiment start. The theory for calculation of the stagnation gradient during excavation acc. to Anagnostou & Kovári (1994) again showed an increase of this gradient as for soil 1-2mm. To characterize the pore pressure distribu­ tion from the RUB tunnelling device, the curves were approximated. Note that only the curves of PWD 4 and 5 are considered for the approximation here due to terminated pressure transfer reformation.

Figure 7. Excess spore pressure distribution during re-penetration in RUB tunnelling device and primary pene­ tration in column test for soil (2) (Zizka, 2019).

Figure 8. Obtained slurry stagnation gradients in soil (2) (Zizka, 2019).

The approximation of pore pressure distribution was divided into two branches. Linear approximation was considered for the area of 6 cm in front of the cutting tools, which approximately corresponds to the extent of slurry penetrated zone during static pri­ mary penetration. From the distance 6 cm onwards, the distribution was approximated by the theory by Bezuijen et al. (2016). Thus the measured pore pres­ sure at the distance of 6 cm was considered as “Excess pore pressure at the interface between pres­ sure transfer mechanism and pure soil”. Subsequently, the pore pressure was input as “Pore pressure head in the excavation chamber” to calculate the non-linear part of the pore pressure distribution. It is possible to see from Figure 7 that the division into two branches fits the measured results well. The comparison of the approximated stagnation gradient from the RUB tunnelling device with the measured in column test shows that during excava­ tion, a lower stagnation gradient was obtained than during static conditions due to existence of increased pore pressure also outside of the slurry penetrated zone. Hence, a fundamentally different behavior was obtained here than for the soil fraction 1-2 mm (1). Based on the pore pressure development obtained for the soil fraction 0.063-4 mm (2), the expected slurry penetration distance assumed in introduction can be revisited. It is expected that the slurry pene­ tration distance during excavation does not change in the device and only the amount of flow through the tunnel face is changing. The obtained experimen­ tal results were obtained with the RUB tunnelling device, hence, they reflect a simplified assessment for the whole tunnel face.

488

6 CONCLUSION AND OUTLOOK

REFERENCES

The significance of the pressure transfer mechanism interaction with cutting tools for the pressure transfer in Case B was investigated in this paper. The column tests were executed for the investigation of the primary slurry penetration and the re-penetration. Re-penetra­ tion tests were conducted while simulating a passing of a single cutting tool by pressure controlled method­ ology. The investigation of the re-penetration behavior with the column tests consequently showed fundamen­ tally different behavior in comparison to primary penetration. Focusing on the re-penetration, Case B-1 (linear pore pressure distribution) delivered higher slurry stagnation gradients during excavation. In contrast, Case B-2 (non-linear distribution) showed decreased slurry stagnation gradients during excavation in comparison to static conditions. The experimental program employed in this paper has certain limitations. The program focused primary on investigations using a poor graded coarse sand (soil 1) from the standard application range of slurry shield. The soil fraction show the linear distribution of pore pressure inside the slurry penetrated zone. The obtained results can be generalized for all soils with such a distribution. Additionally, a well graded sand (soil 2) was investigated. This soil showed non­ linear distribution of pore pressure inside the slurry penetrated zone. Hence, the obtained results for these soils can be generalized for soils with such a distribution. Due to the non-linearity, it is neces­ sary to conduct additional experiments with other well graded sands and with slurry with different con­ centrations of solids in slurry to confirm the expected generalization of the results.

Anagnostou, G.; Kovari, K. (1994). The face stability of slurry-shield-driven tunnels. In Tunnelling and Under­ ground Space Technology 9(2), pp. 164–174. DOI: 10.1016/0148-9062(94)90438-3. Bezuijen, A.; Pruiksma, J. P.; van Meerten, H. H. (2001). Pore Pressures in front of tunnel, measurements, calcu­ lations and consequences for stability of tunnel face. Kyoto, Japan. In Modern Tunneling Science and Tech­ nology: Proceedings of the International Symposium, pp. 27–33. Bezuijen, A.; Steeneken, S. P.; Ruigrok, J. A. T. (2016): Monitoring pressures and analysing pressures around a TBM. Prague, Czech Republic. In Book of abstracts and e-Proceedings of the 13th international conference Underground Construction, pp. 1–8. Broere, W.; van Tol, A. F. (2000): Influence of infiltration and groundwater flow on tunnel stability. Tokyo, Japan. In Geotechnical Aspects of Underground Construction in Soft Ground, pp. 339–344. DIN 4126: 2013–09: Nachweis der Standsicherheit von Schlitzwänden. DIN 4127: 2014–02: Erd- und Grundbau – Schlitzwand­ tone für stützende Flüssigkeiten – Anforderungen, Prüf­ verfahren, Lieferung, Güteüberwachung. Kilchert, M.; Karstedt, J. (1984): Band 2 Standsicherheits­ berechnung von Schlitzwaenden nach DIN 4126. Wies­ baden Berlin: Bauverlag GmbH (Beuth-Kommentare). Küpferle, J.; Zizka, Z.; Schoesser, B.; Röttger, A.; Alber, M.; Thewes, M.; Theisen, W. (2018): Influence of the slurry-stabilized tunnel face on shield TBM tool wear regarding the soil mechanical changes - Experi­ mental evidence of changes in the tribological system. In Tunnelling and Underground Space Technology 74, pp. 206–216. DOI: 10.1016/j.tust.2018.01.011. Xu, T. (2019): Infiltration and Excess Pore Water Pressures in Front of a TBM: Experiments, Mechanisms and Com­ putational Models, PhD thesis, Gent university Zizka, Z.; Schoesser, B.; Thewes, M. (2017): Excavation cycle dependent changes of hydraulic properties of granu­ lar soil at the tunnel face during slurry shield excavations. In 9th Symposium on Geotechnical Aspects of Under­ ground Constructions in Soft Ground (IS-São Paulo). Zizka, Z.; Schoesser, B.; Thewes, M.; Schanz, T. (2018): Slurry Shield Tunneling. New Methodology for Simpli­ fied Prediction of Increased Pore Pressures Resulting from Slurry Infiltration at the Tunnel Face Under Cyclic Excavation Processes. In Int J Civ Eng 15(4), p. 387. DOI: 10.1007/s40999-018-0303-2. Zizka, Z. (2019): Stability of slurry supported tunnel face considering the transient support mechanism during excavation in non-cohesive soil, Doctoral thesis, RuhrUniversitaet Bochum.

ACKNOWLEDGEMENT In this paper, the results from the subproject A6 “Locally transient face support within hydroshields” are presented. The subproject A6 is a part of the “Collaborative Research Centre – SFB 837” at Ruhr-University Bochum in Germany founded by DFG (Deutsche Forschungsgemeinschaft). The authors want to thank to Jakob Küpferle and Arne Röttger who enabled the use of the RUB tunneling device for the investigations presented in this paper.

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Ground movements, interaction with existing structures and mitigation measures

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Interaction between a newly excavated underground ramp and deep existing tunnels A. Afshani, G. Hassan & H. Akagi Department of Civil Engineering, Waseda University, Tokyo, Japan

K. Endou Metropolitan Expressway Co, Tokyo, Japan

ABSTRACT: The close underground excavations inevitably become a common issue in large cities. The proximity of new excavation to existing structures changes the stress-strain regime of soil and can impose a considerable amount of load on nearby structures. A recent excavated case of an underground ramp that eases the access from underground tunnels to aboveground roads provided the chance to study proximity underground excavation and its effect on nearby structures. The ramp was excavated from the ground surface using an earth pressure balanced machine and after passing of a spiral route, it approaches and then connects to a deep existing target tunnel. The lining deformation of the target tunnel monitored before and after the passing of the excavation machine through several monitoring sections along the excavated ramp. The results of the previously performed beam-spring model related to this close excavation are presented. The proximity case of underground constructions is also simulated numerically and verified with measurement results. The verified numerical model is employed to predict the changes in the horizontal earth pressure acts on the target tunnel before and after passing of the machine in the last monitoring section of the ramp when the ramp and target tunnel are almost parallel. The results showed that among the horizontal earth pressure and internal structural forces in the target tunnel, the amount of bending moment changes as large as 35% by close excava­ tion of nearby new tunnel.

1 INTRODUCTION New excavation nearby to existing structures becomes a challenging issue in cities as it changes the stress-strain regime of soil and impose extra internal forces on nearby structures. The interaction between the newly constructed tunnel and the exist­ ing underground structure was studied in past years by numerical and experimental means (Ghaboussi & Ranken 1977; Soliman et al. 1993; Addenbrooke & Potts 2001; Chehade & Shahrour 2008; Wang et al. 2019; Afshani et al. 2020). Chehade and Shahrour (2008) indicated there is no significant interaction effect for distance more than 3D between tunnels which D is tunnel diameter. Wan et al. (2016) stud­ ied field instrumentation of the characteristic of the ground surface response due to tunneling in the proximity of existing structures in London Clay. In their study, vertical and horizontal surface ground displacements measured at an instrumented site in Hyde Park during the passage of two EPBMs have been analyzed and discussed. The extensive inter­ pretation is done by comparing the settlement trough from the field reading to the Gaussian curve. Most of

the intensive field observations are taken using vari­ ous methods, such as precise leveling, total stations, and micrometer sticks. Ng et al. (2015) evaluated the impact of new tunnel construction on nearby existing tunnels by performing a series of three-dimensional centrifuge model tests and numerical calculations in dry sand. They concluded that the stress relief caused by the new tunnel not only led to a reduction in the vertical stress at the invert but it also resulted in substantial stress reduction at the spring line of the existing tunnel. Afshani et al. (2020) reviewed the effect of close underground excavations for tun­ nels having elliptical-shaped cross-section, espe­ cially in coarse-grain soil. In this study, it is demonstrated that the influence distance of close tun­ neling in the case of elliptical cross-section is approximately 4D where D is the tunnel minor diameter. To the best knowledge of authors, there is no inter­ action study between tunnels with clear distance between tunnels less than 0.15D which D is the tunnel diameter. In this paper, excavation of an underground ramp using an earth pressure balance shield machine in a curved alignment in both vertical and horizontal

DOI: 10.1201/9780429321559-64

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directions is discussed. The ramp is excavated from the ground surface along a path with a spiral route and approaches and then connect to a deep existing target tunnel. The lining deformation of a target tunnel moni­ tored before and after the passing of the excavation machine through several monitoring sections along the excavated ramp. The results of the previously per­ formed beam-spring model related to this close exca­ vation is presented in the form of generated internal forces in the target tunnel. This tunneling cases is also simulated numerically and verified with measurement results. The verified numerical model is employed to predict the changes in the horizontal earth pressure and internal forces of target tunnel lining before and after passing of the machine in the last monitoring sec­ tion of the ramp when the ramp and target tunnel are almost parallel.

10.13 m respectively. There are two existing largediameter tunnels (ETL and ETR tunnels shown in Figure 2) with overburden of nearly 50 m, a diameter of 12.3 m and a thickness of 40 cm located in fine sandy soil (Ks). The A ramp connects to the existing tunnel on the left, hereafter called ETL, after section 5 (the connection part has not been shown.). The ramp A mainly passes through the Dc and Ks layers. The in-situ and laboratory tests are done to obtain parameters of each layer. Table 1 lists the geotechnical parameters of each soil layer. 2.2

Measurement data

At the end part of A ramp where it connects to the main tunnel ETL, in addition to the section 5 shown in Figure 1, 10 other monitoring sections

2 MEASUREMENT DATA 2.1

Site introduction

To study proximity excavation, measurement data from an excavation site of a tunneling case in the south of Tokyo is used in this study. It is an under­ ground ramp that creates the access from under­ ground tunnels to aboveground roads. It is a part of a project called north highway ring road which includes highway and underground tunnels. Along the underground tunnels, several exit and entrance ramp were used to increase accessibility between tun­ nels and aboveground highway roads. The ramps were excavated from the ground surface using an earth pressure balanced machine and after passing of a spiral route, it approaches and then connects to a deep existing target tunnel. Here, one of the ramps called A ramp with a length of 459.8 m and a total number of 316 rings is used. Figure 1 shows the top view of the site and location of A ramp. Figure 2 shows the transverse cross-section at the location of section 5 shown in Figure 1. Top surface soil layers in the site are mainly loam (Lm), clay (Dc) and sand (Ds), with standard penetration test (SPT) values between 5 to 15. Under these layers, there are sand (Ks) and sandy mudstone (Kms, Km) deposits with large SPT value. The A ramp has a diameter and thickness of 9.9 m and 35 cm respectively. The ramp was excavated using a shield machine with the length and diameter of 10.845 m and

Figure 1. The general plan of the site.

Figure 2. Details of transverse section 5. Table 1. Soil layer parameters obtained from in-situ and laboratory tests and concrete lining properties. Layers Units Loam (Lm) Clay (Dc) Gravel (Ds) Sandy Mud (Kms) Sand (Ks) Mudstone (Km) Concrete lining A -ramp Concrete lining ETL, ETR

Nspt γ

E

c

ϕ

ν

kPa 54 160 48 1800 75 2000 -

˚ 0 0 34 11 42 7 -

­ 0.45 0.4 0.3 0.3 0.3 0.3 0.17

-

0.17

15 5 8 > 50 > 50 > 50 -

kN/m3 14 15.5 18.5 19 19.5 18.5 26

MPa 7.8 19 20 500 78 450 39000

-

26

33000 -

Nspt is the value obtained from Standard Penetration Test, γ is wet unit weight, E is elastic modulus, c is cohesion, ϕ is internal friction angle and ν is Poisson ratio.

494

are considered to observe the close construction effect. The 10 surveying lines each with three stations of A, B, and C are monitored with sur­ veying instruments. The stations A and C are located in the crown and spring line and station B is located between them. Figure 3 shows the top view of connection part and also a cross-section of survey line 1 when clear­ ance between two tunnels is approximately less than 0.15D (the clearance is 1.413 m and diameter of A ramp is D = 9.9 m). The illustrated numbers along A ramp in Figure 3 are ring numbers (316 rings in total). The advancement of shield machine in the ramp is expressed by ring numbers. The length of rings along ramp changes as the ramp is not straight. The vertical movement at station A in crown and horizontal displacement at station C in the spring line are important and their results are discussed here. Figure 4 shows vertical displacement of station A in the ETL tunnel at survey lines 1, 4, and 8 recorded at the entire of the time when the shield machine moves along the A ramp. Locations of survey lines 1, 4, 8 and also section 5 are demonstrated in Figure 3. It can be seen that upon arrival of the machine in A ramp, the crown of the ETL tunnel (station A) moves down by a maximum of 6 mm. Figure 5 also shows horizontal displacement of the ETL tunnel spring line (station C) at survey lines 1, 4, and 8. At the time of machine arrival in A ramp to the vicin­ ity of survey lines, the spring line of ETL tunnel

Figure 4. Vertical displacement of station A at survey lines 1, 4, and 8.

Figure 5. Horizontal displacement of station C at survey lines 1, 4, and 8.

Figure 3. (Top) Plan view, and (bottom) cross-sectional view of the survey line 1 at end part of ramp A and ETL tunnel.

(station C) displaces toward A ramp by more than 10 mm. By approaching of machine head toward survey lines, no noticeable displacements in the lining of the ETL tunnel are found which indicates that the magnitude of face pressure is equal or less than insitu soil and water pressures. The measurement data indicates that due to the excavation of A ramp, and probably inducing of some percentages of stressrelaxation in soil around A ramp, the soil and lining are drawn toward the gap spaces around excavated area in A ramp. The maximum horizontal displace­ ment at the ETL spring line is larger than vertical crown displacement as the lining spring line is closer to A ramp. Figure 6 shows the schematic of the overall meas­ ured response of the existing main tunnel ETL to the

495

Figure 6. Overall measured response of ETL to tunnel excavation along A ramp.

tunnel excavation along A ramp. Figure 7 shows the shield machine speed and ring number in which face of machine locates (ring 270 ~ ring 316) by date of excavation. According to Figure 7, rate of machine advancement changes between 15 to 20 mm per min­ utes (21.6 to 28.8 m per day) and in the last few rings, it drops to 7.4 to 15 mm per minutes (10.65 to 21.6 m per day). On date 8/27 (Sunday), excavation is interrupted and speed of machine at the start of the next day declines. Figure 8 shows correlation between speed of the shield machine and horizontal displacements of the ETL at stations C of all survey lines (survey line 10 to survey line 1 as illustrated in Figure 3-top). This correlation is expressed in terms of survey line numbers and date of excavation. The direction of excavation is from survey line 10 toward survey line 1. As machine approaches end of the ramp, horizontal displacement of ETL shows increasing trend, and speed of machine decrease. It should be also noted that not only the speed of exca­ vation, but also the distance between ramp A and

Figure 7. Ring number that machine face locates (R 270 ~ R316) and rate of machine advancements by date of excavation.

Figure 8. Measurement data of (left) horizontal displace­ ment of ETL at station C of all survey line, (right) machine speed by date of excavation and numbers of rings along A ramp.

ETL is influential on the horizontal displacement (the net distance between external diameter of ETL and ramp decreases). 3 BEAM-SPRING MODEL The beam-spring model is used for structural calcu­ lation of the tunnel lining in transverse and longitu­ dinal directions. The member forces of segmental rings in the transverse direction is calculated by knowing the amounts of load and characteristics of reinforced concrete lining. The segmental rings are modeled as a beam with rotational and shear springs. Figure 9 shows the schematic of the beam-spring

Figure 9. Schematic of beam-spring model for two neigh­ bor rings.

496

model for two neighbor rings and Table 2 lists the parameters employed in this model. In Table 2, Kv represents axial ground spring, Km shows rotational spring between segment and segment, and Ks is shear spring between ring and ring. The Kv coeffi­ cient is obtained from the value of the Standard Penetration Test of Ks soil in listed in Table 1 and the correlation proposed by Tunnel Structural Design Specification of Metropolitan Expressway (2008). The Ks coefficient is obtained from an equa­ tion recommended by the Railway Technical Research Institute (2006) as follow:

where E = elastic modulus of bolt between ring to ring, I = moment of inertia of bolt section, b = is segment width of ETL (= 2 m). The value for Km is obtained from the literature. The model is used to examine the structural behavior of ETL tunnel lining under the loads of self-weight, soil and water pressure, and the extra loads ori­ ginated from machine operation in A ramp. The self-weight, and soil and water pressures are cal­ culated readily using properties of soil layers and lining specifications. Figure 10 shows the general imposed loads on a target tunnel (ETL) when the machine face and tail passes along a nearby tunnel. It shows that by face passing of machine along new tunnel, lining of target tunnel is pushed toward its center by horizon­ tal pressure of qf, while during machine tail pass, lining is pulled by the effect of horizontal distributed pressure qt. The extra loads due to the machine oper­ ation (qf and qt in Figure 10) along A ramp can be obtained by numerical analyses. The details of the numerical model are given in the next section. According to the field measurements, during the both face and tail pass of the machine, ETL lining is pulled toward A ramp and therefore, in the case of this interaction problem, direction of qf horizontal

Table 2. tunnel.

Beam-spring model parameters for existing left

Parameters

Unit

Value

Tunnel Diameter (Do) Thickness of segment (t) Width of segment (w) Elastic modulus of segment (Ec) Poisson ratio Axial stiffness of spring (Kv) Rotational stiffness of spring (Km) Shear stiffness of spring (Ks)

m m m GPa υ MN/m3 MN.m/rad MN/m

12.3 0.4 2.0 39 0.3 50 15 79

Figure 10. Load system on beam-spring model when (a) machine face passes, and (b) when machine tail passes.

pressure should be opposite of the one shown in Figure 10 (a). To consider the effect of machine face and tail pass, some degree of stress-relaxations is applied. The amount of relaxation is decided based on the amount of deformation in the lining of the ETL tunnel. The amount of stress-relaxation is dis­ cussed in the next section. The values of qf1, qf2, qt1 and qt2 loads are read and used as an input loads in beam-spring model. After performing analyses in the beam-spring model, the maximum and minimum bending moment and corresponding values of axial force per ring are listed in Table 3. Table 3. tunnel.

Results of beam-spring model on existing left

Unit Range of values

497

Axial force

Bending moment

kN/Ring 4710, 5343

kN.m/Ring - 373.3, +376.8

4 NUMERICAL MODELING The plane-strain numerical model is used to study close construction effect between A ramp and ETL tunnel. The soil layer stratigraphy shown in Figure 2 and the soil and concrete lining properties listed in Table 1 are used in the model. The linear elastic with Mohr-Coulomb failure criteria and linear elastic models are used for soil layers and concrete linings respectively. Employment of three-dimensional numerical model is more close to reality to estimate close construction effect, however, as A ramp and ETL tunnel become almost parallel nearby to section 5, here plain-strain model is used. Figure 11 shows the plain-strain numerical model. By using the model, it is intended to stimulate the lining deformation in ETL same as the one that occurred in the field (see Figure 4, 5, and 6) during the face and tail passing of shield machine through the survey line 1. After the creation of an equilibrium state under the self-weight of soil layers, ETL is excavated and its lining is installed. Then, the amount of stress-relaxation for the face and tail pass­ ing of machine are decided. These values are decided by fitting the lining deformation of ETL tunnel with the measurement values shown in Figures 4 and 5. The initial stress in the boundary of A ramp is reduced by 10% to simulated the face passage of the machine and for the tail passage of machine, another more 10% stress relaxation is applied and then lining of A ramp is installed. Figure 12 shows predicted and measured vertical displacement of station A at survey line 1 and Figure 13 indicates predicted and measured horizontal displacement of station A at survey line 1. By using the total stress relaxation of 10% and 20% in numerical model, the overall displacement values of ETL tunnel are close to the values of field measure­ ments. According to Figure 12, the calculated vertical displacement at station A is slightly larger than the measured ones. One possible reason for this might be due to the non-uniform existence of gap and differ­ ent percentage of stress-relaxation around A ramp in the field.

Figure 11. Numerical model.

Figure 12. Predicted and measured values of vertical dis­ placement of station A at survey line 1.

Figure 13. Predicted and measured values of horizontal displacement of station C at survey line 1.

The calibrated model is then used to examine changes of the internal forces in ETL lining and changes of horizontal soil pressure between A ramp and ETL tunnel. Figure 14 shows axial force in ETL tunnel lining at three steps (a) after ETL lining

Figure 14. Predicted axial forces in ETL tunnel for various construction steps.

498

installation, (b) after passing of machine face in A ramp, and (c) after passing of machine tail pass in A ramp. The largest axial force occurs at both left and right spring lines and the biggest changes due to close tunneling happens at left spring line (θ = 270˚). At left spring line, magnitude of axial force changes from 3634 kN to 3904 kN (almost 7% increase in the amount of axial force). Figure 15 shows the dis­ tribution of bending moment in ETL tunnel lining by using the same construction steps explained previ­ ously in this part. Maximum negative and positive bending moment occurs at 45˚ and 85˚ clockwise from crown respect­ ively. The largest changes in the bending moment hap­ pens at the left spring line (θ = 270˚) from +122 to +188 kN.m (equivalent to a 35% rise in bending moment). This considerable amount of change in the bending moment due to close tunneling excavation should be considered during the design stage. The values of bending moments and axial forces listed in Table 3 from the beam spring model entail the values shown in Figure 14 and 15. Figure 16 shows the horizontal stress in the soil between two tunnels. The soil in the pillar between two tunnels experiences large changes due to the close tunnel excavation. The σxx increases from a linear distribution in initial condition into a curved distribution after excavation and lining installation of ETL. The maximum changes occur in the soil near spring line of ETL tunnel. Then, by excavation of A ramp, the amount of horizontal stress reduces by two steps for every 10% relax­ ation in stress at the boundary of A ramp. The amount of reduction in horizontal stress of soil after face and tail passage of machine in A ramp is approximately 8.5%.

Figure 16. Predicted horizontal stress in the pillar between two tunnels.

5 CONCLUSION In this study, the interaction between two under­ ground tunnels with a very small distance from each other was investigated. The new tunnel is an under­ ground ramp with curve shape alignment that was excavated from ground surface using an earth pres­ sure balanced machine and after passing of a spiral route, it approaches and connects to a deep existing target tunnel. Numerical analyses, and results of pre­ viously performed beam-spring model were used to show the lining deformation, internal forces in target tunnel, and horizontal earth pressure distribution between two tunnels in the last monitoring section of the ramp when the ramp and target tunnel are almost parallel. The results of this study are as follow: -

-

Figure 15. Predicted bending moment in ETL tunnel for various construction steps.

499

According to field measurements, the pillar dis­ tance to tunnel diameter in this interaction prob­ lem is less than 0.15D and during machine face and tail pass and lining installation in new A ramp, the spring line of target tunnel is drawn toward new excavation and reaches to final value of 10.3 mm. No pushing in the lining of the target tunnel observed which indicates pro­ gressive relaxation of the stress in the soil around new excavation. During passage of machine face and tail along A ramp, axial force and bending moment increase by 7% and 35% respectively at left spring line of target tunnel. This considerable amount of change in the bending moment due to close tunneling excavation should be taken into account during the design stage.

-

The horizontal stress of soil at pillar between two tunnels reduces by approximately 8.5% after machine pass in A ramp.

REFERENCES Afshani, A., Akagi, H. & Konishi, S., 2020. Close con­ struction effect and lining behavior during tunnel exca­ vation with an elliptical cross-section. Soils and Foundations, 60(1), pp.28–44. Addenbrooke, T.I. & Potts, D.M., 2001. Twin tunnel inter­ action: surface and subsurface effects. International Journal of Geomechanics, 1(2), pp.249–271. Chehade, F.H. & Shahrour, I., 2008. Numerical analysis of the interaction between twin-tunnels: Influence of the rela­ tive position and construction procedure. Tunnelling and Underground Space Technology, 23(2), pp.210–214. Ghaboussi, J. & Ranken, R.E., 1977. Interaction between two parallel tunnels. International Journal for Numerical and Analytical Methods in Geomechanics, 1(1), pp.75–103.

Metropolitan Expressway Co. Ltd. 2008. Tunnel Structural Design Specification (Shield Tunneling). Ng, C.W., Boonyarak, T. & Mašín, D., 2015. Effects of pillar depth and shielding on the interaction of crossing multitunnels. Journal of Geotechnical and Geoenvironmental Engineering, 141(6), p.04015021. Railway Technical Research Institute. 2006. Railway Struc­ ture Design Standard with Commentary report. Soliman, E., Duddeck, H. & Ahrens, H., 1993. Two-and three-dimensional analysis of closely spaced double-tube tunnels. Tunnelling and Underground Space Technology, 8(1), pp.13–18. Wan, M.S.P., Standing, J.R., Potts, D.M. & Burland, J.B., 2016. Measured short-term ground surface response to EPBM tunnelling in London Clay. Géotechnique, 67(5), pp.420–445. Wang, Z., Yao, W., Cai, Y., Xu, B., Fu, Y. & Wei, G., 2019. Analysis of ground surface settlement induced by the construction of a large-diameter shallow-buried twin-tunnel in soft ground. Tunnelling and Underground Space Technology, 83, pp.520–532.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Pile driving interaction with existing tunnel K.J. Bakker Delft University of Technology, Delft & WAD43 bv, IJsselstein, The Netherlands

R. Spruit Delft University of Technology, Delft & Port of Rotterdam, The Netherlands

F.C.M. van Overstraten Kruijsse Delft University of Technology, Delft, The Netherlands

ABSTRACT: The Netherlands is a densely populated country. Any situation where infrastructure needs to be extended there is a probability that this may interfere with existing structures in the vicinity; this may even involve a tunnel. As tunnels in general are not under piled, they may be vulnerable to imposed deformations or loads. Due to the soft soil in the Netherlands, new infrastructure, e.g. bridges will be built on piles; mostly driven piles. Pile driving due to its nature may create deformations and loads, as the soil original soil is dis­ placed from the position of the pile. This volume needs to be absorbed either by compression or kinematically absorbed due to displacement of soil outward from the pile. Near the Thomassen tunnel in Rotterdam, piers need to be built for overbridging of the approach to the Thomassen tunnel in the extension of the Harbour Railway. In order to estimate the deformation effects of pile driving a test was performed on a nearby location. Pile intrusion was monitored with inclinometers at different distances to a series of test piles. The result was applied as input for 3D FEM analysis for the erection of one of the bridge piers. In this paper, test results will be discussed as also the results of the analyses.

1 INTRODUCTION For the extension of the Harbour Railway in Rotter­ dam, to improve transport capacity for the 2nd Maasvlakte Harbour extension, a bridge needs to be built to cross-over the A15 motorway that at this location is seated deep underground as part of the approach to the Thomassen tunnel. This tunnel here passes the Calland Channel next to the site, see Figure 1 and Figure 2 and at this location is seated in a Trench in the Bank. The Tunnel here is located at a depth of about 7.5 m below the average water table (there is a tidal influence in this part of the Harbour), For the foundation, piers will be built on either side of the Trench. These piers will be supported by a large number of piles, that need to be driven into the underground. If in the past piles where moderate in diameter, i.e. in the order of 0.2 – 0.4 m, that didn’t cause too much concern, in the present situation piles are foreseen with an outer diameter of about 0.7 and as such in a formation of 8 x 7, which makes that the effect of driving these in the underground is foreseen to be not negligible. The displaced soil will give rise to sideways displacements that may be felt at larger

DOI: 10.1201/9780429321559-65

501

distances. These deformations if hindered may give rise to imposed loads e.g. here load on the retaining walls of the harbour, the trench and on the tunnel, see also Figure 2. For further identifi­ cation, see also Figure 7, the eastern side founda­ tion of the Arched bridge is indicated as pillar 8.2, and the pier on the Westbank of the trench as pier 8.3. Apart from these main piers add­ itional pier 8.1 and 8.0 may be identified that bear the approach to the arched bridge. In this paper the main attention is focussed on the deformations related to pier 8.2 and its effect on the tunnel. After the driving of the piles, a foundation plat­ form will be built. The platform itself will consist of a concrete plate on top of a configuration of 56 Fundex piles each with a diameter of 711 mm. Due to the ground-penetrating effect of these piles, it was feared that the Sheet pile walls carrying the trench might be displaced and even so that the tunnel itself would be displaced due to the imposed deform­ ations. Further at the other side in this cross section there is the embankment of the Neckar harbour with Pipe-racks of Vopak, Vopak is a company involved with bulk Storage in the area, that are located parallel to the retaining wall. Further there are some water

Figure 1. Artist impression of the Neckar-Harbour, (left) with the new Port Railway. The arch bridge crosses the approach to the Calland-tunnel that lies in a trench in the Bank.

ducts in the underground, see Figure 2, that might be affected due to the pile driving.

Figure 3. Side view pier 8.2.

2 CONSTRUCTION OF THE BRIDGE PIERS For the construction of the pier 8.2 it is foreseen that 56 ground-penetrating screwed piles ⊘ 711 mm will be placed in a grid of 7 x 8 piles, see Figure 3 and Figure 4. The pier dimensions overall are 16 x 17 m. Due to the penetration of the piles, the soil origin­ ally at that location, must be displaced. This “surplus” of material will create sideways displacements in the underground. Only a part of the displaced soil will be absorbed by compression or densification of the soil, the rest may lead to uplift of the soil surface or needs to be kinematically absorbed by lateral displacements. The latter may give rise to unfavourable loads on structures in the vicinity or even to displacement of these. In principle, the upper soil layers here, silty sand and silty clay are poorly compressible; certainly, if foundation elements such as piles are quickly inserted into that ground, the soil at first primarily

Figure 4. Top view pier 8.2; 56 Screw Posts.

Figure 2. Cross section A-A, on pier 8.2, seen from the North, see also Figure 7, the Neckar harbour at the left.

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will react undrained. The obstructed drainage may resist compression and subsequently part of the excess volume will be squeezed-away sideways. Due to geometric effects it is generally assumed that the soil that is driven from its position gives a lateral deformation that diminishes with a ratio of, R0/Ri, where R0 is the diameter of the pile, (sometimes 1.15 R is assumed, where R is the pile radius) and Ri is a coordinate taken with respect to the center of the pile, the assumptions taken for the start of the analysis will be explained in more detail in the next section. Due to the foreseeability of this problem, and because part of the problem may be soil type related the Port Authority, in the preliminary of this project, organized a foundation test on a location nearby of the site, in which five piles have been inserted and soil deformation was measured with inclinometers at different distances of the piles. Based on the test results, as described in the next paragraph, it was established, by approximation, which part of the displaced soil would effectively lead to lat­ eral displacements outside a direct influenced zone of about five times the diameter of the pile. This percent­ age that was derived from the test was used as a starting point for numerical analysis that was per­ formed to calculate the displacement effects on struc­ tures further away. 3 FOUNDATION TESTING A foundation test was carried out where five screw piles with a diameter of ⊘ 711 mm were installed in the ground and horizontal displacements were meas­ ured at different distances from these piles, see Figure 5. The displacements where derived by inte­ gration of inclinometers. The measurements (Table 1) where evaluated in the context of a graduation work at Delft University (Overstraten-Kruijsse, 2019). For this project, draft results of her study were assumed to be characteristic with some conservatism. Since then, no details arose that contradict these assumptions. For the short time it takes to screw a pile in, undrained conditions must be assumed; it only takes a few hours to screw in a pile, which is short com­ pared to the consolidation time of the soil layers.

Table 1.

Figure 5. Test configuration for piledriving in the vicinity of structures, near location Neckar-Harbour.

Furthermore, with silty clay and silty sand, which in general has a low permeability, and therefore it may take weeks to consolidate water overpressures and because porewater is nearly incompressible it must be considered that during pile driving soil must be displaced outwards. Only, in the proximity of the pile, due to non-linear effects, such as plasticity and large displacements and large deformations, soil compression may absorb part of the displaced soil; Further, due to uplift of the soil package another part of the volume may be taken. The remaining volume must be absorbed by kinematic displacement of soil in the outward direction from the pile; this displace­ ment will reduce with a factor of about R0/Ri, as a function of that distance. To evaluate this effect in a simplified way the fol­ lowing explanation may be given; assuming incom pressibility of the soil, the following soil balance must hold; where V ¼ πR0 2 , is the volume per unit length, that needs to be absorbed this volume must balance with lateral displacements, i.e. the volume passes the circumference at a certain distance from the pile. Assuming a circumference at a certain distance from the file of; circi ¼ 2πri , where a volume is

Ground displacement at a distance from the pile, from the foundation test (Overstraten Kruijsse, 2019). Pile 1

Direction Distance Displacements Clay (shallow) Sandy clay(intermediate) Clay (deep)

A m 3.3 mm 12.9 2.0 - 0.5

Pile 2 B

mm 0.0 -0.8 -0.1

A m 3.1 mm -0.1 -0.2 -0.8

Pile3 B

mm -18.6 - 4.4 - 1.0

503

A m 3.9 mm 1.5 0.7 0.4

Pile 4 B

mm 0.1 0.2 -0.1

A m 1.1 mm 26.2 12.1 5.5

Pile 5 B

mm -2.0 -4.3 -3.3

A m 2.5 mm -4.8 -1.8 0.3

B

mm 1.7 1.0 0.1

displaced according to V ≈ ui · circi , where ui is a displacement at a certain radius from the pile, this must balance with the displaced soil and so, rearran­ ging leads to:

Þ

For the outer diameter of a pile, where R0/Ri = 1.0 this leads to:

which is a well-known solution. In order to simplify the evaluation a correction fact or if 1 is introduced according to:

Following this approach, the non-linear effects such as compression, densification and uplift close to the pile will be neglected, but further away from the pile this result matches well. It was assumed here that further away than 8D, the effects are primarily elastic. Here at 3.5 m from the pile (5 times the diameter), the agreement is relatively good or even conservative. This may be attributed to the fact that close to the pile deformations are governed by plastic compres­ sion and densification and further away the displace­ ments are dominated by elastic behaviour. In the present case the objects to be protected all lie further away than 5 times the diameter of the piles and therefore the effects near the pile are of less importance for the object of this study. 4 FOUNDATIONS AT NECKAR-HARBOUR 4.1

Where C1 is a discounting factor 0 < C1 < 1, where it is assumed that C1, depends on the soil type. For the displacement very near to the pile, cave expansion theory might improve the description of the deformation but as here in this paper the focal point is the overall effect of an entire foundation and also for the sake of simplicity, here a more simplified approach was taken. Considering that the test location consists of three distinct types of soil layers, following a best match evaluation with the measurement, it was found that a weighting factor of C1 of 0.25 corrob­ orates best for the Sand and Sandy Clay and a factor of C1 = 0.5 for the clay layers, as indicated in Figure 6.

Interpretation of the soil test to Pier 8-2

The outcome of the previous evaluation, e.g. for the sand would imply that for a centre to centre distance of the piles of 2.25 m, 75 % of the displaced soil volume of a single pile, i.e. a volume of 0.75 x 0.383 = 0.29 m3 per m of shaft would need to be absorbed by uplift of the soil level or by compression of the soil. Uplift may be confined to the soil close to the soil surface, which means that for the deeper soil layers a compression of:

of the soil body around the pile needs to be taken by compression. For sand, Terzaghi & Peck (1948) pro­ vided in article 51 of “Soil Mechanics in Engineer­ ing Practice” guideline values for compaction of “fills” between 4% and 8% densification for sand and/or clay-containing sand.

Figure 6. Fit for ground displacement for Clay; C1 = 0.5, for Sandy-Clay C1 = 0.25; measurements for Clay at an approxi­ mate depth of 5 m below greenfield and for the sandy clay between 6 and 20 m below greenfield.

504

This means that with an average compaction of 6% the material has passed from a loosely packed material up to a densely packing. Which is not unlikely, but also indicates a certain limit to the compaction in this zone. 4.2

ation of the CPT’s JV250, see Figure 8, and JV251, soilparameters were retrieved by Robertson method in conjunction with an interpretation acc. to Table 2b in EC7, NEN-en-9997. The result is given in Table 2. 5 PLAXIS 3D MODEL

Soil investigations and parameter estimation

In the preliminary to this project, CPT tests were performed on the site. An overview of the CPT loca­ tions with respect to the position of the piers to be foreseen is given in Figure 7. Based on an evaluation of about 14 Cpt tests it was found that the stratigraphy consists of a geological deposit originally up to about Dutch Ordnance Level i.e. NAP, overlain by an Antropogeneous Sand layer that raised the greenfield that originally was at about NAP 0.0 to an average height of about 3 to 4 m above NAP. Based on this information an image was built on the ground structure in this area in different layers and with different soil parameters per layer. Overall the soil stratification in this area is rela­ tively homogeneous and based on a further evalu-

5.1

Figure 7. CPT locations (JV250 and JV251).

Figure 8. CPT test result JV 250; indication of stratigraphy in the Underground.

Table 2.

Model description and considerations

For the analysis of the interaction between the exist­ ing structures, i.e. the trench with the tunnel and the retaining wall at the harbour side, a 3D numerical model was established to evaluate the effect of the

Soil parameters for Plaxis calculations, based on CPT JV 250 and JV 251.

Top of soil layer m

Type

CPT, qc MPa

γnat kN/m3

Id [-]

E50ref MPa

c’kar kN/m2

φ’kar °

ψ °

δ °

OCR [-]

5.5 1 -0.5 -1.5 -4.5 -8.5 -11 -12.25 -16 -18.7 -20.5

1 Fill sand firm 2 Silty Sand loose 3 Silty Clay 4 Sand dense 5 Silty Sand loose 6 Sand loose 7 Silty Sand very loose 8 Sand firm 9 Silty sand very loose 10 Clay soft 11 Sand firm

8 2 0.6 9 3.5 6 4 10 5.5 1.6 13

20 16.8 15 19 17.2 18.6 17 19 16.2 16 19.5

0.8 0.25 Na 0.6 0.2 0.4 0.2 0.36 0.20 Na 0.55

30 7 10 35 12 30 10 30 10 6 45

2 2.5 2.5 0.1 0.1 0.1 0.1 0.1 2.5 5 0.1

33 26 29 34 26 33 32 32 28 14 34

2.0 0.0 0.0 3.0 0.0 1.6 1.6 1.6 0.0 0.0 3.2

27.5 25 26 27.5 27.5 27.5 27.5 27.5 20 20 27.5

9.0 3.5 2.0 1.1 1.3 1.3 1.0 1.0 1.0 1.0 1.15

505

pile driving. The model that was used here had pre­ viously been applied to evaluate the effect of loosen­ ing and removing the ground anchors that conflicted with the construction of pier 8.1. To maintain enough stability for the bank, a supporting berm was applied which can be identified in Figure 10 in the corner of the harbour. Evaluation of the predictions and the monitoring afterwards showed a good agreement for the dis­ placement of the bank of the Neckar-Harbour; i.e. a maximum displacement was predicted of about 25 mm, and a displacement was measured of 28 mm; see Figure 9. The good corroboration between predicted and measured displacement gave confidence in the valid­ ity of the underground model. In Figure 10, at the left side, the tunnel may be identified in the trench and on the right side the Neckar harbour with the retaining wall supported by ground anchors. Further, parallel to the bank the Piperack may be identified and the underground the water ducts. Adjacent to the pipe rack there is a minor road and in between the trench and the tunnel the ground level is raised to about NAP + 5.50 m. The trench of the tunnel is supported by props, that in the model are replaced by local supports.

Figure 9. Validation of 3D Plaxis model; comparing predic­ tion and measurement of partial de-activation of ground anchors.

Figure 10. Birdseye view of 3D model seen from the South-East. The tunnel modelled as a lightweight soil body, with construction stiffness.

In the 2D model, that was used to calculate the drift of the tunnel due to the lateral stresses, the props where modelled, as such, see also Figure 15. In the 3D model, the tunnel is modelled as a rigid volume with the effective weight of a tunnel under­ water, i.e. 1050 kg/m3, assuming the design value of the minimum foundation pressure, such as normally considered for immersed tunnels. The pier itself is modelled as a volume with a dimension in the x-y plane of 16 x 17 M. and a depth from ground level to NAP – 24.20 m. Modelling individual piles with volume extension was considered not necessary and with respect to calculation time unrealistically timeconsuming. 5.2 Modelling volume pile driving by volume expansion The pier design aims to build a support of 56 piles of ⊘ 711 mm. Driving these piles would mean oust of a soil volume of 0.397 m3 per m. of pile. For 56 piles under the basement this would add up to 22.23 m3 for every meter of pile. To avoid foreseeable extensive calculation time and difficulties in meshing, the expansion of these piles is modelled here as the expansion of a block volume taken with the size of the pier, i.e. a block of soil with a dimension of 16 x 17 m for the full height of the piles, see Figure 11. If we would model this volume as a cylinder, this would imply a deformation at the outer surface; where L is the circumference calculated ffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi this would imply as a deformation at the outer skirt of the cylinder of about, 22.2/51.8 = 0.42 m. For sand, discounting for the reduction factor C1 of 0.25 this would imply a deformation of 0.25 x 0.42 = 0.11 m and for clay twice this value. In percentage the volume expansion of the soil layers within the layers would expand with 22.2x0.25/(16 x17) = 2.2% for the sand layers and 4.4% for the clay layers, which is assumed in the further calculations as a starting point, further indicated as the 100 % expansion, as indicated in the calculation sequence of Table 3.

Figure 11. View from the North; expansion of the piles modelled as a soil body 16 x 17 m with volume expansion. Indication of groundwater table, Visibility of the soil around the pier deactivated to indicate the topography, see also Figure 2.

506

Table 3.

1 Initial stage 2 Structures 3 Sand fill NAP +3.50 between retaining walls 4 Pre-tension of Anchors 5 Excavation 1 Harbour & Trench 6 Excavation 2 Harbour, adding berm 7 Adding Pipe-rack 8 Loosening of ground-anchors in the corner of the Neckar-Harbour 9 Installing Screw piles; volume expansion 10 Volume expansion 2, 1,5 times

5.3

displacement, i.e. for 100% of volume expansion such as indicated in paragraph 4.1 and 4.2, and in phase 10 one for the situation that this percentage is 50% larger. Here for the Clay layers in the block ΔV of 2.2 % was assumed and for the Clayey sand half of that, 1.1 %. The analysis in the last step, no 10, with a 50 % exceedance of the first step, to verify the influ­ ence of uncertainty in the input.

Construction stages as applied. drained drained drained drained drained drained drained drained

6 CALCULATION RESULTS Undrained Undrained

6.1

Model data

With respect to structural starting points the follow­ ing structural details were taken: Existing Bank at the North of the Neckar harbour – Sheet piling, topsoil level 7.00 M-NAP – Profile AZ36 (S235 GP) ppn 18.50 M-NAP; Sheet piling at Neckar harbour, East bank), – deep AZ36 (S235 GP) ppn 15.00 M-NAP, and AZ13 upcoming with different point levels. – Grouting anchors Type Ø 60.3 x 17.5 steel grad MW450, angles 42.5°/47.5°, anchors c.t.c. 1.26 m, level 3.00 m + NAP; Trench of the Thomassen tunnel – Combi Wall, Top NAP + 4.50, pipe 1620 x 20 c.t. c. 2.52. Point level NAP-27.0; – Pipe Props 1120 x 20 c.t.c. 5.00 m; – Girder 2.0 x 1.0 m Water ducts – 1 x (1200 x 10 mm) 9.00 m from the rear of Neckar harbour bank, at a depth of NAP + 2.0 m. – 1 x (800 x 8 mm) 11.80 from the rear of the Neck arbour bank, at a depth of NAP + 2.0 m Geometrical data

Deformation effects at the banks

In the analysis, the expansion of the ground volume taken by the pier has been calculated and evaluated in steps. In the first step an expansion to 100% of the expected values was applied, as derived in the previ­ ous paragraph. In a second step, the input of the volume expansion was enlarged to 150 % compared to the previous phase, to evaluate the sensitivity to the calculation results. The results presented in this paper mainly refer to the first step, as the latter did not change any of the conclusions drawn. In Figure 12 the horizontal displacement, as pre­ dicted by the model, on the skirts of the soil body are shown, in particular on the bank of the NeckarHarbour more at the front of the picture and in the background of the trench at the Thomassen Tunnel. The displacement on the side of the port, i.e. 0.039 m, see Figure 12, is seen as an upper limit. Due to a modelling restriction, on the side of the Thomassen tunnel, as modelling both sides of the trench was out of the scope of this model, due to an edge effect of the soil body this side of the model, the tunnel will react stiffer than and a larger part of the soil deformation will be squeezed to the port side. Based on the calculation it was estimated that the volume expansion of the construction of Pier 8.2 might lead to an additional loading at the level of the tunnel of about 120 kPa, see Figure 13. It was considered unlikely that the tunnel would be able to withstand such a lateral load without

– Harbour level NAP - 7.00 m-NAP; (Analysis depth, current data indicate a level of NAP -5.35 m) – Underwater embankment running up from 7.00 m-NAP to approx. 1.70 m + NAP; – Bank level top sheet piling NAP +5.50m Water levels – LW: 2.04m-NAP; – HW 2.48m+NAP; – Groundwaterlevel at Ordnance Lever, i.e. NAP; 5.4

Sequence of analysis

The applied construction phasing is shown in Table 3. The calculations include two variants for the

Figure 12. Maximum displacement of the walls, on the side of the Neckar-Harbour (front), approximately 0.039 m and at the side of the tunnel trench (back) of about 0.031 m.

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Figure 13. The effective horizontal load at the tunnel, after boring of the piles may rise from 171 till about 293 KN/ m2, at the bottom of the tunnel.

additional displacements. It must be considered that tunnel bearing on the underground is relatively small so it can’t withstand large horizontal loads. In order to retrieve a better estimate of what would be the effect on the tunnel an additional Plaxis 2D model was developed where the deform­ ation at a certain distance from the tunnel, i.e. at 5 m from the trench was taken as input and applied as a forced displacement in the 2D model. The dis­ placement as applied is indicated in Figure 14, which is the output of the 3D model as described previously.

combi wall of about 5.0 m, at which position the deformation from the 3D model was taken, see Figure 15. The imposed displacement as applied was specified as a forced deformation of 0.05 m. for the vertical, see also Figure 15 In the next step, the reaction of the tunnel on this imposed deformation was calculated, considering the overall effect of the tunnel in the trench, where the interaction of the trench support, including the props was modelled. As a result of this imposed displacement, due to the forced insertion of the screw piles for pier 8.2, a displacement of the tunnel was calculated of about 0.028 m, see Figure 16. In the present situation regarding this tunnel, the owner, i.e. the Ministry of Transport and Public works, did not clearly state an absolute requirement regarding maximum permissible movements. However, based on Euro code NEN-EN-9997, annex H it is generally assumed that maximum relative rotation of 1:500 gives a limit. In addition to that it is not uncommon to assume that 50% of the limit may have developed within a previous phase. This would imply a limit value for relative increase in rotations of 1:1000 or of that order. At a section length of the tunnel of 20 m this would imply a maximum permissible additional dis­ placement of about 0.02 m. Within the project it was recommended to set an alarm value to any additional horizontal displacement of 0.01 m. 6.3 Prop forces along the trench From the 2D model, prop forces are calculated regu­ lar of about (O) 2600 kN/prop, which can reach up

6.2 Plaxis 2D analysis of loads and displacement on the of Thomassen tunnel in reaction of the displacement as retrieved from the 3D model To verify the effect of the horizontal load, in Plaxis 2D a symmetrical model of the tunnel was modelled and a forced displacement was specified as a boundary condition in a cross section at a distance from the

Figure 15. Plaxis 2D modelling immersion trench of the Thomassen tunnel; with imposed displacement.

Figure 14. Horizontal displacement at approx. 5 m from the wall of the Thomassen tunnel.

Figure 16. Forced displacement of the Thomassen tunnel.

508

to about 3600 kN/prop next to the pier. Based on the present data for the props, i.e. tube 1200 x 20 mm length approx. 40 m, a buckling strength is calcu­ lated from (:

The calculated forces, including the increase are well within the margins of admissibility and will not pose a problem. 7 CONCLUDING REMARKS In order to estimate the amount of displacement, Plaxis 3D and Plaxis 2D calculations have been per­ formed. In this case the 3D model had been previ­ ously used to predict displacement for the bank of the Neckar harbour and the corroboration to the measured displacements supports the validity of the soil parameters that were applied. With respect to the input parameter of the imposed deformations due to pile installation

measurements of the test location where evaluated and applied. Regarding the movements at the Thomassen Tunnel, a critical displacement is foreseen. Further it must be mentioned that due to imposed deformations the tunnel could move more than the permissible value of 20 mm, as calculated according to annex H of the Eurocode as limit value. The results of the calculation show that, without additional measures, ground penetrating pile driving would yield deformations to the tunnel that would exceed acceptable limits. In order to limit this effect pre-boring on the pile location is recommended and featured. Given the risks with respect to the tunnel, moni­ toring of the deformations when pile driving was strongly recommended and foreseen.

REFERENCES Overstraten Kruijsse, F.C.M. van. 2019. The installation effects of screwed displacement piles. MSc Thesis. Delft University of Technology. Terzaghi, K. & R.B. Peck. 1948. Soil Mechanics in Engin­ eering Practice. New York: Wiley.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Numerical modelling of framed structures with masonry infills affected by tunnelling-induced deformation and damage D. Boldini Sapienza University of Rome, Rome, Italy

N. Losacco Polytechnic University of Bari, Bari, Italy

A. Franza Universidad Politécnica de Madrid, Madrid, Spain

S.M. Miraei University of Bologna, Bologna, Italy

ABSTRACT: The paper describes a numerical approach aimed at investigating the response of framed structures with masonry infills to tunnel excavation in sand. The numerical model, analysed with the Finite Element method, includes the soil, the tunnel and the framed structure with infills. It proved its capability in reproducing the soil response at different values of the volume loss and the principal soil-structure interaction mechanisms, including sliding and development of a gap at the soil-foundation interface. The results are sum­ marised in terms of angular distortion values for each bay as well as local tensile strains in the masonry infills. Finally, the efficiency of angular distortion models in predicting the maximum tensile strain of the infills is also evaluated.

1 INTRODUCTION This paper deals with the analysis of deformation and damage induced on a framed structure by tunnelling. This soil-structure interaction problem is of par­ ticular interest since, while the effect of tunnel excavation on masonry structures has been thor­ oughly studied (e.g. Potts & Addenbrooke 1997, Pic­ khaver et al. 2010, Burd et al. 2000, Amorosi et al. 2014, Losacco et al. 2014, 2016), the response of framed buildings has been received only a limited attention in the past (e.g., Goh & Mair 2014, Farg­ noli et al. 2015, Boldini et al. 2018). A fully 3D numerical approach based on the Finite Element method was adopted. In order to capture the evolution of the displacement field with increasing volume loss, an advanced constitutive model was employed for the soil, capable of describing the dependency of small-strain stiffness on the mean effective stress and the decay of stiff­ ness with strain level, allowing for early develop­ ment of plastic strains. In contrast, simple isotropic linear elasticity was considered for the structural elements. Both the cases of a bare frame and of a frame with masonry infills were studied, the latter being

DOI: 10.1201/9780429321559-66

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the building components most prone to tunnel­ ling-induced damage. An analysis simulating greenfield conditions was also carried out, to serve as a reference. The Finite Element model was initially validated against available centrifuge data and proved to fairly reproduce the main mechanisms of the soil-structure interaction phe­ nomenon, including the formation of a gap and the relative sliding between the foundation and the soil. 2 THE NUMERICAL MODEL The numerical simulations were inspired by a set of centrifuge tests performed at the Nottingham Univer­ sity focusing on the response of framed buildings on raft foundations to tunnelling (Xu et al. 2019a). The centrifuge apparatus consists of a plane-strain pack­ age including a strongbox and flexible membrane model tunnel. A fine dry silica sand, known as Leighton Buzzard Fraction E, was adopted in all the experiments. Specifically, the test named F2t3b6L, with layout shown in Figure 1 (Xu et al. 2019b), was studied numerically. Considering the acceleration

Figure 1. Layout of the F2t3b6L centrifuge test considered as a reference for the numerical simulations (Xu et al., 2019b).

applied in the Nottingham centrifuge, equal to 68g, the problem is characterised by the follow­ ing dimensions at the prototype scale: tunnel diameter Dt = 6.12 m, depth of the tunnel axis Zt = 11.0 m (thus representing a rather shallow tunnel), frame width B = 31.3 m and height H = 5.4 m, bay length bbay = 5.2 m and height hbay = 2.6 m, thickness of the structural elements (raft foundation, plates and walls) t = 0.22 m. The soil volume is 630 mm wide, 322 m deep and 260 mm thick in the longitudinal direction (corres­ ponding respectively to 43.5 m, 21.9 m and 17.7 m at the prototype scale). For practical reasons, a very small gap exists between the front/back walls of the centrifuge box and the frame; however, plain strain conditions can be practically assumed to be valid for the structure as well. Three-dimensional Finite Element analyses were carried out with the Plaxis 3D software. The differ­ ent ingredients of the interaction problem, i.e. the soil, the frame and the infills, were discretised with solid elements. No plate elements were adopted for structural elements, especially the infills, to have a direct evidence of the tunnelling-induced strain distribution. Given the hypothesis of linear elasti­ city, this did not require the adoption of a very dense mesh discretisation for the infills. A view of the adopted Finite Element mesh is shown in Figure 2. Taking advantage of the plain-strain con­ ditions of the problem, only a slice of the full 3D prototype was considered (Figure 3), with length of 8.772 m parallel to the tunnel axis. This length cor­ responds to the theoretical spacing in the longitu­ dinal direction between masonry infills. The other dimensions were set consistently with those of the F2t3b6L test, but only half of the problem was modelled, considering the symmetry with respect to the tunnel axis. A thickness of 0.25 m was assumed for the infills. Tunnel excavation was simulated by applying incremental displacements at the boundary nodes

Figure 2. 3D Finite Element mesh.

Figure 3. Detail of the infills (a) and Finite Element dis­ cretisation (b).

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to obtain a homothetic contraction of the tunnel cross-section centred at the invert (Boldini et al., 2018). 2.1

Material constitutive laws and properties

The soil behaviour was simulated adopting the SANISAND bounding surface plasticity model originally proposed by Dafalias and Manzari (2004). The constitutive law is capable of reprodu­ cing the mechanical response of sand from small

to medium strains, using a unique set of material constants, regardless of the initial conditions of the soil, and has been successfully used to reproduce the behaviour observed in centrifuge tests of soilstructure interaction due to tunnelling (Xu et al. 2019b, Giardina et al. 2019). The initial relative density of the sand in the centrifuge test was 90%. Accordingly, model parameters were preliminary calibrated against available laboratory and centrifuge tests per­ formed on the same sand in the framework of other experimental campaigns (Visone 2008, Far­ rell 2010). A summary of the adopted material constants is provided in Table 1. For the structural elements, i.e. the frame and the infills, a linear elastic constitutive law was adopted. The frame was assembled by welding two alu­ minium alloy plates and 12 angle sections. A quasi-rigid connection was achieved by welding 60% of the length along the longitudinal direction. On the basis of preliminary bending tests con­ ducted on the frame, Young’s modulus E = 53.8 GPa and Poisson’s ratio of ν = 0.334 were inferred; the unit weight is γ = 27 kN/m3. For the infills, the elastic constants were set equal to E = 4.0 GPa and ν = 0.2 (Cobanoglu et al. 2017); γ = 13 kN/m3 was considered. To replicate a rough soil-foundation interface, a thin layer of sand was glued beneath the structure in the test. This detail was accounted for in the numerical simulations by activating an interface between the frame and the soil, with a coefficient of friction equal to that of the soil at constant volume (i.e. 32°) and null tensile strength. 2.2

Performed analyses

Four different numerical analyses were performed. In the first analysis (labelled “Greenfield”), only greenfield conditions were considered, in order to validate the numerical approach against the experi­ mental evidence. The second analysis (labelled “Frame”), included the bare frame, for which experimental data are also available, for further

Table 1. Material constants of the SANISAND model for the Leighton Buzzard Fraction E (Xu et al. 2019b). Elasticity Critical state Yield surface Plastic modulus Dilatancy Fabric-dilatancy tensor

G0 = 400; ν = 0.05

Mc = 1.287; c = 0.780; λc =

0.00178; e0 = 0.8191; ξ = 2.4352 m = 0.01 h0 = 4.05; ch = 1.1; nb = 2.8 A0 = 0.55; nd = 2.564 zmax = 0; cz = 0

validation. In the last analyses, infills were added to the frame in the central transverse section of the model. In one case, the full stiffness and weight of the infills were activated to reproduce realistic conditions (labelled “Frame with stiff infills”). In this study, maximum tensile strains in the panels were used as indicators of tunnelling­ induced damage. Hence, in order to assess the relative influence of the stiffness and weight of the infill panels on the expected damage, a further analysis in which weightless, fully flexible, “vir­ tual” panels were modelled (labelled “Frame with flexible infills”) was also carried out. The analyses comprised the following stages: 1) gravity activation in the soil domain, assuming a coefficient of earth pressure at rest K0 = 0.5; 2) activation of the frame and the soil-structure inter­ face elements, if included; 3) activation of the infills, if included; 4) deactivation of soil elements inside the tunnel and application of incremental displace­ ments at the boundary nodes. 3 NUMERICAL RESULTS In the following, results are provided in terms of ver­ tical and horizontal displacements at the ground sur­ face, deformed configuration of the frame and, when appropriate, tensile strains within the infills. Larger settlements at the soil surface than at the base of the frame indicate that a gap has formed between the foundation soil and the structure. Similarly, different horizontal displacements between the soil and the foundation at the same location show that slipping has occurred. Results are shown with reference to two values of tunnel volume loss Vlt, defined as the relative differ­ ence between the initial and the deformed crosssectional area of the tunnel: Vlt = 0.4%, representa­ tive of a good tunnelling performance, typically achieved by the use of EPB or slurry shield machines, and Vlt = 1.0%, typical of traditional tunnel excavations. 3.1

Validation of the numerical approach

Figures 4 and 5 show the settlements and horizontal displacements at the ground surface for Vlt = 0.4% and Vlt = 1.0%, respectively. Numerical results for green­ field conditions and for the analysis with the frame only are compared to centrifuge data available in Xu et al. (2019b). The agreement is rather satisfactory for the green­ field case, with a general slight underestimation of settlements and overestimation of horizontal displace­ ments in the numerical solution if compared to the experiment. When the frame is activated, numerical predic­ tions show a reduction of soil settlements beneath

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Figure 4. Comparison between centrifuge results and numerical predictions in the free-field case and for the frame only analysis (Vlt = 0.4%): (a) vertical and (b) hori­ zontal displacements at the ground surface.

Figure 5. Comparison between centrifuge results and numerical predictions in the free-field case and for the frame only analysis (Vlt = 1.0%): (a) vertical and (b) hori­ zontal displacements at the ground surface.

the foundation, while the corresponding values measured during the centrifuge tests are very similar to those obtained in greenfield conditions, for both values of volume loss. In the numerical analyses, a gap is formed between the foundation and the underlying soil sur­ face at the larger volume loss, consistently with the experimental evidence. A good match is observed for the horizontal displacements at the smaller volume loss, with almost null movements of the frame foundation in the transverse direction, indicat­ ing that sliding has occurred. The same feature is observed at Vlt = 1.0%; in this case, though, the comparison between experimental and numerical horizontal displacements at the ground surface is less satisfactory. In summary, a good performance in reproducing the main features of this complex soil-structure inter­ action problem can be assessed for the numerical model.

of vertical displacements at the frame founda­ tion. A gap at the soil-structure interface starts appearing at Vlt = 0.4%, although concentrated in the central part of the structure, and further develops at Vlt = 1.0%. Due to this phenomenon, no significant reduction of soil settlements is observed beneath the frame at large volume loss, with respect to those calculated in greenfield conditions (Figure 8). Horizontal displacements of the frame foundation are nearly zero as in the previous analyses, while those exhibited by the soil at the ground surface are larger, more similar to those obtained in the green­ field case.

3.2

Influence of the masonry infills

Results of the analyses with the masonry infills are shown in Figures 6 and 7. Activation of the infills is associated to a stiffer response in terms

3.3

Deformation and damage parameters

The effects of the frame stiffness and infills on the building deformations (i.e. distortions and max­ imum tensile strains) are evaluated. In order to quantify the deformation of the infills, undergoing shear dominated deformation due to the presence of axially stiff slabs, the average angular distortion β of each bay is used, computed from foundation settlements assuming no tilt (Son and Cording, 2005). Because of the centred position of the

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Figure 6. Comparison between numerical predictions for the bare frame and the frames with infills (Vlt = 0.4%): (a) vertical and (b) horizontal displacements at the ground surface.

Figure 7. Comparison between numerical predictions for the bare frame and the frames with infills (Vlt = 1.0%): (a) vertical and (b) horizontal displacements at the ground surface.

tunnel beneath the structure and the presence of axially stiff floors, β of each bay is given by the slope computed from the bay edges (i.e. the ratio between the differential settlement and the bay length). On the other hand, the maximum tensile strain εt of each infill is adopted as a damage par­ ameter. Tensile strains are estimated with two approaches: 1) directly inferred from the solid finite elements of the infills at the first storey (Figure 9 and 10); 2) from the angular distortion using the expression α × β/2, which was modified with respect to the formula of Son and Cording (2005) by introducing the coefficient α = 2, as sug­ gested by Boone (1996) to account for the slab-to­ column fixity at the edges. In other words, the slab­ to-column fixity gives an elevated nodal stiffness which causes the shear deformations to concentrate in the central part of the panels rather than distrib­ uting uniformly along the infill width, which is not­ able in Figures 9 and 10. Figure 11(a) displays angular distortions induced within the bays at Vlt = 1.0%. As expected, results for the bare frame and the frame with

flexible infills are similar. The bare frame stiffness decreased the distortions of the central bays above the tunnel with respect to the greenfield case, resulting in approximately halved maximum values of β. This was due to the gap formation mechan­ ism, which prevented further settlements at the centreline to be transmitted from the ground to the structure (Franza and DeJong 2019). On the other hand, the presence of stiff infills reduced the build­ ing distortions to an approximately uniform value, similar to the greenfield angular distortion of the external bays. Finally, tensile strain distributions are con­ sidered to evaluate the efficiency of the second (approximated) approach, using β in estimating the tensile strains. Results shown in Figure 11(b) indi­ cate conservative predictions delivered by using α × β/2 for both flexible and stiff infills at the second storey. However, the distribution along the offset of the maximum ε t at the first storey infills (inferred from the direct approach) differ with respect to the angular distortion estimations. This may be due to the first storey infills being in

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Figure 9. Contours of tensile strains in the infills panels (analysis with stiffness and weight activated) (Vlt = 1.0%).

Figure 10. Contours of tensile strains in the infills panels (analysis with fully-flexible panels with no weight) (Vlt = 1.0%).

Figure 8. Comparison between deformed configurations in the frame only analysis and in the analysis with infills (stiffness and weight activated) (Vlt = 1.0%).

contact through the foundation with the soil. Fur­ ther analyses are needed to confirm if these obser­ vations can be generalised to a wider range of considered scenarios.

Figure 11. (a) Angular distortions of bays and (b) max­ imum tensile strains of the panels inferred from angular distortions or solid finite elements of the infills (Vlt = 1.0%).

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4 CONCLUSIONS This paper illustrates the capability of Finite Element numerical modelling, implementing an advanced constitutive law for the soil and a realistic structural model, in predicting the response of framed build­ ings to tunnelling, while providing insights into the structural deformation mechanics. This may help engineers to deliver more effective risk assessments. The main conclusions of this work are: • results for a bare frame simulation assessed the robustness of the adopted numerical approach in replicating experimental outcomes of soilstructure interaction due to tunnelling; • a simulation, accounting for the presence of stiff infills perfectly bound to the frame, confirmed their stiffening action that can lead to a significant reduction in tunnelling-induced foundation settle­ ments. Thus, infills could play an important role in decreasing the risk for damage. However, this effect may depend on the adopted contact law between frame and infills, not investigated here; • for framed structures with both flexible and stiff infills, there was evidence of the impact of gap formation on the soil-structure interaction for volume losses often achieved in the construction of urban tunnels. Consequently, when modelling this problem numerically it is important to include a tensionless soil-foundation interface that allows for the detachment due to the super­ structure induced load redistribution; • the building distortions and damage were quanti­ fied in terms of average angular distortions of bays and maximum tensile strains of infills. Bare frames (or frame with flexible infills) with a semi-flexible response decreased peak distor­ tions associated to the greenfield settlement trough resulting in a parabolic distribution of shear distortions. On the other hand, the nearly rigid frame with stiff infills underwent a low and approximately constant distortion along the entire transverse length of the structure; • finally, the use of angular distortion values, cor­ rected by a coefficient considering the slab edge fixity, allowed for (approximate) estimation the maximum tensile strains of panels.

REFERENCES Amorosi, A., Boldini, D., de Felice, G., Malena, M. & Sebastianelli, M. 2014. Tunnelling-induced deformation and damage on historical masonry structures. Géotech­ nique 64(2): 118–130. Boldini, D., Losacco, N., Bertolin, S. & Amorosi, A. 2018. Finite Element modelling of tunnelling-induced dis-placements on framed structures. Tunnelling and Underground Space Technology 80: 222–231. Boone, S. J. 1996. Ground-movement-related building damage. Journal of Geotechnical Engineering 122: 886–896.

Burd, H.J., Houlsby, G.T., Augarde, C.E. & Liu, G. 2000. Modelling tunnelling-induced settlement of masonry buildings. Proc. Inst. Civ. Eng. Geotech. Eng. 143 (1): 17–29. Cobanoglu, B., Aldemir, A., Demirel, I.O., Binici, B., Canbay, E. & Yakut, A. 2017. Seismic performance assessment of masonry buildings using in situ material properties. Journal of Performance of Constructed Facilities 31(4): 12 pages. Dafalias, Y.F. & Manzari, M.T. 2004. Simple plasticity sand model accounting for fabric change effects. Jour­ nal of Engineering Mechanics 130(6): 622–634. Fargnoli, V., Gragnano, C.G., Boldini, D. & Amorosi, A. 2015. 3D numerical modelling of soil–structure inter­ action during EPB tunnelling. Géotechnique 65(1): 23–37. Farrell, R. 2010. Tunnelling in sands and the response of buildings. Ph.D. thesis, University of Cambridge, UK. Franza, A. & DeJong, M.J. 2019. Elastoplastic solutions to predict tunneling-induced load redistribution and deform­ ation of surface structures. Journal of Geotechnical and Geoenvironmental Engineering 145(4): 04019007. Giardina, G., Losacco, N., DeJong, M.J., Viggiani, G.M.B., Mair, R.J. (2019) Influence of soil modelling on the assessment of tunnelling-induced deformations of structures. Proceedings of the Institution of Civil Engin­ eers - Geotechnical Engineering. [Online] 1–49. Avail­ able from: doi:10.1680/jgeen.18.00127. Goh, K.H. & Mair, R.J. 2014. Response of framed build­ ings to excavation-induced movements. Soils and Foun­ dations 54(3): 250–268. Losacco, N., Burghignoli, A. & Callisto, L. 2014. Uncoupled evaluation of the structural damage induced by tunnelling. Géotechnique 64 (8): 646–656. Losacco, N., Callisto, L. & Burghignoli, A. 2016. Soilstructure interaction due to tunnelling in soft ground, an equivalent solid approach. In: Van Balen, K., Verstrynge, E. (Eds.), Proceedings of the 10th International Conference on Structural Analysis of Historical Constructions: Anam­ nesis, Diagnosis, Therapy, Controls (SAHC 2016, Leuven Belgium). CRC Press, London, pp. 495–502. Pickhaver, J.A., Burd, H.J. & Houlsby, G.T., 2010. An equivalent beam method to model masonry buildings in 3D finite element analysis. Comput. Struct. 88 (19–20): 1049–1063. Potts, D.M. & Addenbrooke, T.I. 1997. A structure’s influ­ ence on tunnelling-induced ground movements. Proc. Instn. Civil Eng. Geotech. Eng. 125: 109–125. Son, M, & Cording E.J. 2005. Estimation of Building Damage Due to Excavation-Induced Ground Movements. Journal of Geotechnical and Geoenvironmental Engin­ eering 131(2): 162–177. Visone, C. 2008. Performance-based approach in seismic design of embedded retaining walls. Ph.D. thesis, Uni­ versity of Naples Federico II, Italy. Xu, J., Franza, A. & Marshall, A.M. 2019a. The response of framed buildings on raft foundations to tunnelling. Journal of Geotechnical and Geoenvironmental Engin­ eering. Submitted. Xu, J., Marshall, A.M., Franza, A., Boldini, D., Amorosi, A. & DeJong, J.M. 2019b. The response of framed buildings on raft foundations to tunnelling: a centrifuge and numer­ ical modelling study. In Haraldur Sigursteinsson, Sigurður Erlingsson and Bjarni Bessason (eds.), Geotech­ nical engineering, foundation of the future; Proc. XVII ECSMGE-2019, Reykjavík, 1-6 September 2019. Ice­ landic Geotechnical Society. 8 pages.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Analytical investigation for the circumferential behavior of the shield-driven tunnel adjacent to a braced excavation H.Z. Cheng, R.P. Chen, H.N. Wu & F.Y. Meng College of Civil Engineering, Hunan University, Changsha, Hunan, P.R. China

ABSTRACT: The excavation adjacent to existing tunnels in close proximity becomes a common recurring problem in urban areas. Due to the presence of joints, both the longitudinal and circumferential flexural stiff­ ness of the shield-driven tunnel are significantly reduced. However, most of the previous studies focus on longitudinal tunnel deformation features. In this study, a simplified method for evaluating the circumferential behavior of the shield-driven tunnel adjacent to a braced excavation is proposed. A continuous ring structure under plane strain condition is used to determine the circumferential behavior of segmental lining. A typical tunnel cross section subjected to a series of circumferential loads is used to study circumferential behavior, including tunnel convergence and lining internal forces. Additionally, a parametric study is performed to develop a better understanding of the governing factors of tunnel behavior, including excavation-tunnel rela­ tive position and the soil condition.

1 INTRODUCTION As urbanization progresses, urban underground space development is becoming increasingly more important. However, the interaction between new underground infrastructures and existing metro tun­ nels has been increasingly serious. New underground construction, including foundation pit excavation and tunnelling and pile driving, will cause significant influence on the stress and deformation state of the surrounding soils and existing tunnels. Then, the ser­ viceability of existing adjacent tunnels may be affected, especially for shield-driven tunnels with lower flexural stiffness due to the presence of the lining joints. In recent years, several cases of tunnel damage caused by nearby excavation have been reported (e.g. Chang et al. 2001, Hwang et al. 2011, Chen et al. 2016). Considering such unfavorable cases, predicting the behaviors of tunnels are important to avoid any damage to existing tunnels. Consequently, both ana­ lytical and empirical approaches were developed to study the mechanism of excavation-soil-tunnel inter­ action. Zhang et al. (2013a) predicted the heave of a tunnel located directly beneath an excavation. The Boussinesq’s solution was employed to analyze the soil stresses and displacement caused by excavation, and the existing tunnel was assumed as an elastic beam. Zhang et al. (2013b) used a two-stage method to study the deformation response for adjacent tun­ nels. The vertical stress relief acting on the existing tunnel was estimated by the Mindlin’s solution

caused by the excavation, and then the deformation response of the tunnel was calculated based on the assumption of the Euler-Bernoulli beam resting on a Winkler foundation. Wu et al. (2015) proposed that the shield driven tunnel was assumed to be a Timoshenko beam model, as the Euler-Bernoulli beam only considered the bending deformation and ignored the shearing deformation of jointed linings. Later, Liang et al. (2017) analysed the tunnelexcavation interaction using the Timoshenko beam model. The existing methods focus on the prediction of the deformation of existing tunnels beneath an excavation. With the increasing of underground space excavation depth, a large number of excavations will be performed beside existing tunnels. The behaviors of tunnels induced by the lateral excavation of a foundation pit are more complicated due to the complex stress and displacement states. Zheng et al. (2018) proposed a simplified empirical method to predict the maximum horizontal tunnel displacement laterally adjacent to excavations based on a series of numerical analysis. Due to the lateral stress relief caused by the adjacent excavation, the existing tunnel would be further horizontally elongated and vertically compressed, and cracks and leakages of the linings may occur. However, a sufficient understanding of the circumferential behavior of the tunnel is still lack­ ing. So far, a simplified method that can be employed to predict the circumferential tunnel responses due to lateral excavation is not available. In this paper, a simplified analytical method is pro­ posed to evaluate the circumferential behavior of the

DOI: 10.1201/9780429321559-67

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shield-driven tunnel adjacent to a braced excavation, including the convergence deformation and the inter force. It combines the closed-form solution proposed by Sagaseta (1987) and the tunnel model set up by Lee et al. (2001a, 2001b), in which the soil is treated as the incompressible materials. A parametric investigation is further performed to develop a better understanding of the governing factors of circumferential behaviors. The effects of tunnel-excavation relative position and the soil resistance are studied. It is expected that the finds obtained from this analysis can provide a quick assess­ ment of the excavation-induced tunnel damage. Add­ itionally, it is believed that the results will help in capturing the factors and parameters of tunnel serviceability. 2 METHOD FOR ANALYSIS The two-stage analysis method is utilized in this ana­ lysis, which is widely employed in solving construc­ tion-induced soil-structure interaction problems. According to the two-stage analysis method, the ana­ lysis of soil-structure interaction problem can be divided into two sequential stages: first, the construc­ tion-induced green soil unloading stress or displace­ ment acting on the existing underground structure is estimated using the available solution; second, the response of existing structures subjected to the associ­ ated construction effect is calculated. 2.1

Figure 1. Schematic presentation for the influence of adja­ cent excavation on the existing tunnel.

the surface of z0. Actually, the ground loss due to the wall deflection is discretized into a series of seg­ ments with a constant height of dz along the vertical direction, and these segments can be seen as a number of point sink in this analysis. Based on the virtual image technique, the horizontal and vertical displacements of subsoil at a point of P(x, z) caused by excavation are predicted as follows, caused by a point sink with an inward radial displacement of a at a depth below the surface ground of z0 2

Þ

Green soil displacement induced by excavation

In general, the excavation-induced ground movements mainly depend on the wall deflection. Therefore, the excavation-induced subsoil displacement is calculated using the analytical solution proposed by Sagaseta (1987). This analytical solution has been widely util­ ized in the determination of the soil movement field caused by underground construction, including tunnelling, pile driving, and braced-excavation. To employ this analytical solution, some basic assumptions are required to be specified in the analytical derivation: (a) soil is considered as a homogeneous and incom­ pressible material; (b) the calculation is conducted under undrained condition; (c) the effect of the enclos­ ure underground structure is not considered; (d) the effect of the surcharge caused by preexisting buildings and vehicles is ignored; (e) the wall deflection is con­ sidered as the ground loss caused by excavation. In the analytical method developed by Sagaseta (1987), the closed-form solution was derived using a virtual image technique. Figure 1 shows a schematic presentation for the virtual image tech­ nique applied in a brace excavation problem. Geo­ metric parameters include excavation depth (he), diaphragm wall depth (h), tunnel diameter (D), tunnel cover depth (C), the horizontal distance between tunnel springline and diaphragm wall (Lt). As shown in Figure 1, δh represents the horizontal displacement of the diaphragm wall at a depth below

where r1 and r2 are the distances to the sink and its image source, respectively, associated with the following explanations

According to the principle of area equivalence, the corresponding area of each segment can be equivalent to a circle using the expression

Therefore, integration of Equations (1) and (2) along the vertical direction, the excavation-induced ground movements can be obtained once the pattern of the wall deflection is determined.

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2.2 Analytical model of the tunnel Underground metro tunnels are usually constructed using shield machines, especially in soft ground cities. Due to the presence of the joint, the shielddriven tunnel is not a continuous ring structure, which significantly reduces the mechanical proper­ ties of the shield-driven tunnel. In general, the tunnel lining structures are subjected to a combination of earth and water pressures. The pressure distribution pattern acting around a lining structure usually employs the definition of Lee et al. (2001a, 2001b), as shown in Figure 2. For simplicity, the total pres­ sures consist of six parts, as described in the following. (1) p1 refers to the total vertical earth pressure at tunnel crown, and can be estimated by:

(2) p2 refers to the reaction pressure at the tunnel bottom, which equals to the summation of p1 and self-weight of tunnel lining:

where t and γc are the thickness and unit weight of the tunnel lining, respectively. (3) p3 refers to the lateral earth pressure at the crown level of the tunnel lining:

where c is the cohesion of the soil, φ is the friction angle of the soil. (4) p4 refers to additional earth pressure developed at the tunnel invert level:

where q1 is the overburden earth pressure at tunnel crown:

(5) p5 refers to averaged self-weight of the tunnel lining: where γi and hi represent the unit weight and the thickness of the soil layer i. n is the number of soil layers above the tunnel crown.q2 is the vertical earth pressure developed at the shoulder regions, which can be approximated by:

(6) p6 refers to lateral soil resistance pressure. For simplicity, p6 is assumed to distribute over the range of 45°-135° with respect to the vertical dir­ ection around the tunnel, and normal to the seg­ mental lining with a parabolic pattern, as defined in the following expression:

where γ is soil average unit weight at the shoulder regions. where ph is the soil resistance at the springline of the tunnel, and ϕ is the angle measured from the ver­ tical direction around the tunnel. Using the Winkler elastic foundation theory, ph can be expressed as:

Figure 2. The circumferential loads on the cross section of jointed shield tunnels (Lee et al., 2001a).

where Ks is the soil resistance coefficient, and Δh is the horizontal displacement of the tunnel lining at the springline level. As demonstrated by Liu et al. (2016), a large stress relief along the horizontal direction is observed due to excavation, in comparison with a relatively smaller vertical stress relief. Then, the existing tunnels would be horizontally elongated. To obtain the convergence deformations of the tunnel and the corresponding lining internal forces (i.e., the

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bending moment and the axial force), two basic assumptions are required. In this analysis, the hori­ zontal convergence value of the tunnel is set to the difference of subsoil horizontal displacement between the two points at the springline of the tunnel based on Sagaseta’s method. As mentioned before, the effect of soil-tunnel interaction is neglected. The second assumption is that the vertical earth pres­ sure is not affected by excavation, considering a smaller vertical stress relief measured. 3 PARAMETRIC ANALYSES To investigate the effects of various factors on the deformations of existing tunnels, a series of paramet­ ric analysis is carried out, including the tunnel pos­ ition and the soil condition. In this section, a simple example is assumed, as shown in Figure 1. The unit weight, cohesion, and friction angle of the soil are set to 18 kN/m3, 17 kPa, and 18.5°, respectively. The groundwater level is assumed to be at the ground surface. Table 1 lists the parameters of the lining of the tunnel. The thickness of the lining of the tunnel is 300 mm, and the tunnel diameter is 6.0 m. It should be noted that the joint flexural stiffness ratio developed by Lee et al. (2001a, 2001b), λ=Kθl/EI, is employed to characterise the relative flexural stiff­ ness of the joint over the rigidity of the lining seg­ ment. EI is the flexural stiffness of the lining segments (per unit length). l is the calculation length and is selected as 1 m to represent a unit length. The joint flexural stiffness ratio is set to a typical value of 0.1. Zheng and Li (2012) demonstrated that the patterns of the wall deflection can be categorized into four types, cantilever-type, convex-type, com­ posite-type, and kick-in-type. The top of the wall was usually restricted for the excavation in soft soils. Then, the pattern of the wall deflection exhibiting a convex type is employed in this study, described by a simple Gaussian curve, as follow:

Table 1.

The magnitude and shape of the wall deflection can be determined using parameters A, B, C, z0. The sum of parameters A and B is equal to the maximum wall deflection δmax, which is located at the level below the surface of z0. The length of the retaining wall is set to 30 m. Parameters A, B, C, z0 are selected as -3, 33, 8 and 15, respectively. 3.1

Effect of the tunnel position

Figure 3 shows the variations of convergence deform­ ations of the tunnel using Δv and Δh with different normalised clear distances between the tunnel and the diaphragm wall. The clear distance is normalized con­ cerning the tunnel diameter. It should be noted that the soil resistance coefficient of Ks is selected to 15000 kN/m3. Δv and Δh represent the final vertical and horizontal convergence deformations, and the ini­ tial values of them, Δv,ini=13.97 mm and Δh,ini =9.91 mm, are also provided in Figure 3. As expected, the convergence deformations decrease with the clear distance at a reduced rate. It is demon­ strated that the excavation-induced tunnel conver­ gence deformations would be effectively reduced increasing the tunnel-excavation clear distance. As show in Figure 3, when the clear distance is larger than 4D, minor convergence deformations can be observed, which can be the guideline to determine the excavation induced the influence zone. Figure 4 show the lining internal forces (i.e., the bending moment and the axial force) of the half lining, for six values of normalised clear distance, Lt/ D, range from 0.5, 1, 2, 3, 4, to 5. The calculated shear forces are not provided, since the effect of the shear forces on displacement is relatively small. Due to the horizontal stress relief, the increases of the mag­ nitudes of the bending moment and the axial force are observed, but the patterns of the lining internal forces are not be influenced. Note that a positive moment corresponds to the case that the inside surface of the lining is subjected to tension. It can be found that the

Parameters of the lining of the tunnel.

Parameter

Value

Tunnel diameter, D (m) Lining thickness, t (mm) Lining width, b (m) Unit weight of lining, γl (kN/m3) Joint number of each tunnel ring Joint position of half structure (°) Circumferential joint flexural stiffness ratio Young’s modulus of segment (kPa) Shear modulus of segment (kPa) Poisson’s ratio of segment

6.0 300 1.0 25 6 30, 90, 150 0.1 3.45e7 1.44e7 0.2

Figure. 3. Effect of normalised clear distance on tunnel convergence.

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maximum positive moment is observed at the tunnel crown. It implies that the inter surface of the concrete surfers the maximum tensile stress, and local concrete may crack when the tensile stress increases to some extent. Besides, the largest opening occurs at joint 2, at the springline of the lining. 3.2

Figure. 4. Effects of normalised clear distance on the lining internal forces: (a) bending moment and (b) axial force.

Effect of the soil resistance

To explore the influence of the surrounding soil on the behaviors of the lining, three soil resist­ ance coefficients (i.e., 7 500, 15 000 and 30 000 kN/m3) are selected in this analysis. Figure 5 shows the variation of convergence deformations of the lining with different soil resistance coeffi­ cients. Note that in the following cases the cover depth of the tunnel is supposed to be 2D, and the clear distance between the tunnel and the wall is supposed to be D. Induced tunnel conver­ gence deformations are found to decrease with an increase in soil resistance coefficient. When soil resistance coefficient is increased up to 30 000 kN/m3, the horizontal and vertical conver­ gence deformations are reduced by up to 66% and 68%, respectively. This is because ground with larger values of soil resistance coefficient possesses stronger stiffness and then it will sig­ nificantly reduce the excavation-induced deform­ ation in the lining. Figure 6 show the lining internal forces (i.e., the bending moment and the axial force) of the half lining for three soil resistance coefficients. From an inspection of the figure, it can be found that with the increase of soil resistance coeffi­ cient, both the magnitudes of the bending moment and the axial force decrease with soil resistance coefficient. It implies that by increasing soil resistance coefficient the lining internal forces in the tunnel would be reduced. Corres­ pondingly, ground improvement can be an effect­ ive way to improve soil resistance in practice, which could alleviate significantly the adverse effects on existing tunnels use to excavation. 4 LIMITATIONS

Figure. 5. Effect of soil resistance coefficient on tunnel convergence.

The proposed analytical method provides an approximate estimation for the deformation and the inter force of the existing tunnel, which could be used to evaluate the tunnel serviceabil­ ity in a preliminary analysis. However, it should be noted that the soil is assumed as a homogeneous and elastic material. The prelim­ inary work does not consider the influence of the soil-tunnel interaction. Besides, the longitudinal behaviors of the tunnel are not involved in this analysis. Therefore, further study should be per­ formed to involve the above-mentioned factors.

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ACKNOWLEDGEMENTS This paper is supported by National Natural Science Foundation of China (No. 41807512).

REFERENCES

Figure. 6. Effects of soil resistance coefficient on the lining internal forces: (a) bending moment and (b) axial force.

5 CONCLUSIONS Based on Sagaseta’ close-form displacement solution, an analytical approach is proposed to predict the behaviors of shield tunnels adjacent to a braced exca­ vation. A parametric analysis is performed to study the effects of the clear distance between the tunnel and the deflection wall, and the soil resistance on the tunnel convergence deformations and the lining internal forces. The following conclusions can be drawn: (1) The proposed analytical method can be used to evaluate the convergence deformations and the lining internal forces of the segmental lining caused by the lateral excavation. Additionally, the analytical solution can be employed in the preliminary assessment of tunnel serviceability. (2) The results suggest that both the horizontal and vertical convergence deformations of the segmen­ tal lining increases dramatically because of the lateral unloading. The convergence deformation and the lining internal forces decrease as the clear distance increased, albeit at a decreasing rate. (3) Increasing the soil resistance will significantly reduce adverse effects in the deformation of shield tunnels, implying that improving ground stiffness can be an effective way to decrease the effect on the existing tunnel due to adjacent excavation.

Chang, C.T., Sun, C.W., Duann, S.W. & Hwang, R. N. 2001. Response of a Taipei Rapid Transit System (TRTS) tunnel to adjacent excavation. Tunnelling and Underground Space Technology 16(3):151–158. Chen, R.P., Meng, F.Y., Li, Z.C., Ye, Y.H. & Ye, J.N. 2016. Investigation of response of metro tunnels due to adja­ cent large excavation and protective measures in soft soils. Tunnelling and Underground Space Technology 58, 224–235. Hwang, R.N., Duann, S.W., Cheng, K.H. & Chen, C.H. 2011. Damages to metro tunnels due to adjacent Excavations. In: Proceeding of TC302 symposium Osaka, International symposium on backwards problem in geotechnical engineering and monitoring of geo­ construction, Osaka, Japan. Lee, K.M. & Ge, X.W. 2001a. The equivalence of a jointed shield-driven tunnel lining to a continuous ring structure. Canadian Geotechnical Journal 38 (3), 461–483. Lee, K.M., Hou, X.Y., Ge, X.W. & Tang, Y. 2001b. Analyt­ ical solution for jointed shielddriven tunnel lining. Inter­ national Journal for Numerical and Analytical Methods in Geomechanics 25(4), 365–390. Liang, R.Z, Xia, T.D., Huang, M.S. & Lin, C.G. 2017. Sim­ plified analytical method for evaluating the effects of adjacent excavation on shield tunnel considering the shearing effect. Computers and Geotechnics 81, 167–187. Liu, G.B., Huang, P., Shi, J.W. & Ng, C.W.W. 2016. Performance of a Deep Excavation and Its Effect on Adjacent Tunnels in Shanghai Soft Clay. Journal of Performance of Constructed Facilities 30(6), 04016041. Sagaseta, C. 1987. Analysis of undrained soil deformation due to ground loss. Geotechnique 37(3): 301–320. Wu, H.N., Shen, S.L., Liao, S.M. & Yin, Z.Y. 2015. Longi­ tudinal structural modelling of shield tunnels consider­ ing shearing dislocation between segmental rings. Tunnelling and Underground Space Technology 50, 317–323. Zhang, J.F., Chen, J.J., Wang, J.H. & Zhu, Y.F. 2013a. Pre­ diction of tunnel displacement induced by adjacent excavation in soft soil. Tunnelling and Underground Space Technology 36, 24–33. Zhang, Z.G., Huang, M.S. & Wang, W.D. 2013b. Evalu­ ation of deformation response for adjacent tunnels due to soil unloading in excavation engineering. Tunnelling and Underground Space Technology 38, 244–253. Zheng, G., Yang, X.Y., Zhou, H.Z., Du, Y.M., Sun, J.Y. & Yu, X.X. 2018. A simplified prediction method for evaluating tunnel displacement induced by laterally adjacent excavations. Computers and Geotechnics 95. 119–128. Zheng, G. & Li, Z.W. (2012). Comparative analysis of responses of buildings adjacent to excavations with different deformation modes of retaining walls. Chin­ ese Journal of Geotechnical Engineering 34(6), 970–977.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Winkler model for axial deformation of compressible piles due to tunnelling

J.J. Crispin Department of Civil Engineering, University of Bristol, Bristol, UK

ABSTRACT: Predicting the influence of tunnelling induced displacements on nearby structures is essential for urban tunnelling projects. However, piled foundations have a reinforcing effect on the soil, influencing the settlement of the associated structure as well as surrounding structures. This means that traditional greenfield settlement prediction methods are not accurate and instead numerical analysis is required to predict ground movements. A simplified, two-stage analysis can be completed by first predicting the greenfield ground move­ ments down the length of the pile and applying these displacements to the end of Winkler springs representing the interface between the pile and the soil. A simple solution is presented for compressible piles in homoge­ neous soil and layered soil. Extension of the solution to inhomogeneous soil layers and to pile groups is discussed.

1 INTRODUCTION Tunnels are increasingly being constructed in urban areas in the vicinity of existing structures, many of which are supported by piled foundations. Design­ ers must ensure that the movements of affected piles are low enough to ensure serviceability of the structure. A variety of methods are available predict the displacement field due to the tunnel in the absence of any structures, referred to as greenfield settlement (discussed in Section 3). However, the presence of the pile will influence this response; piles closer to the tunnel will be pulled down by larger settlements near the base of the pile while piles further from the tunnel will be supported by the soil below, increasing and reducing the surface settlement relative to the greenfield value, respectively. Simple diagrams have been produced by Jacobsz (2002) and Selementas et al. (2005) that can be used to predict whether the pile settles more or less than the surrounding soil, depending on the position of the pile toe. More rigorous analytical solutions are available employing the two-stage analysis method. This involves predicting the greenfield settlement at the pile location, then applying this displacement to the interface between the pile and the soil. This is similar to the approach used to calculate interaction factors for pile groups by Mylonakis & Gazetas (1998). This approach has been employed by multiple authors using different analysis methods and consti­ tutive models. These include finite difference method solutions, e.g. Huang et al. (2009), Franza et al. (2019) (linear-elastic soil model) and

Williamson et al. (2017) (power-law soil model), and boundary element method solutions, e.g. Chen et al. (1999), Loganathan et al. (2001) (linear elastic soil model) and Basile (2014) (hyperbolic soil model). These solutions all require discretising the pile and a numerical approach. Therefore, an analyt­ ical formulation is appealing to engineers for prelim­ inary analysis. In this paper the two-stage analysis method is employed with a compressible pile and the Winkler model. 2 PROBLEM DEFINITION The geometry of the problem is shown in Figure 1. A tunnel of cross-sectional area, At, is driven at depth, H. This results in a displacement field with vertical component uz(x,z) (the horizontal component is not considered here) where x is the horizontal dis­ tance from the tunnel and z is the depth below ground level. The pile of length, L, and cross-sectional area, Ap, located at distance, s, from the tunnel centreline is modelled as a compressible column of stiffness, Ep, supported by distributed Winkler springs and a base spring of stiffnesses k(z) and Kb, respectively. 3 GREENFIELD RESPONSE A review of greenfield settlement prediction methods can be found in Franza (2016). Most methods to predict subsurface displacements due to tunnelling can be broadly grouped in to two

DOI: 10.1201/9780429321559-68

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4 PILE RESPONSE In homogeneous soil (k(z) = k), the response of a pile to an applied head load can be calculated using the well-known equation for pile head stiff­ ness, K0 (Mylonakis & Gazetas 1998).

where λ is a load transfer parameter and Ω is a dimensionless base stiffness constant:

Figure 1. Problem geometry – tunnel excavated under a piled foundation supported by Winkler springs.

However, the displacement field due to the tunnel results in a distributed load applied to the length of the pile and a point load at the pile base. Therefore the settlement of the pile head, w0, is the sum of the contri­ bution by the pile shaft, w0,s and the pile base, w0,b.

main families. The first are based on the observa­ tion that the surface settlement trough can be approximated by a Gaussian distribution (Peck 1969). O’Reilly & New (1982) suggest that as subsurface movements appear to be directed to a “sink” near the tunnel centre and the subsurface vertical displacements can also be described using Gaussian distributions with varying trough width with depth, shown in Equation 1:

4.1

Pile shaft

w0,s is given by summing the incremental displace­ ments of the pile head due to the applied skin friction at each depth, dw0,s(z):

where VL = volume loss in tunnel construction and i(z) is the trough width parameter. i(z) is a function of depth in the form ið zÞ ¼ c1 H - c2 z. Mair et al. (1993) proposed values of c1=0.5 and c2=0.325 for tunnels in clay based on monitoring data and centri­ fuge testing. The second family of methods are based on an analytical solution for tunnels in an incompress­ ible soil using a virtual image technique by Saga­ seta (1987). Verruijt & Booker (1996) generalised the solution to compressible soil, including an additional component to account for ovalisation of the tunnel. This solution was modified by Loganathan & Poulos (1998) to model a larger gap at the top of the tunnel rather than uniform radial ground movement and by González & Sagaseta (2001) to account for nonlinear soil behaviour near the tunnel. A review of these solutions is provided in Pinto & Whittle (2014). In this paper Equation 1 is employed when dem­ onstrating the proposed model, however it can be substituted with another method, numerical results or field test data if necessary.

where tt(z) is the skin friction at depth z due to the is the incre­ tunnel displacement field and mental pile head displacement due to unit skin friction applied at depth . By symmetry of the prob­ lem, this is the same as the displacement of the pile at depth z due to a unit load applied at the pile head. For homogeneous soils this is given by Equation 6.

where K0 is the pile head stiffness from Equa­ tion 2. An analytical solution to the integral in Equation 4 has not been found. Instead, the result can be approximated using Gauss-Legendre quadrature. Employing three points, Equation 4 can be approxi­ mated using Equation 7.

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Pb,P can be calculated by multiply the applied . For homogeneous soil this head load, P, by yields Equation 13, which matches the result in Crispin et al. (2018).

For most problem configurations this approximation gives results with an error of less than 0.1% of the pre­ dicted settlement of a pile directly above the tunnel (s = 0) relative to an adaptive quadrature algorithm. 4.2

Pile base

The contribution of the base response to the pile head settlement is given by:

For homogeneous soil inputting z = L into Equation 6:

is given by

While tension in the Winkler springs along the pile shaft is possible, indicating negative skin friction, the pile base spring cannot be in tension as the stiffness of this spring relies on compression of the soil below. To check this condition has not been violated, the resultant force on the pile base, Pb, can be calculated by sum­ ming the base load due to the applied pile head load, Pb,P, and the base load due to the tunnels influence on the pile shaft, Pb,s, then subtracting the base load due to the tunnels influence on the pile base, Pb,b. Pb,b can be calculated by multiplying Kb by uz(s,L). Pb,s can be calculated using a similar approach to w0,s:

If the resultant load is negative, a (theoretical) correcting load, ΔPb, should be applied to the pile base to bring the resultant load to zero. This reduces the contribution of the base response to the pile head settlement by Δw0,b:

5 ILLUSTRATIVE EXAMPLE Consider a tunnel excavated 30m below ground level in a stiff clay (shear modulus, Gs = 25MPa) under a building supported by a piled raft. The piles are 20m long, 0.6m diameter concrete bored piles (Ep = 20GPa) at various distances from the tunnel. Assuming the piled raft is fully flexible (the piles can freely settle different amounts) and neglecting interaction effects between the piles, the settlement of the piled foundations at different distances from the tunnel can be predicted using the method described above. The concentric cylinder model (Randolph & Wroth 1978) can be used to relate the soil shear modulus variation, Gs(z) to k(z):

where is the incremental pile base load due to a unit skin friction applied at depth For homogeneous soil this is given by:

Similarly to Equation 4, Equation 10 can be approximated using three point Gauss-Legendre quadrature.

where Ds is the pile shaft diameter, νs is the soil Poisson’s ration and rm is an empirical radius beyond which the settlement due to the pile is assumed to be negligible. Equation 16 was calibrated by Randolph & Wroth (1978) using numerical results. For the soil properties and pile dimensions

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being considered, this yields a Winkler spring stiff­ ness, k = 44MN/m2. Kb can be calculated by approximating the pile base as a rigid punch in an elastic half space (Ran­ dolph & Wroth 1978), leading to Equation 17:

where Db is the pile base diameter and η is a factor to account for the depth below the ground surface, taken as unity (Randolph & Wroth 1978). The base stiffness of the piles in this example is therefore Kb = 50MN/m. Inputting these stiffnesses into Equation 3 gives λL = 1.76 and Ω = 0.10. Substituting these into Equation 6 and multiply the result with Equation 1 and the Winkler spring stiffness, k, yields the dis­ placement of the pile head due to the applied skin friction at each depth, dw0,s(z). Equation 7 and Equa­ tion 8 can then be evaluated to get the pile head settlement due to the contribution of the pile shaft and pile base, respectively. The sum of these two contributions gives the total pile head settlement due to the tunnel excavation. In this case the pile base spring is assumed to still be in compression. The different components of the normalised head settlement for piles at varying distance from the tunnel centreline are plotted in Figure 2, compared to the pre­ dicted greenfield settlement. The predicted behaviour matches the expected response; piles close to the tunnel settle more than the surrounding ground and piles further from the tunnel settle less. In this case, this change in behaviour happens at around 12m from the tunnel centreline. Additionally, as the pile is rela­ tively slender and the base is not very stiff, the main contribution to the head settlement comes from the dis­ placement field affecting the pile shaft.

6 LAYERED SOIL This solution can be extended to multiple soil layers with different stiffnesses. Consider the general case of a pile embedded in N soil layers shown in Figure 3. If soil layer n is homogeneous, the stiffness of the length of pile below the top of that soil layer, Kn-1, is given by:

where λn and Ωn are analogous to λ and Ω, respect­ ively (Equation 19), in which kn is the Winkler spring stiffness in soil layer n and Kn is the stiffness of the length of pile below layer n. Therefore, K0 can be cal­ culated by recursively applying Equation 18 for each soil layer starting with layer N and KN = Kb.

The influence of the skin friction on the pile head settlement depends on the stiffness of the section of the pile both above and below where the load is applied. As with the pile head stiffness, the incremental pile head settlement due to unit skin friction at depth z can be calculated by first getting the incremental settlement at the top of the soil layer, wn-1,s/tt (z), then recursively calculating the influence of this displacement on the top of the layer above. This results in Equation 20:

Figure 2. Pile response due to tunnelling, λL = 1.76, Ω = 0.10, L/H = 0.67.

Figure 3. Problem geometry - Pile embedded in N soil layers.

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where n is the layer that the depth z is in. For a homogeneous soil layer, and (zn) are given by Equation 21 and Equation 22, respectively.

where w(z) is the displacement of the pile at depth z. For k(z) described using a power-law with depth, for which a linear function is a specific case, solu­ tions are available to calculate the pile head stiffness (Scott 1981, Guo 2012, Crispin et al. 2018). The method in this paper could be applied to these soil stiffness profiles by extending these solutions to con­ sider loading along the pile shaft, as has been carried out for pile to pile interaction problems (Crispin & Leahy 2019).

8 PILE GROUP RESPONSE Piles are often installed as part of groups. Any piles within a distance of rm from each other will interact (Mylonakis & Gazetas 1998), reducing their differ­ ential settlement. Huang et al. (2009) suggests that a shielding effect occurs due to the presence of other piles between a pile and a tunnel. However, this second-order effect is not limited to a “shielding” pile (closer to a tunnel) reducing the settlement of a “shielded” pile (further away). The further, “shielded” pile will also in turn reduce the settlement of the closer, “shielding” pile by redu­ cing the displacement of the soil in its vicinity. For the same reason, a closely spaced row of piles par­ allel to a tunnel will settle less than an individual pile of the same dimensions at the same distance from the tunnel. As such, this effect may be better described as a reinforcing effect, where the settle­ ment of a pile due to an external displacement field is reduced by nearby piles regardless of the direc­ tion of the source of the displacements. This effect can be modelled using the approach proposed by Huang et al. (2009), which is based on a method proposed by Mylonakis & Gazetas (1998) for calculating interaction factors between piles in groups. Consider a “source” pile (pile 1) at distance s1t from the tunnel and calculate the settlement of the pile due to the tunnel displacement field as a function of depth, w1t(z). This can be carried out using a similar method to the one for single piles above, modified to work for depths other than the surface. The relative displacement of the source pile compared to the applied displacement field is given by:

Note that z0 = 0 and the empty product = 1, which occurs when z is in the top soil layer (n = 1). There­ fore Equation 20 reduces to Equation 6 for a single homogeneous soil layer. Equation 4 can be split into separate integrals for each soil layer and resolved independently using a method similar to that in Equation 7. These contribu­ tions are then summed together with the contribution due to the pile base (Equation 8) to give the total settlement. Following the same approach for the additional base load in layered soil, Equation 11 can be replaced with Equation 23:

where given by:

for a homogeneous soil layer is

Equation 10 can then also be split into separate integrals and solved similarly to Equation 12. Equa­ tions 13 and 14 can be applied as before, substituting Equation 23 for Equation 11. 7 INHOMOGENEOUS SOIL The pile response functions detailed in this paper can be derived, in addition to further results for inhomo­ geneous soil (k(z) variable with depth), by solving the governing differential equation in Equation 25 and applying the boundary conditions of the spring at the pile base and zero shear force at the pile head.

The displacement field created due to this relative displacement can then be calculated using Equa­ tion 27:

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where s1x is the distance of a point from the source pile and ψ(s1x) is the attenuation of settlement from the source pile with the distance of a point from it, s1x. This is the same attenuation function used in the calculation of pile group interaction factors (Mylona­ kis & Gazetas 1998):

This displacement field can then be input into the single pile head settlement prediction method in this paper for all other piles in the group to give the change in their settlements due to the reinforcing effect of this source pile. This should be repeated with all piles in the group (as the source pile) to get the final settlements. Intuitively, these additional displacement fields will in turn produce higher order reinforcing effects from the piles in the group. However, these will very quickly become negligible and therefore the calcula­ tion only needs to be carried out once. 9 CONCLUSIONS A simplified analytical approach has been presented for predicting the response of compressible piles in homogeneous soil due to tunnelling. A numerical solution to the resulting integral and an illustrative example are provided. The solution has been extended to layered soil and inhomogeneous soil layers and pile group response were discussed.

ACKNOWLEDGEMENTS The author would like to thank Prof. George Mylona­ kis and Dr Paul Vardanega for their supervision and guidance as well as Neil Moss for his helpful com­ ments on the manuscript. This work was supported by the Engineering and Physical Sciences Research Council (grant number EP/N509619/1). No new experimental data was col­ lected during this study.

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Bayesian inference for deep excavations

W.J. de Wolf Fugro NL Land BV, Leidschendam, The Netherlands; former Department of Geoscience and Engineering, Delft University of Technology, Delft, The Netherlands

M. Korff Department of Geoscience and Engineering, Delft University of Technology, Delft, The Netherlands; Deltares, Delft, The Netherlands

A. van Seters Fugro NL Land BV, Leidschendam, The Netherlands

J.H. van Dalen Department of Geoscience and Engineering, Delft University of Technology, Delft, The Netherlands

ABSTRACT: In the Observational method Ab Initio approach observational feedback is used to optimize a structural design to field conditions that are found to be uncertain in the design phase. As a deep excavation takes place, the supporting retaining wall deflections can be observed by inclinometers which serve as an extra source of information on the structure’s performance. Although the Observational method can be benefi­ cial for both safety and economic point of view, limited deep excavations have been executed via this design strategy. This is mainly due to the lack of specification on how measurement-processing can be used to sup­ port engineers to assess the structure’s safety. This study introduces a methodology for real-time measurement-processing with the use of Bayesian updating. In this methodology retaining wall deflections are used to update the unique field conditions of the construction site. This methodology is applied to a measurement set gathered at the construction of a deep excavation in Groningen, The Netherlands, to demonstrate its potential to supplement the Observational method.

1 INTRODUCTION The design of a geotechnical structure needs to deal with the uncertainties in the heterogeneous and dynamic field conditions that are unique for each construction site (Hicks & Nutall, 2012). In the Observational method Ab Initio approach a flexible design and construction plan is made to allow anticipation to observational data (Peck, 1969). This data is gathered throughout different construction phases to give feedback on the per­ formance of the geotechnical design. This way the design can be optimized to the in-situ condi­ tions as they appear, which can be beneficial from both safety and economic point of view (Powderham & Nicholson 1996). It is believed that the design method is especially suitable for deep excavations in the soft soil conditions of The Netherlands, because of a staged construction sequence and non-brittle failure mechanisms (Nicholson et al. 1999, Korff et al. 2013).

Despite its potentials, the Observational method Ab Initio approach has few applications. A valuable design approach is presented by the CIRIA guideline C760 on retaining walls from the UK (Gaba et al. 2017). The guideline proposes the use of a traffic light system as a basis for decision making. How­ ever, due to the integral approach of the Observa­ tional method many concerns are raised on how to actually quantify the safety of the structure being built (Patel et al. 2007). This quantification is required to justify interventions and to avoid and reduce the impact of any unforeseen event. In order to do so, it is necessary to subtract crucial informa­ tion from the observational data. In this paper a method is shown for real-time data interpretation with the use of Bayesian updating (Ang & Tang, 2007). In the Bayesian update the pre­ dicted retaining wall performance, as assessed by a computational model in the design phase, is com­ bined with the information obtained from observa­ tions during construction. By describing both sources

DOI: 10.1201/9780429321559-69

529

of information via probability density functions dif­ ferent uncertainties in both the design and construc­ tion phase can be weighted in the outcome of the update. As construction continues a Bayesian update can be performed each time new observations are available. Consequently, the retaining wall behaviour as predicted by the computer model can be re­ assessed throughout the different excavation stages. This paper is structured as follows. First, the prin­ ciples of the Bayesian updating method are intro­ duced. The method is then demonstrated using a case study of a deep excavation in The Netherlands. This is followed by a more general evaluation of the proposed method, to finally conclude on the value of the Bayes­ ian update in the context of the Observational method for the use of deep excavations in soft soil conditions. 2 BAYESIAN UPDATING 2.1

Monitoring wall displacements

Spross et al. 2014 promoted the potential of Bayes­ ian inference by presenting a fictive case of a rock pillar. They illustrated the use of Bayesian updating to find a deformation modulus in accordance with a fictive measurement set. To show general validity of his method, they concluded that more case studies of different geotechnical kinds are needed. This work was used in the underlying study to look into the potential of Bayesian updating for processing retaining wall displacements for deep excavations. Often strict SLS requirements are set on the max­ imum allowable deflection of the retaining wall. Therefore, the performance of the retaining wall can be directly related to the movements that occur during sequential excavation (Gaba et al. 2017). At the beginning of construction there is only one source of information available on the expected deformations of the wall, namely the prediction made by a computational model. This prediction is based on the model input that represents the engineers’ assump­ tions of the initial field conditions. Inevitably, those assumptions carry uncertainty, the magnitude of which depends on site-investigation efforts. The goal of Bayesian updating is to reduce these initial uncer­ tainties by adding a second source of information, which is the monitored wall displacement. Each time measurements are taken, this observational data is combined with the prediction to form a new updated prediction on the expected deformations. This com­ bination is done by the equations of the Bayesian update (Ang & Tang, 2007) that requires the quantifi­ cation of all the uncertainties in both the design, con­ struction and monitoring of the retaining wall. 2.2

model input follows directly from the heterogeneity of soil strength properties and is commonly described by a normal distribution (CUR, 2008). Other input uncertainties can be taken into account as well, for example the natural fluctuations of water levels. The impact of the variance of each input param­ eter can be first accessed by a sensitivity analysis. Each parameter is then assigned with a sensitivity score that represents the impact of its variance on the modelled outcome. Because each construction phase is different, the sensitivity score varies per phase. Hence, during different construction phases, differ­ ent soil parameters can become relevant or irrelevant for the structure’s stability. Consequently, the variance of the prediction on the structure’s performance can be quantified by means of a Monte Carlo simulation. For this simula­ tion only the parameters with significant sensitivity scores should serve as stochastic input. The Monte Carlo simulation then returns an output distribution with properties μMC and σMC representative for the uncertainty in the expectation of retaining wall dis­ placements (Figure 1). For calculations performed with the Spring model, this is typically a lognormal distribution (De Wolf, 2019). To recognize the inaccuracy of the calculation model to predict retaining wall displacements an additional model error is added to the output distri­ bution of the Monte Carlo simulation. For the Spring model, an absolute error of 10% is added in accord­ ance with the Dutch CUR standards. Finally, the properties of the lognormal distribution can be described by mean μd and standard deviation σd . 2.3

Measurement uncertainty

Measurements of the retaining wall displacements have an uncertainty introduced by: (1) The inherent variability between then measure­ ment devices: σ2inh: (2) Error of the measurement device: σ2m:e . Both these factors can be combined via:

Predictive uncertainty

The prediction contains (1) uncertainty in the model input and (2) inaccuracy of the computational model used (De Wolf, 2019). The uncertainty in the

Figure 1. Principle of Monte Carlo simulation.

530

with ζ 2 being the variance of the measurement distribution. Other errors can be added as well to equation (1) (Spross et al. 2014). The mean of the measured dataset is indicated by �x. Note that the natural logarithm in equation (1) is used to transform the normally distributed variances to their lognormal equivalent. This is done in order to combine these measurements with the lognor­ mally distributed prediction in the Bayesian update. 2.4

Bayesian update

Once the first construction phase has been completed the measurement data obtained is added to the ori­ ginal model prediction with initial properties μd , σd . The outcome of the Bayesian update, that follows from applying equations (2) and (3) (Ang & Tang,0 2007), 0 is therefore indicated by the superscripts μd and σd : The Bayesian update can be repeated after 0 the second construction phase has finished: the μd 0 and σd obtained after processing the first measure­ ment set will be updated again with 00a second meas­ 00 urement set, leading to μd and σd . This can be repeated each construction phase such that the model prediction is updated with all the information on hand.

Þ n

2.5

Figure 2. Outline of the methodology.

3 CASE STUDY 3.1

Project description

The methodology is demonstrated by means of a case study that concerns the construction of a 2-layered basement in the north of The Netherlands. Due to the geological history, strong heterogeneity is expected at the construction site. Additional uncertainty is intro­ duced by basing the design on a rather simplified 2D spring model (Figure 3), whereas the asymmetric shape of the deep excavation (Figure 4) might

Calibration

Each updated prediction is associated with a new set of input parameters for the model. This param­ eter set can be found by calibration, which is in principle the opposite of a Monte Carlo simulation (Figure 1): Given a new range of displacements, the calibration is looking for the representative stochas­ tic input distributions. As more observations are added, ideally, the range of possible displacements decreases. This implies that the initially assumed parametric uncertainty decreases as well. Conse­ quently, the calibrated parameter distributions can be used to assess the safety of the build geotech­ nical structure via the principles of Eurocode 7 (Vrijling, 2015). The above described methodology is summarized by Figure 2.

531

Figure 3. Implemented design with stratigraphy.

the field were kept very limited. For cross-section 4, the observed maximum deflection was less than 5 mm. In this case study, the data of the 2 inclinom­ eters of cross-section 4 are analysed with the goal to find representative parameters for loamy clay layer 2B. In the site-investigation no laboratory tests were performed. Instead, Table 2B of the Dutch National Annex NEN9997-1 was used to derive the characteristic soil parameters from CPT data. Therefore, the coefficients of variations (COV) are adopted from the Annex as well to stochastic­ ally describe the uncertainty of these parameters. Their values are stated in Table 1. Figure 4. Top view of the building pit.

3.2

actually lead to a more favourable distribution of strut forces (Fuentes et al. 2018). The implemented design with 3 layers of struts was based on a limit state design and a soil profile taken from the site-investigation (Table 1 and Figure 3). To avoid any installation problems, a Mixed in Place (MIP) wall was selected. During the excavation phases, see Table 2, no difficulties were obtained. In fact, the actual displacements in

Table 1.

Characteristic values for soil layers.

Soil parameters

γsat [kN/m ] φ [� ] δ [� ] c [kPa] OCR [-] k1 [kN/m3] 3

Table 2.

1A.Sand

2A.Clay

2B.Loam/clay

COV

19 30 30 1.2E+04

19 22.5 22.5 4.0E+03

21.5 28 28 2.5 3.0 6.0E+03

5% 10% 10% 20% 20% 20%

Description

1 2

Installation of MIP wall at +7.6 m. Excavation +6.1 m, installation 1st layer of struts. Excavation +3.52 m, installation 2nd layer of struts. Excavation +0.6 m, installation 3rd layer of struts. Excavation: final depth -2.0 m. Installation of concrete floor. Construction of basement floors, stepwise removal of struts.

4 5 6 7

3.3

Bayesian update

Next, the Bayesian update is performed. The inclinometer error σm:e: is set to be 1.36 mm (De Wolf, 2019). This is a rather large standard devi­ ation compared to the actual observations that were less than 5 mm. Therefore the variance of this error in the Bayesian update is reduced for each construction phase via equation 4.

Construction phasing.

Phase #

3

Sensitivity analysis

At first a sensitivity analysis is performed to see what parameters dominate the structural response. To take into account the possible underestimation of the structural force distribution, it was chosen to vary the stiffness modulus EI with 20% alongside the soil strength parameters. The results of the sensitivity analysis are pre­ sented in Figure 5. The friction angle of top layer 1A only has significant influence to the wall deformation for the first part of the excava­ tion. It can be seen that the EI becomes relevant once a certain depth of excavation has been passed. The strength properties of layer 2B con­ tribute to the structure’s performance throughout the whole excavation process. Its friction angle has an increased dominant impact during con­ struction phases 4 and 5.

Figure 5. Result of the sensitivity analysis.

532

Table 3. Mean parameter values for MIP wall and soil layer 2B assumed in design phase (original) and calibrated via Bayesian update in phase 5.

Figure 6 illustrates the Bayesian updates through­ out the excavation phases 2 to 5. It can be seen that the shape of each update gets smaller in the process. As more observational data is added, the incorrect­ ness of the originally assumed parameter input is confirmed as the Bayesian update shifts more towards the measurements. At phase 5 the Bayesian update has fully converged. This means that at that phase the updated prediction coincides with the field observations. The parameters found by the calibra­ tion in phase 5 should thus be representative for the overall observed structural deflections. Figure 7 pre­ sents the results of a forward simulation of the Spring model, performed with these calibrated input parameters (Table 3). 3.4

Mean parameter values: μ

�

2B.φ [ ] 2B.OCR [-] 2B.k1 [kN/m3] EI [kNm2/m]

Original

Calibrated phase 5

33.5 4.0 9.0E+03 8.75E+04

40.4 3.6 50.5E+03 19.19E+04

Calibrated results

Figure 8 illustrates the change of the probability density function (pdf) of the friction angle of layer 2B for different phases. The calibrated results follow the outcome of the sensitivity analysis. It can be seen in Figure 5 that the biggest change in the meanμ is at phase 4 as its sensitivity score strongly increased relative to phase 3. This update in μ can be

Figure 8. Calibrated friction angle of layer 2B.

noticed by the shift of the pdf of phase 4 towards the right. Consequential calibration of phase 5 confirmed this updated μ, leading to a significant decrease of standard deviation σ. Such results could not be found for every input parameter. For both the stiffness EI and the modu­ lus of subgrade reaction k1 of layer 2B the coeffi­ cients of variations remained around 10%. As noticed during the first calibration phases, many mutual combinations were possible between the two parameters leading to the same calculated retaining wall displacement (Figure 9). This means Figure 6. Bayesian update for construction phases 2-5.

Figure 7. Forward simulation with calibrated parameters of phase 5 shows a fit towards the overall observed structural displacements.

Figure 9. Possible parameter input combinations with same displacement as outcome as found for calibration of phase 3.

533

that a slightly higher modulus of subgrade reaction, with a lower stiffness value, leads to the same out­ come. This result makes sense as both parameters have a comparative sensitivity score with a comparable effect on the outcome of the Spring model as can be derived from the sensitivity analysis.

4.2

4 DISCUSSION 4.1

Interpretation of the calibrated solution

The calibrated parameters of phase 5, presented in Table 3, fit the overall measured wall deflection as shown in Figure 7. However, these values are not truly representative for the actual soil conditions: Both the calibrated values for wall stiffness EI and the modulus of subgrade reaction 2B.k1 are unrealis­ tically high. Instead, it should be realized that the results are simply a way to fit the computational model to the observations. Unrealistic calibrated parameters could thus indicate shortcomings in the computational model. The process of the Bayesian update allows to assess the structure’s safety in real-time if, and only if, these results hold till the end of construction. Therefore, the type of computational model is an important choice as it should accurately simulate soil behaviour and soil-structural behaviour in time. For soft soil conditions, this means that it is espe­ cially important that the model is able to recognize the temporary effects of undrained soil behaviour. In the case study however, the long timespan of the overall measurement set indicated that undrained soil behaviour was not relevant during the staged excavation (De Wolf, 2019). Instead, the observed limited MIP wall displacements are most likely a result of more favourable force distributions between the structural elements and the soil than previously assumed. This might have been caused by the following explanations: (1) The cement-soil mixture of the MIP wall has been modelled incorrectly by assuming a too low modulus of subgrade reaction. (2) The stiffness due to the asymmetric shape of the deep excavation might have been more favour­ able than assessed by the 2D Spring model. The contribution of each explanation could not be found due to a lack of additional data. In the project, no strut forces were measured alongside the con­ sidered cross-section. It would have been valuable to have this extra data to be more decisive on the actual force distributions between the structural elements. Additionally, there is little guidance for the selection of input parameters for modelling a MIP wall in existing literature. All in all, it should be realized that limitations of the model and observational data affect the possibilities of engineers to fully explain the site-conditions.

534

Features affecting the performance of the Bayesian update

A valuable feature of the Bayesian update is that uncertainties can be taken into account for both the model and the observations. With this method the prediction on the structure’s performance can be updated anytime during the construction process based on the weight of each uncertainty. According to equations (2) and (3) enough n measurements should be performed in order to converge. Before convergence, the structure’s performance is carefully assessed by not fully rejecting the original predic­ tion. Depending on the certainty of each Bayesian update with regards to this original prediction, the structure’s performance can be reassessed. This can be beneficial in both safety and economic points of view as this methodology can timely reveal flaws in the set-up of the original model prediction. For instance, based on the growing set of measurements in the case study, the decision could have been made to adjust the structure in construction phase 4, as the Bayesian update indicated the incorrectness of the original prediction. Therefore, the Bayesian update and corresponding calibrated parameters can be used to justify the decision to continue with a more eco­ nomical design alternative with two struts instead of three. Typically, the necessary number of measurements in order to converge depends on the specification of the model and measurement errors. However, deter­ mining the magnitude of these errors can still be sub­ jective and case dependent. As demonstrated in this case study, it might be desired to adjust the measure­ ment error as the number of observations grows. However, such choices are not generally specified and might not be easily made in real-time. 5 CONCLUSIONS The case study showed a successful application of the Bayesian update. The value of the proposed methodology is primarily in the ability to determine a set of soil parameters to fit the computational model to the observational data. Consequently, this can be used to justify decisions to adjust the original design. It needs to be emphasized that the calibrated results are not truly representative soil parameters if the constitutive model is not suitable for the siteconditions. Also, additional sources of information might be needed to fully interpret the measured data. The presented methodology with the Bayesian update works with different computational models. In order to use this method for real-time measure­ ment processing in the Observational method it is necessary to investigate some aspects further. The use of different computational models could be tested in order to take into account undrained soil behaviour as relevant for soft soil conditions. It would be valuable to extend this method for an

excavation with adjacent buildings and to combine multiple observed quantities. Therefore, the method­ ology could be extended to add, for example, surface settlements as an additional information source. Finally, the quantification of the uncertainties in both the model and observations play an important role in the outcome of the Bayesian update. Therefore, more case studies should be performed to come up with a more ambiguous formulation of these uncertainties. To conclude, it is believed that the presented methodology with the Bayesian update has the potential to enrich the Observational method in the applications of deep excavations. It could be a starting point for a more active use of real-time measurement interpretation.

REFERENCES Ang, A. H. S., & Tang, W. H. (2007). Probability concepts in engineering: emphasis on applications in civil & environmental engineering (Vol. 1). New York: Wiley. CUR-Commissie, C. 135 (2008) Van onzekerheid naar betrouwbaarheid. CUR-Commissie C, 135. De Wolf, W.J., (2019). The Observational Method for building pits in soft-soil conditions: A study on measurement-processing and feasibility of the Observa­ tional method Ab Initio approach. Repository TU Delft, student master thesis. Fuentes, R., Pillai, A., & Ferreira, P. (2018). Lessons learnt from a deep excavation for future application of the observational method. Journal of Rock Mechanics and Geotechnical Engineering, 10(3),468–485.

Gaba, A, Hardy, S, Doughty, L, Powrie, W, & Selemetas, D. (2017). Guidance on embedded retaining wall design. Report C760, CIRIA, London (pp.203–219). Hicks, M. A., & Nuttall, J. D. (2012). Influence of soil het­ erogeneity on geotechnical performance and uncer­ tainty: a stochastic view on EC7. In Proceedings 10th International Probabilistic Workshop, Universität Stutt­ gart, Stuttgart (pp. 215–227). Korff, M., De Jong, E., & Bles, T. J. (2013, September). SWOT analysis Observational Method applications. In Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris (pp. 2–6). Nicholson, D, Tse, C & Penny, C. (1999). The Observa­ tional Method in ground engineering – principles and applications. Report 185, CIRIA, London Patel, D., Nicholson, D., Huybrechts, N., & Maertens, J. (2007, September). The observational method in geotechnics. In Proceedings of XIV European Confer­ ence on Soil Mechanics and Geotechnical Engineering, Madrid (pp. 24–27). Peck, R.B. (1969). Advantages and limitations of the obser­ vational method in applied soil mechanics. Ninth Ran­ kine Lecture, Geotechnique 19, Ko. 2 (171–187). Powderham, AJ & Nicholson, DP (1996) “The Observa­ tional method in geotechnical engineering” ICE, Thomas Telford, London. Spross, J., Johansson, F., Stille, H., & Larsson, S. (2014, January). Towards an improved observational method. In ISRM Regional Symposium-EUROCK 2014. Inter­ national Society for Rock Mechanics and Rock Engineering. Vrijling, J. K. (2015). Probabilistic Design: Risk and Reli­ ability Analysis in Civil Engineering. Collegedictaat CIE413.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

An underground excavation in Barcelona and its interaction with existing structures A. Di Mariano Geomechanics Group, Centre Internacional de Mètodes Numèrics a l’Enginyeria (CIMNE), Barcelona

A. Varga Department of Engineering Geology and Geotechnics, Budapest University of Technology and Economics (BME), Budapest

A. Gens Department of Civil and Environmental Engineering and Geosciences, Universitat Politècnica de Catalunya (UPC), Barcelona

ABSTRACT: The construction process for the enlargement of an underground railway station in Barcelona required an additional tunnel excavation below existing structures. The paper presents the case study, describ­ ing the details of the excavation method and the support systems used both to guarantee the stability of the tunnel excavation and to limit the effects of the underground constructions activities on the buildings above the alignment. The movements observed in the buildings, once the excavation was completed, are compared with some class A predictions. The construction works had a favourable outcome and no building damage was reported because of the excavation.

1 INTRODUCTION The metropolitan area of Barcelona is densely urban­ ised with a population of over 3 million inhabitants. The public transport system consists of a wide net­ work of tram, metro, city buses and suburban rail­ ways. With the aim to meet the increasing traffic and transportation demand, the metropolitan transport authority ATM (Autoritat del Transport Metropolità) continuously improves the public transport system (Schwarz et al. 2006). Nowadays, 12 metro lines with more than 160 stations compose the Barcelona subway network, covering a considerable part of the city. Provença Station is one of these metro stations, originally built in 1927 (Figure 1). To upgrade its accessibility and evacuation facilities, ATM undertook the enlargement of the station. The construction works included the execution of diaphragm walls and stairs from the platform to street level for new emergency exits, the modification of the original station hall and the excavation of a platform exten­ sion, parallel to the old station. The latter allows the enlargement of one of the platforms of the sta­ tion by means of openings through the adjacent lat­ eral walls of the existing station (Figures 2-4). The excavation of the platform extension, which

involves soft ground materials and is located below existing structures, started in July 2018 and lasted a little over 4 months. In this context, tunnelling­ induced ground movements constituted a major concern and systems to minimize ground deform­ ations were considered beneficial (Gens et al. 2006, Schwarz et al. 2006, Di Mariano et al. 2007, Di Mariano & Gens 2014). The paper presents the case study of the Pro­ vença station enlargement and describes the exca­ vation methods used in the project, along with the techniques employed to support the excavation face. A comprehensive instrumentation system allowed monitoring of the excavation-induced movements in the ground as well as in the existing structures near the tunnel. Before starting the exca­ vation, a number of numerical analyses were car­ ried out by the contractor to provide class A predictions for the performance of the buildings affected by the excavation. The paper shows the comparison of the movements observed in the buildings at the end of the excavation with the numerical results. The comparison shows that the adopted construction method gave rise to very small movements both in the ground and in the sur­ rounding structures. Observed movements were smaller than class A predictions.

DOI: 10.1201/9780429321559-70

536

Figure 3. Plan of the new layout of Provença station.

Figure 1. Map of Barcelona and location of Provença station.

2 LOCATION AND GROUND CONDITIONS Provença Station is located at the city centre of Bar­ celona, beneath Balmes Street, in the area between Rosselló and Provença Streets (Figures 1-2). From the late ‘80s, growth of passenger numbers in one of the platforms of the Station had produced serious difficulties, affecting the security and comfort of all users. The problems of saturation affected mostly the platform where trains head towards the outskirts of the city, especially at peak hours (Figure 4). The high number of passengers led to the enlargement of this part of the Station, whose layout is rectilinear with a length of about 90 m and whose original plat­ forms have a width of about 3 m (Figure 4). The excavation of the new tunnel, parallel to the old sta­ tion and with a total length of 50 m, allows increas­ ing the original platform width up to about 8 m (Figures 3-4). Once the excavation was com­ pleted, the new tunnel was connected to the existing platform by demolishing part of the lateral walls between the piers of the station vault. Geological conditions around the tunnel alignment involve different layers of Quaternary soil deposits, ranging from clayey to gravelly materials. In the

Figure 2. Location of Provença station and project area.

Figure 4. Cross-section of the new layout of Provença Station.

specific area of the station, a superficial layer of fill overlies several lenticular bodies of clay and silt, alternating with layers of gravels immersed in a clayey or silty matrix (Figure 5). Randomly, some bodies of clean gravels or sands appeared in bore­ holes drilled close to the excavation area. The arrangement of materials in the geological profile is due to the double alluvial and colluvial origin of these deposits derived from the presence of a significant watercourse that followed the current layout of Balmes street (Figures 2-3). The new tunnel runs mostly through clayey mater­ ials with variable contents of silt and gravel. At the

Figure 5. Geological profile of the project area. The number above the profile correspond to the street numbers of the buildings above the new tunnel.

537

depth of the tunnel crown, appearance of gravel with a clayey or silty matrix was expected. The low tunnel cover (in the range of 8 to 9 m), the presence of a quite thick fill layer at the ground surface as well as the possible presence of gravel at the tunnel crown led to the selection of a cautious excavation procedure. In the area of the project, the water table is located below the tunnel invert at a depth of about 14 m from the ground surface, corresponding approximately to the depth of the foundation of the old station. 3

EXCAVATION WORKS

Underground excavations in soft ground generally induce soil movements, which can be a risk for nearby structures, particularly in congested urban areas. The specific characteristics of the enlargement of Provença station, with its complex geometry and reduced length of excavation, forced the selection of conventional excavation methods for the tunnel con­ struction. In conventional tunnelling, the stability of the face is generally of critical concern to minimize ground movements, especially in urban environ­ ments. Ground improvement methods are often necessary and, among them, in-tunnel support methods represent an efficient solution. In the case of Provença station, the Forepoling Umbrella System (FUS), sometimes referred to as the Umbrella Arch Method (UAM), was adopted to con­ trol tunnelling-induced ground deformations (Cal­ vello & Taylor 1999, Carrieri et al. 2002, Volkmann & Schubert 2007, Juneja et al. 2010, Yeo 2011, Pinyol & Alonso 2012, Le 2017, Le & Taylor 2017). Steel pipes (also indicated as forepoles) were installed from the tunnel face to provide a roof above the tunnel heading (Figure 6 and 7) and grout was injected through them to form a closed canopy (Le 2017). Forepoles offer immediate ground sup­ port, allowing the excavation to be carried out shortly after their installation (Le & Taylor, 2017). Each steel pipe has a total length (L) of 12 m and their insertion angle (β) varies from a minimum of about 6º to a maximum of approximately 8º, with the embedded length (EL) being 3 m (Figure 6).

Figure 6. Forepoling Umbrella System (Le 2017, Le & Taylor 2017).

Figure 7. Picture of the forepoles injected with grout to form a closed canopy.

The centre-to-centre spacing between forepoles (S) is 200 mm and their diameter is about 140 mm (Figure 7). For safety reasons, the filling angle, α of the FUS (Figure 6) has an exceptionally high value of roughly 85º. Due to the complex geological char­ acteristics of the site and the presence of buildings just above the tunnel alignment along Balmes Street, the excavation process was carried out with very short advance steps, installing the lining almost at the tunnel face. In addition, a comprehensive and accurate monitoring survey was planned to guarantee a good control of the construction works. Tunnelling commenced below the intersection between Balmes and Rosselló streets and progressed towards Provença Street (Figure 2) in the direction from mountain to sea. The 50 m long tunnel runs parallel to Balmes Street except for a short initial section (having a total length of 8 m), near Rosselló Street, that has a slight curvature to adapt to the changing transverse section of the station (Figure 3). The cross-sectional area of the excavation is in the range 18-20 m2 and, as a face support system, the project initially involved the use of a bench-and­ heading approach, with delayed excavation of the central part of the face (ITA/AITES 2007). In add­ ition, the project included the use of confining layers of sprayed concrete, in the excavated part of the tunnel, as well as face fibreglass bolting or grout injections, whenever necessary. The primary lining support involved the placement of steel ribs every half a metre of excavation sealed against the ground by a 28 cm thick shotcrete shell. The final lining, which was installed close to the tunnel face (at a maximum distance of 10 m from the face), has a minimum thickness of 30 cm and consists of cast­ in-situ concrete elements, reinforced with wire mesh. The final invert slab is of the same type as the final tunnel lining and has a thickness of 85 cm. The connection between the new tunnel and the existing platform is ensured by means of four 5 m wide openings through the adjacent lateral walls of Provença station. Piers of 2.2 m in width and 2.0 m in thickness -that are part of the existing struc­ ture of the station- separate these wide openings.

538

These piers actually coincide with the existing reinforcements of the station vault. To improve fur­ ther the resistance of the piers and guarantee the sta­ bility of the new structure, 11 micropiles were installed through each pier from street level with total lengths in the range 27-30 m. In total, 44 micropiles were installed and their tips are located well below the existing foundation of the station. The connection of the new tunnel with the existing station was exe­ cuted after its excavation was completed. 4 CLASS A PREDICTIONS AND ESTIMATION OF MOVEMENTS Before starting the excavation, the contractor carried out different 3D numerical analyses to make class A predictions of the performance of the buildings affected by the excavation and to decide whether additional ground improvement was required. The analyses started considering the behaviour of a typical building in Balmes Street, during its con­ struction. Then, they determined how the subsequent construction of the old Provença Station affected the existing building. Finally, they estimated the ground movements that the construction of the new tunnel would cause to the building. The numerical model also considered the openings that connect the new tunnel with the existing platform (Figure 4 and 8). In order to simplify the characterization and study of the ground, the soil materials were grouped into five main formations. The most superficial one is the fill layer (F), followed by a layer of clay with gravel and silts (CGS) overlying a deposit of clayey sand (SC). Below, a layer of clay with sand (CS) overlies a formation of sands and silty sands (SSS). The constitutive model reproducing the ground behaviour was a simple linear elastic-perfectly plastic model with a Mohr-Coulomb failure criterion. Table 1 summarises the mechanical parameters used in the analyses for each soil layer. Linear elastic constitutive models were adopted to reproduce the behaviour of the old station, the new tunnel and the foundation of the building. On the other hand, an elastoplastic con­ stitutive model with Mohr-Coulomb failure criterion

Figure 8. Scheme of the 3D Numerical model.

Table 1. Geomechanical parameters of the ground materials. Width

γ

c´

ɸ

E

Soil Layer

m

kN/m3

kPa

º

MPa

F CGS SC CS SSS

3.0 6.0 4.0 3.0 3.5

18 20 20 21 21

5 50 10 60 20

28 28 33 29 33

12 16 35 25 40

represented the performance of the building. Numer­ ical analyses did not consider the hydraulic conditions of the site because the excavation took place above the water table and were performed drained. The numerical model was modified a number of times to improve progressively the representation of the real problem, which is quite complex. In general, the results of the different numerical calculations were consistent with one another, attesting the robustness of the analyses and contributing to increase confi­ dence in the reliability of the results. Yet at the time, the numerical model could not be validated with insitu monitoring data. Figure 9 shows the predicted settlements profiles at the base of the building in a representative trans­ verse section of the model. The settlement profiles refer to the three most relevant phases of the numer­ ical analyses, the construction of the building, the excavation of the old station and the construction of the new tunnel. The final maximum predicted total settlement is just above 45 cm (Figure 9). The results obtained in terms of the building deformations due to the new tunnel excavation were associated with very slight damages for the buildings of Balmes Street, with maximum critical tensile strains less than 0.00075 (value suggested by Polshin & Tokar 1957, Burland & Wroth 1974 and Boscardin & Cording 1989 for cracking to first become notice­ able). Therefore, based on the numerical results, Pro­ vença station was not designated as requiring

Figure 9. Predictions of maximum total settlement profiles at the base of the buildings in Balmes Street. First numer­ ical results associated with the originally planned excava­ tion procedure.

539

mandatory ground treatment. Nevertheless, in view of the critical tunnelling conditions, the Contractor proposed a slight modification to the designed exca­ vation procedure. The purpose of such modification was also to avoid the construction constraints associ­ ated with the use of a sequential excavation approach in a tunnel of small dimensions. The modified excavation procedure involved a full-face tunnelling method (Lunardi, 2000, Lunardi & Barla 2014, Barla 2016) with a fibreglass reinforcement at the tunnel face as well as a tunnel cross-section closure after each excavation stage (Figure 10). Temporary steel sets were used for the quick cross-section closure and excavation stages of just half a metre were planned (Figure 10). The tunnel face reinforcement consisted of fibre-glass bolts of the same length as the forepoles (12 m). The Contractor carried out additional 3D and 2D numerical analyses to make new class A predictions of the performance of the buildings affected by the excavation. According to the results of the additional numerical calculations, the new excavation proced­ ure guaranteed the same tunnel face stability as the initially designed one; hence, it was accepted and implemented. Figure 11 shows the new maximum predicted incremental ground settlement profiles due to the tunnel excavation, at the end of the construction pro­ cess (continuous curve). Figure 11 also illustrates, for comparison, the “greenfield” Gaussian normal distribution curve at the ground surface (discontinu­ ous curve), associated with an equivalent circular

Figure 11. Numerical class A prediction of settlement pro­ files at the ground surface, due to the tunnel excavation. The “greenfield” Gaussian settlement trough is also shown for comparison.

tunnel of the same cross-sectional area as the tunnel to be excavated. The empirical normal distribution curve, whose well-known mathematical expression is given by Equation 1 (Peck 1969, Schmidt 1969, Attewell & Woodman 1982, O’Reilly & New 1982, Rankin 1988, Mair & Taylor 1997, Mair 2008), was evaluated as a preliminary assessment to roughly estimate the maximum settlement of the ground, ignoring the presence of buildings and considering a volume loss, VL, of 1.0% (Eq. 3). Volume losses of about 1.0% are frequent in conventional tunnelling and were considered reasonable for the new excava­ tion. The surface settlement s(x) at any distance x from the tunnel centre-line can be expressed as a function of VL (volume of the surface settlement trough per unit length of the tunnel expressed as a percentage of the notional excavated volume) through the integration of Equation 1 and its combination with Equation 2 (Eq. 3).

where s(x) = surface settlement at a distance x from the tunnel centre-line; smax = maximum surface settlement and i = value of x at the point of inflection of the transverse distribution (Figure 11). In most cases, the parameter i (Eq. 1) has a linear relationship (O’Reilly & New 1982, Mair et al. 1993) with the depth of the tunnel axis, H0, (Eq. 2).

Figure 10. Picture of the full-face tunnelling method. Installation of temporary lining and fibreglass reinforcement after one of the excavation stages.

As it is known, the parameter K depends on the soil type and generally increases with depth, z (Mair et al. 1993). In the area of Provença sta­ tion, the tunnel axis is at a depth of about 12 m and K assumes values of about 0.5 at the ground surface (based on previous construction experiences).

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In the Gaussian curve of Figure 11 (Eq. 3), the following values apply, VL = 1.0%, K = 0.5, H0 = 12.0 m and Deq = 5.0 m (equivalent tunnel diameter). Maximum numerical predicted volume losses were below this value and around 0.7% (Figure 11). Comparison between the numerical settlement trough and the “greenfield” distribution curve (Figure 11) shows how the presence of buildings affects the numerical movements’ distribution at the ground surface (Potts & Addenbrooke 1997, Burd et al. 2000, Son & Cording 2005, Franzius et. al 2006, Dimmock & Mair 2008, Farrel 2010, Goh & Mair 2011, Rampello et al. 2012, Giardina et al 2015). The numerical analyses did not consider aspects such as the nonlinear behaviour of the build­ ings, the initial structure damage and/or the torsional behaviour of the buildings that also influence the complex soil-structure interaction problem. Figure 12 shows the class A predictions in terms of the horizontal movements’ distribution along the ground surface (continuous curve). The graph also includes, for comparison, the “greenfield” distribu­ tion of horizontal displacements, u, (discontinuous curve) based on the assumption of O’Reilly and New (1982) that the resultant vectors of ground movement are directed towards the tunnel axis (Burland 2001). According to this assumption, as it is well known, horizontal displacements at the ground surface depend on the settlements’ distribution according to Equation 4.

where s(x) = ground surface settlement at a distance x from the tunnel centre-line (Eq. 3).

Figure 12. Numerical class A prediction of the ground sur­ face horizontal movements in three representative section of the model. The “greenfield” horizontal movements’ dis­ tribution is also shown for comparison.

The presence of the structures on the surface influ­ ence not only the settlements’ but also the horizontal movements’ distribution (Figure 11-12). According to the numerical results, the Balmes Street’s side affected by the excavation, on top of which the buildings are located (right hand side of Figures 11-12), experiences greater movements than the part of the street with no buildings on top (left hand side of Figures 11-12). Numerical class A predictions, associated with the new excavation procedure, indicated settle­ ments at the ground surface with maximum values of about 14 mm (Figure 11), yet much lower hori­ zontal movements (with maximum values of about 4 mm, Figure 12). Numerical results also showed that total settlements were associated with build­ ings’ maximum slopes of about 1/270. According to common damage assessment criteria (Rankin 1988, Boscardin & Cording 1989, BRE 1995, Mair et al. 1996, Burland 2001), these values could lead to possible superficial damage in the buildings, yet they would hardly have structural significance (risk category 2). Once expected movements were evaluated, the excavation started with the awareness that the tech­ nical success of the construction works and the limi­ tation of ground movements would depend fundamentally on the care and quality of the execu­ tion of the adopted modified excavation process. 5 OBSERVED MOVEMENTS AND NUMERICAL RESULTS When tunnelling was completed, the class A predictions were compared with the vertical and horizontal movements observed in the buildings. The monitoring layout included total station prisms on the façade of the buildings along Balmes Street (at different heights from the ground surface) and precision levelling points at ground level (out­ side and inside the buildings). The prisms allowed the measurement of the vertical and horizontal movement of the buildings’ façades at two eleva­ tions, below the first-floor level and on top of the buildings. On the other hand, the precision levelling points enabled to measure not only the settlement of the buildings, at various distances from the tunnel centre-line, but also their tilt with respect to their ini­ tial position. Figure 13 shows a comparison of the observed settlements at the ground level (marker symbols) and the numerical predictions (curves) relative to both the first and second numerical analyses, associated with the two different excavation methods (Figure 9 and 11). Numerical predictions in Figure 13 refer only to the calculation phases relative to the tunnel excavation. The numbers in the graph legend refer to the street numbers of the buildings in Balmes Street, located above the excavation (Figure 5). In general, the observed vertical movements were well below the pre­ dicted values and never greater than about 5 mm.

541

linear stress-strain relationships in the elastic domaincannot reproduce the ground nonlinear behaviour under the effects of variations in stresses (due in this case to an excavation process). This clearly affects the results in terms of tunnel-induced ground move­ ments’ distribution. In addition, numerical models could not be validated before the analyses and model calibration is undoubtedly essential for successful pre­ dictions, even when using simple constitutive models (Negro & de Queiroz 2000).

Figure 13. Comparison of observed and computed vertical movements relative to the tunnel excavation.

Observed volume losses were also well below pre­ dicted values and always less than 0.3%. Figure 14 illustrates the observed horizontal movement of the buildings’ façades (marker sym­ bols) in comparison with the computed horizontal movements at the ground level (continuous curve). Maximum observed horizontal movements at the ground level are never larger than the numerical pre­ dicted values. Observed horizontal movements on top of the buildings (Figure 14) indicated a small tilt of three of the buildings along Balmes Street (build­ ings #124, 128 and 130). The maximum observed tilt refers to building #128 and its value is about 1/ 16300. Thus, observed movements, both in the ground and in the surrounding structures, were never larger than numerical predictions (Figures 13-14). The class A predictions turned out to be conserva­ tive because of the careful excavation process and close attention to construction details. This was essen­ tial due to the critical location of the construction site as well as to the difficult geotechnical aspects of the project. It is certainly difficult to make accurate esti­ mates of damage in the case of old buildings with uncertain histories. In addition, numerical predictions of ground movements are sensitive to many details of numerical analyses, especially constitutive models. The use of simple constitutive models -that adopt

Figure 14. Comparison of observed and computed horizon­ tal movements. Observed movements refer only to the façade of the buildings.

6 SUMMARY AND CONCLUSIONS The construction process for the enlargement of Pro­ vença station required an excavation below existing structures. Geological conditions involved different layers of Quaternary soil deposits ranging from clayey to gravelly materials and the phreatic level was always below the excavation. The complex cross-section of the excavation, its limited length and its small equivalent diameter led to the use of conventional excavation methods with advance construction stages as low as 0.5 m. The forepoling umbrella system contributed to limiting the decompression in the excavation crown, immedi­ ately ahead of the face. Full-face advancements, with a fibreglass reinforcement adjustable to the geo­ logical conditions of the site, proved successful for tunnel face stability. Closing the support along the excavation perimeter, before each new advance stage, helped to limit ground deformations. Before starting the construction works, numerical analyses resulted in conservative class A predictions of the performance of the buildings affected by the excavation. A comprehensive instrumentation system allowed monitoring of the excavation-induced move­ ments in the ground as well as in the existing struc­ tures along the site. The comparison between observed and computed movements proves the favourable outcome of the construction works. Indeed, the careful and welldesigned excavation procedures gave rise to very small movements both in the ground and in the sur­ rounding structures. These procedures were the main factor underlying the success Sof the excava­ tion works. Observed movements were smaller than numer­ ical predictions and, in general, very little buildings displacements were caused by the enlargement of Provença station. Accordingly, there was no reported damage on any of the structures close to the site. This case study reveals the importance of careful and well-designed excavation procedures in diffi­ cult geological conditions, below existing struc­ tures. It also proves the efficiency of the forepoling umbrella system as an in-tunnel measure to reduce the magnitude of ground deformations attributable to volume losses, even in the case of relatively shal­ low tunnels (in this case C/D ≈ 2). It also demon­ strates the success of a full-face excavation method

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in difficult conditions, provided it is combined with proper reinforcements at the tunnel face.

ACKNOWLEDGEMENTS The authors gratefully acknowledge the support of Infraestructures.cat.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Equivalent frame model for the assessment of tunnel-induced damage to masonry buildings D.B. Gulen, S. Acikgoz & H.J. Burd Department of Engineering Science, University of Oxford, Oxford, UK

ABSTRACT: The risk of tunnel-induced damage to existing masonry buildings is typically assessed in cur­ rent practice by either modelling the building as a simplified elastic beam or by more complex procedures involving three-dimensional numerical analysis. This paper describes a preliminary ‘equivalent frame’ (EQF) model, of intermediate complexity, which is currently under development for the analysis of tunnel-induced damage to masonry buildings. The EQF model comprises an assembly of specially-formulated EQF elements. Soil-foundation interaction behaviour is modelled with consistent one-dimensional finite elements, incorporat­ ing a Winkler soil model. The current EQF model employs linear elastic models for the building and the soil, and is limited to two-dimensional (2D) building geometries. A study is described in which the EQF model is applied to an idealized masonry facade founded on a strip footing. Tunnel-induced ground movements are prescribed via empirical models for greenfield displacements. Separate analyses are described, employing the finite element program DIANA, in which the facade is modelled as a mesh of 2D continuum finite elements. The EQF model provides distributions of tensile strain that indicate a good qualitative match with those obtained from the 2D continuum finite element analysis, although the EQF model demonstrates an overall stiffer response.

1 INTRODUCTION The increase in demand for tunnel construction in urban areas generates a need for a proper assessment of tunnel-induced damage in nearby structures, espe­ cially for vulnerable historical buildings. In current engineering practice, risk assessments for tunnel-induced damage to existing masonry buildings are usually based on a phased assessment procedure in which the complexity of the assessment model is selected according to the level of perceived risk (e.g. Mair et al., 1996; Burland, 2001; Harris & Franzius, 2006; Burland et al., 2012). Initially, the building is modelled as a simplified elastic beam and the maximum tensile strains induced by greenfield ground displacements (without considering soilstructure interaction) for a specified tunnel scenario are computed (Burland & Wroth, 1974; Boscardin & Cording, 1989). If the level of building damage is classified as greater than ‘slight’, then more detailed modelling is employed (Mair et al., 1996). In this detailed modelling phase, analyses involving 2D or 3D finite element models of the tunnel-soil-building interactions may be employed (e.g. Burd et al., 2000; Giardina et al., 2013; Yiu et al., 2017). Motivation for the current research is to develop new modelling strategies to form intermediate approaches between simplified elastic beam models

and detailed 3D finite element analyses. Such intermediate-level models provide a means of obtaining damage predictions that are more reliable than those obtained from elastic beam approaches, while avoiding the cost of conducting detailed 3D finite element ana­ lysis. One form of intermediate-level model, ‘equiva­ lent frame’ (EQF) is considered in the current paper. Equivalent frame models are widely used in struc­ tural earthquake engineering practice to investigate the seismic response of masonry buildings (e.g. Magenes & Fontana, 1998; Roca et al. 2005; Bel­ mouden & Lestuzzi, 2009; Lagomarsino et al., 2013; Bracchi et al., 2015; Quagliarini et al., 2017; Siano et al., 2018). This modelling approach involves the idealization of the masonry building into deformable structural elements (such as spandrels and piers) and rigid connections. The idealization of the effective height of the piers (the height of the deformable sec­ tions) mainly depends on the layout of the wall open­ ings (Lagomarsino et al., 2013; Bracchi et al., 2015). Lagomarsino et al. (2013) suggested an equivalent frame idealization criteria, which defines the effect­ ive height of the piers as starting from the corner of the openings with a steepest inclination angle up to the corner of the external walls, floors, footing or the corners of an adjacent opening. Bracchi et al. (2015) compared three alternative equivalent frame ideal­ ization criteria and noted that a consistent

DOI: 10.1201/9780429321559-71

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idealization is obtained when these criteria are applied to walls with regular openings (except at the ends of the wall, where minor differences may occur). The EQF model developed in the current study is concerned with an example masonry building founded on a strip footing (Figure 1). This example building, based on previous work by Yiu et al. (2017), is inspired by masonry buildings in London. Single and twin-tunnels passing underneath the building are considered. The paper is organized as follows. First, the for­ mulation of the EQF model is described. Then the model is applied to example tunnel-building scen­ arios. The analyses are conducted by: i) computing the short-term tunnel-induced ground movements using empirical equations, ii) representing the masonry building as an equivalent frame and iii) capturing the soil-structure interaction by using a one dimensional (1D) finite element Winkler soilfoundation model. The characteristics of the EQF model are compared with a 2D continuum facade model, employing the DIANA 10.3 finite element analysis program. This provides a means of evaluat­ ing the EQF modelling approach.

2 CASE STUDY In this study, the example problem developed by Yiu et al. (2017), for a tunnel (or tunnels) passing beneath a typical London-type masonry building with a strip footing is considered. Geometrical details of the building, foundation and tunnel arrangements employed in the case study are indicated in Figure 1. Single tunnels and parallel tunnels with a spacing of 40m are considered. In the analyses described by Yiu et al. (2017), greenfield tunnel-induced displacements were computed using 3D finite element analysis. Com­ puted values of the settlement volume loss, VL at the ground surface for single and twin-tunnel scenarios were determined in these analyses as 1.65% and 1.64% respectively. The trough width parameter K needed to fit the finite element settlement data with a Gaussian profile is specified by Yiu et al. (2017) as 0.57. For the current modelling, the greenfield ground movements at the foundation base level (where the footing depth is df = 1.0m) are determined using empirical Gaussian distribution models (Peck, 1969; O’Reilly & New, 1982). Yiu et al. (2017) employed a Young’s modulus of Em = 3 GPa and Poisson’s ratio of νm = 0.2 for the masonry. These stiffness parameters are also adopted in the current model. Case study analyses are initially conducted for single and twin tunnels with zero eccentricity (e = 0). These analyses are used to inspect the detailed response of the building. Subsequent ana­ lyses, for eccentricities e = 5m, 10m, 15m, 20m, 25m and 30m are then conducted to explore the vari­ ation of computed tensile strain (considered to be a proxy for damage) with tunnel eccentricity. 3 FORMULATION OF THE EQF MODEL 3.1

Figure 1. Building, foundation, soil and tunnel configur­ ation adopted for example analyses: (a) layout and dimen­ sions of building; (b) dimensions of facade openings and lintels; (c) details of masonry and foundation; (d) single-tunnel scenario; (e) twin-tunnels scenario (Yiu et al., 2017).

Masonry building EQF model

The ‘isolated facade’ modelling approach, (Yiu et al., 2017) with openings for windows and doors, is adopted in the EQF model. The facade is idealized as an assembly of compliant structural elements and rigid connections. Self-weight of the structure is not included in the model. The EQF idealization is illustrated in Figure 2, where the bold black sections indicate rigid connec­ tions and the thinner black lines indicate compliant structural elements. Determining the appropriate arrangement of the rigid and compliant components is an important part of the EQF modelling process; the current arrangement is based on geometric modelling criteria proposed by Lagomarsino et al. (2013). The footing-soil interaction is represented by a linear elastic Winkler model, connected to the EQF elements at the base of the facade. Tunnel-induced ground movement are applied to the model via equiva­ lent forces applied to the base of the EQF model.

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by

The internal and external displacements are related where the transformation matrix T is,

Figure 2. Equivalent frame model (EQF).

3.2 Finite element formulation of the EQF elements The EQF model comprises an assembly of 1D EQF elements as shown in Figure 3. The elements are for­ mulated in terms of a local coordinate, x, aligned along the element. The EQF model incorporates both spandrel (horizontal) and pier (vertical) elem­ ents. For the piers, an appropriate coordinate trans­ formation needs to be applied to the stiffness equations so that they are consistent with the global axis directions. The EQF elements consist of a central flexible section (Section F, length LF ) which deform in bend­ ing according to 2D Euler-Bernoulli beam theory. Transverse and axial displacements, v and u respectively, are considered. Section F is elastic, with Young’s modulus, E, second moment of area, I, and cross-section area A. Rigid sections (indicated as Sections 1 and 2 in Figure 3) are connected to each end of the flexible section. These rigid sections pro­ vide a means of modelling the connections between individual piers and spandrels. The EQF element is formulated (Eqs 1-11) in terms of the displacements and rotations at the exter­ nal nodes (indicated as filled circles in Figure 3). These displacements/rotations are contained within the element displacement vector Ue where,

The displacements and rotations at the two internal nodes (indicated with open circles in Figure 3) are contained in the internal displacement vector Ui where,

The element stiffness matrix is determined by the considering the deformations in the flexible section. In this section the axial and lateral displacements are,

where u is the local displacement vector and NF is the shape function matrix,

where NiL and NiH are conventional linear Lagran­ gian and Hermitian shape functions (see Appendix, Eqs 31-36) expressed in terms of the mapped coordinate,

where The generalized strain vector is εa is the axial strain and κ is the curvature (Burd, 2019). The generalized strain vector is related to the nodal displacements/rotations by,

The first row of the matrix BF is,

where the prime indicates a derivative with respect to αF . The second row of BF is,

Figure 3. Diagrammatic representation of the EQF elem­ ent, showing the flexible section (F) and the two rigid sec­ tions (1 and 2) (modified from Sekulovic & Salatic, 2001).

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where the double prime indicates a second deriva­ tive with respect to αF. The stiffness matrix for the element is determined from, Gapping and/or sliding at the soil-structure inter­ face is not allowed in the current form of the model. The local displacements are determined from the nodal displacements by u ¼ NUe where N is a shape function matrix that is appropriate for each section in the element. For Section 1 the shape function matrix is,

where,

where,

Equation 10 is evaluated by Gauss integration. 3.3

EQF Winkler soil-foundation model

The soil is represented by 1D finite elements, as shown in Figure 4, that are formulated (Eqs 12-29) to have displacements u that are compatible with the EQF elements employed in the equivalent frame. Shear and normal tractions, t and p, act on the element such that,

where p is the local traction vector the local displacement vector the soil stiffness matrix,

For Section 2 the shape function matrix is,

where,

, u is and Ds is

The parameters kh, and kv defining the soil stiffness in the horizontal and vertical directions respectively. The vector u represents the displacements at the base of the equivalent frame. When tunnel-induced displace­ ments are applied to the model, the reference configur­ ation (at which p ¼ 0) is u ¼ uG where uG are the specified greenfield displacements at the base of the equivalent frame. In this case, Equation 12 becomes,

Figure 4. The EQF Winkler soil-foundation model.

For section Section F, the shape function matrix is where NF is given in Equation 5. The indi­ vidual stiffness matrices arising from Section 1, K1, Section 2, K2, and Section F, KF, are,

The stiffness matrix for the soil element is deter­ mined by summing the three stiffness contributions to give,

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The vector of equivalent nodal forces, F, for the element is determined by summing the individual contributions of each section, as follows:

Figure 5. Isolated facade 2D continuum model in DIANA.

4 2D CONTINUUM MODEL A separate model, in which the facade is represented as a mesh of 2D continuum linear elastic plane stress finite elements, has been developed in the finite element program DIANA 10.3 (Figure 5). This 2D continuum model provides a means of assessing the veracity of the equivalent frame model. The facade, lintels and the footing are modelled as 6-node tri­ angular elements and the openings are generated as voids (Burd et al., 2000; Yiu et al., 2017). The mesh size is selected as 0.5 m. Thus, 2568 meshed elem­ ents are generated for the whole facade geometry (excluding the lintel elements). Each triangular plane stress element has 3 integration points. The soil-structure interaction is represented in the model with elastic structural-line zero-thickness inter­ face elements. These 6-noded elements (3 nodes on the upper faces and 3 nodes on the lower face) are avail­ able as built-in models in DIANA. Stiffness parameters for the interface were selected to be the same as those employed in the EQF foundation model. Greenfield horizontal and vertical displacements are imposed on the lower face of the interface elements as prescribed displacements.

These equivalent nodal forces are the form of,

where FU represents the equivalent forces due to the displacement at the base of the equivalent frame and FG is an additional term associated with the pre­ scribed greenfield displacements. The vector FG is determined from the sum of the contributions from the three sections as,

5 RESULTS 5.1

Greenfield displacements

Greenfield settlements, Sv are specified in the ana­ lyses at the level of the base of the foundation on the basis of the conventional empirical equation pre­ sented in Equation 30 as (Peck, 1969),

In the current implementation of the model, the greenfield displacements are imposed via the appli­ cation of nodal forces to the base of the equivalent G G frame corresponding to -FG (i.e. –(FG 1 þ F2 þ FF )). For the current calculations, the soil stiffness coefficients are chosen as kh ¼ 1:69 x 107 N/m2 and kv ¼ 2:7 x 107 N/m2. These data were selected to provide (approximate) comparability with the data reported by Yiu (2018) for strip footings embedded in Terrace gravel.

where: D is the tunnel diameter, x is the transverse distance from the tunnel centre line and i defines the location of the inflection point (i = Kz0, where z0 is the depth of the tunnel centre). Horizontal displace­ ments are determined by assuming that ground movements are directed towards the tunnel centre line (O’Reilly & New, 1982). Ground movements for the twin tunnel case are determined by superposition. Greenfield vertical displacements determined with this empirical approach are plotted in Figure 6.

549

Figure 6. Greenfield settlements at foundation base level induced by tunnelling: (a) single tunnel; (b) twin tunnels.

5.2

Figure 7. Vertical displacements at foundation base level induced by tunnelling with e = 0: (a) single tunnel; (b) twin tunnels.

Building response

The vertical and horizontal displacement responses computed at the foundation base level, for the zero eccentricity case, for both the EQF and 2D con­ tinuum models are presented in Figures 7, 8 respectively. Both models demonstrate a flattening of the greenfield vertical displacement profile, as a result of soil-structure interaction (Figure 7a, b). In addition, both models indicate a relatively large displacement gradient at the location (x = +-10m) corresponding to the door openings. The computed horizontal displacements at the base of the foundation, shown in Figures 8a, b indicate that the induced displacements in the building foundation are significantly lower in magni­ tude than the corresponding greenfield displacements. Overall, the EQF results demonstrate a response that is stiffer than is obtained from the 2D continuum models. This observation is consistent with the recent work of Malomo et al. (2019), which demonstrated a similar tendency of equivalent frame models to be over-stiff in the context of seismic analysis. According to the building damage classifications (Burland & Wroth, 1974; Boscardin & Cording, 1989), the tunnel-induced damage in masonry build­ ings is associated with the magnitude of induced ten­ sile strains. Thus, the tensile strains are determined

in the EQF and 2D continuum models to characterize the level of expected damage. The tensile strains are computed in the EQF ana­ lysis by applying beam theory to the flexible sections in the model. The maximum tensile strain occurring in the model is then determined. For the 2D continuum models, following Yiu et al. (2017), a ‘characteristic strain’ approach is adopted to characterize the magni­ tude of the induced tensile strains. In this approach, the strains computed at the Gauss points are used to determine the characteristic strain, εt99, which is not exceeded in 99% of the facade area. This approach, in which 1% of the facade area with the highest tensile strains is discounted, avoids spurious results associated with strain concentrations around the internal corners. In Figure 9, the maximum principal strain distribu­ tions are compared for the EQF and 2D continuum models. Due to symmetry, only the left half of the models are shown. For the single tunnel case, the EQF model (Figure 9a) is able to capture the shear dominant response mechanisms of the first storey pier elements, as determined by the 2D continuum model (Figure 9b). For the twin tunnel case, the EQF model (Figure 9c) also captures the concentration of strains at span­ drel elements near the top edge of the facade as com­ puted by the 2D model (Figure 9d).

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Figure 8. Horizontal displacements at foundation base level induced by tunnelling with e = 0: (a) single tunnel; (b) twin tunnels.

The maximum tensile strain (computed from the EQF) model and characteristic strain, εt99 (from the 2D continuum model) are presented in Figure 10 for a range of tunnel eccentricities, e. Single and twin tunnel scenarios are shown in Fig­ ures 10a, b respectively. Figure 10 demonstrates that the characteristic strains determined with the 2D con­ tinuum model are the order of 600 με for the two zero-eccentricity cases, with a tendency for the char­ acteristic strain to reduce with increasing tunnel eccentricity. These data may be compared with equivalent results in Yiu et al. (2017) for a nonlinear facade model in which the characteristic strain for the zero eccentricity cases is the order of 1000 με. The larger characteristic strains reported by Yiu et al. (2017) are consistent with the use of a nonlinear masonry model in the Yiu et al. (2017) analyses. The EQF models indicate maximum tensile strains that are significantly greater in magnitude than the characteristic strains obtained from the 2D continuum model. This is in spite of the fact that the induced deformations in the EQF models are lower in magni­ tude than in the 2D continuum models (Figure 7, 8). It appears that the system adopted in the current work to characterize the maximum induced tensile strain in the EQF models is inconsistent with the characteristic strain procedure adopted in the 2D analysis.

Figure 9. Major principal strain distributions induced by single and twin tunnels with e = 0: (a) EQF model, single tunnel; (b) DIANA-2D continuum model, single tunnel; (c) EQF model, twin tunnels; (d) DIANA-2D continuum model, twin tunnels.

6 CONCLUSION The EQF model is applied to an example problem consisting of the response of a two-storey masonry building to nearby tunnelling works. The accuracy of the EQF model is evaluated by comparing the computed displacements and strains with those determined from a 2D continuum model. The EQF results are seen to be consistently stiffer than those obtained from the continuum analysis. This apparently over-stiff behaviour of the EQF model will be addressed in future work. The incorp­ oration of shear deformations in the EQF elements (e.g. via Timoshenko beam theory) provides a possible route for future developments. Additionally, the EQF models are seen to predict values of tensile strain that significantly exceed those that are determined using the 2D model employing a characteristic strain approach. This lack of

551

Hermitian shape functions

REFERENCES

Figure 10. Variation of maximum tensile strain (for the EQF model) and characteristic strain, εt99, (for the 2D con­ tinuum model) with tunnel eccentricity, e: (a) single tunnel; (b) twin tunnels.

comparability between the two modelling procedures signifies a limitation of the current EQF modelling approach. In future work, procedures will be devel­ oped to ensure that strain predictions employing the EQF model are consistent with those that are obtained when more sophisticated models are employed. Appendix Shape functions along a line of length, Le , in terms of the variable α (0 � α � 1) are specified below. Lagrangian shape functions

Belmouden, Y., & Lestuzzi, P. 2009. An equivalent frame model for seismic analysis of masonry and reinforced concrete buildings. Construction and Building Materials 23(1): 40–53. Boscardin, M. D. & Cording, E. J. 1989. Building response to excavation-induced settlement. Journal of Geotech­ nical Engineering 115(1): 1–21. Bracchi, S., Rota, M., Penna, A., & Magenes, G. 2015. Consideration of modelling uncertainties in the seismic assessment of masonry buildings by equivalent-frame approach. Bulletin of Earthquake Engineering 13: 3423–3448. Burd, H. J. 2019. Formulation of 1D soil structure inter­ action models. Private communication. Burd, H. J., Houlsby, G. T., Augarde, C. E. & Liu, G. 2000. Modelling tunnelling-induced settlement of masonry buildings. Proceedings of the Institution of Civil Engin­ eers-Geotechnical Engineering 143(1): 17–29. Burland, J. B. 2001. Assessment methods used in design. In J. B. Burland, J.R. Standing & F.M. Jardine (eds), Building Response to Tunnelling: Case Studies from Construction of the Jubilee Line Extension, London, volume 1: the pro­ ject: 23–43. London, UK: Thomas Telford. Burland, J. B., & Wroth, C. P. 1974. Settlement of build­ ings and associated damage. In A.C. Meigh (ed.), Pro­ ceedings of the conference on settlement of structures: 611–654. London, UK: Pentech Press. Burland, J. B., Chapman, T., Skinner, H. D. & Brown, M. (eds) 2012. ICE manual of geotechnical engineering London, UK: ICE Publishing. Giardina, G., van de Graaf, A. V., Hendriks, M. A. N., Rots, J. G., & Marini, A. 2013. Numerical analysis of a masonry façade subject to tunnelling-induced settle­ ments. Engineering Structures 54: 234–247. Harris, D. I., & Franzius, J. N. 2006. Settlement assessment of running tunnels–a generic approach. In K. J. Bakker, A. Bezuijen, W. Broere & E. A. Kwast (eds), Geotech­ nical aspects of underground construction in soft ground: 225–230. London, UK: Taylor & Francis. Lagomarsino, S., Penna, A., Galasco, A., & Cattari, S. 2013. TREMURI program: An equivalent frame model for the nonlinear seismic analysis of masonry buildings. Engineering Structures 56: 1787–1799. Magenes, G., & Fontana, A. D. 1998. Simplified non-linear seismic analysis of masonry buildings. Proceedings of the British Masonry Society 8: 190–195.

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Mair, R. J., Taylor, R. N. & Burland, J. B. 1996. Prediction of ground movements and assessment of risk of building damage due to bored tunnelling. In R. J. Mair & R. N. Taylor (eds), Geotechnical aspects of underground construction in soft ground: 713–718. Rotterdam, the Netherlands: Balkema. Malomo, D., Morandini, C., Penna, A., & DeJong, M. J. 2019. Assessing the reliability of the equivalent-frame idealisation of URM façades with irregular opening layouts by comparison with discrete micro-models, SEDEC 2019 Conference, Greenwich, London, 1–9 September. O’Reilly, M. P. & New, B. M. 1982. Settlements above tun­ nels in the United Kingdom- their magnitude and predic­ tion. In M. J. Jones (ed.), Proceedings of Tunnelling’ 82: 173–181. London, UK: Institute of Mining and Metallurgy. Peck, R. B. 1969. Deep excavations and tunnelling in soft ground. Proc. 7th Int. Conf. Soil Mech. Found. Engng: 225–290. Quagliarini, E., Maracchini, G., & Clementi, F. 2017. Uses and limits of the equivalent frame model on existing

unreinforced masonry buildings for assessing their seis­ mic risk: A review. Journal of Building Engineering 10: 166–182. Roca P., Molins C. & Mari A. R. 2005. Strength capacity of masonry wall structures by the equivalent frame method. Journal of Structural Engineering ASCE 131 (10):1601–1610. Sekulovic, M., & Salatic, R. 2001. Nonlinear analysis of frames with flexible connections. Computers and Struc­ tures 79: 1097–1107. Siano, R., Roca, P., Camata, G., Pelà, L., Sepe, V., Spacone, E., & Petracca, M. 2018. Numerical investi­ gation of non-linear equivalent-frame models for regu­ lar masonry walls. Engineering Structures 173: 512–529. Yiu, W. N. 2018. Finite element analysis of short-term and long-term building response to tunnelling. PhD thesis. University of Oxford. Yiu, W. N., Burd, H. J., & Martin, C. M. 2017. Finite-element modelling for the assessment of tunnel-induced damage to a masonry building. Gѐotechnique 67(9): 780–794.

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Use of underpinning, horizontal jet grouting and ground freezing for ground stabilization to control settlement of existing MRT tunnels during construction of a link-way and railway tunnels for a new MRT line in Singapore Kaoru Hashida Taisei Corporation, Japan

Tadashi Hashimoto Geo-Research Institute, Japan

Yong Kwet Yew National University of Singapore, Singapore

Ramesh Nair Land Transportation Authority, Singapore

John Busbridge Golder, Canada

ABSTRACT: For construction of the new Thomson-East Coast Line (TEL) project in Singapore, twin bored tunnels in stacked configuration and a pedestrian link-way are being constructed directly under two operating MRT lines. Ground conditions consist of thick deposits of soft (under­ consolidated) marine clay overlying Old Alluvium (OA), which at this location is relatively cohe­ sionless. The construction scheme devised to enable construction of the TEL tunnels without inter­ ruption to the existing MRT lines, consisted of ground stabilization to allow underpinning of the existing lines and construction of the new tunnels by the Sprayed Concrete Lining (SCL) method. Both vertical and horizontal Jet Grouting Pile (JGP) as well as ground freezing were employed in the ground stabilization. The presence of existing MRT lines limited the coverage that could be achieved by conventional vertical JGP from the surface alone; as a result, Horizontal JGP was adopted to improve marine clay under the existing tunnels and allow driving of Adits by means of an Open Faced Rectangular Shield Machine. vertical JGP from the Adits was used to stabilize the lower zone of marine clay. Because of the density and potentially flowing characteristic of the under­ lying OA, the JGP could not be relied upon to stabilize this deposit and ground freezing was adopted to stabilize the zone at the interface of the marine clay and the OA. This was the first time that ground freezing had been applied for an underground railway project in Singapore. 2D and 3D FEM Analysis were carried out to predict the movement of the existing tunnels caused by each con­ struction activity - Shield Excavation, ground freezing and SCL Tunnels excavation. Construction control was achieved based on actual monitoring data compared to predicted values, and the work was completed successfully without breaching the Work Suspension Level. The paper describes the impact assessment of the existing live MRT tunnels including overall integrity and stability due to the TEL construction.

1 INTRODUCTION 1.1 Project site The location of the Tunnels within this Project is shown in Figure 1.

Tunnelling work consists of the construction of an underground pedestrian link-way which is a connecting passage from the new Marina Bay Sta­ tion extension to the existing Marina Bay Station as well as mining of the two TEL railway tunnels. All of these tunnels are constructed directly underneath

DOI: 10.1201/9780429321559-72

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longitudinal soil profile along the construction of sta­ tion extension and tunnels is shown in Figure 3. OA (C, D, E) which is located at the interface with Kal­ lang Formation has comparatively higher permeability. Two railway tunnels align vertically as shown in Figure 4. marine clay layer was improved before­ hand by JGP (Jet Grouting Pile) and the upper tunnel was excavated all the way in the JGP treated marine clay. The lower tunnel was excavated in treated marine clay at the tunnel crown and in the original OA layer with water cut off wall using artificial ground freezing. Figure 1. Location of tunnel in T226.

2 CONSTRUCTION & DESIGN CONDITIONS 2.1

the two existing ‘live’ MRT Lines (NSL and CCL) and the construction of two Railway Tunnels which are located below the link-way. Figure 2 shows the 3D image of structures after the tunnelling works are completed. 1.2

Soil condition

The subsurface conditions at site consist of hydraul­ ically reclaimed sand fill followed by Kallang For­ mation. This formation includes upper marine clay (M), estuarine clay (E), fluvial clay (F2) and lower marine clay (M). This is underlain by Old Alluvium (OA) with various weathering grades such as OA(E) – SPT ‘N’ Value 1) this is the area where plastic failure might be expected. There is an obvious correlation between the plastic zone of the undrained model and the area where suctions occur shown in Figure 7. Similar remarks can be made of the consolidation model with the lowest soil perme­ ability (keq=3.16x10-10 m/s) which exhibits a slightly smaller plastic zone compared with the undrained case but a similar magnitude of the maximum plastic strains, hence, explaining the suctions observed in those areas (see Figure 7). The consolidation case with keq=15.8x10-10 m/s shows a plastic zone almost confined to the crown and invert where the plastic strains do not reach the magnitudes of the two cases previously mentioned. Finally, the consolidation case with keq=31.6x10-10 m/s does not indicate plastic fail­ ure. These results reveal a relationship between the soil permeability and the size of the plastic region, the latter becoming smaller as the soil permeability increases. This can be explained by the different level of unloading required in the four cases to achieve the target volume loss, as indicated by the different mag­ nitudes of radial pressure applied on the tunnel boundary, as discussed in Section 2.4. In particular, larger stress magnitudes were required for larger model permeability and this effectively translates in smaller unloading of the soil prior to the installation of the lining and hence, less perturbation of the initial stress field. 3.3

Permeability anisotropy ratio

The results of the parametric study presented in the section above correspond to models where the

Figure 7. Contour plot distribution of the pore water pressure (kPa) at the end of the tunnel construction for undrained and coupled consolidation analyses (compressive pore water pressures are shown as positive).

602

Figure 8. Contour plot distribution of the volumetric plastic strain (%) at the end of the tunnel construction for undrained and coupled consolidation analyses (dilative strains are shown as negative).

equivalent soil permeability keq was varied keeping the permeability anisotropy ratio kh/kv equal to 10. It has been shown above (see Figure 7) that the permeability anisotropy has a marked effect on the pore water pres­ sure distribution around the tunnel, which is more clearly observed in the cases with larger soil permeabil­ ity. The analyses presented in this section aim to illus­ trate the influence of the permeability anisotropy ratio kh/kv on the ground response by considering ratios of 1/10, 1 and 10. In order to isolate the effect of this par­ ameter, the equivalent soil permeability keq is kept con­ stant at 31.6x10-10 m/s adopting the appropriate combinations of kh and kv. Figure 9 depicts the surface settlement trough of the three additional analyses con­ sidered in this section, along with the undrained result, the empirical curve is not included for the sake of clar­ ity. There are no significant differences among the three analyses, the surface settlement with kh/kv equal to 1/10 and 1 are very similar, and slightly shallower than the case with kh/kv equal to 10. Regarding the development of plastic strain, Figure 10 shows contour plot distributions of the volumetric plastic strain around the tunnel cavity at the end of the excavation

Figure 10. Contour plot distribution of volumetric plastic strain at the end of the excavation – varying kh/kv (dilative strains are shown as negative).

for the three analyses under evaluation. The contours show a small plastic region at the crown and invert locations for the anisotropy ratio kh/kv=1/10 and 1 whereas the case with kh/kv =10 does not indicate any plastic failure, note that the latter corresponds to the analysis labeled as keq =31.6x10-10 m/s in the previous section. Following the argument presented above, the analyses under evaluation develop very little or no plasticity owing to the significant portion of the volume loss being caused by the consolidation itself. The negligible differences, in ground movements, between these analyses highlight that the compression pattern around the tunnel caused by differences in the pore water pressure distribution and therefore volume change around the tunnel (not shown for reasons of space) does not significantly influence the shape of the trough. 4 DISCUSSION AND CONCLUSIONS

Figure 9. Surface settlement trough – influence of the per­ meability anisotropy ratio kh/kv.

The effect of modelling consolidation during tunnel construction in London Clay was investigated through a number of plane-strain analyses. A reasonable

603

construction time determined from field monitoring data and permeability values within those measured in London Clay were employed. It was shown that mod­ elling consolidation along with tunnel construction does influence the short-term ground movements. The results including consolidation present a surface settle­ ment trough that is deeper and narrower than that obtained under undrained conditions which results in an improvement with respect to the empirical Gaussian settlement, although there are still significant discrep­ ancies with the latter. As expected, differences with the undrained case become more pronounced as the adopted soil permeability is larger. It was noted that as the permeability is increased, the ground losses coming from soil compression are more significant, which in turn means that less unloading of the tunnel boundary is required to achieve a given volume loss. Conse­ quently, the plastic region around the cavity is reduced with larger permeability. Additionally, the impact of varying the permeabil­ ity anisotropy ratio was investigated. The results indicate that the ground response is less influenced by the anisotropy ratio than it is by the permeability magnitude. It should be highlighted that the modelling presented in this paper assumes that the tunnel section under consideration behaves as a drain, even after the lining is installed. Although this condition is not unreasonable for certain tunnelling methods, the results presented might not be applicable if during construction the excavation face is impermeable and/or waterproofing meas­ ures are considered around the tunnel lining. To conclude, the approach adopted in this paper assumes that the consolidation that would pro­ gressively develop as the tunnel face advances can be modelled in a plane-strain analysis. In order to apply this methodology with confidence, the results should be validated with a 3D analysis.

ACKNOWLEDGEMENTS This research is funded by the Engineering and Physical Sciences Research Council (EPSRC) through a Doctoral Training Grant (Ref: EP/ R512540/1) awarded to the first author.

REFERENCES Addenbrooke, T. I., Potts, D. M. & Puzrin, A. M. 1997. The influence of pre-failure soil stiffness on the numer­ ical analysis of tunnel construction. Géotechnique, 47 (3): 693–712.

Avgerinos, V. 2014. Numerical investigation of tunnelling beneath existing tunnels. PhD thesis, Imperial College London. Avgerinos, V., Potts, D. M. & Standing, J. R. 2016. The use of kinematic hardening models for predicting tunnelling-induced ground movements in London clay. Géotechnique, 66(2): 106–120. Franzius, J., Potts, D. & Burland, J. 2005. The influence of soil anisotropy and K 0 on ground surface movements resulting from tunnel excavation. Géotechnique, 55(3): 189–199. Gasparre, A., Nishimura, S., Minh, N. A., Coop, M. R. & Jardine, R. J. 2007. The stiffness of natural London Clay. Géotechnique, 57(1): 33–47. Gonzalez, N. A., Rouainia, M., Arroyo, M. & Gens, A. 2012. Analysis of tunnel excavation in London Clay incorporat­ ing soil structure. Géotechnique, 62(12): 1095. Hight, D. W., Gasparre, A., Nishimura, S., Minh, N. A., Jardine, R. J. & Coop, M. R. 2007. Characteristics of the London Clay from the Terminal 5 site at Heathrow Airport. Géotechnique, 57(1): 3–18. Jurecic, N., Zdravkovic, L. & Jovicic, V. 2013. Predicting ground movements in London Clay. Proceedings of the Institution of Civil Engineers-Geotechnical Engineering, 166(5): 466–482. Mair, R. J. 2008. Tunnelling and geotechnics: new horizons. Géotechnique, 58(9): 695–736. Mašín, D. 2009. 3D modeling of an NATM tunnel in high K0 clay using two different constitutive models. Journal of geotechnical and geoenvironmental engineering, 135 (9): 1326–1335. Measham, P. G., Taborda, D. M. G., Zdravkovic, L. & Potts, D. M., 2014. Numerical simulation of a deep excavation in London Clay. Proceedings of the 8th European Conference on Numerical Methods in Geotechnical Engineering, Delft, The Netherlands, 2014: Taylor & Francis Nyren, R. 1998. Field measurements above twin tunnels in London Clay. PhD thesis, Imperial College London. Peck, R. B., 1969. Deep excavations and tunnelling in soft ground. Proc. 7th ICSMFE, Mexico, 1969: Sociedad Mexicana de Mecanica de Suelos Potts, D. M. & Zdravkovic, L. 1999. Finite element ana­ lysis in geotechnical engineering: theory, London, Thomas Telford. Potts, D. M. & Zdravkovic, L. 2001. Finite element ana­ lysis in geotechnical engineering: application, London, Thomas Telford. Shin, J. H. & Potts, D. M. 2002. Time-based two dimen­ sional modelling of NATM tunnelling. Canadian geo­ technical journal, 39(3): 710–724. Standing, J. R. & Burland, J. B. 2006. Unexpected tunnelling volume losses in the Westminster area, London. Géotechnique, 56(1): 11–26. Taborda, D. M. G., Potts, D. M. & Zdravković, L. 2016. On the assessment of energy dissipated through hyster­ esis in finite element analysis. Computers and Geotech­ nics, 71: 180–194. Wongsaroj, J., Soga, K. & Mair, R. J. 2013. Tunnelling­ induced consolidation settlements in London Clay. Géo­ technique, 63(13): 1103.

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Semi-coupled modelling of soil-structure interaction during tunnel construction: Two case studies from Bank Station Capacity Upgrade A. Luciano, M.N. Pascariello & E. Bilotta Department of Civil, Architectural and Environmental Engineering, University of Napoli FedericoII, Naples, Italy

S. Acikgoz Department of Engineering Science, University of Oxford, Oxford, UK

R. Mair Department of Engineering, University of Cambridge, Cambridge, UK

ABSTRACT: Detailed site investigation and monitoring data gathered during underground construction projects creates an opportunity to evaluate and improve the techniques used to model the influence of tun­ nelling on existing buildings. This paper discusses the application of the semi-coupled modelling tech­ nique to two historic buildings affected by the construction of new tunnels for the Bank Station Capacity Upgrade in London, UK. Using the semi-coupled modelling approach, the buildings were idealised as equivalent elastic anisotropic solids, while high fidelity models were used to model nonlinear behaviour of the soil and tunnel excavation process. These analyses highlight the flattening of displacement profiles, and the important influence of tunnel advance on skew facades. However, comparison of modelling results to levelling data, highlight some limitations of using equivalent solid models to represent masonry buildings.

1 INTRODUCTION The effects of underground construction on nearby buildings must be taken into account during design. In particular, the presence of architectural heritage may present a severe constraint for urban tunnelling design and construction projects (Viggiani et al. 2006; Ram­ pello et al. 2012; Amorosi et al. 2014; Bilotta et al. 2017). This is particularly the case for heritage masonry structures which may experience cracking due to ground movements. To evaluate tunnelling-induced damage in heritage masonry structures, it is often not sufficient to use conventional empirical assessment techniques. These techniques simplify the problem and neglect soilstructure interaction. Finite-element based numerical methods are increasingly used to analyse the tunnel­ soil-structure interaction problem. These tools offer the ability to model both the excavation and the build­ ing while accounting for three-dimensional effects. Past experience has shown that accurate ground characterization and suitable constitutive models for soil are necessary for capturing tunnelling-induced ground movements (Addenbrooke et al. 1997). On the other hand, the geometric layout and the mechan­ ical behaviour of the building need to be modelled in detail to capture soil-structure interaction (Giardina

DOI: 10.1201/9780429321559-79

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et al. 2013). Studies which feature detailed models of both the soil and structure exist (Yiu et al. 2017) but are difficult to use in a consultancy setting due to the associated computational expense. Therefore, some simplifications are typically adopted in the analyses. The monitoring data collected during construction projects can be used to critically evaluate numerical models and assess the suitability of the simplifying assumptions used during the design stage. In this paper, numerical analyses are carried out to model a tunnel-soil-structure interaction problem from an ongoing project. Two Grade-1 listed build­ ings (St. Mary’s Abchurch and the Mansion House) were affected recently by the construction of new tun­ nels for the Bank Station Capacity Upgrade (BCSU) project in London (UK). Prior to construction, detailed soil investigations were carried out, and during construction, the buildings were monitored. These aspects, discussed in Section 2, presented an opportunity to perform state of the art numerical ana­ lyses and assess their accuracy. Two types of analyses were conducted (see Section 3): first, greenfield then, soil-structure interaction ana­ lyses were performed. For the former analyses, detailed model of the shotcrete lining process was adopted with detailed soil models. The comparison of settlement pre­ dictions from the analyses and the levelling data

helped determine the key parameters required to cap­ ture greenfield profiles. For the soil-structure inter­ action analyses, simplifications were required due to the complexity of the models. The semi-coupled approach proposed by Pickhaver et al. (2010) and Losacco et al. (2014) was used to idealise the buildings as embedded equivalent elastic anisotropic solids. Comparing the modelling results to levelling data enabled a critical evaluation of the impact of the sim­ plifications. These results are discussed in Sections 4 and 5, respectively for St Mary’s Abchurch and Man­ sion House; Section 6 summarises the findings. 2 BANK STATION CAPACITY UPGRADE 2.1

Urban environment

Bank Station is an underground station in London, which connects the Central, Northern and Waterloo & City Lines and provides a terminus for the Dockland Light Railways. Bank Station is connected via under­ ground tunnels to the nearby Monument station, which serves the District and Circle Lines. The Bank Monu­ ment Station Complex is the fourth busiest interchange on the London Underground Network and one of the world’s most complicated subterranean railway sta­ tions. Due to the overcrowding, congestion and passen­ ger delay in the Northern Line, the station is being upgraded to handle the increasing passenger demand and support the expected economic growth of the area. The existing southbound platform for the Northern Line will be converted into a passenger concourse and a new southbound running and platform tunnel will be located to the west of the existing platform. In addition, new passenger exchange tunnels, such as the Moving Walkway tunnel, will be constructed. The urban environment affected by the BSCU pro­ ject is part of the Bank Conservation Area. The under­ ground construction works affect two Grade I listed buildings: St. Mary’s Abchurch and the Mansion House. St Mary’s Abchurch (Figure 1a) is a 17th cen­ tury church, designed by Sir Christopher Wren. It is a single storey rectangular brick building with stone dressings, and a hipped roof. Its four facades are con­ structed of 0.75 m thick double leaf brick masonry with rubble infill. At the north-western corner of the church is a bell tower. This paper will focus on the south and east façade walls of the church which are in close proximity to tunnelling works (see Figure 1b). Trial pit investigations have shown that the south and west walls are founded on cemented rubble and chalk blocks at depths of between 2.4 m and 3.5 m below churchyard level. During the 20th century, the church appears to have been affected by nearby construction activity and a World War II bomb drop. Existing vis­ ible cracks are likely due to these events. The Mansion House (Figure 2a) was built between 1739 and 1752, as the official residence for the Lord Mayor of London. This is a five-storey load bearing

Figure 1. A) A photograph of St. Mary’s Abchurch; b) plan view of the site and the location of buildings and new tunnels. The circle markers on the ground refer to the loca­ tions of levelling pins.

brick building. The stiff facades are about 1.7 m thick (on average) and clad with stones. The internal struc­ tural layout is complex and irregular and features a vaulted basement. Roofs are a combination of timber trusses and timber joists. A condition survey carried out in 1985 indicates that the Mansion House sits on a shallow foundation. The east façade of this building is founded 7m below ground level. The ana­ lyses in this paper will focus on this specific façade, which was affected by the construction of Northern Line Running Tunnel and Moving Walkway Tunnel. 2.2

Ground conditions and tunnelling works

The strata levels encountered in the BSCU ground investigation are, from the ground surface: Made Ground (about 6 to 10 m thick); River Terrace

606

Figure 2. A) Point cloud model of the east façade of Man­ sion House; b) plan view of the site and the location of buildings (highlight the building in the figure), new and old tunnels. The circle markers on the ground refer to the loca­ tions of levelling pins.

Deposits (up to 5 m thick); London Clay (unit B about 17 m thick; unit A3 about 14 m thick; unit A2 about 12 m thick); Lambeth Group (about 18 m thick). Below these layers, Upnor Formation, Thanet Sand and Chalk are encountered, generally well below the tunnel depth. The ground conditions beneath the two buildings and the tunnels’ position are shown in Figure 3. The groundwater table is located about 7 m below the ground level, at the top of the River Terrace Deposit layer. Due to a downward seepage to the Chalk deep aquifer, a sub-hydrostatic piezometric profile is pre­ sent in the London Clay. In the case of St. Mary’s Abchurch the pore pressure reaches a value of 330kPa at the bottom of the soil layer, whereas in the Lambeth group the water pressure decreases until a value of 200kPa. Similar conditions occur beneath the Mansion House. The tunnels were constructed following conven­ tional heading with shotcrete lining in London Clay. Relevant tunnel dimensions for the two buildings are shown in Figure 3a and b. The Northern Line Tunnel is constructed first as a pilot tunnel and later enlarged to 9.8m under St Mary’s Abchurch. Under Mansion House both tunnels (Northern Line and Moving Walkway Tunnels with diameters 5.3m and 8.2m) are constructed following an excavation

Figure 3. Ground conditions and tunnel position: (a) beneath St. Mary’s Abchurch; (b) beneath the Mansion House.

sequence proceeding from top heading to invert. A single shotcrete lining 0.15 m thick has been used in the pilot tunnel, while double lining, 0.3 m thick each, has been used in the enlargement tunnel. The geometry of the tunnels is uneven, but it is modelled as circular shape adopting an equivalent diameter.

607

2.3

Monitoring

Table 1.

While only negligible impact is expected in each building, there are significant uncertainties regard­ ing the behaviour of the ground and the building during the works, thus requiring accurate monitoring. A monitoring system was installed and main­ tained by Geocisa UK. The position of the manual levelling points (LP) installed at the ground level along Abchurch lane and around the church is shown in Figure 1b, with circle mark­ ers. Those around the Mansion House are shown in Figure 2b. In general, levelling points were placed in close vicinity of the buildings, and recorded the displacements experienced by the buildings. This was evident when data from nearby building sockets and levelling points were compared; they revealed near-identical response. In places where no nearby buildings exist (e.g. see the church yard in Figure 1b), levelling points measured greenfield settlements. 3 NUMERICAL ANALYSES Finite element analyses were carried out using Plaxis 3D (Brinkgreve et al. 2013). The software allows the tunnel excavation process to be simulated. Analyses conducted in this paper assume that the ground level and soil layers are horizontal. A simplified bi-linear profile of pore pressure has been assumed in the ground according to Section 2.2. This bilinear line is shown on the left-hand side of Figure 3a. Undrained conditions were assumed for the fine-grained layers. Made ground, River Terrace deposit and Lambeth group were modelled using an elastic- perfectly plas­ tic model (Mohr-Coulomb). There was limited data concerning these layers in the Geotechnical Investi­ gation report (URS 2014), and the use of more detailed soil models could not be justified. Nonethe­ less, the mechanical parameters of the MohrCoulomb model were calibrated against the result of available Standard PenetrationTests). The Hardening Soil-Small Strain Model (Benz et al. 2009) was used to characterize the London Clay behaviour. Results of Pressuremeter, Triaxial and Oedometer tests con­ ducted on samples of this material were available. These were used to choose soil models’ parameters, presented in Table 1. The utilized symbols are not explained in this paper for brevity but they are con­ sistent with the Plaxis software manual (Brinkgreve et al. 2013). In the following sections, the analyses carried out for St. Mary’s Abchurch (Section 4) and the Man­ sion House (Section 5) are described separately. The results concerning the greenfield reference condi­ tions and the soil-structure interaction analyses are compared to the displacements measured during construction.

Soil parameters.

Parameter

MG

LC RTD Unit-B

γ’ [kN/m3] γSAT[kN/m3] cu [kPa] c’ [kPa] φp’ [°] ψ [°] ν’ [-] ν’ur [-] K0 [-] K0nc [-] OCR E’v[MPa] Eu[MPa] pref [kPa] m [-] Eu50ref[MPa] E’50ref[MPa] Eoedref[MPa] Gurref [MPa] Eurref [MPa] G0ref [MPa] γ0,7 [-]

9 19 20 0 34 0.2 0.45 0.44 5 -

9 19 0 39 17,5 0.2 0.5 0.37 70 -

9.5 19.5 106 5 23.7 12.5 0.2 1.3 0.60 4.73 200 1 6.5 4.9 4.9 3.22 7.72 61.76 1.1E-04

LC UnitA3

LC UnitA2

LG

9.5 19.5 187 5 24 12.5 0.2 1 0.59 2.84 200 1 17.7 13.3 13.3 16.58 39.79 86.05 1.1E-04

10.5 20.5 337 5 28.2 12.5 0.2 1 0.53 3.59 200 1 17.7 13.3 13.3 16.58 39.79 86.05 1.1E-04

12 22 500 25 25 14 0.2 ­ 1 0.58 ­ 375 500 ­ ­ ­ ­ ­ ­ ­

Note: MG = Made Ground; RTD = River Terrace Deposit; LC = London Clay; LG = Lambeth Group.

4 ST. MARY’S ABCHURCH 4.1

Greenfield analyses

The area around the church is densely urbanized. However, in the church yard, greenfield conditions can be assumed. Data from the church yard (Figure 1b), was used to evaluate greenfield finite element analyses. To model greenfield displacements, a mesh was constructed. This mesh had a length of 146 m along the tunnels alignment and a transverse width of 90 m. The tunnel construction process (conventional heading with SCL) was modelled in detail. The pilot tunnel front was excavated entirely and then the elas­ tic lining was activated in the next analysis step. The enlargement of the pilot tunnel proceeded from heading to invert in 1 m ring segments. This process was modelled, with elastic elements for the lining, which gain their full stiffness with each analysis step. 4.2

Soil-Structure interaction analyses

With the purpose of investigating how the existing masonry structure was affected by the excavation, a series of 3D soil-structure interaction analyses have been performed taking into account the structural

608

arrangement of the building and its orientation rela­ tive to the tunnel axis. Following Losacco et al. (2014), a semi-coupled approach was adopted which simplified the structure to a series of equivalent solids. The properties of the equivalent solids were chosen to capture the stiffness and weight of each façade. Fixing the height equal to the embedded part of the building (Figure 4), each façade was modelled using the evaluated properties. The equivalent selfweight γ*, the E* and G* modulus have been calcu­ lated reproducing the flexural stiffness EI and the shear stiffness GA of each façade:

where Areal is the area of the façade without the openings, A* the cross-sectional area and I* the second moment of area of the equivalent solid. The constitutive relationship chosen for the elas­ tic equivalent solids was the Jointed Rock (Brink­ greve et al. 2013) which allows defining orthotropic properties for the material. According to Losacco et al. (2014), a modifying factor α was used, which is a function of the aspect ratio H/L, the opening ratio Wr and the geometric factor L/Heq (L being the length of the façade, H total height of the build­ ing, Heq height of the equivalent solid). Hence, multiplying E* and G* to the factor α, the elasticity terms E2 and G2 for the equivalent solid are deter­ mined. Table 2 summarises the adopted values for the equivalent solid parameters.

Table 2.

Structural parameters of St. Mary’s Abchurch.

Façade

γ* [kN/m3]

α [-]

E2 [kPa]

G2 [kPa]

East South North West

81,28 74,72 56,59 183,54

0,49 0,465 0,7 0,43

6,58E+07 5,74E+07 1,04E+07 1,81E+08

1,08E+07 7,60E+06 2,79E+05 1,49E+07

4.3 Comparison between numerical results and the monitoring data The results of the greenfield analyses for the advancements of the front excavation face are shown in Figures 5 and 6, as surface settlement profile along Abchurch Lane for different tunnel advances. These tunnel advances are ys=-46m for the pilot tunnel and -16m, -46m for the enlarge­ ment. ys=-16m indicates that the tunnel heading is directly under the northeast corner of the building while ys=-46m indicates that the tunnel trough has completely developed under the church. The tunnel advance coordinate ys is shown in Figure 1b and the tunnel position is indicated inset in Figures 5 and 6. In the same figures, the FEM results are compared with the monitoring data from levelling points.

Figure 5. Greenfield settlement trough Pilot tunnel: advance -46 m.

Figure 4. 3D view of the FE mesh showing the tunnels and the equivalent solid of St. Mary Abchurch.

Figure 6. Greenfield settlement trough Enlargement tunnel: advance -16 m and -46 m.

609

In Figure 5, a good agreement was achieved between the simulations and the LP data shown in Figure 1b. The simulations are for the final advance (-46m) of the pilot tunnel when the trough is completely developed. Similarly, in Figure 6, a reasonable agreement is observed between the enlargement tunnel simulation and the field data. In both figures, a good fit was reached on the left side of the trough where pure greenfield points were located in the church yard (for points located between -15 and -25 on the x-axis). On the other side, higher settlements were obtained in the field due to the influence of other buildings on the measured data, which was not considered in greenfield analyses. The comparison between the results of the soil-structure interaction analyses and the levelling point settlements, are plotted in Figure 7 and Figure 8 respectively for the east and south façades of St Mary’s Abchurch. As expected, the soil-structure interaction reduces the curvature observed in greenfield simulations. While the displacements are similar, the simulated response appears stiffer than the actual structure. More detailed measurements have demonstrated that existing damage and an interaction with an adjacent building has altered the response. These indicate the need to use more detailed building models to capture the soil-structure interaction response. 5 THE MANSION HOUSE 5.1 Greenfield analyses Following the same approach adopted for St. Mary’s Abchurch, first greenfield analyses were carried out, as

Figure 8. SSI settlement of South Façade - Enlargement tunnel: advance -16m and -46m.

a reference condition to be compared to the soilstructure interaction analyses. The ground conditions were modelled similar to Section 4, with minor differences. The Mansion House is directly affected by the excavation of the new southbound Northern Line running tunnel and of the adjacent Moving Walkway tunnel. Both were constructed with conventional method and they have a sprayed concrete lining (external equivalent circular diameter of 5.3m the former, 8.2 m the latter). The relative position and depth of these tunnels are shown in Figure 3b. The running tunnel was constructed with a full-face exca­ vation while the Moving Walkway tunnel was con­ structed with the top heading and bench excavation method. The tunnelling process was modelled accordingly. 5.2 Soil-Structure Interaction with partially uncoupled approach Each external façade of Mansion House was mod­ elled as an equivalent solid. This solid model is shown in Figure 9, and its approximate properties are given in Table 3. Since only the east façade response is explored in detail, the parameters γ* and α have been determined for this façade, other facades are assumed to have the same parameters.

Table 3.

Figure 7. SSI settlement of East Façade - Enlargement tunnel: advance -16m and -46m.

Structural parameters of Mansion House.

Façade

γ* [kN/m3]

α [-]

E2 [kPa]

G2 [kPa]

East

52.94

0,90

3,27E+07

5E+06

610

Figure 9. Structural elements included in the finite element models: running tunnel, moving walkway tunnel, equivalent solid.

Figure 10. Mansion House East façade measured vs com­ puted settlements after running tunnel (RT) excavation.

weight increases the maximum settlement, as observed in other works (Franzius et al. 2004; Bilotta et al. 2017). The flattening effect of the stiff building on the settlement trough was also well captured. Figure 11 shows the settlements of the East façade after the MW excavation. The predicted volume loss is close to the measured one, but the simulated soil-structure response appears flatter than the measured response. This indicates differences between the real stiffness of the building and equiva­ lent solid. These differences may be due to openings or nonlinearity in the structural response. Both aspects are not considered by the adopted elastic for­ mulation of the equivalent solid. Finally, the effect of the interaction between the two tunnels can be observed in Figure 12. In this figure, the dashed line presents the superposed results from the independent analyses of RT and MW excavations. In contrast, the solid line presents the results of the sequential RT and MW excavations. It is clear that the analysis which neglects the excavation of the running tunnel before the excavation of the walkway tunnel (dashed line) underestimates the computed settlement trough calculated considering the whole process (solid line). Although the first tunnel (RT) has a smaller diameter than the second one (MW) the interaction between the construction of the two tunnels is not

As shown in Figure 9, two different tunnels were excavated under Mansion House. These are the Run­ ning Tunnel (RT), and the larger Moving Walkway (MW) tunnel. The interaction between these tunnels were expected to alter the displacement field that would be caused by tunnelling operations. Therefore, several 3D models were analysed to highlight the effect of tunnel-soil-structure interaction on the prob­ lem. The results presented in this section refer to the settlement trough upon finishing the construction of the entire RT and MW tunnels. In the first model, the soil structure interaction problem was considered by modelling the RT exca­ vation and the Mansion House as an equivalent solid. Since the mechanical characteristics of the masonry of Mansion House were unknown, prelim­ inary analyses were carried out with a range of suit­ able parameters, suggested by Frischmann et al. (1994). Hence, this model (RT excavation only) was used to tune the characteristics of the solid in order to reproduce the measured displacements. The final values of the parameters adopted for the equivalent solid are shown in Table 3. Once the properties of the equivalent solid were determined, another model was constructed, includ­ ing the Moving Walkway (MW) tunnel excavation. In this case the ground movements induced by the excavation of the walkway include the effects of interaction with the previous excavation of the run­ ning tunnel. In order to assess the importance of such interaction, a latter model was analysed where only the walkway was excavated and the excavation of the running tunnel was neglected. 5.3 Comparison between numerical results and the monitoring data Once all models were completed, the numerical results were compared with the monitoring data from levelling points. Figure 10 shows the displace­ ments induced on the east façade of the Mansion House by the RT excavation. Both greenfield and soil-structure interaction analyses are presented. The analyses demonstrated that the effect of the building

Figure 11. East façade measured vs computed settlements after running tunnel (RT) and moving walkway (MW) excavation.

611

REFERENCES

Figure 12. East façade computed settlements: effect of interaction between the two tunnels (RT and MW) excavation.

negligible. This is not surprising, due to the small dis­ tance between the two tunnels; this distance is lower than two times the lower diameter. In order to evaluate the forces imposed on the building, such interaction effects must be considered. 6 CONCLUSIONS This paper discussed the application of the semicoupled modelling approach to the numerical modelling of the effects of tunnelling on two historic build­ ings in London: St. Mary’s Abchurch and the Mansion House. These buildings were affected by the construction of new tunnels for the Bank Station Capacity Upgrade. The semi-coupled analyses provide a convenient way to account for the nonlinear interactions between tunnels, soil and the buildings. The results demonstrated that with detailed soil models and staged excavation analyses, accurate greenfield pre­ dictions are possible. The results further indicated the importance of interaction between adjacent tun­ nels. The soil-structure interaction analyses indicated the flattening of displacement profiles due to struc­ tural stiffness. However, for both buildings, the equivalent solids resulted in stiffer response than the response indicated in the field data. The discussions highlighted potential reasons for these differences, such as the presence of existing damage, openings and nonlinearity, and the interaction between adja­ cent buildings.

Addenbrooke, T. I. Potts, D. M. & Puzrin, A. M. 1997. The influence of pre-failure soil stiffness on the numerical ana­ lysis of tunnel construction. Géotechnique 47, 3, 693–712. Amorosi, A. Boldini, D. de Felice, G. Malena, M. & Sebastianelli, M. 2014. Tunnelling-induced deformation and damage on historical masonry structures. Géotech­ nique 64(2), 118–130. Benz, T. Vermeer, P.A. & Schwab, R. 2009. A small-strain overlay model. Int. J. Numer. Anal. Meth. Geomech. 33 (1), 25–44. Bilotta, E. Paolillo, A. Russo, G. & Aversa, S. 2017. Dis­ placements induced by tunnelling under a historical building. Tunnelling and Underground Space Technol­ ogy, 61, 221–232. Brinkgreve, R.B.J. Swolfs, W.M. & Engine, E. 2013. PLAXIS 3D User’s Manual. PLAXIS bv, The Netherlands. Franzius, J.N. Potts, D.M. Addenbrooke, T.I. & Burland, J. B. 2004. The influence of building weight on tunnelling-induced ground and building deformation. Soils Found. 44 (1), 25–38. Frischmann, W.W. Hellings, J.E. Gittoes, G. & Snowden, C. 1994. Protection of the Mansion House against damage caused by ground movements due to the Docklands Light Reilway extension. Proc. Instn. Civ. Engng. 1994, 107, 65–76. Giardina, G. van de Graaf, A. V. Hendriks, M.A.N. Rots, J. G. & Marini, A. 2013. Numerical analysis of a masonry facade subject to tunnelling-induced settlements. Engng Struct. 54, 234–247. Losacc o, N. Burghignoli, A. & Callisto, L. 2014. Uncoupled evaluation of the structural damage induced by tunnelling. Geotechnique 64(8), 646–656. Rampello, S. Callisto, L. Viggiani, G.M.B. & Soccodato, F. 2012. Evaluating the effects of tunnelling on historical buildings: the example of a new subway in Rome. Geo­ mech. Tunnelling 5(3), 275–299. Pickhaver, J. A. Burd, H. J. & Houlsby, G. T. 2010. An equivalent beam method to model masonry buildings in 3D finite element analysis. Comput. Struct. 88, 19–20, 1049–1063. URS. 2014. Revision 1.0. Geotechnical Investigation Report – Bank Station Capacity Upgrade. URS-8798­ RPT-CIV-002190. Viggiani, G.M.B. Soccodato, F.M. & Burghignoli A. 2006. A study of the interaction between the new C line of Roma underground and the Aurelian Wall. Geotechnical Aspects of Underground Construction in Soft Ground, pp. 687–693. Yiu, W. N. Burd, H. J. & Martin, C. M. 2017. Finiteelement modelling for the assessment of tunnel-induced damage to a masonry building. Géotechnique,67, 9, 780–794.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Centrifuge tests on tunnel-building interaction in liquefiable soil G. Miranda, V. Nappa & E. Bilotta Department of Civil, Architectural and Environmental Engineering, University of Napoli Federico II, Naples, Italy

S.K. Haigh & S.P.G. Madabhushi Department of Engineering, University of Cambridge, Cambridge, UK

ABSTRACT: Relatively shallow and light underground structures, such as urban tunnels, may run through liquefiable sand deposits. In urban area they are likely close to the foundations of buildings and easily inter­ act with them during earthquakes. The reciprocal influence of a tunnel and an adjacent building in presence of soil liquefaction has been investigated in this work. For this purpose, centrifuge tests were carried out at the Schofield Centre of the University of Cambridge (UK) on a reduced scale model of box-type section tunnel. In a first test a model tunnel was embedded in a liquefiable layer of sand. In the second one, a model building was added and founded in the proximity of the tunnel. The study contributes to the wider topic of the resilience of urban environment to natural hazards, and to earthquake-induced soil liquefaction specifically.

1 INTRODUCTION Shallow tunnels in urban area are likely located close to the foundations of buildings and they easily interact with them during earthquakes. The recipro­ cal influence of a tunnel and an adjacent building has been investigated in the literature (Pitilakis and Tsi­ nidis, 2014; Tsinidis et al., 2015; Dashti et al., 2015; Hashash et al., 2017; Loncarevic et al., 2019). Nevertheless, the influence of soil liquefaction on the tunnel-building interaction has not received attention to date. The project STILUS, carried out within the framework of the European funded net­ work SERA (Seismology and Earthquake Engineer­ ing Research Infrastructure Alliance for Europe) intends to investigate this problem through a series of centrifuge tests. This problem appears rather important consider­ ing the rapid extension of the built environment, both above- and underground, to areas prone to liquefaction. In fact, relatively shallow and light structures buried in liquefiable deposits may suffer large uplift during earthquakes. Similar events have involved sewer pipes or large underground struc­ tures, as for instance open-cut railway tunnels (Koseki et al., 1997a; Yasuda and Kiku, 2006). The uplift behaviour of underground structures caused by liquefaction is often studied by phys­ ical models: for instance 1-g shaking table models of completely-buried box structure, semiburied roads and sewer manholes and pipes and possible mitigation measures (Koseki et al.,

DOI: 10.1201/9780429321559-80

613

1997b, Otsubo et al., 2014, Watanabe et al., 2016) or centrifuge models of tunnel of different shapes, embedded in sand layers of different density, with several overburden and groundwater level (Chou et al., 20010; Chian and Madabhushi, 2012; Chian et al., 2014). Experimental evidences indicate that the width of the underground structure and the depth of the liquefied layer below it have a large influence on the uplift displacement. This is triggered as far as high excess water pressures develop, as those induced by strong motions. Chou et al. (2010) have identified and analysed experimentally four uplift mechanisms for an underground structure undergo­ ing ground liquefaction: (i) ratcheting mechanism associated with cyclic movement of tunnel relative to backfill; (ii) migration of pore water mechanism; (iii) bottom heave mechanism resulting from shear failure in deeper soft soil; (iv) viscous flow of liquefied soil. This paper investigates through physical modelling the interaction of such mechanisms and the dis­ placements induced on aboveground structures that may be founded nearby. 2 EXPERIMENTAL LAYOUT Two models were created at reduced scale and tested accordingly at 60-times increased g-level in the Turner Beam Centrifuge at the Schofield Centre of the University of Cambridge.

Table 1.

Physical properties of Hostun Sand.

Soil Hostun Sand HN31

Gs

emax

emin

2.65

1.041

0.555

γ (kN/m3) 16.8

The ground layer consisted of homogenous Hostun sand at a relative density of about 45%-50%. Table 1 reports its properties as known in literature. The sand was dry pluviated in thin layers through an automatic hopper system. To achieve a specific relative density it needs to be poured from a particular height (700mm during these tests) and at a particular flow rate. The sand is placed in a hopper suspended above the model container. A 10mm nozzle was placed at the bottom of the hopper to control the flow rate and the drop height was con­ trolled through the program used to control the equipment. The sand pourer was stopped at desired locations to allow placement of instruments. During the model preparation, arrays of miniature pore pres­ sure transducers (PPTs), piezoelectric and MEMS accelerometers are deployed at the desired locations. The piezoelectric accelerometers are calibrated before use in a centrifuge test using a specially designed calibrator. A calibration factor for the accelerometer is obtained in the units of g/V. The piezoelectric accelerometers measure acceleration in the direction of the thread. Displacement transducers (LVDTs) were used to measure the settlements at different locations. Prior to use, the LVDTs were calibrated by applying known displacements from a screw gauge and its output was measured. To allow a comparative numerical analysis to be conducted, knowledge about the stiffness of the sand within the models is required. A method of charac­ terizing the soil models in the centrifuge is by meas­ uring the shear wave velocity Vs in the test. Shear wave velocity Vs was obtained by using a miniature air hammer which operated at strains around 0.03%. The small strain shear modulus G0 was then obtained using the equation:

To achieve this, an air hammer was installed at the bottom of each model. Through a set of valves, air is supplied to one end of the air hammer which pushes a metal pellet within the air hammer causing it to impact on the opposite end of it. This induces a shear wave in the model. Accelerometers placed at different but know elevation recorded the arrival times of the shear waves. Hence the shear wave vel­ ocity between adjacent accelerometers can be deter­ mined. In the first test a model tunnel (Figure 1) is

Figure 1. Test #1 model (up) and instruments layout (down).

embedded in the sand layer. The rectangular model tunnel is made using an extruded section of alumin­ ium alloy. Rough dimensions and properties are pro­ vided in Table 2. The rectangular tunnel may be representative of a section of a metro station tunnel that can accommodate two separate platforms. The soil cover above the tunnel corresponds to an embed­ ment ratio C/HT = 1.1. In the second test a linear-elastic sway frame (SDOF) made of aluminium is founded in the sand layer. Rough dimensions and properties are provided in Table 3. Figure 2 shows the centrifuge model including the sway frame (i.e. the building), that is located beside the tunnel. The sway frame is fitted with accelerom­ eters to capture its horizontal sway as well as rock­ ing behaviour. Two vertical accelerometers are positioned on the base of the structure to enable measurement of rocking angles. A displacement Table 2.

Tunnel dimensions and properties.

Height (m) Lenght (m) Cover Test #1 (m) Cover Test #2 (m) Mass (kg/m)

614

Model Scale

Prototype Scale

0.076 0.153 0.084 0.082 7.6

4.56 9.18 5.04 4.92 27 360

Table 3.

Structure dimensions and properties.

Height (m) Frequency (Hz) Natural period (s) Damping (ξ) Mass (kg) Area (m2) Stiffness (kN/m) Bearing pressure (kN/m2)

Model scale

Prototype scale

0.01 160.55 0.0062 0.1761 0.97 0.02 996 0.95

4.8 2.68 0.37 0.18 209 520 73.8 59 772 57.09

Table 4.

Model earthquakes, Test #1.

PGA Signal Frequency (Hz) Cycles (g)

Arias

Intensity (cm/s)

1 2 3 4 Kobe

0.159 0.405 5.071 7.798 0.74

1 -

Table 5.

Figure 2. Test #2 model (up) and instruments layout (down).

transducer is located at the base of the frame to measure vertical settlement. The sand layer needs to be saturated before testing. To avoid the incompatibility between the dynamic and diffusion time scaling laws, a high viscosity aque­ ous solution of hydroxypropyl methylcellulose (HPMC) is used to saturate the sand layer, with a viscosity 60 times larger than water. Considering the need to study the displacement field around the tunnel, high speed photogrammetry is used in the tests. Hence stems the choice to use a transparent side container. A rigid container with

10 15 -

0.058 0.100 0.310 0.339 0.167

Model earthquakes, Test #2.

PGA Signal Frequency (Hz) Cycles (g)

Arias

Intensity (cm/s)

1 2 3 4

0.139 0.372 5.280 8.128

1

10 15

0.067 0.105 0.369 0.416

a Perspex window is used. It is known that this type of model container may cause boundary effects affecting the response, particularly when liquefaction is reached. Therefore Duxseal® is used on the walls, to minimize the boundary effects: it has been showed that it can reduce the stress wave reflections by about two-thirds. A servo-hydraulic earthquake actuator is used to apply near-sinusoidal earthquake motions to the cen­ trifuge model. The amplitude of the signal will be increased during the test, until soil liquefaction is achieved. After preparation, the models were loaded on the centrifuge basket, connecting the devices to the onboard acquisition system. At each test start the centrifuge was spun up to 60g. The features of the earthquakes fired are summarized in Table 4 and Table 5. Time histories of acceleration and pore pressure in the ground are recorded during shaking, along vertical and horizontal arrays. Similarly, displace­ ment time histories at a few points at ground sur­ face (settlement) and on the sway frame (settlement, horizontal displacement and tilt) are monitored. Particle Image Velocimetry (PIV) Photogram­ metry enables a deep insight on the triggering of uplift and the evolution of the mechanism. Comparing the time histories measured in cen­ trifuge model #1 and #2 enabled to highlight the influence of tunnel-building interaction on the displacements field induced by soil liquefaction.

615

3 EXPERIMENTAL RESULTS 3.1

Excess pore pressure

The Excess Pore-Fluid pressure ratio (ru) was obtained by dividing the Excess Pore Fluid Pres­ sure (Δu) by the value of the initial effective verti­ cal stress σ'v,0. The soil was considered to be liquefied when the value of ru reached 0.8. During the first earthquake, the largest value of ru was reached close to the soil surface in both tests, being respectively equal to 0.28 and 0.08. The second earthquake almost produced liquefaction in the first test, where in free field conditions (PPT 1) ru=0.75. Nevertheless, during the second test, the higher confining stress beneath the foundations kept the soil far from the condition of liquefaction (ru,max=0.20). EQ3 finally led the soil to liquefaction: in the first test all the PPTs from 1 to 6 showed that ru,max > 0.90. The soil beneath the tunnel got very close to liquefaction, being ru,max=0.8. In the second test, the volume of liquefied soil was reduced by the stress increase given by the foun­ dation. Figure 3 shows a comparison between the excess pore pressure ratio recorded by PPT 2 and PPT 3 between test 1 and test 2. It is clear the effect due to the presence of the structure, lead­ ing to a value of ru,max ≃ 0.4 below the founda­ tion while in the test 1 the pore pressure ratio is equal to 1. PPTs n. 7 and n. 8 recorded the behaviour of the soil underneath the tunnel. During the first 3-5 cycles of the earthquake, pore pressure builds up. Later, when the soil around the tunnel liquefies, the tunnel itself is pushed upwards by the buoyant force, the stress state of that volume of soil is subjected to a considerable reduction,

Figure 3. Comparison between the pore pressure ratio obtained from test 1 and test 2 – PPTs placed close to the ground surface.

and the pore pressures tend to reduce – on aver­ age - while the tunnel heaves. Figure 4 shows the counterphase of the pore pressure build-up and dissipation cycles recorded by PPT 7 and PPT 8, which are located symmetrically with respect to the tunnel. This behaviour represents the evidence of the rocking motion combined with the heave. When the tunnel motion stops, the pore pres­ sures rise again for a few seconds, and then they start dissipating very slowly, as the water flow is obstructed by the tunnel invert. This behaviour is more evident during Test #1, where the uplift is larger. 3.2

Accelerations

Response spectra were obtained by the acceleration time histories recorded by the piezo accelerometers placed within and outside the model. Figures 5 and 6 show how the seismic signal is altered when it reaches the soil surface. Although the PGA at the surface is larger than in the input signal, the PSA is much lower within the range of the fundamental fre­ quency of the signal, due to the liquefied soil that limits the propagation of shear waves to the surface. This behavior is not observed below the foundation (TEST 2, ACC 3), where liquefaction does not occur, and the response spectrum is always higher than the input response spectrum, within all ranges of periods. 3.3

Displacements

The data recorded by the LVDTs and the PIV showed that during the first two earthquakes the displacements were negligible. The third and the fourth earthquakes induced liquefaction of the soil, that caused settlements of the structure and uplift of the tunnel (Figure 7). In both tests, the displacements are larger during EQ3 than EQ4, even though the latter is stronger; this is due to the density increase at the end of EQ3, that reduces the development of liquefaction in the following earthquake. The uplift motion of the tunnel is paired with a rocking behavior, as seen in Figure 7, and is more evident in the first test rather than in the second. The final displacements show that both the tunnel and the structure are subjected to differential displacements, where the tunnel rotates clockwise, and the structure rotates anti­ clockwise. The structure settlement is larger on the side closer to the tunnel, as well as the tunnel uplift is larger on the left-hand side, where there is no structure limiting its heave.

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Figure 4. Pore pressure ratio obtained from the earthquake 3 for the test 1 and test 2 – PPTs placed beneath the tunnel.

Figure 5. Comparison between response spectra obtained by records close to the surface, Test #1, eq #3.

Figure 6. Comparison between response spectra obtained by records close to the surface, Test #2, eq #3.

Through the PIV also the soil displacements around the tunnel and beneath the structure were measured. When the structure is absent (first test), the ground settles beside the tunnel, and the tunnel cover is pushed upwards by the

tunnel itself. In the second test, the ground settlements extend to part of the tunnel cover and the contour map shows the interaction zone between the left foundation and the right side of the tunnel.

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4 CONCLUSIONS The project STILUS, within the framework of the European funded network SERA intended to investigate the problem of the reciprocal influence of a tunnel and an adjacent building in the pres­ ence of soil liquefaction through a series of cen­ trifuge tests. The paper described the results of the first two centrifuge tests carried out within the pro­ ject. In a first test a model tunnel was embedded in a liquefiable layer of sand. In the second one, a model building was added and founded in the proximity of the tunnel. The results clearly show how the distribution of excess pore pressure induced by shaking is affected by the presence of the tunnel and of the building. They also show that the tunnel uplift is driven by the increase of pore pressure below the tunnel invert and the concurrent reduction of effective stresses (and shear resistance) in the cover. The calculated distribution of excess pore pres­ sure induced around the tunnel and the building basement may also be useful to identify where miti­ gation techniques that may locally reduce porepressure build-up would be most effective against the effects of soil liquefaction and should be imple­ mented in the tests.

ACKNOWLEDGEMENTS The activity presented in the paper was carried out within the research project STILUS (StructureTunnel Interaction in LiqUefiable Sand). The project was partly funded within the framework of the Euro­ pean funded network SERA (Seismology and Earth­ quake Engineering Research Infrastructure Alliance for Europe, grant agreement No. 730900).

REFERENCES

Figure 7. Vertical displacements fields at the end of shaking.

Chian, S.C., Madabhushi, S.P.G., 2012. Effect of buried depth and diameter on uplift of underground struc­ tures in liquefied soils. Soil. Dyn. Earthq. Eng. 41, 181–190. Chian, S.C., Tokimatsu, K., Madabhushi, S.P.G., 2014. Soil liquefaction- induced uplift of underground structures: physical and numerical modeling. J. Geotech. Geoen­ viron. Eng. 140, 04014057. Chou JC, Kutter BL, Travasarou T, Chacko JM, 2010. Cen­ trifuge modeling of seismically induced uplift for the BART transbay tube. J Geotech Geoenviron 137 (8):754–765 Hashash, YMA, Musgrove, M, Dashti, S, Cheng, Ph, 2017. Seismic Performance Evaluation of Underground Struc­ tures – Past Practice and Future Trends. In: Proceedings of PBD-III Performance Based Design in Earthquake Geotechnical Engineering, Vancouver, 2017. Paper305. Koseki, J., Matsuo, O., Ninomiya, Y., Yoshida, T., 1997a. Uplift of sewer manholes during the 1993 Kushiro-Oki earthquake. Soils Found. 37 (1), 109–121.

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Koseki, J., Matsuo, O., Koga, Y., 1997b. Uplift behavior of underground structures caused by liquefaction of surround­ ing soil during earthquake. Soils Found. 37 (1), 97–108. Otsubo, M., Towhata, I., Taeseri, D., Cauvin, B., Hayashida, T., 2014. Development of structural reinforce­ ment of existing underground lifeline for mitigation of liquefaction damage. In: Geotechnics of Roads and Rail­ ways: Proceedings of the XV Danube-European confer­ ence on geotechnical engineering. Vol. 1, 119–125, Wien Pitilakis, K., Tsinidis, G., 2014. Performance and Seismic Design of Underground Structures. Geotechnical, Geo­ logical and Earthquake Engineering, 28, pp. 279–340.

Tsinidis G., Pitilakis K., Madabhushi G., and Heron C. Dynamic response of flexible square tunnels: centrifuge testing and validation of existing design methodologies Géotechnique 2015 65:5, 401–417 Watanabe K., Sawada R., Koseki J., 2016. Uplift mech­ anism of open-cut tunnel in liquefied ground and simplified method to evaluate the stability against uplifting, Soils and Foundations, Volume 56, 3, 2016, 412–426 Yasuda, S., Kiku, H., 2006. Uplift of sewage manholes and pipes during the 2004 Niigataken-Chuetsu earthquake. Soils Found. 6 (46), 885–894.

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The effect of deep excavation on existing railway tunnel M. Mitew-Czajewska Warsaw University of Technology, Warsaw, Poland

ABSTRACT: The paper presents a case study of a deep, extensive excavation executed in the downtown of Warsaw in a very close proximity to the existing railway tunnel. The tunnel, containing 10 railway tracks, is located at the outlet from the main Warsaw railway station. The construction of the excavation may affect the track geometry and cause track settlement or heave depending on the stage of construction. Prior to the construction, the numerical 2D FEM analysis were performed in order to evaluate the impact of the excavation to the railway tunnel. After the project implementation, a comparative analysis of the results of numerical analyzes and measurements of vertical displacements of railway tracks carried out during construction was performed. It occurred that the theoretical displacements of the rail obtained in the analysis were underestimated. A discussion of numerical analysis in relation to measurement results is presented.

1

INTRODUCTION

2 DESCRIPTION OF THE CASE

Nowadays, due to a growing urbanization and rapid population growth in cities, one can observe a significant increase in urban density especially along the newly built metro lines. New structures are constructed in close proximity to the existing buildings, often prone to settlement. In such con­ ditions, in dense urban area, the excavation design must also include verification of the impact of the excavation on surrounding struc­ tures (Amorosi et al. 2014, Bogusz & Godlewski 2018, Haibin et al. 2014, Kadlicek et. al. 2016, Łukasik et al. 2014, Mitew-Czajewska 2015, Mitew-Czajewska 2018, Mitew-Czajewska 2019, Phien-wej et. al. 2012, Romani et. al. 2012, Sha­ fiku & Al-Ameri 2018, Superczyńska et al. 2016, Totsev 2012, Yu & Jia 2012). This is made usu­ ally basing on finite element analysis, using pro­ found numerical and geotechnical knowledge including e.g. the choice of proper material con­ stitutive models. In the paper, a case study is presented describ­ ing extensive analysis of the influence of a deep and very large excavation located in the down­ town of Warsaw on the existing railway tunnel containing 10 railway tracks. Imposed limits on the movements of the tracks closest to the exca­ vation were the reason to carry out the FE ana­ lysis of the excavation impact prior to the construction.

DOI: 10.1201/9780429321559-81

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2.1 Location of the excavation The analyzed office building is a part of a larger office complex consisting of three buildings - 2 buildings approx. 90 m high and a tower more than 150 m high. The entire complex is located at Chmielna St. in Warsaw in the vicinity of the 10 railway tracks located in the tunnel. Location of the entire investment in rela­ tion to the street layout and Central Station (the main railway station in Warsaw) is shown in Figure 1. Gen­ eral plan view of the tracks, the site and location of calculation cross-sections is shown in Figure 2. 2.2 Geometry of the excavation and the structure of the tunnel The analyzed excavation is a 4 level underground car park of the huge investment consisting of 3 inde­ pendent office buildings. The size of the excavation is 78.5 m in direction perpendicular to the railway tunnel and approx. 3 times that size along the rail­ way. Maximum depth of the excavation amounts to 18.9 m in the central part and 17.9 m along the exca­ vation walls. It was designed that all 3 underground parts of the excavation should be constructed separ­ ately, one-by-one, using top&down method with 31 m deep, 0.8 m thick peripheral diaphragm wall and underground slabs supporting the wall in the temporary excavation state. The buildings are founded on a foundation slab and d-wall type piles.

2.3

Geotechnical conditions

The area of the investment is located at Chmielna St., which is, according to the Detailed Geological Map of Poland, located on the moraine plateau. The ground level is at a height of approx. 35­ 36.5 m above the level “0” of the Vistula river (local altitude system). The geological crosssection (including the numbers of geological layers) along the analyzed section location is shown in Figure 4. In general, in the investment area, below the ground surface there are artificial, human-made soils to the depth of from 2 m, locally in the vicin­ ity of the railway tunnel even to 10.30 m bgs (below ground surface). These soils are composed of fine, medium, clayey sands and locally clay mixed with brick rubble. Below this layer lies a discontinuous layer of medium density fine and silty sands. In some locations, instead of sands, from the depth of 8.50 m to 17.5 m bgs there are silty clays and silts. This layer covers a discontinuous, highly eroded layer of boulder clay, occurring at a depth of 11.80 to 16.0 m bgs. All cohesive soils are stiff. Below the

Figure 1. Location of the construction site.

Figure 2. Plan view of the construction site in relation to the railway trucks.

The tunnel is a reinforced concrete frame structure consisting of 7 spans of about 8 to 12 m. The walls of the tunnel are 0.5 m thick, while reinforced concrete columns supporting the upper slab have a square cross-section of 0.7 m. Beam roof slab has a height of 2 m. Walls and columns are set on 0.8 x 1.40 m benches. The track surface is laid on crushed stone ballast. The details of the analyzed cross-section are pre­ sented in Figure 3.

Figure 4. Geological cross-section.

Figure 3. Analyzed cross-section.

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cohesive soils layer, to a depth of 27 to 35 m bgs there is a layer of hydrated very dense fine and medium sands. For the analysis purpose, the interpretation of geological conditions was made and a geotechnical profile including 5 geotechnical layers was created as shown in Figure 5. Geotechnical parameters of these layers taken to the numerical analysis are given in Table 1. Two underground water tables were found in the area of the investment. The first, non­ continuous water level, was spotted from 1.5 to 9 m bgs. The second, continuous aquifer is located within the sandy sediments of geological layers VI and VII and its table stabilizes at about 10 m bgs. In the analysis the ground water level was adopted as continuous, at an ordinate 24.05 m above “0” of Vistula river (i.e. 12.3 m bgs).

2.4

Construction stages

The construction of the underground part of the building and the final state were modeled assuming the following construction stages: – stage 1 - initial conditions including existing struc­ tural elements of the tunnel and railway trucks, activation of geostatic stresses in the ground, terrain elevation 36.35 m above 0 of the Vistula river, – stage 2 – execution of a shallow, preliminary exca­ vation, construction of the 0.80 m thick, 31 m deep diaphragm wall, embedded min. 2 m in the imper­ meable soil layer (geotechnical layer VIII), – stage 3 - excavation to the ordinate 31.05 m above 0 of Vistula river (i.e. 5.3 m bgs), – stage 4 - construction of the 0.30 m thick -1 slab, – stage 5 - excavation to the ordinate 24.05 m above 0 of Vistula river (i.e. 12.3 m bgs), lowering the level of groundwater table to the ordinate 16.50 m above 0 of Vistula river (i.e. 19.85 m bgs), i.e. below the level of the bottom of the final excavation, – stage 6 - construction of the 0.30 m thick -3 slab, – stage 7 - excavation to the target elevation, i.e. 18.45 m above 0 of Vistula river (i.e. 17.9 m bgs) with several local trims to an ordinate 17.45 m above 0 of Vistula river (i.e. 18.9 m bgs), – stage 8 – construction of the 2.3 m thick founda­ tion plate, – stage 9 - construction of the building and oper­ ational load (varying from 190 to 410 kPa).

3 NUMERICAL ANALYSIS 3.1

Figure 5. Geotechnical cross-section.

Table 1 . Geotechnical parameters of soil layers con­ sidered in the numerical analysis. Layer

ɣ kN/m3

E MPa

Eur MPa

K0 -

ϕ’ °

c’ kPa

Layer I - Fill Layer IVc - Cl Layer VI - FS Layer VII - MS Layer VIII - Cl

18 20.5 20 20.5 21

25 45 120 150 80

75 135 360 450 240

0.577 0.515 0.470 0.426 0.500

25 29 32 35 30

0 5 0 0 1

Basic assumptions

A numerical model (Figure 6) was made for the cross-section located in the middle of the investment area, closest to the railway tracks (Figure 3 - geom­ etry, Figure 4 – geotechnics) in order to predict the theoretical, both vertical and horizontal displace­ ments of the railway tracks and the tunnel structure. The analysis was made using the finite element method in 2D space, using the GEO5 FEM software (Fine Ltd 2019), taking into account the elasticplastic analysis of the two-phase medium, assuming a flat deformation state. With respect to the soil body, a modified model with the Coulomb-Mohr plasticity condition was used, without isotropic and kinematic reinforcement, but allowing to define unloading-reloading modulus. An assumption was made about the unconstrained flow. In reference to reinforced concrete structures, i.e. for diaphragm walls, underground slabs, foundation slab, tunnel structure etc. the isotropic linear-elastic model was used. The finite elements mesh (Figure 6) was made of isoparametric, six-node triangular elements, interface and beam type elements.

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The reinforced concrete structures of the building and the tunnel were modeled using beam elements assuming their appropriate stiffness. At the contact between the structure and the soil, zero-thickness contact elements were used. They allow modelling of adhesion and slip phenomena in accordance with the Coulomb-Mohr plasticity condi­ tion. The parameters of contact elements were in accordance with the parameters of the adjacent soils. The reduction factor was considered, R=0.67. The finite element mesh of the numerical model covering both the entire, 78.5m wide, underground part of the building and the 75.4m wide railway tunnel is shown in Figure 6. The whole model including also ~80m of surrounding soil body is therefore very large. Detailed data of the numerical model is as follows: – dimensions – 74 x 230 m – number of finite elements – 16156 – number of nodes - 26989 – number of beam elements – 1448 – number of contact elements – 4344. At the time the analysis was performed, during design stage, the construction stages were planned as mentioned in Section 2.4. 3.2

Results of the numerical analysis

As a result of the numerical analysis performed prior to the construction, at the design stage, following issues were to be verified and compared to permis­ sible and safe values: displacements of the track, dis­ placements (and differential displacements) of tunnel walls and foundations. Vertical and horizontal displacements were obtained for the whole model. Vertical displacements of the model in stages 5 and 7 (intermediate and final excavation stages) are shown in Figure 7. Vertical and horizontal displacements of tunnel walls, foundations and tracks closest to the excava­ tion wall are presented in Figure 8, Figure 9 and Figure 10, in stages 5, 7 and 9 respectively. The results of the analysis proved that:

– the value of the horizontal displacement of the railway tunnel wall directly adjacent to the exca­ vation does not exceed 5 mm in the stage of the deepest excavation (Figure 8) and is 11.5 mm after final structure construction and loading (Figure 9), – values of the horizontal displacements of the remaining tunnel walls are gradually decreasing as the distance from the excavation wall increases and amount to about 0.3 mm in the last, the most distant wall (distance from the excavation edge is 82 m), – values of vertical displacements of the tunnel walls foundations in sections 1 and 2, closest to the excavation wall, are 4.9 mm and 1.9 mm respectively (upwards) in the final excavation stage (Figure 9), and 27.8 mm and 15.9 mm in the last stage after final loading, – in further sections/walls, values of vertical settle­ ments gradually decrease, reaching a value of approximately ±1 mm, – track settlements/heave in excavation stages are within 1 to 2 mm, except for 1st track closest to the excavation, where, in stage 7, an uplift of max­ imum 3.4 mm occurs (Figure 8), – the difference in the values of settlement of the neighboring tunnel walls foundations does not exceed 12 mm (over min. 10 m span) in all con­ struction stages, – maximum value of the settlement of the track clos­ est to the excavation in the final stage is 25.1 mm. This value will be approx. constant along the track and will slowly increase overtime during construc­ tion of next levels of the building. In such a case, it should not affect the train operation. Basing on the observations listed above, it can be concluded that both the theoretical values of settle­ ments and horizontal displacements of the tunnel construction does not exceed the permissible values required by the tunnel owner. Calculated values of maximum horizontal dis­ placements of the tunnel walls were compared with

Figure 6. Finite elements model mesh.

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Figure 7. Results of numerical analysis, plots of vertical displacements in the model, stages 5 and 7.

Figure 8. Vertical displacement of the track closest to the excavation in the 5th stage of construction.

Figure 10. Vertical displacement of the track closest to the excavation in the 9th stage of construction.

the permissible values given in the PN-85/B-03010 standard. It is stated that: The permissible value of the ratio f3/h = 0.004. f3=h ¼ 0:001150:004

ð1Þ

where: h - wall height equal to 10.13 m; f3 - hori­ zontal displacement of walls equal to a maximum value of 11.5 mm in stage 9. The numerical analysis described in this section are the prediction made prior to the beginning of the

Figure 9. Vertical displacement of the track closest to the excavation in the 7th stage of construction.

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construction, based on the assumptions set by the contractor at that stage of the investment. 4 COMPARISON OF THE RESULTS WITH ON SITE DISPLACEMENTS MEASUREMENTS During construction the stages of construction of the underground part of the investment were modified significantly by the contractor. As designed - individ­ ual excavations were to be carried out separately (Figure 11), one after the other, beginning from the middle one, then the left one and the right one at the end. While in the revised version of the design, imple­ mented during construction, all three underground parts were excavated simultaneously with the middle part a bit ahead of the others. As a result of the removal of such a large mass of soil at once a significant relaxation of the whole ter­ rain around the excavation occurred. After the 5th stage of construction, i.e. the excavation till 12.3 m bgs, maximum measured displacement (uplift) of the first track, closest to the excavation, reached 10 mm (Figure 12), while the maximum pre­ dicted displacement for the final excavation stage was 3.4 mm (Figure 9). That means the actual uplift exceeded the expected values three times while the target excavation depth was still not achieved.

5 FURTHER ANALYSIS Basing on the experimental data presented in the article (Mitew-Czajewska 2019) it was stated that due to the execution of 10 to 20 m deep excava­ tions protected by diaphragm walls supported by underground slabs in temporary and permanent state metro structures located in its vicinity may experience up to 10 mm vertical displacement (uplift) caused by unloading. Having that in mind and considering the extraordin­ ary large dimensions of the excavation it may be esti­ mated that the displacement limit given above should be increased by 50 - 80 percent so the final uplift of the tunnel structure and tracks should not exceed 15 – 18 mm. This is of course only an estimate. In order to be certain, 3D FEM model in Plaxis is currently being built (SiemińskaLewandowska et al. 2019). Extensive information on the construction stages implemented during construction is now being analyzed, creation of the 3D model including structure and geology simplifications is in progress. Figure 13 shows the first approximate numerical model of the ana­ lyzed case. After completion of the model, first calculations will be made. The results, i.e. theor­ etical vertical displacements of the railway track closest to the excavation in the 5th excavation stage, will be compared with the measured values of the real vertical displacements in that stage. Figure 12 shows. On this basis, the model will be calibrated and the appropriate material model of the soil will be selected. Then final analysis will be completed and the prediction of the final displacements of the tunnel and the track in all construction stages will be made.

Figure 11. Location of the 3 parts of the excavation in rela­ tion to the railway track and positions of geodesic measure­ ment points.

Figure 12. Values of vertical displacement of the track closest to the excavation measured on-site after the 5th stage of construction.

Figure 13. 3D FE model created in Plaxis 3D (top), typical cross-section (bottom), calibration in progress.

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6 CONCLUSIONS In the paper a case study of the impact of a very extensive and deep excavation on existing railway tunnel and tracks is presented. The construction of the investment is now in progress. The real vertical displacements of the track closest to the excavation is being monitored. Values of the track uplift in the 5th excavation stage exceeded the theoretical values estimated by means of 2D FE analysis made during design stage. The reason for this discrepancy is the significant change of the stages of construction of the underground part of the investment. In the revised version of the design, implemented during construction, all three underground parts are exca­ vated simultaneously with the middle part a bit ahead of the others instead of being excavated one after another, after loading the foundation soil with the foundation plate and underground structure to compensate the load of the excavated soil. It was decided that, in order to estimate the real impact of the final excavation in such a case, the 3D FE analysis is necessary. The creation and calibra­ tion of the 3D model basing on the values of the geo­ desic measurement results is in progress. The results and conclusions including comparison of 2D and 3D models will be presented in nearest future.

REFERENCES Amorosi A., Boldini D., de Felice G., Malena M., di Mucci G. 2014. Numerical modelling of the interaction between a deep excavation and an ancient masonry wall, in: Yoo, Park, Kim & Ban (Eds), Geotechnical Aspects of Underground Construction in Soft Ground: 245–250. Korean Geotechnical Society, Seoul, Korea. Bogusz W., Godlewski T. 2018. Predicting the impact of underground constructions on adjacent structures as an element of investment risk assessment, in: Special issue: XVI DECGE 2018 Proceedings of the 16th Danube-European Conference on Geotechnical Engin­ eering, ce/papers, vol. 2, Issue 2-3: 281–286. Ernst&Sohn a Wiley Brand. Fine Ltd. 2019. GEO5 User’s manual. Fine Ltd. Prague. Haibin H., Hao C., Jubo Z. 2014. The influence of founda­ tion excavation on the existing metro tunnel in compli­ cated environment, EJGE 19:3377–3385.

Łukasik, S. & Godlewski, T. et al. 2014. Technical require­ ments for designed and constructed investments, which could influence existing metro structures. Warsaw: Building Research Institiute. In polish. Kadlicek T., Janda T., Šejnoha M. 2016. Applying hypoplastic model for soft soils to the analysis of anchored sheeting wall, Acta Geodyn. Geomater, Vol. 13, 2 (182): 125–136. Mitew-Czajewska, M. 2015. Evaluation of deep excavation impact on surrounding structures a case study. In: C. Madryas, A. Kolonko et al. (eds), Underground Infrastructure of Urban Areas 3: 161–172. CRC Press/ Taylor & Francis Group: Balkema. Mitew-Czajewska, M. 2018. Parametric study of deep excavation in clays, Bulletin of the Polish Academy of Sciences, Technical Sciences, Vol. 66, No. 5: 747–754. Mitew-Czajewska, M. 2019. Displacements of structures in the vicinity of deep excavation, Archives of Civil and Mechanical Engineering, Vol. 19, 2:547–556. Phien-wej N., Humza M., Zaw Aye Z. 2012. Numerical modeling of diaphragm wall behavior in Bangkok soil using hardening soil model, in: Viggiani (Ed), Geotech­ nical Aspects of Underground Construction in Soft Ground: 715–722. Taylor & Francis Group, London. Romani E., Sorge R., Guiducci G., Lucarelli A., Furlani G. 2012. Deep excavation of Malatesta Station in Rome: Design, construction and measures, in: Viggiani (Ed), Geotechnical Aspects of Underground Construction in Soft Ground: 301–308. Taylor & Francis Group, London. Shafiqu Q.S.M., Al-Ameri A.A.S. 2018. Effect of deep sup­ ported excavation on the adjacent deep foundation, in: Abu-Farsakh M., Alshibli K., Puppala A. (eds), Advances in Analysis and Design of Deep Foundations: 265–287. Springer, Cham. Superczyńska, M. & Józefiak, K. & Zbiciak, A. 2016. Numerical analysis of diaphragm wall model executed in Poznań clay formation applying selected FEM codes. Archives of Civil Engineering Vol. LXII, 3: 207–224. Siemińska-Lewandowska, A. et al. 2019. 3D numerical analysis of the impact of three deep excavations on the existing railway tunnel in Warsaw. Warsaw: Internal document. Institute of Roads and Bridges Warsaw Uni­ versity of Technology. In polish. Totsev A.E. 2012. Deep excavation in Bulgaria – compari­ son of measured and computed performance, in: Vig­ giani (Ed), Geotechnical Aspects of Underground Construction in Soft Ground: 807–812. Taylor & Fran­ cis Group, London. Yu X., Jia B. 2012. Analysis of excavating foundation pit to nearby bridge foundation, Procedia Earth and Planet­ ary Science 5:102–106.

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Simplified modelling of the transient response of underground structures due to dynamic loads generated from underground tunnels L. Pelecanos University of Bath, UK

K. Senthil & S. Rupali Dr BR Ambedkar NIT Jalandhar, India

ABSTRACT: Underground tunnels are widely used nowadays in major cities resulting in a significantly con­ gested underground space. In many cases they are located close to underground structures, such as foundation piles, retaining walls etc. and civil engineers need to assess the interaction between tunnels and such structures. Although static tunnel-structure interaction has been widely studied over the past decades, there are still aspects of dynamic tunnel-structure interaction that still need to be clarified. Such dynamic interaction may arise in cases of vibration from high-speed railways close to buildings or in cases of explosions in tunnels. Both of these two cases involve transient dynamic loads but they differ in their frequency content, duration and amplitude. This study aims to assess the transient dynamic response of underground structure due to dynamic loads gener­ ated in underground tunnels. A simplified two-dimensional plane-strain finite element model, with a simple tunnel and buried wall geometry, is developed using a dynamic time-domain formulation. Two different input dynamic loading regimes are considered and the dynamic response (accelerations and displacements) of the top of the structure is evaluated. Different aspects of the model are discussed, including the effect of the modelled dynamic boundary conditions and the most critical time of the structural response.

1 INTRODUCTION Underground tunnels in congested cities are built close to other existing infrastructure, such as piles, retaining walls, buildings, utility pipes etc. Static tunnel-pile interaction is nowadays fairly wellestablished (Loganathan & Poulos, 1998; Nip & Pelecanos, 2019). On the other hand, dynamic loads generated in these tunnels may propagate rapidly and affect these kinds of infrastructure and probably cause unfavourable distress. Different approaches exist that can analyse the dynamic response of these structures to dynamic loads from underground tun­ nels. The most commonly used method is the Dynamic Finite Element (FE) method, which may be employed in different ways, such as full 3D ana­ lysis, simplified 2D plane-strain models (Senthil et al., 2017), or even coupled FE-Boundary Element approaches. The dynamic FE method has been employed over the years in different dynamic load problems, such as impact (Senthil et al., 2013, 2016, 2018), earthquakes (Pelecanos et al., 2013, 2015, 2016, 2018) and train vibrations (Coulier et al., 2013). This paper examines the case of a simplified 2D plane strain analysis of the dynamic response of an

underground wall structure due to dynamic loads gen­ erated in underground tunnels. Dynamic time-history analysis is performed for two different kinds of dynamic loads, i.e. harmonic train vibrations and tran­ sient blast load. Moreover, issues related to the bound­ ary conditions of the model are discussed and their effect on the predicted dynamic wall behaviour is evaluated. 2 NUMERICAL MODEL 2.1

Geometry

The numerical model considered in this work is a 2D plane-strain FE model and the FE mesh is shown in Figure 1. The tunnel is a 2x2m square-boxed tunnel placed at 10m below the ground surface, whereas the wall is 4m deep and placed at 10m away from the tunnel. All materials are assumed to be linear elastic. The material parameters adopted are as follows: the soil has a shear modulus, G=200MPa, Poisson’s ratio ν=0.3 and unit weight, γ=20 kN/m3, whereas the wall has a Young’s modulus, E=30000MPa, Pois­ son’s ratio ν=0.2 and unit weight, γ=24 kN/m3.

DOI: 10.1201/9780429321559-82

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The external boundary conditions (BCs) of the bottom and lateral side boundaries were varied in order to examine the effect of the truncation bound­ aries. Three cases were considered and their results are compared in the following sections: • Case 01: all boundaries have the viscous BC (i.e. horizontal and vertical linear dashpots) • Case 02: all boundaries have full fixity (i.e. zero horizontal and vertical displacements) • Case 03: the bottom boundary is fixed, whereas the lateral boundaries are viscous.

Figure 1. Geometry of the considered FE model.

2.2

To specify the Standard Viscous BC (Lysmer & Kuhlemeyer, 1969), a series of linear horizontal and vertical dashpots are used, i.e. normal (CN ) and tan­ gential (CT ) to the external boundary. Their viscos­ ities, where added in the damping matrix, ½C], in the global force equilibrium equation, and they are spe­ cified by the following equations (Potts & Zdravko­ vić, 1999, 2001; Kontoe, 2006):

Formulation

The problem is of a dynamic nature and therefore global force equilibrium is satisfied by the following equation (Chopra, 1995; Cook, 2007):

Where, ½ M ], ½C], ½K ], are the mass, damping and stiffness matrices, whereas f ug, fu_g, fug and fF g are the global acceleration, velocity, displacement and external force vectors. An implicit dynamic time-domain formulation using the constant average acceleration version (β ¼ 1=4; γ ¼ 1=2) of the Newmark-β (Newmark, 1959) method is employed for the solution of the previously-mentioned global force equilibrium, for each increment, i, adopting a small time step, Δt:

Where, A is the boundary area of influence of the dashpot, whereas, ρ, Vs and Vp are the material dens­ ity and shear and p- wave velocities of the soil medium material, respectively. 3 APPLICATIONS Once the model is developed and setup, two cases of dynamic loads of different nature are considered. The first is a harmonic load from train traffic of monochromatic periodic nature, whereas the second is a transient blast load from an explosion. 3.1

The dynamic load is applied as nodal forces at the internal nodes of the tunnel. In the case of harmonic train loads, two vertical (downward) point forces are applied at two nodes at the bottom of the tunnel, whereas in the case of blast load, vertical and hori­ zontal (outward from the tunnel) point forces are applied at all the nodes of the tunnel (Figure 2).

Figure 2. Application of dynamic loads: (a) train harmonic axle load, (b) transient blast explosion load.

Dynamic harmonic load from train traffic

In this case, the input load is applied as a pair of ver­ tical point forces, F(t), at the base floor of the tunnel, which are harmonic and described by: 5Þ Where, F0 ¼ 0:1MN and ω ¼ 63 rad=sec are the amplitude and circular frequency of the applied force. The horizontal, ax , and vertical accelerations, ay , along with the horizontal, ux , and vertical displace­ ments, uy , are monitored at the top of the wall. These values represented the “input” vibrations from the ground on a superstructure on top of the ground. Figure 3 shows the predicted accelerations and dis­ placements at the top of the wall for the 3 cases of BCs considered and discussed in section 2.2. It is shown that, in general, case 01 (black line), with the all

628

viscous BC (that represents semi-infinite compliant soil) shows consistently the smallest fluctuation of values of both acceleration and displacement. This is expected, as in this case the dynamic waves propagate out of the problem in all directions and are “absorbed” by the viscous BC, as if they were propagating to infin­ ity in a real problem. In particular, when this is compared with case 03 (red line), that is of bottom fixity (that represents the presence of an infinitely-stiff bedrock) one may observe that there are larger values of acceleration and displacement fluctuations which are generated due to reflected waves from the bottom fixed bound­ ary (bedrock) back to the problem. This is more pro­ nounced in the figures of the vertical accelerations and displacements. 3.2

Transient blast load from an explosion

In this case the input load is applied as concentrated point forces on all the internal nodes of the square tunnel towards the outer direction (i.e. towards the soil). The point forces are calculated from the expected blast normal pressure time-history, pðtÞ, which is defined by the following equation (Fried­ lander, 1946) and shown graphically in Figure 4.

Where, the following parameter values are tA ¼ 0:001 sec, adopted: pso ¼ 100 kPa, t0 ¼ 0:05 sec, β ¼ 1:2, according to Friedlander (1962). Again, the horizontal, ax , and vertical accelerations, ay , along with the horizontal, ux , and vertical displace­ ments, uy , are monitored at the top of the wall. These values represented the “input” vibrations from the ground on a superstructure on top of the ground.

Figure 3. Harmonic load application: (a) horizontal accel­ eration, (b) vertical acceleration, (c) horizontal displace­ ment, (d) vertical displacement.

Figure 4. Applied blast load time history.

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Figure 5 shows the predicted accelerations and displacements at the top of the wall for the 3 cases of BCs considered and discussed in section 2.2. It is shown that, in contrast to the previous example, there is no distinct and clear pattern of differences between the 3 different BCs con­ sidered. More specifically, there is no clear evi­ dence that case 01 (complete viscous BC, that is a semi-infinite soft soil) results in smaller fluctuations. However, it is interesting to note that, in this case, the most critical time is in the short term (t 1/50

< 10 10 – 50 50 – 75 > 75

negligible slight moderate high

in Table 1. As a result, so-called impact categories are derived for each building. 2.5 Comments on building vulnerability and risk mapping The impact categories above provide valuable informa­ tion to evaluate the risk of building damage due to groundworks. However, details about the building vul­ nerability, which is characterised by different factors including the building geometry, the foundation scheme, the material and the condition of the building, likely influence these predictions. Table 2 shows a qualitative assessment of building vulnerability according to Dzegniuk et al. (1997). This point scoring system was developed to assess building vulnerability in mining areas and is currently refined for the building stock in Norway. After validation, this qualitative assessment of building vulnerability will be incorpor­ ated into a more accurate and reliable risk mapping. Further details of this methodology are provided else­ where (Piciullo et al., 2020.). Table 2. Building vulnerability assessment according to Dzegniuk et al. (1997). The number of points for each par­ ameter are shown in parenthesis. Building length in m: ≤10 (4); 11-15 (7); 16-20 (11); 21-25 (16); 26-30 (22); 31-35 (29); 36-40 (37); >40 (42) Building solid shape: Regular, compact (0); Little dismem­ bered (3); Well dismembered (6); Regular, vast (6); Dis­ membered, vast (8) Building foundation: On flat level, buildings with or with­ out basement (0); On uneven elevation, surface (3); On uneven elevation, surface with partial basement (6); As above but with a passage gate (8) Building ground foundation: Rigid (0); Low-Rigid (4); Non-Rigid (8) Existing protection for mining operation effects: Bolting (0); Fractural bolting (4); None (6) Technical condition of the building: Good (0); Average (4); Bad (12)

Figure 4. Building deformation parameters adopted for impact assessment: (a) maximum settlement and (b) slope.

Total score: Vulnerability class

634

≥48 C1

37-47 C2

28-36 C3

21-27 C4

≤20

C5

3 CASE STUDY This section describes the adopted case study close to Oslo, Norway, which was used to apply the impact assessment procedure introduced above. 3.1 Project overview Between 2000 and 2004, the Norwegian National Rail Administration, Bane NOR, built a new double track railway between Sandvika and Asker. Figure 5 visualises the case study area during construction. The railway line connects to existing railway tracks (see top of Figure 5). For the first 700 m a cut-and-cover tunnel was built (Phase 1 and 2 in Figure 5). Subsequently, a rock tunnel was constructed. The final structure was either directly founded on bedrock or on steel core piles that were drilled into bedrock. Figure 5 shows that various buildings are near the eastern part of the railway line (i.e. Phase 2). 3.2 Ground conditions The ground conditions and depth to bedrock varied considerably along the cut-and-cover section. The soil profile consists of a thin dry crust, underlain by deposits of soft sensitive clay. Beneath the clay not­ able moraine layer exists above bedrock. Towards the West, the thickness of the moraine layers

decreased; notable moraine layers with varying thickness were, however, still registered. The sensitive clay was characterised as normally consolidated and as a typical quick clay. Undrained shear strength, cu, values between 5 – 15 kPa were measured. For the encountered clay, the following deformation properties were derived: tangent moduls M = 5 MPa, modulus number m = 15 – 22 and a consolidation coefficient cv = 2 – 10 m2/year. A detailed discussion of strength and deformation properties of Norwegian clays is provided elsewhere (Karlsrud & Hernandez-Martinez 2013). 3.3 Excavation support and construction procedure The excavation depth varied between 9 to 13.5 m below the ground surface. Sheet pile walls at a horizontal distance of about 17 m were used as support. Tie-back anchors were drilled into bedrock and used to laterally support the sheet pile wall. A maximum of four anchor levels, depending on the excavation depth, were installed. However, during the construction of Phase 1 extensive reduction of the original ground water surface was recorded. In Phase 2, the final lateral support level was carried out with internal struts, as can be seen in Figure 6. 3.4 Field instrumentation Pore pressure data at bedrock level was obtained at 12 locations surrounding the deep excavation. The initial ground water surface was approximately 2 m below the surface. Hydrostatic pore water pres­ sure was observed. Levelling bolts were installed on buildings that were within 100 m of the deep excavation. This resulted in about 100 buildings that were monitored. Biweekly measurements were carried out which was reduced to a single reading per month as settlement rates decreased.

Figure 5. Case study area (adopted from Karlsrud et al. 2015).

Figure 6. Cross-section for Phase 2.

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3.5

Observations

A maximum pore pressure reduction of approxi­ mately 10 m in pressure head was monitored throughout the construction works. This decrease in pore pressure was caused by (i) leakage between the casing of the tie-back anchors and the sheet pile wall, (ii) leakage along the outside of the casing for steel core piles and (iii) leakage at the toe of the sheet pile wall. Extensive grouting was carried out to mitigate leakage caused by the tie-back anchors and the steel core piles. A toe beam was cast to stop the leakage at the toe of the sheet pile wall. In addition, infiltration wells were installed to increase the pore pressure levels. The observed effects resulted in significant settle­ ments surrounding the deep excavation. Throughout the construction period a maximum vertical displace­ ment of approximately 135 mm was measured. These settlements increased with time due to consolidation and had a considerable impact on the surrounding buildings. Braaten et al. (2004) provide a detailed discussion of this case study including a likely explanation of the observed pore pressure reduction and excava­ tion-induced settlements. Additionally, Brendbekken et al. (2004) summarises the geotechnical design.

Figure 7. Building polygons and deep excavation polygon (blue) in case study area.

4 IMPACT ASSESSMENT The impact assessment procedure described in Sec­ tion 2 was applied to the case study discussed above. In the following, the required inputs are first dis­ cussed, after which short- and long-term settlement predictions are compared to field measurements. 4.1

Inputs

As mentioned above, the described impact assessment methodology has been implemented in the commer­ cial software ArcGIS Pro. The associated code was written in Python and will be made publicly available after further validation. The required inputs to per­ form an impact assessment are the shapefiles of the building polygons, a user defined polygon that defines the location of the excavation, the depth to bedrock model (Section 2.1), the excavation depth and con­ solidation properties of the clay layer. Figure 7 shows the case study area in ArcGIS including both the building and deep excavation polygons. The crucial input parameters related to the short-term and long­ term settlement calculations are discussed below. 4.2

Short-term predictions

The building and deep excavation polygons (Figure 7) were used to compute the distance between the sheet pile wall and each corner point of a building polygon. In addition, the input of the excavation depth was required to employ the solid design curve shown in

Figure 7. This design line was applied, because it rea­ sonably well describes settlements for excavations sup­ ported with internal struts and tie-back anchors (Figure 2). For this case study, an excavation depth of 13.5 m was adopted. This procedure allowed one to determine the short-term settlements for each building corner point. Figure 8 shows the results of the short-term ana­ lysis. For each corner point and building wall, the computed settlements or slope values are compared with the adopted impact categories (Table 1). Figure 8 depictures symbols used to visualise different impact category options for a single building. The overall short-term impact category can either be based on the maximum settlement of a corner point or the max­ imum slope of a building wall. This can be defined by the user. 4.3

Long-term predictions

Having discussed the short-term predictions, further user input is required to compute the long-term settlements. As described above, a combination of the empirical observations of pore pressure reduction caused by excavation works (Figure 3), the depth to bedrock model and the modulus concept proposed by Janbu (1963) was adopted. Table 3 lists the input values to calculate the consolidation settlements, which is again conducted for each building corner point. Note that the used soil properties are in line

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Figure 8. Short-term impact predictions. For symbols and colour code refer to Figure 9.

Figure 9. Visualisation of different options of impact cat­ egories based on maximum settlement, slope and overall impact category. The overall impact category shown is based on the maximum slope of a building wall.

Table 3. User defined input values for calculating long-term displacements. Variable

Unit

Value

Dry crust thickness Groundwater depth Groundwater reduction Clay specific weight Overconsolidation ratio Janbu’s reference pressure Janbu’s modulus number

m m m kN/m3

1 2 11.5 18 1 0 18.5

kPa

Figure 10. Total impact prediction and depth to bedrock. For symbols and colour code refer to Figure 9. The shown overall impact category is based on the slope.

with the soil investigation data of the case study area (Section 3.2). The computed long-term settlements are added to the calculated short-term settlements to obtain the total settlements at each building corner point. Figure 10 shows the predicted impact categories for the total settlements. Similar to above, the overall impact cat­ egory in Figure 10 is based on the maximum slope value of a building. The same colour code as visualised in Figure 9 is adopted. As expected, considering also consolidation settlements often increases the impact category compared to the short-term predictions (com­ pare Figure 8 with Figure 10). This is particularly obvious for the buildings south to the excavation where the depth to bedrock is considerable (i.e. ≥ 20 m). 5 EVALUATION OF IMPACT ASSESSMENT The historical settlement data acquired for the dis­ cussed case study allowed an evaluation of the con­ ducted impact assessment. Precise levelling data of 26 buildings next to Phase 2 of the excavation (Figure 2) were available. The monitoring was con­ ducted in the period between October 2000 to Octo­ ber 2017. The final levelling data was obtained approximately 13 years after the construction works finished. Figure 11 employs the maximum settlement of a building corner to obtain impact categories. A comparison between Figure 10 and Figure 11 reveals that for the same buildings the impact cat­ egories based on the slope are often lower than the

637

Figure 13. Prediction accuracy index (PAI) for impact cat­ egories based on the maximum slope of a building.

impact assessment methodology provides often accurate predictions. For 54 % of the analysed buildings the predicted settlement impact classes were identical to the measured ones. Conservative predictions by one impact category (i.e. PAI = 1) were obtained for 19 % of the structures. 15% of the buildings had a PAI value of 2, while 12% of the buildings were overpredicted by three impact categories (i.e. PAI = 3). Underpredictions were not obtained. Figure 13 compares the predicted and actual impact categories using the slope of a building wall as the assessment criteria. Again, the PAI concept is employed. The predictions were often of conservative nature, but PAI values of 3 were not obtained. This indicates that degree of overprediction is likely lower when adopting the slope as decision criteria (compare Figure 12 and Figure 13). 38% of the predicted impact categor­ ies employing the slope as the decision criteria agreed with the observations, whereas for 12% of the buildings the impact category was underpre­ dicted by one impact category (PAI = -1). PAI values of 1 and 2 were obtained for 38% and 12% of the buildings, respectively.

Figure 11. Comparison of predicted and observed impact categories based on maximum settlement of a building. The colour of a building area indicates the prediction, while the colour of a circle shows the actual impact category based on the maximum measured settlement.

Figure 12. Prediction accuracy index (PAI) for impact cat­ egories based on maximum vertical settlements.

6 CONCLUSIONS

predictions based on maximum settlements. In add­ ition, Figure 11 shows a comparison between pre­ dicted and observed impact categories. Figure 12 provides a more direct comparison using the prediction accuracy index (PAI) as defined by Schuster et al. (2009). The PAI for a building is calculated by taking the difference between the predicted impact category and the observed impact category. For example, if the pre­ dicted impact category is 2 and the observed impact category is 1, the respective PAI is 1. This indicates that negative PAI values are derived for unconser­ vative predictions and positive PAI values for overpredictions. From Figure 12 it is apparent that this

An impact assessment framework that provides the first step towards a reliable procedure to map the risk of building damage due to excavation-induced dis­ placements was described. This approach considers both short-term and long-term soil displacements by adopting empirical settlement observations and obser­ vations of pore pressure reduction. Additionally, a representative soil stratification model was derived using soil investigation data and applied to compute consolidation settlements. For each building, impact categories were then derived for short-term settlements and a combination of short-term and long-term settle­ ments (i.e. total settlements). It was shown that the pre­ diction based on maximum building settlement result in higher impact categories compared to using the slope of building walls. Comparisons were presented

638

between impact predictions and impact categories based on monitoring data of a case study. This data indicates that the impact assessment framework results in accurate predictions with the required conservative­ ness. Further studies could refine these predictions by incorporating more details of the excavation such as corner effects. So far, the focus of this prediction tool has been on computing maximum settlements and slope values. Compared to building distortion parameters such as the deflection ratio, the angu­ lar distortion and the horizontal strain, the max­ imum settlement and slope cannot be directly linked to building strains and damage. While this is a limitation, the used parameters often enable an evaluation with historical case studies. This is more complicated with building distortion param­ eters, which require more detailed monitoring data to be derived. Such detailed monitoring data is often not available in practice. After validation of this impact assessment produce with a wider set of case studies, the adopted methodology can be further developed to compute also building distortions. As briefly outlined in this contribution, further work is currently carried out to incorporate a building vulnerability assessment. This work aims to consider geometrical and structural details about buildings affected by excavation-induced subsid­ ence. It is envisioned that coupling the presented impact assessment with a representative building vulnerability assessment will increase the accuracy and reliability of future predictions.

ACKNOWLEDGMENTS The authors are thankful to the Norwegian Research Council and the partners of the research projects BegrensSkade I and BegrensSkade II/Remedy for funding this research.

REFERENCES Braaten, A., Baardvik, G. & Vik, A., 2004. Observed effects on the pore pressure caused by extensive founda­ tion work and deep excavations in clay. In EDS XIV Nordisk Geoteknikermøte, Ystad, Sweden: H119–H132 (in Norwegian). Brendbekken, G., Baardvik, G., Braaten, A. & Vik, A., 2004. Geotechnical design of deep excavation in soft quick clay with instrumentation of sheet pile wall and struts to control its functionality. In EDS XIV Nordisk Geoteknikermøte, Ystad, Sweden: A103–A144 (in Norwegian). Burland, J.B., Mair, R.J., Standing, J.R., 2004. Ground per­ formance and building response due to tunnelling. In Jardine, Potts & Higgins (eds) Advances in Geotech­ nical Engineering: 291–342. London: Thomas Telford.

Clarke, J.A. & Laefer, D.F., 2014. Evaluation of risk assessment procedures for buildings adjacent to tunnelling works. Tunnelling and Underground Space Tech­ nology, 40:333–342. Devriendt, M. D., Palmer, E., Hill, R. & Lazarus, D., 2013. Historic and non-historic building impact assessment methodology for major tunnelling infrastructure pro­ jects. In Bilotta, Flora, Lirer & Viggiani (eds), Geotech­ nical Engineering for the Preservation of Monuments and Historic Sites: 335–341. CRC Press. Farrell, R.P. & Mair, R.J., 2012. Centrifuge modelling of the response of buildings to tunnelling. In Viggiani (eds) Geotechnical aspects of underground construction in soft ground: 361–370. Rome: CRC Press. Janbu, N., 1963. Soil compressibility as determined by oed­ ometer and triaxial tests. In EDS Proceedings of the 3rd European Conference Soil Mechanics: (1) 19–25, Wies­ baden, Germany. Karlsrud, K. & Hernandez-Martinez, F.G., 2013. Strength and deformation properties of Norwegian clays from laboratory tests on high-quality block samples. Canad­ ian Geotechnical Journal, 50:1273–1293. Karlsrud, K., Langford, J., Lande, E.J. & Baardvik, G., 2015. Evaluation of damage and deformations caused by drilling of tie-back anchors and bored piles for deep excavations. BegrensSkade I report 1+2.4 (in Norwegian). Langford, J., Karlsrud, K., Lande, E.J., Eknes, A.Ø. and Engen, A., 2015. Causes of unexpectedly large settle­ ments induced by deep excavations in soft clay. In M. G. Winter, D.M. Smith, P.J.L. Eldred & D.G. Toll (eds), Proceedings of the XVI ECSMGE Geotechnical Engin­ eering for Infrastructure and Development: 1115–1120, ICE publishing. Langford, J., Karlsrud, K., Lande, E.J., Baardvik, G. & Engen, A., 2016. BegrensSkade – Limitation of damage caused by groundworks. In Grundläggningsdagen GD2016. Stockholm, Sweden. Mair, R.J., Taylor, R.N. & Burland, J.B., 1996. Prediction of ground movements and assessment of risk of building damage due to bored tunnelling. In Mair & Taylor (eds), Fourth International Symposium on Geotechnical Aspects of on Underground Construction in Soft Ground: 713–718. London: Balkema. Mason, J. & Hansray, M., 2005. Crossrail Line 1, Assessment of Settlement Impacts on the Built Heritage, Vol. 1 of 3, Technical Report, Alan Baxter & Associates. Matheron, G., 1963. Principles of geostatistics. Economic geology, 58(8):1246–1266. Piciullo L., Ritter S., Kydland Lysdahl A.O., Kahlström M., Langford J. & Nadim F., 2020. Assessment of building damage due to excavation-induced displacements: the GIBV method, Tunnelling and Underground Space Tech­ nology, accepted for publication. Rankin, W. J., 1988. Ground movements resulting from urban tunnelling: predictions and effects. Geological Society, London, Engineering Geology Special Publica­ tions, 5(1):79–92. Schuster, M., Kung, G.T.C., Juang, C.H. & Hashash, Y.M., 2009. Simplified model for evaluating damage potential of buildings adjacent to a braced excavation. Journal of geotechnical and geoenvironmental engineering, 135 (12):1823–1835.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

A new approach for compensation grouting in highly permeable gravel

M. Sailer & J. Fillibeck Center for Geotechnics, Technical University of Munich, Germany

S. Geuder Southern Bavarian Motorway Office, Germany

ABSTRACT: The Oberau Tunnel is located in southern Bavaria near Garmisch-Partenkirchen. A sector of the tunnel crosses the highly permeable gravel of the Giessenbach Valley. The shotcrete driving of the two tunnel tubes is carried out in this area with only a small overburden under settlement-sensitive buildings. In order to reduce the expected settlements of several centimetres to a tolerable level, compensation grouting was carried out underneath the buildings. This paper deals with the problem of compensation grouting in coarse-grained soils and presents a concept of compensation grouting, which was successfully applied in the highly permeable gravel of the Giessenbach Valley. In addition to the findings from the compensation grout­ ing on a test field, the uplift of a 6000 m² industrial hall is shown. The successful execution of the compensa­ tion grouting could be demonstrated on the basis of geotechnical deformation measurements in the subsurface, real-time measurements on the structure and the efficiency of the compensation grouting derived from them.

1 INTRODUCTION The A95 motorway from Munich to GarmischPartenkirchen ends approx. 3 km before the village Oberau and leads to the B2 federal highway. With an average traffic volume of approx. 26,000 vehicles daily, on weekends and holidays of up to 45,000 vehicles, there is considerable tailback on the main road through Oberau. The extension of the federal highway and a tunnel through the village will enable the traffic volume to be handled in the future. For the tunnel drive the method of shotcrete tunnelling was used. In addition to two sections in solid rock, the route crosses the loose soils of the Giessen­ bach Valley over a length of approx. 530 m. In this section, the tunnel passes beneath several buildings of an industrial park. The foundations of the existing buildings are located here only approx. 11 m above the top of the tunnel. Due to the predicted settlements of approx. 6 cm, a considerable damage of the settle­ ment sensitive buildings had to be assumed. In order to ensure a damage-free tunnelling underneath the buildings, compensation grouting was carried out. Compensation grouting is a special grouting tech­ nique which can be used to uplift buildings or to pro­ tect them against settlements during tunnelling (Chambosse & Otterbein 2001). Further general information and case studies are given e.g. in Burland et al. 2001 or Essler et al. 2000.

A special challenge in Oberau was the execu­ tion of the compensation grouting on an area of approx. 6000 m² in the highly permeable gravel to uplift an industrial hall. In the context of this paper the associated problem as well as the suc­ cessfully applied compensation grouting concept will be discussed. 2 GEOLOGICAL AND HYDROGEOLOGICAL CONDITIONS IN THE GIESSENBACH VALLEY In the course of the tunnel planning extensive ground investigations were carried out. In the area of the Giessenbach Valley, 19 core drillings and 14 dynamic penetration tests were carried out. In 17 of those boreholes groundwater gauging stations were installed. The resulting geological longitudinal sec­ tion is shown in Figure 1. The tunnelling section in the area of the Giessen­ bach Valley consists of alluvially deposited layers of loose soil. A gravel layer was encountered from the ground surface to a depth of approx. -50 m. Below this are the glacial deposits of a ground moraine. The gravel layer can be divided into an upper and a lower gravel layer. The boundary of this layer was largely explored at a depth of approx. 15 m to 22 m below the ground surface. The tunnel runs mainly in the upper

DOI: 10.1201/9780429321559-84

640

Figure 1. Geological longitudinal section in the Giessenbach Valley.

gravel, only in some areas the excavation of the bench and the invert is executed in the lower gravel. The upper gravel has a comparatively higher pro­ portion of stone and fine grains and is heavily inter­ spersed with stones and blocks in sections. These were found with edge lengths of up to 50 cm. In some cases also small fine-grained intermediate layers were found. The lower gravel consists mainly of sandy, weakly stony gravel. Locally both loose gravel layers without any content of fine grain, as well as thin sand and silt layers were explored. According to the results of the borehole dynamic probing, the upper and the lower gravel predomin­ antly have a medium to very dense packing. Accord­ ing to DIN 18130-1, the upper gravel with a k-value of approx. 1.5 x 10-4 m/s to 2.0 x 10-3 m/s can be classified as highly permeable. The water level is usually about 2 m below the tunnel floor. In spring, however, when the snow melts, the water level can rise to the level of the tunnel axis. 3 SHOTCRETE TUNNELLING METHOD IN THE GIESSENBACH VALLEY

injection tube spiles with a length of 6 m to 8 m were used as advance supporting measure (Figure 2) (Filli­ beck et al. 2017). 4 COMPENSATION GROUTING IN THE GIESSENBACH VALLEY 4.1 General explanations on the method of compensation grouting Settlements on buildings can be corrected with the method of compensation grouting by creating a local heave of the ground. For this purpose, several rows of sleeve pipes (also known as tubes á manchettes/ TAMs) are installed under the building. The TAMs have injection openings at a distance of 0.50 m, which are covered by a rubber sleeve. Via the numer­ ous sleeves, a cement-containing grout is injected into the soil (Figure 3, top). The high injection pres­ sure leads to a displacement of the grains and to the development of grout filled fractures in the soil (frac­ ture grouting), causing the building to heave. Due to

The two tunnel tubes (the east and the west tunnel) were excavated by means of top heading at an axial distance of 30 m. The east tunnel followed the west tunnel at a distance of approx. 100 m. The crosssection of each tunnel was approx. 112 m². In order to ensure a safe tunnelling in the area of the existing buildings, face anchors and up to 72 self-drilling

Figure 2. The heading concept and the support systems under the buildings.

Figure 3. Principle of compensation grouting (top) and injection technique (bottom).

641

the large number of installed sleeves and the multiple injection of small quantities of grout, the releveling of the building can be very specifically adapted to the requirements. The deformations occurring on the structure are monitored by real-time measurements by means of a liquid levelling system. In the course of the installation, the TAMs are inserted in the cased boreholes. The casing is removed and the annular gap around the TAMs is filled with a sleeve grout (Figure 3, bottom). With the aid of an injection pipe with double packers, the grout is pressed into the soil. Once the specified quantity of grout has been reached (approx. 10 to 80 l) (Raabe & Stockhammer 1995), the injection pipe is relocated to the next sleeve. In each injec­ tion pass the injections are applied on every second sleeve. In the following injection pass the remain­ ing sleeves are applied. Thereby an injection round is finished. After all planned sleeves have been applied, the process can be repeated and the next injection round begins. How effective the grout is leading to a correspond­ ing heave at the ground surface can be assessed on the basis of the efficiency of the compensation grouting. The efficiency of the compensation grouting describes the ratio of the volume of heave generated to the amount of grout injected. 4.2

Problem definition

The execution of the compensation grouting is technic­ ally challenging and cost-intensive. The crucial cost factors are the installation of the TAMs, the required amount of grout and the time for the injections. Com­ pensation grouting in highly permeable gravel is par­ ticularly challenging because, in contrast to finegrained soils, the grout can flow in the large pores without any heave effect. Therefore, in earlier projects in coarse-grained soils, the pores of the soil were filled in advance by additional pore injections at the TAMs of the compensation grouting or from additional TAMs above and below the level of the compensation grouting. This results in a considerable cost increase compared to the compensation grouting in fine-grained soils. In order to reduce the costs, a different compensa­ tion grouting concept was pursued in the highly per­ meable gravel of the Giessenbach Valley: By injecting a grout as viscous as possible, an uncon­ trolled and inefficient spreading of the grout into the gravel should be prevented. Additional levels of TAMs or a scheduled pore injection prior to the com­ pensation grouting were not intended.

Figure 4. Overview of the test field.

Similar to the compensation grouting under the industrial hall, the TAMs were installed at a depth of 7.5 m below the ground surface. For this purpose, six cased boreholes, each 18 m long, were drilled from a shaft with a gradient of 3%. In each borehole a row of steel TAMs (Ø 60.3 x 3.60 mm, 0.5 m sleeve spacing) was installed. The TAMs had a maximum distance of approx. 2.15 m as planned. Three different variants of grout mixtures were tested in the test field. The aim was to use a grout as viscous as possible that is rich in solids, easy to pro­ cess, does not significantly change its properties in the storage tank within a few hours and is pumpable. A finished product of cement, fillers, latent-hydraulic and puzzolanic substances was used for all mixtures. The properties of the finally selected grout mixture (variant 3) are summarized in Table 1. In contrast to the pore injections, relatively small grout quantities of 30 to 45 l per sleeve per injection pass were injected according to the novel grouting concept described here. A total of 15 measuring points of the liquid levelling system were installed to measure the deform­ ations occurring on the ground surface. Due to the deep location of the TAMs, it was expected that the heave generated in the subsoil would not occur to the full extent on the ground surface. However, in order to be able to recognize as early as possible whether the intended heave of the subsoil is achieved by the new grouting concept, the subsoil deformations were additionally measured directly above the TAMs (1 m distance) using a horizontal inclinometer. The deformation measurements after the second and fourth injection of all the sleeves are depict in Figure 5. Already after the second round, a heave of 4 mm was measured with the horizontal inclinometer. This shows that the grout used does not penetrate the pore Table 1.

4.3

Preliminary examinations on the test field

The test field (Figure 4) was used to demonstrate that the novel grouting concept described in Chapter 4.2 was suitable for carrying out the compensation grouting in gravel and to select an appropriate grout mixture.

Properties of the selected grout mixture variant 3.

Grout mixture

W/s Marsh- Yield ratio Density time point kg/l s N/m²

Variant 3

0,7

642

1,60

42

54

Temperature °C 15-22

Figure 5. Measurements of the horizontal inclinometer and the liquid levelling system on the test field.

space without effect, but leads to the desired displace­ ment of the soil at an early stage. After completion of the fourth injection round, the deformation at the hori­ zontal inclinometer increased to up to 18 mm. As expected, the heave at the ground surface was somewhat lower due to the large soil overburden of 7.5 m. Nevertheless, a surface heave of approx. 2 mm and 5 mm was achieved in the second and fourth injection round respectively. In order to investigate whether the necessary heave of the industrial hall to prevent settlement damage could be achieved, additional injection rounds were carried out. As a result, up to 13 injections were exe­ cuted in certain areas. Thus, heaves at the ground sur­ face of approx. 19 mm at the edge and approx. 70 mm in the middle of the test field were achieved. The results of the test field showed that the pro­ posed grouting concept is suitable for the application of compensation grouting in such permeable gravel. With the chosen viscous, grout mixture with high solid content, both a desired early heave and heaves with a sufficient magnitude were achieved.

Compensation grouting was carried out on an area of approx. 6000 m². The level of compensation grout­ ing was arranged in the middle between the tunnel and the footings of the industrial hall. For the installation of the TAMs, starting from four shafts, a total of 7730 m were drilled with individual lengths of up to 53 m, with a gradient of 3% and a maximum centre distance of 2 m. In addition, horizontal inclinometers were installed above the TAMs of the shafts L1, L3 and L4. The hall construction consists mainly of a precast reinforced concrete construction. A total of 75 pillars are located in the settlement area. On the basis of a static-constructive assessment of the permissible settlement differences of the industrial hall, alarm, intervention and limit values were determined for monitoring the building. The resulting tangent inclinations between two adjacent individual foot­ ings of the pillar construction are listed in Table 2. The pillar deformations were automatically recorded every minute by a liquid levelling system and com­ pared with the permissible settlement differences.

4.4 Execution of the compensation grouting under the industrial hall

Table 2. Defined alarm, intervention and limit values.

4.4.1 General information The two tunnel tubes crossed the industrial hall in a depth of 11 m and over a length of approx. 88 m (Figure 6).

Inclination between footings

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Alarm value

Intervention value

Limit value

1/500

1/350

1/250

horizontal inclinometers, the local distribution of the grout quantities and the efficiency of the compensa­ tion grouting will be discussed.

Figure 6. Scheme of horizontal inclinometers and TAMs under the industrial building.

During the compensation grouting, the deformations were also monitored by the construction site personnel. The deformations could be viewed at any time via an online portal. If the permissible values were exceeded, an automatic notification was sent. For stability reasons (stability of the working face and the fresh shotcrete lining), no compensation grouting was carried out in an area of 20 m ahead and behind the working face. Since the top heading had already exceeded the permissible settlement dif­ ferences, it was necessary to compensate a part of the settlements by a preliminary heave of the build­ ing. For this purpose, 2/3 of the predicted settlements of the top heading were applied as a mirror image of the settlement trough (negative settlement trough) as a preliminary heave. The shape of the predicted settlement trough and the scheduled preliminary heave are shown in Figure 7. Analogous to the execution of the top heading with a preliminary heave of the industrial hall, an intermediate heave of the building was necessary prior to the excavation of the bench and the invert. The following evaluations refer to the more demand­ ing execution of the preliminary heave, since com­ paratively larger heaves had to be achieved and the first injection rounds in the highly permeable gravel are classified as more critically. In the following, the findings from the deformation measurements of the

Figure 7. Scheduled preliminary heave and the predicted settlements due to the top heading [mm].

4.4.2 Evaluation of the soil deformations in the vicinity of the compensation grouting Figure 8 depicts the measurements of the horizontal inclinometer installed at shaft L3 at a specific date, the injected amount of grout per square metre and the number of injections per sleeve. Per injection, 30 to 45 l of grout per sleeve were injected at a rate of 10 l/min. The horizontal inclinometer is located approx. 2.5 m above the level of the compensation grouting at shaft L2 and shaft L3. In the first injection round, the injections were carried out at all the sleeves. Already during the second injection round, heaves occurred along the entire section of measurements on the horizontal inclinometer. Although the injected amount of grout and the achieved heave varied to a certain range, on average approx. 75 l/m² were neces­ sary to achieve heaves of 3 to 5 mm (section of meas­ urements 14 to 41 m). To uplift the ground near the shaft (1.5 to 6.5 m), proportional grout quantities were required (approx. 175 l/m² with approx. 11 mm heave). The advantage of the applied grouting concept is illustrated through the attainment of an initial heave from the injection of a relatively small quantity of grout. This is in contrast to the design variation with additional pore injections prior to the compensation grouting, where significant quantities of grout are required before any heave is observed. Assuming an even distribution of the grout, a pore injection of 75 l/ m² would only fill the pores of a 30 cm thick, 1 m² large soil layer (porosity approx. 25 %) without achieving any heave. With such a small thickness of the injected soil, it cannot be excluded that the frac­ tures developing from the subsequent compensation grouting (section 4.1) develop through the entire thin layer of improved soil. Thus, the grout will also pene­ trate into non-injected soil areas and surface level heave greater than that of the applied grouting con­ cepts can not be achieved. 4.4.3 Evaluation of the achieved efficiency of the compensation grouting In order to achieve the scheduled heave at the industrial hall, a total of approx. 1,285,000 l grout were injected over the course of 178 days. Figure 9 on the left shows the area of compensation grouting (approx. 6000 m²) with the amount of grout per quadrant (2.5 m x 2.5 m). Depending on the size of the scheduled heave, the required grout quantities also varied. The efficiency of the compensation grouting was determined in order to compare zones of different heaves within the industrial hall. Not only the injections directly under the isolated foot­ ings lead to an uplift of the corresponding pillar, but also the injections within a certain radius around the isolated footing. By comparing the

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Figure 8. Sleeves and their respective number of injections (top). Measurements of the horizontal inclinometer and grout quantities (bottom).

Figure 9. Total efficiency [%], achieved column uplift [mm] and required grout per quadrant [l] (left). Development of the efficiency (%) (right).

deformations that occurred with the injections car­ ried out during this period, it was possible to deter­ mine the maximum radius around the footing at which an injection point (sleeve) still has an influ­ ence on the uplift of the pillar. It is about 3 m for the majority of the investigated measuring points. All injections within this radius were considered, to determine the efficiency. In order to estimate the achieved volume of heave on the ground surface, only the measured values of

the liquid levelling system at the pillars were avail­ able. Due to the relatively large distance between the pillars (6.25 m x 22.5 m), to simplify matters, it was assumed that within the radius of 3 m the same heave was achieved as at the corresponding pillar. Figure 9 (left) depicts the total efficiency (ratio of total volume of heave to total volume of grout each within the radius) at selected pillars after completion of the preliminary heave. While a total efficiency of approx. 7 % was achieved at the edge of the hall, it

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ation measurements with horizontal inclinometers it could be shown that the first heaves of the soil were already achieved after the injection of small amounts of grout. The alternative design variation including a pore injection was found to be inefficient.

was approx. 15 % in the middle of the hall. Since, in the middle of the hall, the surrounding injections make it more difficult for the grout to flow away, a comparatively higher heave can be achieved there. Furthermore, the efficiency increases with the increasing number of injection rounds. Figure 9 on the right shows the development of the efficiency regard­ ing the grout quantities and the corresponding heave of each injection round. Already after the second injection round, a significantly higher efficiency is achieved in the middle of the hall than at the pillars along the edge. With an increasing number of injection rounds this dif­ ference becomes more and more obvious. While in the last injection round the efficiency in the middle of the hall was approx. 30 to 50 %, the maximum efficiency at the edge of the hall was only approx. 10 to 20 %.

REFERENCES

5 SUMMARY The completion of the Oberau bypass will signifi­ cantly improve traffic conditions along the B2 fed­ eral highway. The construction of the tunnel section in the Giessenbach Valley was particularly demand­ ing. In this area the tunnelling was carried out in highly permeable gravel with an overburden of approx. 11 m to the existing buildings. Due to the compensation grouting, it was pos­ sible to compensate the tunnel induced settlements without causing any damage to the buildings. A grouting concept was developed, with which the otherwise very time consuming and cost-intensive compensation grouting measure in gravel could be greatly reduced. By means of geotechnical deform­

Burland, J. B., Standing, J. R. & Jardine, F. M. 2001. Build­ ing response to tunnelling: case studies from construc­ tion of the Jubilee Line Extension, London. London: CIRIA and Thomas Telford. DIN 18130-1: 1998-05,Bestimmung des Wasserdurchläs­ sigkeitsbeiwerts Teil 1: Laborversuche (DIN 18130-1: 1989). Chambosse, G. & Otterbein, R. 2001: State of the art of compensation grouting in Germany. Proc. 15th Int. Conf. on Soil Mechanics and Foundation Engineering, Istanbul, Turkey: 1511–1514. Balkema: Rotterdam. Essler, R. D., Drooff, E. R. & Falk, E. 2000. Compensation Grouting, Concept, Theory and Practice. Proceedings, Advances in Grouting and Ground Modification. ASCE Geo-Institute Special Publication No. 104, Denver, CO: 1–15. Reston, VA: ASCE. Fillibeck, J., Klinger, A., Geuder, S. & Willberg, U. 2017. Geotechnische Herausforderungen beim Bau des Tun­ nels Oberau. Beitrag zum 22. Symposium Felsmechanik und Tunnelbau. Fachsektionstage Geotechnik der DGGT, Würzburg 2017. Raabe, E.W. & Stockhammer, P. 1995. Einsatz von Soil­ crete und Soilfrac im Tiefbau – Möglichkeiten und Grenzen beider Injektionstechniken. Innovationen in der Geotechnik; Beiträge zum 10. Christian Veder Kollo­ quium. Institute for Soil Mechanics and Geotechnical Engineering, Graz University of Technology: 127–144.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

A case study on the effects of anchor drilling in soft, low sensitive clay and sandy, silty soils T. Sandene & S. Ritter Norwegian Geotechnical Institute, Oslo, Norway

E.J. Lande Department of Civil and Environmental Engineering, Norwegian University of Science and Technology, Trondheim, Norway

ABSTRACT: This paper presents monitoring results from a deep supported excavation in central Oslo, Norway. Sheet pile walls and tensioned tieback anchors drilled into bedrock supported the excavation in soft low sensitive clay to a depth of about 9 m. In two specific areas, relatively large settlements were observed, coinciding with drilling of anchors for the sheet pile wall. The immediate subsidence during drilling was fol­ lowed by a period of ongoing settlements at a lower rate. These ongoing displacements were not expected as pore pressure reduction alone could not explain the rate or extent of displacements measured. Immediate sub­ sidence during drilling was largest in areas with sandy and silty layers and is related to erosion and loss of soil volume due to excessive flushing and remoulding of soil during drilling. The observed mechanisms are characteristic for deep excavations in Norway and the collected data inform further work to reduce ground displacements caused by excavation works.

1 INTRODUCTION

2 OVERVIEW OF CONSTRUCTION PIT AND SURROUNDINGS

This paper focuses on ground displacements observed during the excavation of the construc­ tion pit for the new National museum of Norway in central Oslo. The construction of the museum, which will be open to the public in 2020, is currently taking place at the site of the decommissioned westbound train station in Oslo (“Vestbanen”). The old station building was trans-formed into the Nobel Peace Center museum and exhibition centre. The groundworks for the new National museum included to remove the remaining train station, to remove and relocate other subsurface infrastructure, to establish construction pits for the basement and to construct the foundations for the new structure. The main objective of this paper is to address excess ground displacements caused by drilling of cas­ ings for strand anchors to bedrock. This paper first pro­ vides an overview of the presented case study, after which widely observed settlement causes of deep exca­ vations in Oslo are introduced. Then, the monitoring programme and conduct-ed excavation works are described before a detailed discussion of the observed displacements is provided. Finally, conclusions are drawn.

DOI: 10.1201/9780429321559-85

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Figure 1 shows an overview of the construction site including the sheet pile walls (bold lines) and varying excavation levels. The sheet pile walls were installed very close to neighbouring structures and existing infrastructure to maximize the available area. The building ground is restricted by Dokkveien to the southwest, the Nobel Peace Center to the east, Dron­ ning Mauds gate (with underlying parking garage) to the northeast and access roads to the E18 highway to the northwest. In addition, Figure 1 shows the location of cross sections A-A and B-B which will be discussed in detail below and indicates the location of piezo­ meters (PZ) that were installed near the excavation pit. Prior to construction the area was relatively flat at an elevation of about +2.3 to +2.5 (Oslo local eleva­ tion reference). Dokkveien inclines from the south towards northwest up to about +6.5 where it crosses over the tunnel entrance for the E18 ramps. The ramps incline from +1.2 near the tunnel entrance under Dokkveien to Dronning Mauds gate at +8.9. Figure 1 shows the general level of excavation in different areas of the main excavation pit. The deep basement, which covers an area of about 130 by 85 m, required an excavation until a level of -6.40. This resulted in a general excavation depth of 8.5 to 9.0 m.

Figure 1. Localization of construction pit, neighbouring structures and cross sections A-A and B-B. Figure 2. Illustration of different contributions to ground displacements (Langford et al. 2015).

3 NEIGHBORING STRUCTURES AND INFRASTRUCTURE The four lane E18 highway crosses directly under­ neath the construction site in two separate tunnels (Figure 1). Above the E18 tunnels is a sewage tunnel. The rock overburden from the top of the E18 tunnel to the bottom of the excavation is approxi­ mately 18 m, while the sewage tunnel has about 5-6 m of rock cover after the excavation is complete. Dokkveien, which is serviced by trams, lays on an embankment of crushed rock. It is partly supported by a 50 m long retaining wall towards the construc­ tion site. A joint in the retaining wall indicated dif­ ferential settlements in the order of 7-8 cm prior to any construction activity. This is probably caused by the weight of the embankment in combination with large variability in depth to rock. The E18 ramps exit the tunnel on a 16 m long waterproof concrete “trough” which is founded on end-bearing concrete piles to rock. The next 45 m lays on rockfill embankment with increasing height. After that a 75 m long inclining bridge lifts the ramps up to Dronning Mauds gate. The bridge is founded on pillars and piles to rock. Other buildings or bridges in the area are founded on end bearing piles to rock or directly on rock.

ground displacements caused by deep excavations in typical Norwegian ground conditions. The effect of drilling is often difficult to predict, while the effects of drainage may cause damaging settlements several hundred meters from the excavation. Due to the very low permeability of the clay, pore pressure lowering occurs first in permeable layers beneath the clay and bedrock or due to fractures in the bedrock. Subsequently, the pore pressure reduction spreads into the clay layers which increases the effect­ ive stresses in these layers and results in consolidation settlements. Water infiltration wells provide effective means to preserve the ground water level. This miti­ gation measure is often required, because large exca­ vations with a considerable number of potential leakage points are difficult to make completely water tight. The effects of settlements from drilling in clayey materials are believed to be caused by mainly local remoulding of soil around the casing including removal of excess material during drilling (Langford et al. 2015). When drilling in predominantly granular material (i.e. sand and silt), erosion of soil may occur and cause significant subsidence (Konstantakos 2004, Kullingsjö 2007, Lande et al. 2020). These effects from drilling are important to document, as they are usually not quantified when performing risk analysis of possible groundwork damage in Norway.

4 GENERAL CAUSES FOR DISCPLACEMENTS INDUCED BY DEEP EXCAVATIONS

5 GROUND CONDITIONS

It is well known that deep excavations in soft soils can cause displacements and settlements on sur­ rounding ground, structures and infrastructure (Peck 1969, Mana & Clough 1981, Karlsrud & Andresen 2008). Studies reported by Langford et al. (2015) clearly indicate that initial and secondary effects from drilling for tieback anchors and piles from inside an excavation may cause excessive ground settlements. Figure 2 illustrates typical causes of

Ground conditions at the construction site are typical to lower areas in central Oslo. Historically, the area was a shallow bay which has been reclaimed for building purposes. The top 2.0-2.5 m of soil is characterized as urban fill material including sand, gravel, rock and material from old demolished buildings (bricks, con­ crete, etc.). The clay starts at around +0.0 masl, which is 2-3 m below ground surface. The top 2-4 m of clay has some degree of weathering, and from around -4.0

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– Monitoring of anchor forces at 4 locations – Measurement of ground displacements and dis­ placements of adjacent buildings and bridges; a total of 193 monitoring points at a maximum distance of approximately 450 m from the excavation The following mitigation measures to prevent water leakage into the excavation and to limit pore pressure reduction were carried out: – Water infiltration through 20 to 50 m long bore­ holes in rock at 4 locations – Systematic rock grouting of in 10 m depth beneath the toe of the sheet pile wall – Casting of concrete beams at the sheet pile toe where the toe was exposed – Pre-grouting of anchors and piles if leakages were observed – Polyurethane injection of sheet pile interlocks and anchor casings if significant leakages were observed

Figure 3. Active undrained shear strength cuA with depth.

masl and deeper it is normally consolidated (NC). Figure 3 shows the design value of the active undrained shear strength profile compared with calcu­ lated NC-profile (0.3 times the in situ vertical effective stress σv0ʹ). In deeper areas and in local ravines, moraine deposits may be found between clay and rock. Such a moraine deposit was encountered between the construction pit and the E18 ramps to the northwest. In this area silty, sandy deposits were registered between a depth of more than 8 to 10 m below surface Depth to bedrock varies from around 2 m to over 20 m. The rock surface is hilly, and in some places has nearly vertical cliffs. Figure 4 presents measured pore pressure levels on the outside of the excavation pit prior to the construc­ tion works. The piezometers were installed close to the bedrock at all locations and to approximately 5 m soil depth in locations PZ1001 and PZ1003 (Figure 1). The results show that the pore pressure at bedrock is about 2 m lower compared to the hydrostatic pressure in PZ1001 and PZ1002, and about 1 m lower in PZ1003 and PZ1006. This is most likely caused by drainage through cracks, fissures and faults in the rock to the underlying tunnels. 6 MONITORING PROGRAM AND EXCAVATION 6.1

Monitoring program and mitigation measures

The monitoring program for the excavation pit for the new museum consisted of: – Pore pressure measurement at 8 locations (4 loca­ tions in addition to those shown in Figure 1) – Inclinometer readings at the sheet pile wall at 8 locations

The water infiltration wells were installed with 3 wells on the north and north-eastern side of the exca­ vation, and one between the excavation and the E18 ramps (near section B-B in Figure 1). Attempts were made to install wells on the southern side, near PZ1001, but none of the holes proved suitable for water infiltration in rock. 6.2

Excavation method

The excavation was supported by sheet pile walls driven to rock with strand anchors as lat­ eral support of the sheet pile walls. Strand anchors are commonly installed by drilling of a steel casing through soil and 1-2 m into rock. An open hole is drilled into the rock beneath the casing which is subsequently grouted to secure the strands. Once the grout has cured the anchors are post-tensioned. For this project, cas­ ings were drilled using an eccentric down-the­ hole drilling system, which uses water and com­ pressed air for flushing and compressed air for driving the hammer. In total around 470 strand anchors were installed. The sheet pile toe was horizontally fixed by steel dowels into rock before excavation. The dowels were installed through casings welded on the sheet piles. In areas where the toe was exposed and rock beneath the toe of the sheet pile was excavated, a concrete beam with rock anchors was installed. This measure secured the toe of the sheet pile wall and provided additional sealing against groundwater leakage. Due to the varying depth to bedrock both soil and rock was excavated. The total excavated volume was approximately 110 000 m3; approximately 20 000 m3 was excavated rock. The varying depth to bedrock affected also the foundation solution. About 1/3 of the museum is founded directly on rock, while the rest is founded on end bearing micro piles (steel core piles).

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Figure 6. Section A-A with excavation support system and settlement measurement points.

Figure 4. In-situ pore pressure with depth before construction.

6.4 6.3

Section A-A Dokkveien

Figure 5 shows a closer plan view of section A-A, which is shown in Figure 1. The sheet pile wall (SPW) is shown together with the retaining wall, which supports Dokkveien and the tram tracks. The numbered points indicate points where settlements were measured before, during and after excavation phases. Each tram track and the kerbstones on both sides were monitored with about 1-2 mm accuracy. Figure 6 shows cross section A-A including the local soil stratification, rock and ground surface and excavation level. Section A-A was located in a local deep rock zone about 10 m wide at the base. The rock inclined steeply on both sides of this zone. Each of thefive anchor rows (R1 to R5) are identified with dis­ tance (cc) between anchors, the casing diameter (Ø) which was used when drilling through soil and posttension load.

Section B-B E18

Figure 7 shows a plan view of the area surrounding section B-B. The section is covering two different phases: 1. Excavation of the main pit with drilling of piles outside the sheet pile wall 2. Drilling of a permanent retaining wall (RD-wall) with permanent anchors close to E18 and excava­ tion between the main pit and E18 Figure 8 shows a section view of the first phase. The distance between the two-pile groups (in leftright direction on Figure 7) was 7.2 m. For the first phase, only settlement points 804, 808 and 810 showed any response on groundwork activities. Figure 9 gives an overview of the second excava­ tion phase which was close to the E18 ramps. The drilled RD-wall consists of 406.4 mm diameter steel piles with welded-on interlocks to create a continuous retaining wall. In the area on the left side of the E18-tag on Figure 7 there is one row of anchors, while there are two rows on the right side as illustrated in Figure 9. Because the wall and anchors are permanent, casings are larger than used at the other sections. Measurement points 804, 808 and 810 were destroyed during preparation works before drilling the RD-wall and were replaced by 821 to 824. Points 802, 803, 806 and 807 are places on the center strip between on and off ramps. 7 GROUND DISCPLACEMENTS DURING CONSTRUCTION 7.1

Figure 5. Plan view of the area around section A-A.

Section A-A Dokkveien

Figure 11 shows measured settlements against time for the monitoring points near section A-A (Figure

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5) as well as pore pressure head relative to sea level measured at bedrock in PZ1001. The time periods when anchor drilling took place are also indicated. Excavation to the next anchor row was usually per­ formed within one or two weeks after drilling. All points were measured every 3 to 6 months from March 2014 to April 2015. From April 2015 to Sep­ tember 2016 the points were measured every other week, before intervals were gradually increased again as construction activities affecting the sur­ roundings were finalized. The results show that total displacements were larger closer to the excavation with 8.0 to 8.5 cm at points 523 and 524. Total settlements were in the order of 4.0 cm at points 519 and 520. Although the measurement intervals were long until April 2015, settlements were triggered during early phases of drilling and excavation. The settlements were ongoing until May-June 2016 when they stopped for a few months before accelerating again around Decem­ ber 2016. The rate of settlements later decreased again, and no further settlements were measured after April 2017. Pore pressures at PZ1001 indicated an in-situ pore pressure head near bedrock varying between -1.5 to 0 masl before construction started. The results show that drilling of anchor rows R4 and R5 caused instant drops in pore pressure, followed by a gradually decrease down to -3.6 masl in Febru­ ary 2016. Significant leakages of ground water through and around anchor casings were observed and were probably the main cause of pore pressure reduction. Polyurethane grouting was performed but was only partly successful in stopping leakages. Pore pressures later increased as basement structures were completed, and groundwater level within the excavation was raised when the self-weight of the new structure could handle the uplift. The ground­ water level within the excavation had to be tempor­ ary reduced again in July and December 2016 to perform repair and injection of leaking cracks in the concrete basement walls. The measured settlement rates between March 2016 to April/May 2017 indicate that the pore pressure at PZ1001 had to be below -3.0 to -3.5 masl before any significant consolidation settlements occurred. This is particularly evident in the period after April 2016 and can be further highlighted in the period from December 2016 to April 2017. In this time period, the rate of settlements increased instantly during the temporary groundwater drawdown and decreased immediately when pore pres­ sure levels increased to -3.0 m. The following conclusions can be drawn: – Until October 2015 the settlements were most likely caused by displacements of the sheet pile wall and local erosion of soil around the anchor casings from excessive flushing during drilling.

Approximately 60 % of the total displacements were measured at this point. – In October 2015, when pore pressures first are lowered enough, the rate of settlements were 2-3 mm/month, which is significantly higher than the rate later caused by pore pressure reduction alone. This was most likely caused by reconsoli­ dation of remoulded clay around the casings, in combination with redistribution of stresses after excessive flushing of clay during drilling 7.2

Section B-B E18 first phase

Figure 12 shows measured settlements near cross sec­ tion B-B for the first excavation phase (main excava­ tion pit). The figure also shows the measured pore pressure near bedrock at PZ1002 and PZ1003, and progress of drilling anchor rows and piles in the area between the sheet pile wall and the road embankment. The displacement trend at section B-B in Figure 12 is comparable to the trend measured for section A-A during drilling of the first two rows of anchors. During drilling of rows R3 and R4 there is however a sharp increase in settlement rate, which decreases after drilling of anchors is completed but continues at a constant rate of nearly 7-8 mm/month for as long as the measurement points survived. Total sub­ sidence measured by January 2016 were around 12­ 13 cm at points 804 and 808. Pore pressure measurements at PZ1002 and PZ1003 show a draw down following the completion of anchor row R4. The draw down effects are however less dra­ matic than those measured at Dokkveien due to the water infiltration wells installed in the area. Total draw down in the period when displacement measurements were active was about 1.5 to 2.0 m compared with the 2.5 to 3.0 m at Dokkveien, where it also was con­ cluded that around 3.0 m drawdown was required before any significant displacement rate could be measured. There is however some distance between the area where settlements were measured and the two piezometers, and it is possible that the local pore pres­ sure reduction was larger. The total amount and rates of settlement is however far larger than the 1-2 mm/ month measured at Dokkveien due to pore pressure reduction alone, and the reduction was less than at Dokkveien. The dramatic increase in settlements rate during drilling of R3 and R4 was most likely caused by dis­ turbance effects in the silty sand layer above bedrock (Figure 8). It was observed during drilling that com­ pressed air through neighbouring casings, dowel cas­ ings on the sheet pile wall and even through the soil underneath the sheet pile wall. It is therefore likely that the air flushing eroded and transported granular mater­ ial on its way through the soil deposits, this likely explains the sudden 5-6 cm of settlements measured during drilling of R3 and R4. Drill rig operators

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Figure 7. Plan view of the area around section B-B.

Figure 9. Section B-B with excavation support system ­ Second excavation phase.

observed ground settlements. These values are signifi­ cant but are also comparable to other studies (Konstan­ takos 2004). The transport and displacement of soil particles and water with air and how air moves through and affects the soil is probably more complicated than a pure volumetric loss. Such simulations will require further studies. Figure 10 shows results from inclinometer readings on the sheet pile wall near section B-B. Positive values are deformation towards the excavation side. The figure reveals that while the wall moves towards the excavation during the first excavation phases, it moved out­ wards to a total of 55 to 60 mm during the final phases of excavation. The simulations with loss

Figure 8. Section B-B with excavation support system and pile locations – First excavation phase.

reported that drilling resistance was high when going through the silty sand layer, and this routinely led to more extensive use of air flushing. The rate of subsid­ ence after drilling continued at a near constant rate until measurement points were destroyed, which can be explained by reconsolidation and stress redistribu­ tion in disturbed zones. A notable amount of ground water leakage into the excavation pit was later observed. This leakage was particularly visible through pipes and channels eroded into the excavation pit by the air flushing. Polyurethane grouting was performed in casings, sheet pile interlocks and soil, but was only partly successful. Preliminary FEM simulations where the loss of soil was simulated by applying negative volumetric strain in the anchor areas has been carried out. An average soil volume loss in the order of 0.4 to 0.5 m3 pr. anchor were indicated in order to replicate the

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Figure 10. Inclinometer measurements on sheet pile wall at section B-B.

Figure 11. Measured settlements, anchor drilling progress and measured pore pressures at Dokkveien.

Figure 12. Measured settlements, anchor drilling progress and measured pore pressures at E18 ramps first phase.

of soil volume also resulted in the same type of behavior of the sheet pile wall, meaning that the outwards movement could be partly explained by erosion during drilling. These inclinometer readings indicate however that horizontal dis­ placements of the sheet pile wall probably have a minor or no effect on the vertical ground sur­ face displacements 10-15 m behind the excavation. 7.3

Section B-B E18 second phase

Figure 13 shows the measured ground settlements near section B-B and the respective pore pressure data against time. In addition, the period while

drilling the RD-wall and the anchor drilling for the second phase is indicated. Some minor settle­ ment of around 10 mm were measured before drilling of the RD wall at points in the middle of the ramps (802 to 807). Another 5-10 mm of settlements were measured at these points during drilling of the RD wall itself. The most obvious settlements were measured during and shortly after drilling of the anchors. Immediate settle­ ments ranged from 10 to around 40 mm at point 807. Even more subsidence was measured at point 822, but the point and was lost during drill­ ing works. Settlements continued at a lower rate at all measurement points for 6-7 months after drilling.

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Figure 13. Measured settlements, anchor drilling progress and measured pore pressures at E18 ramps second phase.

Pore pressure measurements at PZ1002 were lost after April 2016 because the piezometer malfunc­ tioned. Piezometer PZ1003 indicates that the pore pressure level because of infiltration measures was relatively identical at the period of the RD-wall drill­ ing and before the construction works started (spring 2014). PZ1002, however, indicated a total pore pres­ sure reduction of approximately 1.0 to 1.5 m before it malfunctioned. It is obvious from Figure 13 that pore pressures were reduced from July/August 2015, while no sig­ nificant ground displacements were measured before drilling of the RD-wall in May 2016. For this reason, pore pressure reduction likely does not explain the measured displacements. 8 DISCUSSION Inclinometer readings were performed near cross section A-A but are not reported in this paper. The reasons for this is that firstly the contractor failed to perform a good enough zero-reading before excava­ tion started, and that secondly the inclinometer casing was hit by an excavator which caused unreli­ able results. The overall trend from the measure­ ments however suggest a behavior similar to section B-B (Figure 10). The RD wall in section B-B was not instrumented by inclinometers. Calculated displacements sug­ gested a maximum inwards movement of 10-15 mm, which is significantly lower than the theoretical value required to obtain settlements of 40-50 mm. Moreover, it is worth to point out that in other loca­ tions of sheet pile walls outwards movements were observed. The acquired data shows that settlements decline rapidly with distance from the excavation. By contrast, ground water drawdown and consolida­ tion effects often affect a considerable area

surrounding a deep excavation. It can therefore be followed that the observed subsidence is caused by local disturbance from anchor drilling. The two locations where the ground settlements were measured did not have any critical infrastruc­ ture or structures nearby which were particularly damaged by settlements (only a bump in the road). Results from the measurements are however still interesting for other cases where such structures may be present. 9 SUMMARY AND CONCLUSIONS This paper provided displacement data measured during drilling of casings for anchors for three cases at two locations during construction of the new National museum. When comparing the displace­ ments with the concurrent pore pressure and inclin­ ometer measurements, it is concluded that the drilling activities themselves are the main contribu­ tor to most of the subsidence measured outside the excavation pit. Settlements was measured to be both immediate during drilling, but also had longer lasting effects for several months after drilling. While the immediate settlements are caused by excessive flush­ ing and erosion of soil, the lasting effects are believed to be caused by reconsolidation of remoulded soils and redistribution of stresses follow­ ing the losses of soil volume. The full effects how­ ever will require further studies.

REFERENCES Karlsrud, K. & Andresen, L. 2008. Design and perform­ ance of deep excavations in soft clays. Proc. 6th Int. Conf. on Case Histories in Geotechnical Engineering, Arlington, Virginia, Paper no.12.

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Konstantakos, D.C., Whittle, A.J., Regalado, C. & Scharner, B. 2004. Control of ground movements for a multi-level-anchored diaphragm wall during excavation. Proc. 5th Int. Conf. on Case Histories in Geotechnical Engineering. New York, Paper no 5.68. Kullingsjö, A., 2007. Effects of deep excavations in soft clay on the immediate surroundings - Analysis of the possibility to predict deformations and reactions against the retaining system. Doctoral thesis. Chalmers Univer­ sity of Technology, Göteborg, Sweden. ISBN 978-91­ 7385-002-5. Lande, E.J., Karlsrud, K., Langford, J. & Nordal, S. 2020. Effects of drilling for tieback anchors on surrounding ground - results from field tests. Journal of Geotechnical

and Geoenvironmental Engineering. DOI: 10.1061/ (ASCE)GT.1943-5606.0002274. Langford, J., Karlsrud, K., Lande, E.J., Eknes, A.Ø., & Engen, A., 2015. Causes of unexpectedly large settle­ ments induced by deep excavations in soft clay. Pro­ ceedings of the XVI ECSMGE. Edinburgh: ICE Publishing, 1115–1120. Mana, A.I. & Clough, G.W. 1981. Prediction of movements for braced cuts in clays. Journal of Geotechnical and Geoenvironmental Engineering, ASCE. 107, GT6: 759–777. Peck, R.B. 1969. Deep excavations and tunnelling in soft ground. Proc. 7th Int. Conf. on Soil Mechanics and Foundation Engineering. 225–290. Mexico City.

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Prediction of long-term settlement in shield tunnel using GA-BP neural network Yi-Ming Shen, Dong-Mei Zhang & Jie Zhang Department of Geotechnical Engineering, Tongji University, Shanghai, China

Dong-Mei Zhang & Jie Zhang Key Laboratory of Geotechnical and Underground Engineering of Minister of Education, Tongji University, Shanghai, China

ABSTRACT: In recent years, the hotspot problem in the tunnel has been shifted from how to better con­ struct the tunnel to how to effectively maintain and ensure the safety of the tunnel, among which the long­ term settlement prediction of tunnel is of great interest. Through the existing monitoring data to quantitatively predict the long-term settlement of the tunnel in the future will help us to make effective judgments and take corresponding active control measures in advance to guarantee the safety of the tunnel. As a time-series data, the monitored long-term settlement of the tunnel can be highly fitted by Genetic Algorithm optimizing Back Propagation neural network. However, as a simple mathematical algorithm, the neural network does not con­ sider the longitudinal deformation characteristics and mechanical mechanism of the tunnel when performing the fitting prediction. Based on the traditional Genetic Algorithm optimizing Back Propagation neural net­ work, this paper proposes a classification prediction strategy that can overcome the shortcomings of disregard­ ing the structure. In order to verify the superiority of the proposed strategy, the model is trained and predicted using the monitoring data from the Oriental Sports Center to the Lingzhao Xincun in Shanghai Rail Transit Line 8. Compared with the traditional Genetic Algorithm optimizing Back Propagation model, the improved strategy can reduce the prediction error of tunnel settlement by 20%. In addition, the predicted settlement using the proposed strategy is closer to the monitoring value. The Mean Square Error between them reaches 0.457 and the THEIL Inequality Coefficient is only 0.028, proving the validity of the method.

1 INSTRUCTIONS With rapid development of the metro system, the maintenance of existing tunnels has become a big challenge but vitally significant mission. As an important index of the indicator monitoring system, the long-term settlement of tunnel reflects the safety status of the shield tunnel to a certain extent (Li et al. 2017, Fu et al. 2017). Large displacements and differential settlement appear will result in the longi­ tudinal joint opening and then tunnel leakage to threaten the operation safety of the tunnel (Shao et al. 2016), and even cause the tunnel collapse for the most serious case (Elbaz et al. 2016). Therefore, many researchers have studied the long-term tunnel settlement using various methods. Based on the con­ formal mapping of complex variable methods, Zhang et al. (2013) derived the analytical expression of long-term settlement of shallow buried tunnel during stable seepage considering the effect of tunnel lining. Wongsaroj et al. (2007) studied the long-term ground response due to the tunnel leakage

based on the case study of St James’s Park using soil–fluid coupled three-dimensional finite element simulation. Huang et al. (2013) investigated the tunnel displacement caused by nearby deep excava­ tion using finite element method. However, the per­ formance of long-term settlement of shield tunnels is coupled with various factors. It is unreasonable to study long-term settlement of the shield tunnel simply considering the unilateral factors. Instead, as first-hand information, the on-site settlement moni­ toring data shows the true response of the tunnel, which can be used to judge the safety status of the tunnel and make reasonable predictions quantita­ tively. Once the tunnel settlement exceeds the thresh­ old, active control measures will be taken in advance to ensure the safety of the tunnel. Theoretically, the observed tunnel settlement with the same time interval can be regarded as a time series model, which means that it can be used to pre­ dict the settlement in the future in a short time. As a mathematical method for prediction based on exist­ ing data, the Genetic Algorithm (Holland & John,

DOI: 10.1201/9780429321559-86

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1992) optimizing Back Propagation (GA-BP) neural network not only optimizes the weights and thresh­ olds of Back Propagation (BP) neural networks, pre­ venting them from falling into non-optimal solutions, but also improves the convergence speed and fitting of the network, which is widely used in various fields. Li et al. (2015) found that the GA-BP model has a high accuracy for predicting the lateral displacement of the diaphragm wall of deep excava­ tion. Xue and Liu (2017) concluded that the Genetic Algorithm and Particle Swarm Optimization is reli­ able to predict soil liquefaction although the lique­ faction potential is difficult to assess due to the high uncertainty of the seismic environment and soil properties. Despite the advantages mentioned above, the mathematical algorithm only fits the data from the view of mathematics and can’t consider the struc­ tural characteristics and mechanical mechanism of the tunnel. In this paper, a classification prediction model which is called the improved strategy later is proposed based on GA-BP to overcome the short­ coming. The result of case study shows that the improved strategy greatly improves accuracy and provides a new idea for settlement prediction when compared to traditional GA-BP neural network. 2 BASIC THEORIES OF GA-BP MODEL AND IMPROVED STRATEGY 2.1

Introduction of GA-BP model

As an optimized algorithm, GA-BP neural network overcomes the shortcomings of BP neural networks. It uses the global optimization ability of the Genetic Algorithm to provide the optimal combination of weight and threshold for BP neural network to avoid the shortcomings of falling into the local optimal solution. Meanwhile, it can accelerate the conver­ gence speed of the neural network and improve the prediction accuracy as well. Figure 1 shows the

Figure 1. Flowchart of the optimization algorithm.

flowchart of the optimized algorithm. The basic prin­ ciples and training procedures of BP neural networks have been described in many literatures (E.Rumel­ hart et al. 1986, Zhang et al. 2018, Wang et al. 2007), and will not be repeated here. Instead, a detailed introduction to the key processes of opti­ mizing BP neural networks using Genetic Algorithm will be provided. Firstly, the population is encoding automatically, in which each individual in the population will be encoding in real numbers. Figure 2 shows the top­ ology diagram of the neural network. The weights and thresholds of each individual in the topology diagram of the neural network, including the weight of the input layer to the hidden layer, the weight of the hidden layer to the output layer, and the thresh­ old of each neuron in the hidden layer and the output layer, will be arranged in an “chromosome”. The “chromosome” length will be determined by Equa­ tion 1.

where L is the total length of each “chromosome”, A1 is the total number of input layer neurons in the topology diagram, A2 is the total number of hidden layer neurons in the topology diagram, A3 is the total number of output layer neurons in the topology diagram. Secondly, the fitness value is calculated. The qual­ ity of each individual in the population will be judged by the magnitude of the fitness value. The

Figure 2. Topology diagram of the neural network.

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initial weights and thresholds of each individual con­ tained in the “chromosomes” encoded in the first step (Equation 1) will be input as parameters to the BP neural network. Using the training data, the BP neural network is trained to predict the output. The error between the predicted output and the expected output will be used as the fitness value of each indi­ vidual. The individual fitness value can be calculated using Equation 2.

where F is the individual fitness value, k is the coef­ ficient, n means the total number of neurons in the output layer, ei is the expected output of the i-th neuron in the output layer of BP neural network, ai is the pre­ dicted output of the i-th neuron in the output layer of BP neural network. Thirdly, selection of a better individual. During this step, the error between the predicted output and the expected output is taken as the individual fitness value, which is calculated by Equation 2. Therefore, the smaller the fitness value of a “chromosome”, the higher the quality of the “chromosome”. The roulette algorithm is taken to determine the probability that each individual will be selected for the next round. The selected probability of each individual can be presented in Equations 3 and Equations 4.

¼

where Fi means the fitness value of the individual i of the population, k is the coefficient, pi is the probabil­ ity corresponding to the population individual i, N is the number of populations. Fourthly, performing gene crossover on chromo­ somes. Using Equation 3 and Equation 4, the indi­ viduals selected in the previous step will form a new population, among which two individuals will be selected arbitrarily and be crossed at a certain prob­ ability, while the positions of the intersections are selected randomly as well. The operation method that the m-th chromosome and the n-th chromosome used to intersect at the j position is shown in Equa­ tions 5.

where b 2 ½0; 1].

Finally, performing gene mutation on the chromo­ somes. In this step, an individual is randomly selected from the population generated by Equation 5 and mutated according to a certain probability. The location of the mutation is also randomly deter­ mined. The method for mutating the j-th gene of the i-th individual is shown as Equation 6.

where amax is the upper bound of the gene aij, amin is the lower bound of the gene aij, f(g)=r2(1-g/Gmax)2, r2 is the random number, g is the current number of iterations, Gmax is the maximum number of evolutions, and r is the random number between 0 and 1. With this procedure, the individuals with the best fitness in the population are selected. When the fitness of the individual obtained using Equation 2 satisfies the target requirement, the information contained in the chromosome will be given as an initial value to the BP neural network. If not, another cycle starts. 2.2

Improved strategy

Using the GA-BP neural network mentioned above, the prediction of tunnel settlement can be carried out based on the observed data monitored with the same time interval. However, such a prediction method only fits the monitoring data from the view of mathematics and can’t consider the differences in settlement along the alignment. In order to fully grasp the relationship between the settlement and the structural characteristics of the tunnel, a large amount of monitored data has been gathered and observed carefully. In engineering, a large proportion of the monitored data of the shield tunnels in Shanghai are collected by the Wireless Sensor Net­ work (WSN), which have been widely introduced in earlier publications. (Luca et al. 2010, P. J. Bennett et al. 2010). Among the massive data, two typical sections of Line 11 and Line 8 are selected here to illustrate the improved strategy. Figure 3 shows the sketch map of Line 11 from Oriental Sports Center to Sanlin Road. The tunnel settlement of this section with a time inter­ val of one year is shown in Figure 4. Figure 5 shows the sketch map of Line 8 from South Xizang Road to China Art Museum. The settlement of this section is graphed in Figure 6, covering the time interval from May 20, 2018 to April 18, 2019. It can be found that the settlements varied significantly along the tunnel because of the different buried stratums, impaction of adjacent engineering, and longitudinal deformation characteristics of tunnel. The distribution of the long-term settlements of the two typical sections exhibit fluctuation along the tunnel axis, while the overall development of settlement with time gradually gets stable. Therefore, the entire section

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Figure 3. Sketch map of Line 11 from Oriental Sports Center to Sanlin Road.

Figure 6. Settlement of metro Line 8 from South Xizang Road to China Art Museum.

should be divided into sub-sections according to the characteristics of the observed deformation. The improved strategy suggests that each part corresponds to an undulation interval of the shield tunnel settlement and the settlement trend of each part is predicted independently. 3 CASE STUDY Figure 4. Settlement of metro Line 11 from Oriental Sports Center to Sanlin Road.

In order to verify the feasibility and superiority of the improved strategy, a practical case has been given to prove. 3.1

Background

Shanghai Rail Transit Line 8 was opened to the public in 2009. Total 273 monitoring points were installed in up line of the section of Oriental Sports Center-Lingzhao Xincun. Figure 7 presents the

Figure 5. Sketch map of Line 8 from South Xizang Road to China Art Museum.

Figure 7. Sketch map of Line 8 from Oriental Sports Center to Lingzhao Xincun.

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In order to further demonstrate the superiority of the improved strategy, the same data is used to train by the GA-BP neural network, among which all the param­

Figure 8. Settlement data for each monitoring point every six months.

sketch map of Line 8 from Oriental Sports Center to Lingzhao Xincun. Figure 8 shows the settlements with the time interval of six months. Based on the settlement distribution as shown in Figure 8 and the improved strategy, the section is divided into five parts: from monitoring point S750 to monitoring point S820; from monitoring point S821 to monitor­ ing point S889; from monitoring point S890 to monitoring point S939; from monitoring point S940 to monitoring point S989 and from monitoring point S990 to monitoring point S1022. During the training process, the neural network adopts a three-layer structure of 6 input neur­ ones-5 hidden neurones-1 output neurone. The parameters of the Genetic Algorithm are set as follows: the evolutionary algebra is taken as 300 times; the crossover probability is taken as 0.2; the mutation probability is taken as 0.1. The population size is taken as 15. 3.2

Result

The data used by the improved strategy is monitored from September 13, 2018, to April 21, 2019. In the training sample, the data monitored of the first six months are used as the input data, while the data of the seventh month are used as the output data to train the neural network. In the predicting sample, the data of the second to seventh month are used as the input data, while the data of the eighth month are used as the expected output data. The error between the expected output data and the actual data for the eighth month is used as a basis for judging the quality of the neural network. Figure 9 shows the neural network training results for each part based on the longitudinal deformation characteristics of the shield tunnel, from which we can clearly see that the prediction result of each part is quite close to the monitored results. Except for a few specific points, the error of each monitoring point does not exceed 10%, which satisfies the engineering control standards.

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Figure 10. Comparison of absolute error between two pre­ diction methods and true values.

prediction methods and true values. Compared to the GA-BP model, the improved strategy performs better in terms of approximation to the monitoring values. According to statistics, all the results are tabu­ lated in Table 1. The calculated result shows that the absolute prediction error is controlled within 0.5 mm for the 149 sets of monitored points using the improved strategy, while it is only 97 sets using the GA-BP model. Thus, the prediction accuracy is improved by about 20%. Only 8 sets of data points has an absolute error which exceeds 1.5mm using the improved strategy, accounting for 3% of the total monitoring points, which is in accordance with engineering regula­ tions. The forecast data of the improved strategy is more able to approximate the true value since The Mean Square Error of the improved strategy Figure 9. Comparison of trained results and monitored results based on longitudinal deformation characteristics of is 0.457. THEIL Inequality Coefficient (THEIL IC) is defined as a coefficient varying from 0-1 shield tunnel. to describe the difference between the prediction value and the true value. The smaller the coeffi­ eters are set consistent. The predicted results using the cient value, the higher the prediction accuracy. GA-BP model and the improved strategy are compared The calculation shows the THEIL IC is about with the monitored values respectively. Figure 10 0.028 using the improved strategy, while it’s shows the comparison of absolute error between two 0.033 using the GA-BP model.

Table 1.

Analysis of prediction results of two training methods. Absolute error between predicted value and true value

Training method

1.5

Mean Square Error (MSE)

THEIL IC

GA-BP model Improved strategy

97 149

109 83

59 33

8 8

0.638 0.457

0.033 0.028

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4 CONCLUSION

is taken place at the third position of the chromo­ some. Then a brand-new individual, Chromosome B:11311111, can be obtained.

Based on the neural network optimized by Genetic Algorithm, this paper proposes a new classification and prediction method considering the longitudinal deformation characteristics of the shield tunnel. The following conclusions can be drawn.

ACKNOWLEDGMENTS

(1) GA-BP neural network is proved to be a useful method widely used in various fields for non­ linear predicting. (2) The observed long-term settlement is fluctuating along the tunnel axis but will gradually get stable with time in long-term. Therefore, the simple use of the GA-BP neural network to fit the monitoring data is not able to reflect the mechanical mechanism of the longitudinal deformation of the shield tunnel. Based on the longitudinal deformation characteristics of shield tunnels, a classification prediction model of the improved strategy is proposed. (3) Compared with the GA-BP neural network model, the improved strategy can improve the prediction accuracy by about 20%. The predicted results using the improved strategy is less fluctu­ ating with an error coefficient of 0.028 in THEIL IC. What’s more, only 3% of the total data, has an absolute error over 1.5mm.

NOMENCLATURE Chromosome: an ordered series. Take an example: there are eight weights and thresholds in a neural network with the same value of 1, then the ordered series is 11111111, which we name chromosome because of the much same structure. Chromosome length: The digits in a chromosome. Take an example: the ordered series of the chromo­ some is 11111111, then the chromosome length is 8. Roulette algorithm: a mathematical algorithm to select better individuals from a large number of groups. The whole groups are arranged in a circle while a better individual corresponds to a sector with a greater angle. Spin the pointer and the individual is selected to the next round once its corresponding sector is pointed by the pointer. Repeat the action until enough individuals have been selected (An individual can be selected many times). Gene crossover on chromosomes: Take Chromo­ some A:11111111 and Chromosome B: 22222222 as example. Suppose that exchange is taken place at the fifth position of the chromosomes. Then two brandnew individual, Chromosome C:11112111 and Chromosome B: 22221222, can be obtained. Gene mutation on the chromosomes: Take Chromo­ some A:11111111 as example. Suppose that mutation

This study is financially supported by the National Natural Science Foundation of China (Grants No. 41772295, 51978517), Innovation Program of Shanghai Municipal Education Commission (Grant No. 2019-01-07-00-07-456 E00051) and key innov­ ation team program of innovation talents promotion plan by MOST of China (No. 2016RA4059).

REFERENCES BENNETT, P. J., KOBAYASHI, Y., SOGA, K. & WRIGHT, P. 2010. Wireless sensor network for moni­ toring transport tunnels. Proceedings of the Institution of Civil Engineers, 163, 147–156. ELBAZ, K., SHEN, J. S., ARULRAJAH, A. & HORPIBULSUK, S. 2016. Geohazards induced by anthropic activities of geoconstruction: a review of recent failure cases. Arabian Journal of Geosciences, 9, 708. FU, L., HUANG, Z., HUANG, H., ZHANG, J. & YIN, G. 2017. Health diagnosis method of shield tunnel structure based on cloud theory. Chinese Journal of Engineering, 39, 794–801. HOLLAND, JOHN, H. 1992. Adaption in Natural and Arti­ ficial Systems. Ann Arbor MI: The University of Mich­ igan Press, 211. HUANG, X., SCHWEIGER, H. F. & HUANG, H. 2013. Influence of Deep Excavations on Nearby Existing Tunnels. International Journal of Geome­ chanics, 13, 170–180. Li, X., LIN, X., ZHU, H., WANG, X. & LIU, Z. 2017. Condition assessment of shield tunnel using a new indi­ cator: The tunnel serviceability index. Tunnelling & Underground Space Technology, 67, 98–106. Li, Y., XUE, Y., YUE, L. & CHEN, B. 2015. Displace­ ment Prediction of Deep Foundation Pit Based on Genetic Algorithms and BP Neural Network. Chinese Journal of Underground Space and Engineering, 11, 741–749. MOTTOLA, L., PICCO, G. P., CERIOTTI, M., GUNA, S. & MURPHY, A. L. 2010. Not all wireless sensor net­ works are created euqal: a comparative study on tunnels. ACM Transactions on Sensor Networks, 7(2). RUMELHART, DE, HINTON, GE & WILLIAMS, RJ 1986. Learning representations by back-propagating errors. NATURE, 323, 533–536. SHAO, H., HUANG, H., ZHANG, D. & WANG, R. 2016. Case study on repair work for excessively deformed shield tunnel under accidental surface surcharge in soft clay. Chinese Journal of Geotechnical Engineering, 38, 1036–1043. WANGZHUO, JIALI-MIN, QINYONG & WANGYAN-HUI 2007. RAILWAY PASSENGER TRAFFIC VOLUME PREDICTION BASED ON NEURAL NETWORK. Applied Artificial Intelligence, 21, 1–10. WONGSAROJ, J., SOGA, K. & MAIR, R. J. 2007. Modelling of long-term ground response to tunnelling under St James’s Park, London. Géotechnique, 57, 75–90.

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XUE, X. H. & LIU, E. L. 2017. Seismic liquefaction potential assessed by neural networks. Environmental Earth Sciences, 76, 15. ZHANG, D., LIU, Y. & HUANG, H. 2013. Leakageinduced Settlement of Ground and Shield Tunnel in Soft

Clay. JOURNAL OF TONGJI UNIVERSITY(NATURAL SCIENCE), 41, 1185–1190+1212. ZHANG, L., WANG, F., SUN, T. & BING, X. 2018. A constrained optimization method based on BP neural network. Neural Computing & Applications, 7, 1–9.

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Tunnelling through a piled foundation: Interaction effects Davor Simic Head of Geotechnical Area, Ferrovial-Agroman, Spain

Belén Martínez-Bacas Geotechnical Area, Ferrovial-Agroman, Spain

ABSTRACT: Many authors have discussed interaction mechanism between the tunnel excavation and the existing pile. This paper addresses the changes on the pile axial forces, focusing mainly on the differences in behaviour when the tunnel is excavated below pile toes compared to the cases where the pile is resting deeper than tunnel level. For this investigation large scale numerical models have been employed to account the non­ linear soil/pile stress-strain behaviour. First, an isolated pile has been modelled under the axial load conditions imposed by the Osterberg cell test, obtaining the interaction parameters. Next, a detailed 3D model including the piles and the sequential excavation of the tunnel has been calibrated using settlement data and horizontal displacements obtained from monitoring during construction. Conclusions from the model interpretation show the important changes in load distribution of the piles depending on the tunnel relative situation and the pile lengths.

1 INTRODUCTION As it is widely known the ground movements induced by the tunnel excavation are the result of a complex interaction between the soil and the EPB operation (Figure 1 from Potts et al., 2001). In this sense, the following sources of ground deformation can be distinguished: – A face loss due to the stress relaxation of the soil ahead of the cutterhead, partially compensated by the pressure in the chamber of the EPB. – A shield loss due to the diameter of the cutting wheel being larger than the shield diameter. This ground loss is partially compensated by the ben­ tonite injection through the shield ports, in the machines where this feature is active. – The post-shield loss due to the gap between the excavated surface and the external diameter of lining ring. Two sections of the post-shield loss are distinctively present: a part adjacent to the tail of the shield where the grout is still fluid exerting a pressure that partially compensates for the ground convergence and the farthest part where the grout has already set, so the ground forces act rigidly against the lining ring. These ground movements develop sequentially as the TBM advances creating a transient 3D deformation field that interacts with the nearby pile foundation. The aim of this paper is to analyze by means of a numerical model how the pile resistance components

(shaft friction and end bearing) are progressively modi­ fied with the EPB advance. To this purpose, two com­ plex interaction mechanisms have been introduced in a numerical model, i. e. the EPB operation effects as the machine advances and the soil/pile stress changes during the tunnel excavation. A Plaxis3D model was developed introducing the aforementioned effects and, finally, to limit the model complexity, the superstruc­ ture stiffness has been simplistically introduced by means of an equivalent beam whose parameters were derived from a separate structural simulation not described in this paper (Simic and Bacas, 2019). Two types of behavior have been established in the technical literature (Dias and Bezuijen, 2014; Mair and Williamson, 2014): 1. The pile tip is located above the tunnel crown, where the pile settlement induced by the tunnel is larger than the ground surface settlement 2. The pile tip is situated below the tunnel, in which case the ground deformation induced by the tunnel generate negative skin friction forces on the pile. In both cases the axial load distribution along the pile is modified when compared to the initial state and, depending on the pile connection to the super­ structure, leads to a force redistribution in the build­ ing and foundations that can have an import impact in their respective factors of safety. The numerical modelling is useful to understand the evolution of pile displacements and the axial force distribution with the tunnel advance, previously

DOI: 10.1201/9780429321559-87

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Figure 1. Sources of volume loss. Example of the TBM (Potts and Zdravković, 2001).

investigated by Franza et al. (2017, 2019), Hong et al. (2015), Lee (2012), Lee et al. (2012), Soomro et al. (2017) among others. 2 DESCRIPTION OF THE PROJECT. STRATIGRAPHY AND GEOTECHNICAL PARAMETERS The railway connection tunnel to the Barcelona Air­ port has been excavated with an EPB TBM with an excavation diameter of 10.6 m, below the existing terminal building with an overburden varying between 10 m to 15 m. The Project is located in the soft deltaic sediments of Llobregat river close to the coastline with a shallow water table. The terminal building is resting on a mixed foundation, consisting of slabs and piles of different lengths, see Figure 2: – 6 m bored piles of 0.65 m diameter, which support the part of the building of airside (Rambla) – 8 m long square driven piles of 0.4 m side, which support the façade of the terminal (Hall) – 50 m long square driven piles of 0.4 m side, sup­ porting the main building. Finally, part of the superstructure above the tunnel (connection box) rests on a slab foundation having a thickness of 1.20 m. Geologically the soil consists of soft deltaic Holo­ cene sediments to a depth of about 47 m where the

Figure 3. Transversal section: soil layers, tunnel and length of the piles.

Pleistocene fluvial rounded gravels are found, that form the base of the delta. Boring and CPTU tests were performed and the following units have been identified, see Figure 3: – – – –

0m-2m: UG0: fill 2m-4m: UG1: clayed silt 4m-14m: UG2: sand 14m-48m: UG3: silt and grey clay interbedded with sand bands – >48 m: UG4: sand and gravel The water level is found at 3 m depth. The geotechnical parameters have been derived from the triaxial tests and calibrated simulating the triaxial test calibration tool of with the finite element software (Figure 4), see Table 1 and Table 2. 3 CALIBRATION OF THE SOIL/PILE INTERACTION PARAMETERS

Figure 2. Pile foundations of the Terminal building.

To study the pile-soil interaction an Osterberg cell test was available in the same ground. For the inter­ pretation the pile with Osterberg cell was 3D mod­ elled using a Plaxis3D software. The numerical model was 20 m (width) x 20m (length) x 74m (deep) with the soil layers shown in Figure 3 and soil parameters shown in Table 1 and Table 2. The UG3 was calibrated using the triaxial soil lab tests tool as shown in Figure 4. The pile with a diameter of 1.2m, was simulated with two embedded beams

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Table 2.

Parameters Mohr Coulomb soil model.

Parameters

UG0

UG1

UG4

Condition γunsat (kN/m3) γsat (kN/m3) E’(MPa) ν ϕ’(˚) c’ (kPa) k0

Drained

Undrained 15.3 19.5 5 0.35 29 2 0.49

Drained 19.0 21.0 160 0.33 38 1 1

Table 3.

Designed tip and skin resistance.

Layer Depth (m) Figure 4. Comparison between simulated triaxial test and numerical model of this test.

UG0 0-2 UG1 2-4 UG2 4-10 10-14 UG3 14-20 20-37 37

Table 1. Parameters of the Hardening soil with small strain soil model (HSsmall). Parameters

UG2

UG3

Condition γunsat (kN/m3) γsat (kN/m3) E50ref ’(MPa) Eoedref MPa) Eurref(MPa) G0.7(MPa) γ0.7 νur m ϕ’(˚) c’ (kPa) pref (kPa) k0NC OCR

Drained 16.70 20.50 15 15 45 56 0.2e-3 0.2 0.5 30 1 100 0.47 1

Undrained 15 19.2 3.3 3.3 9.9 33 0.1e-3 0.2 1 22 1 100 0.53 1

elements: the upper part from ground level to a ­ 23 m depth and the bottom part from 23 m to a ­ 37 m depth. The skin friction of the upper and bottom sections of pile were preliminary obtained from the CPTU tests as shown in Table 3. The end bearing of 2 MPa corresponds to the CPTU tip resistance at a depth of 37 m (Figure 3). The pile was modelled under the axial load conditions imposed by Osterberg cell test, consisting of eleven load cycles from 500 kN to 5000 kN. The load cycles were simulated applying two points load at each section of the pile, respectively.

20.4 5 0.4 21 9 0.64

Tip resistance (MPa)

Skin friction (kPa) 30 30 30 72 100 40

2

Figure 5. Comparison between the Oserberg-cell test and the numerical model simulation.

Figure 5 Presents vertical displacement versus load cycle, it is observed a good calibration between the numerical model and the Osterberg-cell test.

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4 TUNNEL EXCAVATION BELOW THE TERMINAL BUILDING 4.1

Description of the model

Figure 6 presents the geometry of the 3D model for the back-analysis. The dimensions are 200 m (transverse to the tunnel direction, x-direction) x 220 m (longitudinal tunnel direction or y-direction) x 74 m (depth) and the number of zones 433000. The soil layers are the same as in Figure 3 and the soil parameters are presented in Table 1 and Table 2. The superstructure, modelled as an equivalent beam, is resting on the 238 piles and the base slab. The TBM excavates through the layer UG3 (Figure 3). Table 4 shows the tunnel data. The advancement of the TBM is 1 m in each calculation step. The tunnel was simulated considering the gaps between the soil and shield, the grout pressure and the face pressure, as well as the jack forces of the TBM machine. Figure 7 shows the chamber earth pressure and grout pressure during tunnelling. During excavation the earth pressure was measured with 7 pressure cells placed inside the TBM cham­ ber. The grout injection pressure was applied at 6 injection ports. Input values of chamber pressure and grout pressure were introduced in the model corresponding to ring 1037, halfway through base slab. Therefore, the input face pressure was 240

Figure 6. 3D model for the back-analysis of ground dis­ placements induced by the tunnel.

Table 4.

Tunnel data.

Figure 7. Face and grout pressures applied to the TBM.

kPa with an increment with depth of 15 kPa/m, and the input grout pressure was 300 kPa with and increment with depth of 10 kPa/m. The TBM machine, and the base slab and were simulated with structural plate elements. Table 5 pre­ sents the parameters. The lining rings of the tunnel were simulated as solid elements having the properties of the precast concrete material (model linear elastic, non-porous and Young modulus 31 MPa). As explained, the foundation piles have differ­ ent lengths, see Figure 8. The piles belong the footings from 1 to 5 and from 6 to 19 are 8 m and 6 m long piles, respectively, so their tips rest on the UG2 layer. The piles belong the footings from 20 to 39 are 50 m long piles and their tips rest on the UG4 layer, below the tunnel springline. The piles were simulated with struc­ tural embedded pile elements. Table 6 presents the pile structural and interface parameters. The tip resistance and skin friction values were obtained from the Osterberg cell calibration as described above. The construction stages of the model were the following:

Tunnel shield Excavation diameter (m) Tail diameter (m) Volume loss (%) Rings Number of segments per ring Ring thickness (m) External diameter (m) Internal diameter(Dt) (m) Ring width (m) Depth of springline (m)

10.60 10.51 0.5 7 0.32 10.24 9.60 1.60 19

Table 5.

Plate data.

Plates parameters

Base slab Connectionbox

Thickness(m) 1.2 Density (kN/m3) 25 Material type Concrete-Elastic Behavior E(MPa) ν12

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Isotropic 31500 0.2

TBM 0.17 247 SteelElastic Isotropic 200000 0.3

4.2 Comparison of the model displacements with the monitoring data Figure 8 shows the location of settlement survey monuments (P1, P2 and P3) and the inclinometer located inside the terminal building which were monitored during construction. Figure 9 shows the movement of survey monu­ ments placed next to the base slab during tunnelling. The monitoring values are close to the model results. Figure 10 compares the ground settlements along 2 sections: section 1 through the slab and section 2

Figure 8. Plan view of the 3D model. Pile caps, survey monuments (P1, P2 and P3) and inclinometer localization.

Table 6.

Pile data.

Pile parameters Skin friction (kN/m)

Tip resistance (kN) E’(MPa) Diameter(Dp) (m) Length (m) Position relative to the tunnel

LP1,

LP2,

LP3 Depth (long (m) piles) 0-8 8-10 10-14 14-48 48-50

SP1 (Short piles)

SP2

(Short piles) Figure 9. Comparison of the survey monument settlement during the tunnel advance with the model results.

160 160 160 62 236 740

15.3 19.5 5 0.35 384.8

19.0 21.0 160 0.33 798

36300 0.45 50 adjacent

36300 0.45 8 above

28600 0.65 6 above

– Initial stage: the base slab and all footing and piles foundation are activated. On the other hand, it was applied the vertical load from the superstruc­ ture acting on the footings, corresponding to every column of the building. – Advancement stage of 1 m for the TBM. During tunneling has been considered the face pressure, the conicity of the machine, the grouting and jack thrust. To simulate the complete advance of the TBM the model was run for 114 construction stages.

Figure 10. Comparison of measured ground settlement to the model results.

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through pile foundation. It is to be noticed that section 1 experienced a heave, probably due to the low over­ burden existing in the tunnel section below the slab. Figure 11 compares the monitoring inclinometer measurements with the model data, showing good correlation as well. As conclusion, the aforementioned calibration proves the adequacy of the model to simulate the behavior of the piles during the tunnel excavation. In the next paragraphs the variation of the force distribu­ tion along the different piles is analyzed through the model results. 5 BEHAVIOUR OF THE PILES 5.1

Pile displacements

The pile foundation settlements and deflections during the tunnel advance have been obtained from the model. The sketch in Figure 12 shows the piles selected for this analysis:

Figure 12. Deformed mesh (scale up to 150 times). Max­ imum lateral displacement is 14 mm in the LP1.

– Two short piles of different length: SP1 8 m long and SP2 6 m long – Three 50 m long piles (LP1, LP2 and LP3), stand­ ing at different distances to the tunnel. Figure 13 shows the evolution of the head pile settlements during the tunnelling. Short piles above the tunnel have larger settlement than long piles. The settlement of SP1 and SP2 is 5% of the pile diameter (Dp), which correspond to 20 mm and 30 mm, respect­ ively. This result has sense considering that SP2 is shorter than SP1. The piles LP1 and LP2 are at an offset of 7 m, equivalent to 0.7 of the internal tunnel diameter (Dt)). These piles have a settlement equivalent to 1%

Figure 13. Evolution of the pile settlement during tunnelling.

Figure 11. Comparison of inclinometer horizontal move­ ments and model results.

Dp (4 mm). The settlement of the long pile LP3, placed at 2.3Dt (22 m), is smaller, 0.3%Dp (1 mm). Also, Figure 13 indicates the zone of influence of the tunnelling. The x-axis 0 corresponds to the loca­ tion of the pile, negative values indicates that the tunnel face has not arrived the pile, positive values the tunnel face is getting ahead. It can see that when distance of TBM is 3 times the internal tunnel diam­ eter farther from the pile location, the increment of the settlements is negligible. The settlements increase substantially between -1Dt to 1Dt. These results are in line with Dias and Bezuijen (2015), Franza et al. (2019), Lee (2012), Lee and Ng. (2006), Soomro et al (2017) and Willianson (2014). On the other hand, Figure 14 shows the deflec­ tions along the piles in two directions: perpendicular to the tunnel axis (transversal tunnel direction) and parallel to axis direction (tunnel direction). The larger deflection occurs in transversal direction reaching 14 mm for the pile LP1. However, the deflection of the LP2 is 10 mm due to this pile is sur­ rounded by different pile groups. The pile LP3,

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Figure 14. Final deflection of the piles. Figure 16. Force evolution of the long pile LP2.

farther from the tunnel, has a maximum value of 5 mm. Similarly, the movement of the short piles depends whether the pile is in the border or it is sur­ rounded by another group of piles. Thus, the behav­ iour of the SP1 pile is similar the upper part of the LP1 and the behaviour of the SP2 is similar the behavior of LP2. 5.2

Pile force distribution

Another important conclusion of the model is the variation of the pile force distribution during the tun­ nelling process. The set from Figure 15-18 shows, for each of the selected long and short pile, the change of axial force distribution for different situ­ ations of the EPB face relative to the pile. To make this behavior more comprehensive, the unit shaft frictions are also represented in the same tunnelling stages. The piles have markedly different behaviour whether they have their tips below the tunnel or not. As many authors have remarked (Dias and Bezui­ jen, 2014, Franza et al. 2019, Lee, 2012; Mair and

Figure 15. Force evolution of the long pile LP1.

Figure 17. Force evolution of the long pile LP3.

Williamson, 2014, Soomro et al. 2017), for the behavior of the long piles, having the tip below the tunnel springline (LP1, LP2, LP3), when the tunnel face is close to the pile (≥-1Dt), the part of the pile above the springline increases the axial force due to the mobilization of the negative skin friction. How­ ever, the part of the pile below the springline, the axial force decreases due to the soil moves upward while pile settles, causing positive skin friction (see Figures 15-17). For the piles closer to the tunnel (2Dt) the maximum value of the axial force occurs when the TBM has passed (≥3Dt), indicating that the pile forces are not affected by the grout pressure and ring placing (Figure 17).

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piles depending on the tunnel relative situation and the pile length:

Figure 18. Force evolution of the short pile SP1.

If the pile is surrounded by other piles the development of the negative skin friction is minor. Figure 18 shows the axial force distribution for the short pile. When the TBM passes below the SP1 tip (stage 0Dt and 1Dt) the negative skin friction around the tip induces a tensile axial force in the lower part of the pile. For the stage ≥3Dt, due to the effect of the grout pressure and the ring placing the axial force is compressive along all the pile reaching its maximum value. 6 CONCLUSIONS Many authors have discussed interaction mech­ anisms between the tunnel excavation and the existing pile. This paper addresses the changes on the pile axial forces, focusing mainly on the differences in behavior when the tunnel is exca­ vated below pile toes compared to the cases where the pile is resting deeper than tunnel level. For this investigation 3D numerical models have been employed to account for inter­ action effects, incorporating the non-linear soil/ pile stress-strain behaviour both for shaft friction and end bearing. To this purpose, first an iso­ lated pile behaviour has been modelled under the axial load conditions imposed by the Oster­ berg cell test, obtaining the model parameters that match the load-settlement results of the test. The EPB excavated tunnel passes close to an array of piles having their toes above and below the tunnel. A detailed 3D model has been cali­ brated using settlement data and horizontal dis­ placements obtained from monitoring during construction. Conclusions from the model interpretation show the important changes in load distribution of the

– Long piles are overloaded in compression. The effect starts to be noticed when the tunnel face is one diameter distance and increases progressively reaching it maximum value when the tunnel face is at pile position. Beyond that the maximum axial force drops slightly because of the grout pressure for piles located at less than one diameter of the tunnel (< 1 diameter of the tunnel). For piles located farther the maximum axial force occurs when the TBM passes beyond two tunnel diameters. – The long piles develop negative skin friction in a substantial section of their shaft, up to the springline level, causing the increment of axial force. – The short piles close to and above the tunnel are partially in tension due to the negative skin fric­ tion effects when the tunnel face is below, but finally reach a compressive state due to the effect of the grout pressure.

REFERENCES Dias, T. G. S. & Bezuijen, A. 2014. Pile-Tunnel Interaction: A conceptual analysis. In C. Yoo, S. Park, B. Kim, H. Ban (eds,), Geotechnical Aspects of Underground Construction in Soft Ground; Proc. intern. symp., Seoul, 25-27 August 2014: 251–255. Leiden: Balkema Dias, T. G. S. & Bezuijen, A. 2015. Data Analysis of Pile Tunnel Interaction. Journal of Geotechnical and Geoen­ vironmental Engineering. Vol. 141 (12):1–15 Franza A., Marshall, A. M., Haji T., Abdelatif A. O., Carbonari S. & Morici M. 2017. A simplified elastic analysis of tunnel-piled structure interaction. Tunnelling and Underground Space Technology 61:104–121. Franza A., Jimenez R. and Marshall A. M. 2019. Elastic analysis of tunnelling beneath capped pile groups. In H. Sigursteinsson, S. Erlingsson & B. Bessason (eds.), Geotechnical Engineering foundation of the future. Proc. of the XVII ECSMGE-2019, Reykjavik, 1-6 Septem­ ber 2019. Icelandic Geotechnical Society: www. jtfi.net. Hong, Y., Soomro, M.A. & Ng, C.W.W. 2015. Settlement and load transfer mechanism of a pile group due to side-by-side twin tunnelling. Computers and Geotech­ nics. Vol. 64, 105–119. Lee, C. J. 2012. Numerical analysis of the interface shear transfer mechanism of a single pile to tunnelling in wea­ thered residual soil. Computers and Geotechnics. Vol 42: 193–203 Lee, S. W., Choy, C. K. M., Tse, S.C., van Gool, F.R., Cheang, W. W. L. & Brinkgreve R. B. J. 2012. 3D Nu­ merical modelling of tunnelling intersecting piles. In Viggiani (ed.), Geotechnical Aspects of Underground Construction in Soft Ground. Proc. of the 7th Inter­ national Symposium. Rome, 17-19 May 2011:919–925. Lee, G. T. K. & Ng, C. W. W. 2006. Three-dimensional numeri-cal simulation of tunnelling effects on an exist­ ing pile. In Bakker, Bezuijen, Broere and Kwast (eds), Geotechnical Aspects of Underground Construction in

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Soft Ground Proc. of the 5th intern. Symp., Amsterdam, 15-17 June 2005:139–144 Mair, R. J. & Williamson M. G. 2014. The influence of tunnelling and deep excavation on piled founda­ tions. In C. Yoo, S. Park, B. Kim, H. Ban (eds.), Geotechnical Aspects of Underground Construction in Soft Ground; Proc. intern. symp., Seoul, 25-27 August 2014:21–30. Leiden: Balkema. Potts D. M. & Zdravković L. 2001. Finite element analysis in geotechnical engineering application. Vol. 2. London: Thomas Telford.

Simic D. & Martinez-Bacas B. 2019. Soft ground tunnelling below a mixed foundation building. In D. Peila, G. Viggiani and T. Celestino (eds.), Tunnels and under­ ground cities: engineering and innovation meet archae­ ology, architecture and art; Proc. WTC2019., Naples, 3-9 May 2019.Vol. 1. Leiden: Balkema. Soomro, M.A., Ng, C. W.W., Liu, K. & Memon, N.A. 2017. Pile responses to side-by-side twin tunnelling in stiff clay: Effects of different tunnel depths relative to pile. Computers and Geotechnics. Vol. 84: 101–116 Williamson M. G. 2014. Tunnelling effects on bored piles in clay. PhD thesis, University of Cambridge, U.K.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

The use of protective structures to reduce tunnelling induced damage to buildings G. Song, A.M. Marshall & C.M. Heron Faculty of Engineering, University of Nottingham, Nottingham, UK

ABSTRACT: Excavation of tunnels in urban areas causes ground movements that could damage existing nearby structures. The interaction between tunnel construction and piled structures has attracted considerable attention from researchers; however the study of protective measures used to reduce the effect of tunnelling on the structures has received less attention. Piled walls are sometimes used in practice to reduce the effect of tunnelling on buildings, however detailed experimental data related to this problem are limited. In this paper, results from three centrifuge tests are presented which aim to quantify the effect of a protective wall in redu­ cing piled structure damage caused by tunnelling. A hybrid testing technique (coupled centrifuge-numerical modelling) is adopted, where a numerical model is used to solve the structural domain (building and founda­ tion loads, including redistribution of loads due to ground movements caused by tunnelling), and the complex non-linear soil, soil-wall interaction, and soil-pile interaction behaviour is modelled within the centrifuge domain. This paper focuses on settlement data obtained from the tests, with results used to demonstrate how the length of the protective wall significantly affects the soil movements on the building side of the wall, with subsequent impacts on pile settlement.

1 INTRODUCTION Tunnel excavation in urban areas causes ground movements which could damage existing nearby assets (e.g. buried pipelines, building foundations). The tunnelling-induced ground movements, as high­ lighted by Mair and Taylor (1997), are mainly due to (1) soil stress-relief at the tunnel excavation face, (2) over cutting from the tunnel excavation face, and (3) variation of ground stresses close to the tunnel lining (long-term effect due to consolidation). Tunnel-pile­ structure interaction (TPSI) problems have drawn con­ siderable attention from researchers (Mair and Taylor 1997; Standing and Potts 2008; Jacobsz et al. 2004; Franza et al. 2019; Lee 2012; Loganathan et al. 2000). The construction of a tunnel close (adjacent) to a piled structure has the potential to cause differential settlements and threaten the serviceability of a structure. To protect structures from tunnelling­ induced damage, several remedial measures have been adopted in practice: (1) compensation grouting, (2) structure jacking, and (3) the use of protective structures or curtain walls. The purpose of the remed­ ial measures are to reduce the uneven structure settle­ ments caused by tunnelling. There are several examples of the application of protective structures applied as a barrier between a tunnel and an adjacent piled foundation (Ledesma and Alonso 2017; Di Mar­ iano et al. 2007), however there are very few

DOI: 10.1201/9780429321559-88

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experimental studies which have considered this prob­ lem. Bilotta (2008) conducted centrifuge tests to investigate the effect of a diaphragm wall on soil movements caused by tunnel volume loss in overconsolidated clay; the study did not explicitly include a structure or foundation system. To the authors’ knowledge, no centrifuge modelling studies have yet been conducted which include all the interacting com­ ponents; i.e. tunnel, protective structure, foundations, and structure. In order to achieve more sustainable cities and enable more efficient and cost-effective pro­ tective measures to be developed, it is necessary to improve our understanding of these complex inter­ action problems. In this paper, results from three centrifuge tests are presented which study the performance of protective walls in reducing the effect of tunnelling-induced ground movements on structures. Two model protect­ ive walls with different embedded depths were stud­ ied; a `short’ wall where the toe of the wall was located at the tunnel springline, and a `long’ wall where the toe was located below the tunnel invert. This paper focuses on tunnelling induced movements of the soil, the protective walls, and the piles. In the centrifuge tests, axial loads and bending moments were measured in the piles and protective walls, using fibre Bragg grating strain sensors (Song et al. 2019), however these data are not presented here due to space limitations. The effect of the protective wall

on reducing pile settlement is quantified and com­ pared against data from a test with no protective wall. 2 EXPERIMENTAL SYSTEM AND SETUP The centrifuge tests were conducted on the University of Nottingham Centre for Geomechanics (NCG) 2 m radius, 50 g-tonne geotechnical centrifuge at an acceleration of 80 times gravity (i.e. 80 g). Figure 1 shows the layout of the tests, along with the definition and detail of various geometric and material parameters. The coupled centrifuge-numerical modelling (CCNM) technique originally developed by Franza and Marshall (2019) was used in this study, but with a modified pile loading system. For these tests, the CCNM technique allows the tunnel, soil, (depending on the test) protective wall and piles to be modelled in the centrifuge, whereas the framed building is simu­ lated in a numerical model, with pile head displace­ ments (vi ) and loads (Pi ) being transferred between the centrifuge and numerical models using a real-time interface developed by Idinyang et al. (2019). The CCNM technique enables a realistic replication of the load redistribution (due to the characteristics of the structure) that occurs when piles are affected by tun­ nelling, thereby providing a better simulation of the global interaction problem than the case where, for example, constant dead loads are applied to the pile heads. A benchmark test, labelled CCNM3, was con­ ducted where only the piled structure and tunnel were included (i.e no protective wall). Tests with protective walls are labelled CCNM-R-S and CCNM-R-L, where the ‘R-S’ and ‘R-L’ refer to tests with a ‘short’ wall (length Lw ¼ 207 mm; toe of the wall located at tunnel springline) and

Figure 1. Test layout in model scale.

a `long’ wall (Lw ¼ 297 mm; toe of the wall half a tunnel diameter below the tunnel invert), respectively. The protective walls were located 55 mm away from the tunnel centreline (giving a clear distance between the tunnel and pile of 4.5 mm) and were coated with sand to give a rough surface (to replicate, for example, nondisplacement secant pile walls). 2.1

Centrifuge model

The centrifuge strong box has inner width x height dimensions of 700 mm x 500 mm, had a depth of 150 mm, and the `front’ face equipped with transpar­ ent acrylic. The model tunnel had an initial diameter of 90 mm and was located 175 mm above the strong box base, giving a clear distance of 130 mm between the bottom of the tunnel and strong box base. Tunnel volume loss was implemented using an eccentric rigid boundary mechanical model tunnel (Song et al. 2018). The model tunnel enables non­ uniform radial displacements around the tunnel lining (typical for shallow tunnelling), with max­ imum soil displacement at the tunnel crown and no displacement at the tunnel invert. In practice, a 0.8 m diameter concrete pile has an axial stiffness EA ¼ 10 - 14 x 103 MN (assuming concrete has a Young’s modulus E ranging from 20-28 GPa. To match the diameter of the pile (centrifuge acceleration is 80 g), a 10 mm diameter aluminium hollow tube was used. The thickness of the model tube was 1 mm, which gives an axial rigidity EA ¼ 19:4 x 103 MN in prototype scale (slightly higher than the 0.8 m full-scale concrete pile). The piles were driven into the soil at 1 g which in effect replicates a bored pile scenario. To increase the inter­ face friction of the model piles, sands was bonded to the surface and tip of the model pile, which gives a final pile diameter of 11 mm. A 10mm thick aluminium plate with a width of 148 mm was used in the centrifuge tests (the width of the strongbox is 150 mm) to model a protective wall, which gives a flexural rigidity EI ¼ 34:8 x 103 MNm2 in prototype scale. Using a geometric scaling factor, the aluminium plate represents a 0.8 m thick wall at prototype scale. However, a 0.8m thick concrete wall has a flexural rigidity of EI ¼ 10 - 14 x 103 MNm2 (assuming same range of E as previously). There­ fore, in terms of flexural rigidity, the aluminium plate represents a concrete wall that has a thickness of between 1 and 1.2 m. To reduce the interface friction between the protective walls and the strongbox front/back walls, Polytetra­ fluoroethylene (PTFE) strips were placed on the front/back face of the protective walls. Two cameras were used to take images during the tests, and geoPIV-RG (Stanier et al. 2015) was used to track subsurface soil and protective wall movements.

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2.2 Numerical (structural) model The numerical model for the structure was devel­ oped within ABAQUS (Hibbitt 2002) and simulated a five-storey steel frame building with no groundfloor slab (see Figure 1). A linear elastic constitutive model was used for the steel frame. The dimensions of the steel frame building as well as column and beam size, in prototype scale, are given in Figure 1. The pile head loads were calculated based on a five storey buildings for storage purpose, where the per­ manent load is 3 kN=m2 , and the variable load is 7.5 kN=m2 . This gives a total load of 2364 kN for the two inner piles and 1630 kN for the two outer piles (prototype scale). 2.3 Model preparation The soil was prepared at 1 g using the sand pluvia­ tion method. Prior to sand pluviation, the strong box was placed with the front acrylic window facing downwards, allowing the sand to be poured in the direction of the longitudinal tunnel axis. The alumin­ ium plate (protective wall) was placed on the acrylic window (front window), and two temporary supports were located inside the strong box to prevent the sand from spreading during sand pluviation. A thin layer of dyed sand was placed uniformly on the sur­ face of the acrylic window to increase the `tracking’ ability of geoPIV-RG. The sand was then prepared according to a methodology calibrated to achieve a relative density of Id ¼ 90%. After sand pouring, the back wall was bolted to the strong box, and the strong box was rotated to its upright position. The temporary support was then disassembled. To repli­ cate non-displacement piles, the model piles were pushed into the sand sequentially, starting from pile 1 and proceeding to pile 4; a support frame was used to ensure the piles were kept vertical during installation.

each tunnel volume loss increment, pile displace­ ments measured from the centrifuge domain are passed to the ABAQUS model via the real-time interface, and new pile head loads calculated from ABAQUS are then passed back to the pile load con­ trol system in the centrifuge. Once a stable condition is achieved, another increment of tunnel volume loss is initiated and the process is repeated. Meanwhile, images were taken by the cameras to measure the soil and protective wall displacements. Note that the CCNM program was used in all tests (CCNM3, CCNM-R-S, and CCNM-R-L), hence an accurate simulation of load redistribution onto the pile foun­ dations was simulated for each case, depending on the amount of pile displacement caused by tunnelling (which varied depending on the length of the protective wall). 3 RESULTS 3.1 The effect of the protective walls on soil displacements Figure 2 shows the ground surface vertical and hori­ zontal displacements for tests CCNM3, CCNM­ R-S and CCNM-R-L at Vl;t ¼ 2%. It is worth noting that the vertical and horizontal displacements discussed in this subsection were not measured under planestrain conditions due to the existence of the piles.

2.4 Testing procedure A 5 N constant load was applied to the pile (load controlled) during centrifuge spin-up to 80 g to ensure minimum relative displacement between the soil and piles. Three stabilisation cycles were then performed (going from 80 g to 10 g and back to 80 g); the intention of these cycles is to achieve a uniform stress distribution within the soil body to obtain better consistency between centrifuge tests. The piles were then loaded to the designated work­ ing load (225 N for outer piles, and 370 N for inner piles) in 50 N stages, starting from pile 1 and moving sequentially to pile 4 (refer to pile number­ ing in Figure 1). The CCNM real-time data exchange interface program was then activated, followed by the initiation of ABAQUS (simulating the frame structure). The tunnel volume loss (Vl;t ) process was then started, with Vl;t increased up to ≈3% in incre­ ments of ≈0:1%. Within the CCNM program, for

Figure 2. Measured surface displacement at Vl;t ¼ 2%: (a) vertical settlement, (b) horizontal displacement (positive values are downwards and to the right).

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The piles were located in the middle of the strongbox (halfway between the front and back walls) whereas displacements were measured at the acrylic window. For test CCNM-R-S, the soil on the retained side of the wall (x=Dt 40:7) shows a broadly similar magnitude of vertical settlement to the soil in refer­ ence test CCNM3 (with no protective wall), suggest­ ing that the length of the wall in test CCNM­ R-S was not sufficient to reduce tunnelling-induced surface soil settlement at the location of the building. The short wall, with its toe at the tunnel springline, is seemingly ineffective at blocking the progression of ground displacements, and the wall follows the settlement of the surrounding soil. In addition, for test CCNM-R-S, the surface soil horizontal displace­ ments close to the wall (x=Dt between 0 and 1) shows less displacement towards the tunnel than the soil in test CCNM3. These results suggest that a `short’ protective wall could be used to reduce sur­ face horizontal displacements to some degree, but

that this wall in ineffective at reducing surface settlements. For the vertical settlement in test CCNM-R-L, the amount of settlement behind the wall (x=Dt 40:7) is minimal. Moreover, surface soil settlements located ‘in front’ of the wall (x=Dt 50) are greater than in tests CCNM3 and CCNM-R-S. For test CCNM-R-L, horizontal soil displacements towards the tunnel in the area behind the wall are less than in test CCNM­ R-S. The results demonstrate that the `long’ wall (with its toe half a tunnel diameter below the tunnel invert) provided a greater level of reduction of sur­ face soil displacements (vertical and horizontal) than the ‘short’ wall. Figure 3 presents the vertical and horizontal soil displacement contours for all three centrifuge tests at a tunnel volume loss of Vl;t =2.0 %. For Test CCNM3, a reasonably symmetric `chimney-like’ zone of relatively high vertical settlements is observed above the tunnel crown, despite the fact

Figure 3. Contours of vertical and horizontal displacement (mm) at Vl;t ¼ 2:0% for all tests (positive values are downwards and to the right).

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that the piled structure is located on the right-hand side of the tunnel. The loaded piles (located within the centre of the strongbox) did have some effect on ground displacements at the plane of the acrylic wall, where surface settlements are noted to be greater to the right of the tunnel than to the left. For test CCNM-R-L (`long’ protective wall), the soil located at the structure side the wall (x=Dt 40:7) shows considerably less settlement than the equiva­ lent soil on the opposite side of the tunnel (x=Dt 5 - 0:7). For test CCNM-R-S, the soil at the structure side the wall does not show as significant a reduction in soil settlements as CCNM-R-L. For protective wall tests, the `chimney-like’ soil settle­ ment mechanism above the tunnel crown tends to `tilt’ towards left in Test CCNM-R-L, with vertical settlements being reduced nearer the protective wall. For horizontal displacement in test CCNM3, as was the case for vertical settlements, the piles located to the right side of the tunnel tended to reduce tunnelling-induced soil movements in the region around the piles. As a result, the boundary line (distinguishing the soil displacement directions) is tilted slightly towards the right. For tests CCNM­ R-S and CCNM-R-L, in general, the horizontal soil displacements behind the wall are less than the soil displacements in test CCNM3. For test CCNM-R-S, the major zone of horizontal displacement was located close to the tunnel springline, which will encourage the toe of the protective wall to move towards the left. A notable mechanism is observable in the horizontal displacement data on the right side of the protective wall for test CCNM-R-S, indicating that the wall was shifted to the left and rotated by the action of the tunnel volume loss (the wall dis­ placements are examined in more detail in the next section). For test CCNM-R-L, the toe of the wall was located below the tunnel invert. Data from Zhou (2015) (greenfield tunnelling) showed that the soil around the bottom half of the tunnel undergoes very little to no displacements. Therefore, the soil dis­ placements around the toe of the long protective wall are expected to be very small. A zone of hori­ zontal displacements is observable behind the long wall at a depth of z=zt = 0.4-1.2, which is indicative of a bending mechanism of the wall (examined in the next section). 3.2

The response of the protective walls

Figure 4 shows the horizontal displacement of the walls for tests CCNM-R-S and CCNM­ R-L. Some sand particles intruded into the inter­ face between the PTFE strips placed along the front face of the wall and the acrylic window, near the toe of the wall in both tests. As a result, geoPIV-RG displacement data close to the toe of the wall was not used, and the horizontal dis­ placements were estimated based on the deformed shape of the wall above the toe (upper and middle portions of the walls).

Figure 4. Horizontal displacement and bending moment along the wall with tunnel volume loss: (a) horizontal dis­ placement, (b) bending moment.

Figure 4 demonstrates that the deformed shape of the `short’ and `long’ protective walls is very differ­ ent. For test CCNM-R-S, the toe of the wall was located at the tunnel springline (z=zt = 1). Figure 3 demonstrated that the major zone of horizontal dis­ placements towards the tunnel was located close to the tunnel springline (z=zt = 0.8); these tunnelling­ induced soil movements tended to move the toe of the wall towards the tunnel. The displacement at the upper portion of the wall (close to the soil surface) is minimal. For test CCNM-R-L, the toe of the wall is located below the tunnel, where the surrounding soil experiences limited movement during tunnelling. With an increase in tunnel volume loss, the middle portion of the wall (above to the tunnel springline at z=zt =0.8-1.0) shows the most horizontal displace­ ment towards the tunnel, whereas the upper portion of the wall moved towards the right (z=zt =0-0.4). The fixity of the wall beneath the tunnel level causes wall bending at a depth of z=zt = 0.8; the soil located below the upper portion of the wall (z=zt =0.4-0.6) provides the resistance against the horizontal move­ ment of the wall towards the tunnel and the soil located at the upper portion (z=zt =0-0.4) provides little resistance (low horizontal confining stress),

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therefore can not prevent the wall from bending towards the structure side. As discussed previously, the flexural rigidity of the model wall in prototype scale is greater than that of a 0.8 m thick concrete wall, therefore under the same tunnel volume loss, a greater level of wall deformation is expected for a ­ 0.8 m concrete wall. To summarise, the displaced shape of the protect­ ive wall highly depends on the length of the wall with respect to the depth of the tunnel. For CCNM­ R-S, the tunnelling-induced ground movements caused the toe of the wall to move towards the tunnel, with little horizontal displacement occurring near the ground surface. For CCNM-R-L, the toe of the wall located below the tunnel experienced limited displacement. With an increase in tunnel volume loss, soil movements towards the tunnel occurred at a depth of z=zt = 0.8, resulting in the wall bending at this depth. 3.3 The effect of the protective walls on pile settlements Figure 5 shows the pile head settlement with tunnel volume loss for tests CCNM3, CCNM-R-S and CCNM-R-L; note the scale of the settlement for each pile is different. For test CCNM-R-L, because the wall effectively reduced soil movements immediately behind the wall (see Figure 2, 3), the settlement of pile 1 was signifi­ cantly reduced when compared with the other tests.

Figure 5. Pile head settlement with tunnel volume loss.

Pile 1 in test CCNM3, where the protective wall was not used, experienced the greatest amount of settle­ ment with tunnelling. For pile 2, test CCNM­ R-L shows the least pile settlement, followed by tests CCNM-R-S and CCNM3. For pile 3, the amount of pile settlement with tunnelling is generally less than piles 1 and 2 in all three centrifuge tests, test CCNM­ R-L provided the least settlement, and tests CCNM­ R-S and CCNM3 show similar magnitude of settle­ ment. For pile 4, as the location of the pile is furthest away from the tunnel and protective wall, the effect of the protective wall on pile head settlement is minimal, and all tests presented a similar settlement with tunnelling. From the above observation, it can be summarised that the use of protective walls can reduce tunnelling­ induced pile settlement. In addition, the efficiency of the protective walls in terms of pile settlement can be quantified using Equation 1. The maximum efficiency of a protective wall is achieved if the tunnelling­ induced pile settlement is fully prevented (zero pile displacement, giving ηrs Vl;t ¼ 100%), whereas the other extreme case is that the protective structure provides zero protection to pile settlement (i.e. the wall does not exist, equivalent to test CCNM3 in this case and resulting in ηrs Vl;t ¼ 0). Note that CCNM3 here does not represent an assumption that the piles moved according to a greenfield value (as done using empir­ ical analysis methods of tunnel-pile interaction such as Devriendt and Williamson (2011)) since it is experimentally obtained using the CCNM technique, hence it implicitly accounts for the effects of initial pile safety factor (which can be achieved using the analytical method of Franza et al. (2020))and load redistribution amongst piles due to the building stiff­ ness. Centrifuge test results from CCNM-R-S and CCNM-R-L should fall within these two extreme cases.

where S is pile settlement, the subscript Vl;t indi­ cates the tunnel volume loss, the superscript rs refers to a case where a protective wall is considered, and the superscript s refers to the case without the use of a protective wall. Figure 6 presents the variation of efficiency par­ ameter ηrs Vl;t with tunnel volume loss for the four model building piles in tests CCNM-R-S and CCNM-R-L. For test CCNM-R-S (`short’ protective wall), at lower values of tunnel volume loss (Vl;t ¼ 0:5 - 1:5%; note that settlement data below Vl;t ¼ 0:5% is unavailable because of the very small magnitudes), the efficiency parameter for all four piles decreased as tunnel volume loss increased. With further increase in tunnel volume loss (Vl;t 41:5%), the efficiency parameter increased slightly. Piles 1, 2, and 3 show similar values of

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settlements to structures caused by tunnelling. The surface and subsurface soil displacements, deform­ ation of the protective walls, and the settlement of the piles with tunnel volume loss were presented. The use of a `long’ protective wall, where the toe of the wall extends below the invert of the tunnel (by a tunnel diameter in the tests pre­ sented here) can prevent vertical and horizontal soil displacements more efficiently than a `short’ wall where the toe is located at or above the tunnel springline. The length of the wall has a significant effect on the resulting ground and wall deformation mechanisms. During tunnel volume loss, the use of protective walls can prevent tunnelling-induced pile head settlement. The efficiency (ηrs Vl;t ) of the protective walls in reducing tunnelling induced pile settlement depends mainly on the length of the wall and the relative location of the pile with respect to the tunnel (x=Dt ); it is less affected by the magnitude of tunnel volume loss. The outcomes from these tests will help improve the understanding of the complex tunnel, soil, pro­ tective wall, piled structure interaction problems. The work presented in this paper could help engin­ eers to refine their design of protective walls in prac­ tice, and enable more efficient and cost-effective protective measures to be adopted.

Figure 6. Efficiency parameter ηrs Vl;t with tunnel volume loss.

efficiency (on average ≈10%), whereas pile 4 (fur­ thest from the tunnel) shows values of efficiency close to 0 (negative values are obtained for Pile 4, but these are due to very small magnitude displace­ ments, therefore the data above Vl;t ≈1:0% is not con­ sidered; see grey dashed line in Figure 6). For test CCNM-R-L, pile 1 shows a steady increase of efficiency parameter with tunnel volume loss. For piles 2, 3, and 4, the trend is similar to that from test CCNM-R-S, where efficiency decreases from about Vl;t ¼ 0:5 - 1:5%, then increases or stays steady for higher volume losses, however the magnitude of efficiency parameter is considerably larger for all piles, especially those closest to the tunnel/protective wall. In general, the variation of efficiency parameter for all four piles is not very sen­ sitive to tunnel volume loss; wall length as well as the distance between the pile and tunnel (x=Dt ) have the dominant role. 4 CONCLUSIONS Results from three centrifuge tests were presented to investigate the effect of protective walls on reducing

REFERENCES Bilotta, E. (2008). Use of diaphragm walls to mitigate ground movements induced by tunnelling. Géotechnique 58(2), 143–155. Devriendt, M. & M. Williamson (2011). Validation of methods for assessing tunnelling-induced settlements on piles. Ground Eng, 25–30. Di Mariano, A., J. M. Gesto, A. Gens, & H. Schwarz (2007). Ground deformation and mitigating measures associated with the excavation of a new metro line. In Proc. XIV European Conference on Soil Mechanics and Geotechnical Engineering, ECSMGE, pp. 1901–1906. Franza, A. & A. M. Marshall (2019). Centrifuge and real-time hybrid testing of tunnelling beneath piles and piled buildings. ASCE Journal of Geotechnical and Geoenvironmental Engineering 145(3), 04018110. Franza, A., A. M. Marshall, & R. Jimenez (2020). Non­ linear soil-pile interaction induced by ground settle­ ments: pile displacements and internal forces. Géotechnique, In press. Franza, A., A. M. Marshall, & B. Zhou (2019). Greenfield tunnelling in sands: the effects of soil density and rela­ tive depth. Géotechnique 69(4), 297–307. Hibbitt, K. (2002). ABAQUS/Explicit User’s Manual: Version 6.3. Hibbit, Karlsonn & Sorensen. Idinyang, S., A. Franza, C. M. Heron, & A. M. Marshall (2019). Real-time data coupling for hybrid testing in a geotechnical centrifuge. International Journal of Phys­ ical Modelling in Geotechnics 19(4), 208–220. Jacobsz, S. W., J. R. Standing, R. J. Mair, T. Hagiwara, & T. Sugiyama (2004). Centrifuge modelling of tunnelling near driven piles. Soils and Foundations 44(1), 49–56.

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Ledesma, A. & E. E. Alonso (2017). Protecting sensitive constructions from tunnelling: the case of world heritage buildings in barcelona. Géotechnique 67(10), 914–925. Lee, C. J. (2012). Numerical analysis of the interface shear transfer mechanism of a single pile to tunnelling in weathered residual soil. Computers and Geotechnics 42, 193–203. Loganathan, N., H. G. Poulos, & D. P. Stewart (2000). Cen­ trifuge model testing of tunnelling-induced ground and pile deformations. Geotechnique 50(3), 283–294. Mair, R. J. & R. N. Taylor (1997). Theme lecture: Bored tunneling in the urban environment. Proceedings of the fourteenth international conference on soil mechanics and foundation engineering (Hamburg, 1997), Balkema, 2353–2385. Song, G., A. M. Marshall, & C. M. Heron (2018). A mechanical displacement control model tunnel for simulating eccentric ground loss in the centrifuge. In Proceedings of the 9th International Conference on

Physical Modelling in Geotechnics (ICPMG 2018), July 17-20, 2018 , London, United Kingdom, pp. 455–460. CRC Press. Song, G., A. M. Marshall, & C. M. Heron (2019). Load redistribution of piles affected by tunnelling: hybrid centrifuge tests using fibre Bragg grating. In Proceedings of the XVII European Conference on Soil Mechanics and Geotechnical Engineering, ECSMGE-2019, Reykjavik, Iceland. DOI:10.32075/ 17ECSMGE-2019-0236. Standing, J. R. & D. M. Potts (2008). Contributions to Géo­ technique 1948-2008: Tunnelling. Géotechnique 58(5), 391–398. Stanier, S. A., J. Blaber, W. A. Take, & D. White (2015). Improved image-based deformation measurement for geotechnical applications. Canadian Geotechnical Jour­ nal 53(5), 727–739. Zhou, B. (2015). Tunnelling-induced ground displacements in sand. Ph. D. thesis, University of Nottingham.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Evaluation method on ground movement using continuum ground model M. Sugimoto, J. Chen, P.T. Anh & K. Manabe Nagaoka University of Technology, Nagaoka, Japan

L.G. Lam Can Tho University, Can Tho, Vietnam

S. Chaiyaput King Mongkut’s Institute of Technology Ladkrabang, Bangkok, Thailand

ABSTRACT: To analyze the ground movement and the influence of shield tunneling above tunnel struc­ tures, the continuum ground model with segmental lining is in use. The analysis region in the transverse section should confirm that the side boundaries do not give significant effects on the analysis results. On the other hand, the lower analysis region is selected based on the empirical value to fit the analysis result for the measured data in practice. The effects of tunnel excavation in the numerical analysis are classified into two parts: one is the unloading effect because of the removal of the soil weight at the tunnel section, and the other is the effect of stiffness reduction inside the tunnel. The former heaves the ground around the tunnel, of which the magnitude increases as the distance from the tunnel invert to the bottom of the ana­ lysis region increases. This influence increases as the ground stiffness decreases. This is because the numerical analysis is based on the stress–strain relationship and the displacement comes from the distance multiplying with strain in the case of the elastic model. The latter one causes the upper part of the ground around the tunnel to move toward the tunnel, which decreases the ground reaction. The analysis results include both effects. This study shows both effects on ground displacements separately, taking the lateral region size W and the lower region size d as parameters, and discusses the method to evaluate the analysis results.

1 INTRODUCTION Tunnel structures require very high standards of safety, not only from the aspect of the construction process, but also from the viewpoint of their service­ ability. Soil–structure interaction is of utmost importance in the analysis of the response of tunnel structures. Two main models evaluate the response of shield tunnel structures and the effect of grounds: the beam-spring model and the continuum ground model. The beam-spring model (Yamaguchi 1978, ITA 1999) simulates two parallel segment rings, where each ring consists of beam elements representing segments. Rotational spring elements with a rotational stiffness coefficient kθ represent the radial joints which connect the segments in a longitudinal direction, and shear spring elements with a shearing stiffness coefficient ks represent the ring joints which connect the segment rings. The support of the ground around the tunnel is modeled by springs with a coefficient of ground reaction kr in a normal direction, which has no stiffness on the

tension side. As the ground is only represented by springs, the analysis cannot provide any information with regard to the settlement at the ground surface and to the possible stress and deformation behavior of the ground. The beam-spring method is a fast and simple calculation method. Therefore, it is often applied even though it has limited potential for inter­ pretation with respect to the real situation because of the many simplifications made. It is often used to determine the thickness and required reinforcement of the supporting circular ring following the results of a more sophisticated calculation method. The continuum ground model (Galli 2004, Kasper 2004, Oriol 2012, Ngoc-Anh 2014) takes the tunnel structure and the surrounding ground as a whole under stress and deformation according to the con­ tinuum mechanics. Not only the internal forces and deformation of the lining structure are calculated, but also the stress and deformation of the surround­ ing ground are simulated. It fully reflects the inter­ action of the surrounding ground and the tunnel structure. However, because the surrounding ground and the simulation of the interaction between the

DOI: 10.1201/9780429321559-89

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ground and structures are complex, the continuum ground model is still in the development stage. Com­ pared with the beam-spring method, the continuum ground method fully considers the interaction between the tunnel structure and the surrounding ground. Combined with the specific construction process, it can fully simulate the internal forces of the tunnel structure and the deformation of the sur­ rounding ground in each construction phase, which is more consistent with the engineering practice. At present, the continuum ground method is mainly achieved by the finite element method. The analysis region in the transverse section should con­ firm that the horizontal boundaries have no significant effect on the pend on the size of the analysis region below the tunnel (Tamura 2002, Hisatake 2002). In this paper, effects of the shield tunneling are clas­ sified into two parts. One is the unloading effect because of the removal of the soil weight at the tunnel section, and the other is the effect of the stiffness reduction inside the tunnel because of the replacement of the soil inside the tunnel with the segmental lining. The former one heaves the ground around the tunnel, of which the magnitude increases as the increase of the distance from the tunnel inverts to the bottom of the analysis region. This influence increases as the ground stiffness decreases. The latter one causes the upper part of the ground around the tunnel to move toward the tunnel, which decreases the ground reaction. The ana­ lysis results include both effects. This study shows both effects on ground displacements separately, taking the lateral region size W and the lower region size d as parameters, and discusses the method to evaluate the analysis results. 2 NUMERICAL ANALYSIS 2.1

3D continuum ground model

In this paper, the effects of the analysis region size were studied using a 3D FEM continuum ground model adopting displacement boundary conditions on the excavation surface. The 3D model was implemented in the finite element code DIANA 10.2. The outline of the continuum ground model is shown in Figure 1. Table 1 shows the summary of the element types in the proposed model. The ground was modeled by an 8-node isoparametric solid element. 2.1.1 Lining model The tunnel lining is performed by two or more rings of connection along the tunnel alignment, shown in Figure 2. A ring is composed of eight segments in the circumferential direction. The composition of the segmental lining model was evaluated by 4-node quadrilateral isoparametric curved shell elements. The input parameters of the segmental properties were as follows: Young’s modulus, E; Poisson’s ratio, v; and density, γ.

Figure 1. Boundary condition of the 3D FEM model. Table 1.

Element types in 3D model.

Components

Element

Ground

8-node isoparametric solid element 4-node quadrilateral isopara­ metric curved shell element 8-node plane quadrilateral inter­ face element 2-node torsion spring element

Lining Interaction

Joint

Segment joint Ring joint

Radial direction 2-node Tangential direction translation spring Axial direction element

2.1.2 Joint model The radial joint between segments in a ring was modeled by torsion spring elements with two nodes. The torsion spring generates a moment around the tunnel axis because of the relative rotation of the segments on both sides. The spring constant of the segment joints, kθ, was determined by

Figure 2. Schematic 3D tunnel lining and joint.

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where k'y is a coefficient based on joint type (1 is used here), E is Young’s modulus of the segment, IZ is the moment of inertia, and r is the tunnel radius (ACTEC 1999). The circumferential joint is between rings, shown in Figure 2. Each joint is represented by the shear spring in radial and circumferential directions and the translation spring in axial direction. The spring constant of the ring joint in radial direction, ksr, and circumferential direction, kst, can be expressed as

normal direction against the lining, σn , is equal to the earth pressure at rest, σn0 , when the ground dis­ placement in the normal direction, un , is equal to zero. This means that σn0 acts on the lining at the start of the analysis without deformation. Here, the required displacement to generate σn0 in the interface element is defined as

where kn is the normal stiffness coefficient of the interface element. 2.2

where E and G are Young’s and shear modulus of the segment, respectively; It is the moment of inertia for the area of one joint; Lj is the length between the consecutive joints; b and h are the segment width and height, respectively; and v is Poisson’s ratio (RTRI 1997). 2.1.3 Interaction between ground and lining The interaction between the segment lining and the ground was modeled with the zero-thickness threedimensional interface element Q24IF, which is com­ patible with the solid and shell elements of the model. The Q24IF element is an interface element between two planes in a three-dimensional configur­ ation. The element is based on linear interpolation. Newton-Cotes integration scheme is applied. The characteristics of this interface element were set, as shown in Figure 3, so that the earth pressure in the

Figure 3. Characteristics of the interface element.

Phase analysis

The numerical modeling of the tunnel lining was performed by the 3D FEM with a displacement boundary condition to describe the initial ground dis­ placement, including the gap between the excavation surface and the segmental lining. To clearly separate the unloading effect due to tunnel excavation and the installation of the lining, phase analysis was carried out with three calculation phases, that is, the initial stress analysis, unloading, and stiffness reduction inside tunnel, as illustrated in Figure 4. Phase 1: Initial stress analysis. Before excavation, the self-weight of the ground is loaded to establish in situ stresses in the soil profile concurrent with no displacement. Phase 2: Unloading. The unit weight of the soil inside the tunnel γs is changed to the equivalent unit weight γred, which is defined as the difference between the equivalent unit weight of the selfweight of the lining and the unit weight of water. The displacement as produced by Phase 2 is shown as

Figure 4. Schematic of sequential analysis.

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Table 3. Properties of the tunnel lining and the ground condition.

Phase 3: Stiffness reduction inside tunnel. The shell elements representing the lining model and the interface element between the ground and the lining are installed. The characteristics of the interface element in Figure 3 are adopted by applying enforced displacement u0 to the excavation surface. The displacement as produced by Phase 3 can be expressed as

It is noted that the results of the ordinary method to analyze tunnel excavation and lining installation simultaneously are equal to the results produced by Phase 2 and Phase 3 because of elastic analysis. The comparison between the ordinary method and the proposed two-phase method to simulate the tunnel excavation and lining installation is shown in Table 2.

Component Segment Radius (m) Width (m) Thickness (m) Young’s modulus (GN/m2) Poisson’s ratio Density (kN/m3) Joints Segment J. spring const. (MN-m/rad/m) Ring J. radial spring const. (MN/m/m) Ring J. tangential spring const. (MN/m/m) Ring J. axial spring const. (MN/m/m) Ground Vertical earth pressure at tunnel crown (kN/ m2) Water pressure at tunnel crown (kN/m2) Submerged density (kN/m3) Water density Coefficient of earth pressure at rest Kh0 Coefficient of ground reaction kn (MN/m3) Young’s modulus (MN/m2)

value

4.0 1.2 0.376 33.0 0.2 28.0 43.9 511 1080 166 342.27 300.8 5.5 10.0 0.5 10 30.11

3 EFFECT OF REGION SIZE ON SIMULATED GROUND DISPLACEMENTS To study the effect of region size on simulated ground displacements above the tunnel, sensitivity analysis is conducted by changing the values of region width W and depth d under the tunnel. 3.1

Analysis parameters

Table 3 shows the properties of the lining and the ground conditions for the analysis, which were set based on a site data. The concrete segmental ring has a diameter of 8 m, width of 1.2 m, and thickness of 0.376 m. The following empirical equation (RTRI 2002) was used as the relationship between Young’s modulus of the ground in the 3D FEM continuum model, E, and the coefficient of ground reaction, kn.

where α is the factor for the test method of E, BV is the equivalent diameter of the tunnel, and r is the radius of the tunnel. 3.2

Figure 5 and Table 4 show the 3D FEM model and simulation cases for the effect of region size. 4 RESULTS AND DISCUSSION OF THE EFFECT OF REGION SIZE 4.1

Table 2. Comparison between the ordinary method and the proposed two-phase method.

Ordinary Excavation Lining Phase 2

Unloading

Phase 3

Stiffness reduction

Load

Stiffness

− Self-weight of soil + Self-weight of lining Sum of ordinary method No change

− Removing soil + Segment installation No change Sum of ordinary method

Analysis model

Effect of lateral region size

By changing the lateral region size W from 1H to 5H, the displacements of ground surface ΔU2 and ΔU3 during the stage of unloading and the stage of stiffness reduction inside the tunnel are simulated, respectively, which are shown in Figures 6 and 7. Figures 8 and 9 show the displacements of excava­ tion surface ΔU2 and ΔU3 (+: outward of tunnel) during the stage of unloading and the stage of stiff­ ness reduction inside the tunnel, respectively. Ground displacements at points of ground surface and excavation surface above the tunnel central line are shown in Figures 10 and 11. In these two figures,

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Figure 5. Size parameters of the analysis region.

Table 4.

Analysis cases for the effect of region size. Width of lateral analysis region W (×H) 1

Depth of lower analysis region d (×D)

1 2 3 ○ 4 5

2

○

Figure 8. Displacement of excavation surface (ΔU2).

3

4

5

○

○ ○ ○ ○

○

○

Figure 9. Displacement of excavation surface (ΔU3).

Figure 6. Displacement of ground surface (ΔU2).

Figure 10. Ratio of ground displacement (ΔU2).

Figure 7. Displacement of ground surface (ΔU3).

the ratio of ground displacement is defined as the relative ratio of ground displacement versus the ground movement under the condition of W = 5H.

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Figure 11. Ratio of ground displacement (ΔU3).

Figure 12. Displacement of ground surface (ΔU2).

Based on the results, the ground surface and the excavation surface moved up due to the reduction of the weight inside the tunnel without stiffness changing during the Phase 2 of unloading. The shape of the tunnel ring becomes flat in the horizontal direction during the Phase 3 of stiffness reduction inside the tunnel. This is because the coefficient of lateral earth pressure is less than one and the tunnel crown moves downward. Furthermore, the difference of simulated ground displacements both at the ground surface and at the excavation surface when W is larger than 3H is less than 4% compared with the condition of W = 5H. This indicates that the effect of lateral region size is negligible when W is larger than 3H.

Figure 13. Displacement of ground surface (ΔU3).

4.2

Effect of lower region size

By changing the lower region size d from 1D to 5D, displacements of ground surface ΔU2 and ΔU3 during Phase 2 and Phase 3 are simulated, respect­ ively, which are shown in Figures 12 and 13. Figures 14 and 15 show the displacements of excavation sur­ face ΔU2 and ΔU3 (+: outward of tunnel) during Phase 2 and Phase 3, respectively. Ground displace­ ments at points of the ground surface and excavation surface above the tunnel central line are shown in Figures 16 and 17. In these two figures, the ratio of ground displacement is defined as the relative ratio of ground displacement versus the ground movement under the condition of d = 5D. Clearly, based on the results, vertical rebound deformation under tunnel invert increases with the increase of d. The ground surface and the excavation surface move up due to this rebound deformation during Phase 2. The displacement of the ground sur­ face in vertical direction is almost proportional to d, and the increment of the displacement gradually decreases as d increases. This is because of the effect of the stiffness reduction inside the tunnel due to the replacement of the soil inside the tunnel with the seg­ mental lining. The ground surface and the excavation surface remain unmoved with the increase of

Figure 14. Displacement of excavation surface (ΔU2).

d during Phase 3. 7 The difference of simulated ground displacements both at the ground surface and at the excavation surface when d is larger than 3D is less than 4% compared with the condition of d = 5D. These indicate that the unloading effect greatly

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4.3

Setting of recommended region size

For the above reasons, it is recommended that the lateral region size W and the lower region size d can be adopted as 3H and 3D for FEM analysis, respectively. 5 CONCLUSION This study developed a 3D FEM continuum model with radial joints, circumferential joints, and nontension boundary between the ground and the lining. Furthermore, the ground displacements are simulated by its model to confirm the effect of lateral and lower region size in the case of staggered building. The following conclusions can be made: Figure 15. Displacement of excavation surface (ΔU3).

1. The effect of lateral region size is negligible when the width of analysis model W is larger than 3H, where H is the depth from the ground surface to the bottom of the tunnel. 2. Vertical rebound deformation under the tunnel invert increases with the increase of lower region size d. The ground surface and the excavation surface moved up due to this rebound deform­ ation during the Phase 2 of unloading. The ground surface and the excavation surface remain unmoved, even increasing d during the Phase 3 of stiffness reduction inside the tunnel. The effect of lower region size is negligible when the lower region size of analysis model d is larger than 3D, where D is the tunnel diameter. 3. To avoid the impact of lower region size, ground movement ΔU2 during the Phase 2 of unloading should not be taken over to the Phase 3 of stiff­ ness reduction inside tunnel.

Figure 16. Ratio of ground displacement (ΔU2).

For future research, validating the proposed model using site data on the ground displacement around the tunnel is recommended.

REFERENCES

Figure 17. Ratio of ground displacement (ΔU3).

influences the ground movement below the tunnel and the stiffness reduction inside the tunnel mainly influences the ground movement above the tunnel. Furthermore, to avoid the impact of lower region size, ground displacement ΔU2 during Phase 2 should not be taken over to Phase 3.

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RTRI (Railway Technical Research Institute) 2002. Design Standards for Railway Structures and Commentary (Urban Mountain Tunnel). Tokyo: Maruzen. (in Japanese) Tamura T. & Adachi T. 2002. On the domain allocation of finite element method for tunnel excavation, Journal of JSCE 2002(701):231–242. (in Japanese) Yamaguchi, Y., Kawada, H. & Yamazaki, M. 1978. Com­ parison of some segment design methods. Technical Documents for Structure Design 55: 3–8. (in Japanese)

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

The response of the Europier Terminal Building to the excavation of the T2B basement at Heathrow Airport G.R. Taylor Mott MacDonald, London, UK

ABSTRACT: The new Terminal 2 Concourse B (T2B) is a satellite pier which forms part of the remodelling of the Eastern Campus at Heathrow Airport. The T2B concourse includes a 15m deep, 280m long, 100m wide (max­ imum) basement. The substructure was constructed using both top-down and bottom-up techniques with excava­ tion taking place between one metre thick diaphragm walls embedded between 5m and 7m below formation level. The diaphragm wall retained cuts were supported at a single high level by either a stiff ground slab (top-down) or temporary multi-strand ground anchors (bottom-up). The 3-storey Europier Terminal Building was constructed in 1995 and is of steel frame construction. It is a repetitive modular building, approximately 280m in length in 39 equal bays, using a minimum number of standardised components including glazing. It contains four structural movement joints, approximately equally spaced throughout the length of the structure. The new works were located in close proximity to the Europier Terminal Building with the diaphragm walling taking place within 10m of its foundations; the building was to remain operational throughout basement construction. The ground condi­ tions are typical of the London Basin comprising Made Ground overlying River Terrace Deposits (the Taplow Gravel), which is underlain by the London Clay Formation. The Lambeth Group and Chalk underlie the London Clay Formation at depth. At Heathrow, the groundwater level is typically 2-3m below existing ground level. This paper presents the response of the Europier Terminal Building to the excavation of the T2B Basement, compares the observed movements to those anticipated and discusses the risk mitigation strategy adopted and its efficacy. The T2B underground works were undertaken with minimal impact on the Europier Terminal building which remained operational throughout; the measured movements were less than anticipated in the impact assessment and the sensitive features, the large glass panel glazing units, were unaffected.

1 INTRODUCTION Terminal 2 Concourse B (T2B) is a new satellite pier which forms part of the remodelling of the Eastern Campus at Heathrow Airport. The pier building occu­ pies a strip of north-south land between the two exist­ ing runways and to the east of the current T1 and T2 Terminal buildings (Figure 1). Construction of the T2B building was split into two phases to allow for the most efficient use of the overall stand layout at the air­ port (Figure 2). T2B Phase 2 is an extension of T2B Phase 1, which was completed at the end of 2009. The T2B Phase 2 concourse, which was con­ structed between 2010 and 2013, includes a 15m deep (2-3 levels), 280m long, 100m wide (max­ imum) basement. The substructure was constructed using both top-down and bottom-up techniques with excavation taking place between one metre thick dia­ phragm walls embedded between 5m and 7m below formation level. The diaphragm wall retained cuts were supported at a single high level by either a stiff ground slab (top-down) or temporary multi-strand ground anchors (bottom-up).

This paper presents the response of the adja­ cent Europier Terminal Building to the excavation of the T2B Phase 2 basement, compares the observed movements to those anticipated and dis­ cusses the risk mitigation strategy adopted and its efficacy. 2 GROUND AND GROUNDWATER CONDITIONS The ground profile at the site is typical of the London Basin, comprising Made Ground overlying River Terrace Deposits (the Taplow Gravel), which in turn is underlain by the London Clay Formation (Table 1). The Lambeth Group and Chalk underlie the London Clay Formation at depth. The Made Ground is typically between one and two metres thick, generally consisting of airport apron/pave­ ment, whilst the underlying River Terrace Deposits are up to six metres in thickness. The London Clay extends to a depth in excess of 60m below existing ground level at the site.

DOI: 10.1201/9780429321559-90

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3 EXISTING BUILDINGS

Figure 1. Terminal 2 Concourse B Site Location.

Figure 2. Terminal 2 Concourse B and Europier Building (from Shanghavi et al, 2017).

Table 1.

Strata thicknesses. Thickness

Strata

m

Superficial Deposits River Terrace Deposits London Clay Lambeth Group

1-2 up to 6 > 60 Not proven

The hydrogeology of the near surface deposits in the London Basin is related to the River Terrace Deposits that overlie the London Clay. These deposits contain a few metres depth of water perched on the London Clay with the upper part of the gravels typic­ ally unsaturated. This water forms a minor perched aquifer, which is commonly referred to as the Upper Aquifer. At Heathrow, the groundwater level in the Upper Aquifer is typically two to three metres below existing ground level. The pore pressure within the upper part of the London Clay is generally controlled by the presence of this perched water and typically exhibits a hydrostatic pressure distribution from the standing water level in the Upper Aquifer.

As shown in Figure 1 the T2B site is located to the east of the existing Terminal 2 and Queen’s buildings at Heathrow Airport. The Europier Terminal build­ ing is situated to the northwest and extends into the northern portion of the T2B Phase 2 site (Figure 2). That part of the Europier Terminal building situated immediately adjacent to the T2B Phase 2 structure was to be demolished during T2B Phase 2 construc­ tion; the remainder of the Europier Terminal build­ ing was to remain operational during site development (although ultimately the Europier Ter­ minal structure was to be completely demolished). The remainder of the T2B Phase 2 site comprised operational taxiways/apron and the interface between T2B Phases 1 and 2. The Europier complex consists of the Eurolink which extends from Terminal 1, Eurolounge which sits central to and underneath the link and Europier which branches at approximately 45 degrees north east, from the end of the link. At the time of the T2B Phase 2 works the Europier Terminal building was of the order of 15 years old. It consists of 12 Airbridges accessed via an Arrivals over Departures floor plate configuration and various ramp level accommodations. Loading equipment is stored in a temporary structure area created between the Europier complex building wings. The Europier complex is also served by numerous apron level roadways. The 3-storey Europier Terminal building was con­ structed in 1995 and is of steel frame construction with longitudinal horizontal bracing systems at roof level and vertical bracing systems at Apron to Departures levels (Figure 3 and Figure 4). Lateral roof bracing was provided at the ends of the building only, with internal sections of the building between movement joints being unbraced laterally. This sug­ gests that the columns from Departures level to roof were designed as cantilevers to resist wind loads and other notional horizontal loads. The ‘As-built’ draw­ ings indicate that the main circular columns are con­ crete filled up to 600mm above the structural floors that they support. The slab at Departures level was of composite metal deck and in situ reinforced concrete construc­ tion and acted as a diaphragm to transfer wind and notional horizontal loads to the lower level bracing systems. The corrugated steel-clad roof was supported on curved structural steel rafters which in turn were fixed to splayed feature ‘tree’ structures on the top of the main columns (Figure 4). The Arrivals level structure and bridges to the Stand Fixed Links were also suspended from these feature ‘tree’ structures. The Arrivals level slab was of similar construction to the Departures slab but was not believed to con­ tribute to the stability of the building. The Europier Terminal building was clad with a proprietary cladding/glazing at upper levels; the glazed panels were considered very sensitive to

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length of the structure. The new works were located in close proximity to the Europier Terminal Build­ ing with the diaphragm walling taking place within 10m of its foundations; the building was to remain operational throughout T2B Phase 2 basement construction. 4 POTENTIAL IMPACT ASSESSMENT

Figure 3. Europier Building – Elevation adjacent to pro­ posed Terminal 2 Concourse B Phase 2 Building.

The potential impact of the T2B Phase 2 excavation on the T2B Phase 1 and Europier Terminal buildings was generally carried out in accordance with the widely accepted three-phase approach to potential damage assessment (Mair et al., 1996). The damage potentially caused to the adjacent Terminal buildings by the excavation-induced ground movements was assessed through a staged process in which detailed consideration was given to: (a) The overall structural stability of the build­ ings, and (b) The potential weakness of sensitive features (for example the glazed façades).

Figure 4. Europier Building – splayed feature ‘tree’ structures.

movement/deflection. Each cladding unit consisted of 14 double height, approximately (2x4) m, glass panels in aluminium frames. The support services accommodation at Apron level cladding was con­ structed of stack-bonded blockwork. The steel columns of the building superstructure were founded on a grid of reinforced concrete pad foundations; the ground floor slab consisted of ground-bearing reinforced concrete on crushed con­ crete sub-base constructed on the original Apron slab. The Stand Fixed Link structures were of steel frame construction and appeared to be independently braced but connected to the main Europier Terminal building. The Europier Terminal building was a repetitive modular structure, approximately 280m in length comprising 39 equal bays, using a minimum number of standardised components including glaz­ ing, cladding, air bridges, ceiling panels and roof coverings. It contained four structural movement joints, approximately equally spaced throughout the

The design and implementation of protective measures and associated instrumentation and moni­ toring, as well as corresponding trigger levels, were developed on the basis of the results of this impact assessment. Consideration was given to the likely modes of deformation of the buildings, and to the determin­ ation of the ‘greenfield’ strains likely to result from the box excavation. The methodology adopted for ground movement estimation was based on the case history data presented in CIRIA Report C580 whilst the simple elastic beam theory proposed by Burland & Wroth (1975) was used to determine the potential ‘limiting’ tensile strain that could be developed and the consequent potential damage category through consideration of a two-dimensional transverse sec­ tion taken through the Terminal building structure. Both the wished-in-place case and a simplified construction sequence were considered when modelling the impact of the excavation-induced ground movements caused by the box excavation. When each construction stage was considered, all the exca­ vations completed prior to that stage were included in the analysis for that stage, i.e. the results repre­ sented the cumulative movements induced. An integral part of the impact assessment process is damage categorisation. The damage classification system proposed by Burland et al (1977) and subse­ quently developed upon by Boscardin & Cording (1989) and Burland (1995) was used to assess the impact on the Europier Terminal building and to evaluate the overall structural stability in relation to the effect of excavation-induced ground movement. Detailed consideration was given to the structural nature, current condition and fabric of the building,

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potential lines of weakness and sensitive features. Additionally, cognisance was taken of the building foundations (nature and configuration) and any pre­ vious movement that the structure had undergone. Even if the impact upon the Europier Terminal struc­ ture was classified as slight or very slight overall, the glazed façades may still be susceptible to cracking. Due to the fragile nature of the glass panel clad­ ding, these features are very sensitive to minor movement/deflection, and the possibility of some sections fracturing or cracking a detailed assessment of the glass panels was carried out; both longitudinal and transverse movement was considered. In the absence of allowable movement criteria for the glass cladding panels of the Europier Terminal building, a movement tolerance equal to 50% of that advised for the glass cladding units of the Terminal 2 Concourse B Phase 1 Building was assumed (i.e. ±12.5mm in-out of the longitudinal building plane and ±7.5mm right-left along the longitudinal building plane); the glass panels of the Europier Terminal build­ ing were considered potentially more sensitive to movement/deflection given the less robust nature of the frames to the glazing. In addition, and in the absence of a pre-construction condi­ tion survey of the glazing units, an allowance of 25% of the movement tolerance was made for the original installation and subsequent inservice movement. The Europier Terminal Buildings, including the associated glass panels, were assumed to respond as perfectly rigid bodies to the

differential movements induced by the box exca­ vation. In terms of overall stability potential damage categories from negligible to slight were suggested by the impact assessment. In critical areas, for example at the pinch point between the T2B Phase 2 box excavation and the Europier Terminal Building, it was considered pru­ dent to locally tape windows and internal/external glass walls prior to construction. 5 INSTRUMENTATION AND MONITORING A comprehensive monitoring system was designed to measure the ground and structure movements induced by the T2B Phase 2 construction works, and thus confirm design expectations, allow comparisons of actual movements with trigger levels as well as enable timely implementation of pre-planned mitiga­ tion measures, as appropriate for the trigger levels reached. Both manual and remote data capture moni­ toring systems were incorporated within the layout (Figure 5). The monitoring system to measure the impact of the excavation-induced ground subsidence on the T2B Phase 1 and Europier Terminal buildings comprised: a) 3D prisms located at two different levels on the façades of the buildings that were to be manually read using Total Stations. This system provided detailed building deflection monitoring data.

Figure 5. Instrumentation layout adjacent to the Europier and T2B Phase 1 Terminal Buildings.

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b) Water Levelling System (T2B Phase 1 building only): a series of level gauges located within the ground floor of the building to provide differen­ tial settlement monitoring data. A precise levelling monitoring point was attached to each adjacent level gauge to provide a secondary system. One precise levelling point per gauge was used to enable absolute movement of the system to be established. c) A grid of precise levelling points located at appropriate intervals along the facades of the buildings, thus enabling a check on the vertical displacements measured using the Total Stations and the Water Levelling System; the intervals were based on the nature of the proposed works and adjacent existing infrastructure as well as the layout of the Total Stations and the Water Levelling System. In addition to this instrumentation, which was installed on the buildings, inclinometers were installed both within the diaphragm walls of the T2B Phase 2 basement and the surrounding ground; these instruments extended to the toes of the diaphragm walls, some 15-20m below existing ground level. The purpose of these instruments was to measure the movement of the embedded retaining walls and the ground behind. Movement trigger levels were established for implementation during the construction works in the vicinity of the T2B Phase 1 and Europier Terminal buildings. The trigger level criteria were developed from an understanding of the time it takes to implement the recovery cycle, including discovery, review and decision-making times as well as the modification implementation period (Table 2). When assessing the impact of the ground movements on the existing Terminal Buildings, the results of regular visual inspec­ tions of the buildings were also taken into account as well as the trigger levels specified. On the T2B Phase 2 project the green-amber-red traffic light system was used to represent the condi­ tion of the various parts of the works during construction. Prior to the commencement of any construc­ tion work buildings and infrastructure located within the predicted zone of influence were monitored in advance of the works (background monitoring), to establish base readings and

identify naturally occurring background move­ ments (both seasonal and otherwise), to detect inconsistencies in surveying techniques and to identify potential on-going trends from the ori­ ginal construction. 6 CONSTRUCTION AND BUILDING RESPONSE The embedded retaining walls of the 280m long, 100m wide (maximum), 15m deep basement to the T2B Phase 2 Terminal Building comprised one metre thick diaphragm walls toed 5-7m into the London Clay Formation. Diaphragm wall panel width varied from 3.1m to 7.2m; the smaller width panels were used in areas where the embedded retaining walls were close to existing buildings such as the Europier Terminal and T2B Phase1 buildings. Diaphragm wall installation commenced with those panels closest to the Europier buildings in the fourth quarter of 2010. In terms of construction methodology, a mix of top-down and bottom-up sequences was adopted throughout the footprint of the T2B Phase 2 base­ ment. As shown in Figure 6 a top down construc­ tion methodology was adopted immediately adjacent to the T2B Phase 1 building and the north­ ern part of the Europier building, the diaphragm wall retained cuts being supported at high level by a single stiff ground slab. A bottom-up construction methodology was adopted adjacent to the southern part of the Europier Terminal building with support during construction being provided by temporary multi-strand ground anchors (a Single Bore Mul­ tiple Anchorage system). The 100-200mm nominal diameter SBMAs were installed between 2 and 4m below existing ground level at declination angles of between 30° and 40° to the horizontal; there were two anchors per 3.1m wide wall panel, installed

Table 2. Trigger levels for the Europier Terminal Building. Trigger Value (mm) Direction

Amber

Red

Right/Left In/Out

±5 ±8

±7.5 ±12.5

693

Figure 6. Construction sequences adopted within T2B Phase 2 basement footprint (from Shanghavi et al, 2017).

1.75m apart. Each ground anchor had a total length of the order of 40m with a corresponding free length of the order of 10m. The base slab in the northern section of the T2B Phase 2 basement footprint was groundbearing to resist uplift (hydrostatic) pressure and heave of the London Clay Formation; large diam­ eter (1.5-1.8m) bored tension piles were required as part of the solution. This arrangement contrib­ uted to a reduction in the excavation-induced ground movements and thus the impact on the adjacent existing buildings. In the central and southern sections of the T2B Phase 2 basement footprint where the control of excavation-induced ground movements was of less importance a suspended slab solution including a collapsible void former and underslab drainage (pressure relief pipes incorporated into the internal drainage system of the basement) was used. In the potential impact assessment, the topdown construction methodology was considered to provide a high stiffness excavation support system whereas the bottom-up construction methodology was considered to provide a low stiffness excavation support system. The T2B Phase 2 basement substructure is shown in Figure 7. The time displacement profiles for a number of Precise Level Monitoring Points fixed to the front façade of the Europier Terminal building are presented in Figure 8. As evident from Figure 8 excavation in the vicinity of the Europier Terminal building commenced during the first quarter of 2011; movement had essentially ceased by the end of the first quarter of 2012. The pattern of behaviour exhibited over time by these monitoring points is consistent with their position relative to the T2B Phase 2 construction works. A maximum settlement of approximately

Figure 8. Time displacement profiles for Precise Level Monitoring Points installed on the Europier Terminal building.

10mm is indicated by these measurements. This compares to an anticipated settlement in the potential impact assessment of up to 21mm. The ten most critical differential settlements across the footprint of the Europier Terminal building expressed as a percentage of the amber trigger level are shown in Figure 9. The nature of the data presented in Figures 8 and 9 reflects the position of the Terminal building to the T2B Phase 2 basement excavation works. In Figure 10 the movement vectors for the 3D prisms installed at upper and lower levels on the facades of the Europier Terminal building are presented together with those for the T2B Phase 1 Terminal building. These vectors of overall movement are in general agreement with the dif­ ferential settlement data shown in Figure 9. The data suggests that the Europier Terminal build­ ing is tilting slightly towards the T2B Phase 2 excavation.

Figure 9. Differential settlements from Precise Level Monitoring Points installed on the Europier Terminal building.

Figure 7. T2B Phase 2 basement substructure (from Shanghavi et al, 2017).

694

ACKNOWLEDGEMENTS The contributions of colleagues at MM, Balfour Beatty Ground Engineering and Geotechnical Observations (Dr Andrew Ridley) on the T2B project are gratefully acknowledged. The views expressed in this paper are those of the author and do not necessarily represent those of either MM or HAL (Heathrow Airport Limited).

REFERENCES

Figure 10. Retro target vector plot for points installed on the Europier Terminal building.

7 CONCLUSIONS This paper has presented the response of the Europier Terminal Building to the excavation of the T2B Phase 2 basement at Heathrow Airport, comparing the observed movements to those anticipated and discussed the risk mitigation strategy adopted including construction method­ ology (the adoption of a combination of topdown and bottom-up construction sequences together with the use of ground anchors to restrain the embedded retaining walls during basement excavation) and its efficacy. The T2B Phase 2 works were undertaken with minimal impact on the Europier Terminal build­ ing which remained operational throughout; the measured movements were less than anticipated in the impact assessment and within the move­ ment trigger values set; the sensitive features, the large glass panel glazing units, were largely unaffected as a result.

Boscardin, M.D. and Cording, E.J. 1989. Building response to excavation-induced settlement. In Proceedings of the American Society of Civil Engineers, Journal of Geotech­ nical Engineering, Volume 115, No.1 January 1989, Paper 23066, 1-21. Burland, J.B. 1995. Assessment of risk of damage to build­ ings due to tunnelling and excavations. Invited Special Lecture to IS-Tokyo ‘95: 1st International Conference on Earthquake Geotechnical Engineering, In Ishihara (ed.), Proceedings 1st International Conference on Earthquake Geotechnical Engineering: 1189–1201. Rot­ terdam: Balkema. Burland, J.B. and Wroth, C.P. 1975. Settlement of buildings and associated damage. State of the Art Review. In Pro­ ceedings of the Conference on Settlement of Structures: 611–654. Pentech Press. Burland, J.B., Broms, B.B. and de Mello, V.F.D. 1977. State of the Art Report: Behaviour of Foundations and Structures. In Proceedings of the 9th International Con­ ference on Soil Mechanics and Foundation Engineering, Volume 2, 495–546. Gaba, A.R., Simpson, B., Powrie, W. and Beadman, D.R. 2003. Embedded retaining walls – guidance for eco­ nomic design. CIRIA Report C580. London: CIRIA. Mair, R.J., Taylor, R.N., and Burland, J.B. 1996. Prediction of ground movement and assessment of risk and building damage due to bored tunnelling. In Mair, R.J. and Taylor, R.N. (eds), Geotechnical Aspects of Underground Construction in Soft Ground: 713–718. Rotterdam: Balkema. Shanghavi, H., Straw, J., Patel, R., and Winsor, D. 2017. Heathrow Terminal 2B: delivering the biggest airside basement at Britain’s largest airport. In Proceedings of the Institution of Civil Engineers Civil Engineering, Paper 1600017, 1–6.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Longitudinal structural deformation of shield tunnels induced by overlying excavation H.N. Wu, S. Chen, R.P. Chen, Y. Liu, H.Z. Chen & F.Y. Meng College of Civil Engineering, Hunan University, Changsha, Hunan, P. R. China

S.L. Shen College of Engineering, Shantou University, Shantou, P. R. China

ABSTRACT: This paper investigated the longitudinal deformation behavior of a shield tunnel caused by over­ lying excavation based on a case study from Shenzhen, China. A soil-tunnel interaction model, which combines the structural model proposed by Wu et al. (2015) with the Winkler foundation, was proposed to analyze the deformation behaviors of the longitudinal structure as well as the circumferential joints of the tunnel. By compar­ ing the predicted tunnel settlement with the measured results in the case history of Shenzhen, the effectiveness of the proposed soil-tunnel interaction model was verified. The calculated deformation of the circumferential joints was compared with the leaking location of the groundwater observed in the inner surface of the shield tunnel. The results indicated the groundwater leakage generally occurred between the location with the maximum open­ ing of joint and the location with the maximum shearing dislocation. This further indicated the longitudinal deformation of the shield tunnel was accumulated by both bending opening and shearing dislocation in the joints, rather than pure opening of joints that commonly used in the previous studies and the waterproofing design.

1 INTRODUCTION The shield-driven method has been widely used in the construction of metro tunnels. It is commonly observed that shield tunnels have the problem of dif­ ferential settlement in longitudinal direction during operation (Mair 2008; Shen et al. 2014). The differ­ ential settlement has leaded to a significant longitu­ dinal deformation of the tunnel, resulting in a series of problems such as segment cracks, groundwater leakage, and distortion of the track. The mechanism of longitudinal deformation and taking its effect into account in the design stage has become great issues of concern. One of the key problem involved falls on the proper modelling of soil-tunnel interaction in the longitudinal direction. Many attempts have been made to establish a soil-tunnel model for longitudinal analysis during the last three decades. One of the major difficulties lies with the accurate simulation of the structural behavior of the segmental lining. The shield tunnels are lined with precast concrete segments that con­ nected by bolts to form a lining ring. In the longitu­ dinal direction, the tunnel is composed of a series of segmental rings, with a staggered or non-staggered installation pattern (Shen et al. 2014). Since the joints between rings have a lower stiffness than the segmental rings, joint deformation is the main

element causing tunnel deformation in longitudinal direction. According to field investigation and previ­ ous studies (Wang 2009; Wu et al. 2015), such longi­ tudinal deformation occurs in the combination of bending mode, in which the segmental rings rotate rigidly with opening of the joints, and dislocation mode, in which the two adjacent rings are staggered. The existence of the circumferential joints not only affect the deformation behavior of a tunnel but also cause a different internal force distribution compared with a continuous tubular structure. The effects of the joints should be taken into account in a tunnel model. One way to deal with this issue is to model the seg­ mental rings and joints by individual elements (Koi­ zumi et al. 1988). This method is seldom used due to its computational complexity. The other way is to consider the tunnel lining as a continuous beam or tube with a discounted stiffness, by which simple solutions can be obtained (Shiba et al. 1988; Liao et al. 2008). The tunnel is considered as a homogenous Euler-Bernoulli beam (hereafter called EB beam) with a discounted stiffness by applying a reduction factor, η, to the bending stiffness (EI)eq of the tunnel lining (Bogaards and Bakker 1999; Talmon and Bezuijen 2013). Since an EB beam is in a pure bending condition, the tunnel based on this presents a deformation mode of rotational bending, which fails to capture the actual structural behavior of shearing

DOI: 10.1201/9780429321559-91

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dislocation. Wu et al. (2015) proposed a Timoshenko beam simplified model, which idealized the tunnel as a continuous Timoshenko beam with equivalent flex­ ure and shear stiffness. Since the Timoshenko beam theory takes into account shear deformation and rota­ tional bending effects, the model proposed by Wu et al. (2015) presents the deformation behavior of shearing dislocation of segmental tunnels reasonably. This paper aims to combine the longitudinal struc­ tural model proposed by Wu et al. (2015) and the classical Winkler foundation for soil-tunnel interaction analysis. Based on the soil-tunnel interaction model, the longitudinal deformation behavior of a shield tunnel caused by overlying excavation was investi­ gated based on a case study from Shenzhen, China. 2 METHOD OF ANALYSIS 2.1

Formulation of longitudinal structural model

Figure 1 gives an illustration of the longitudinal model proposed by Wu et al. (2015). The tunnel was idealized as a continuous Timoshenko beam. In order to consider the effect of circumferential joints, Wu et al. (2015) used an equivalent flexural stiffness (EI)eq and an equivalent shear stiffness (κGA)eq for the longitudinal structure, which were derived based on some geometric relationships. The calculation unit is composed by one segmen­ tal ring and the adjacent joint. Bending equivalence and shear equivalence can be conducted based on the following geometric relationships:

where, θ = equivalent rotation angle of the unit; θj = rotation angles of the joint; θs= rotation angles of the ring; u= shear displacement of the unit; uj= shear displacement of the joint; us= shear displace­ ment of the segmental lining.

Figure 1. Illustration of Timoshenko beam simplified model (Based on Wu et al. 2015).

The rotation angles of the joint and the segment lining can be calculated based on the following equa­ tions proposed by Liao et al. (2008):

where, lb= length of the bolt; Es = Young’s modulus of concrete segment; Is = area moment of inertia of the cross-section; Kf is the coefficient of rotational stiffness for the circumferential joint, , where ψ=location of neutral axis angle (Liao et al. 2008); λ= influence factor of circumferential joints; ls= length of the ring. Accord­ ing to Eq. (1), Eq. (3a), and Eq. (3b), the equivalent flexural stiffness of the unit (EI)eq can be obtained (Liao et al. 2008):

where, η = reduction factor of flexural stiffness. Based on the equivalence principle of Eq. (2), the equivalent shear stiffness was derived by Wu et al. (2015). The shear displacement of the joint and the ring can be calculated as follows.

where, γb = shear distortion of bolt; κb= Timoshenko shear coefficient of bolt, and for a circular section, κb = 0.9; Gb= shear modulus of bolt; Ab = cross sec­ tional area of bolt; γs= shear distortion of the ring; κs = Timoshenko shear coefficient of segmental ring, and for an annular cross section, κs = 0.5; Gs= Shear modulus of segmental ring; As= cross sectional area of segmental ring. Substituting Eq. (5a) and Eq. (5b) into Eq. (2), the equivalent shear stiffness (κGA)eq can be expressed by the following equation:

where, ξ is a modification factor take account of the effects of the friction between segments, the sealing gasket, and the tongue and groove joint.

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The internal force and the overall deformation of the tunnel can be obtained using the Timoshenko beam theory. For a Timoshenko beam simplified tunnel, the equilibrium equation, the compatibility condition, and the material law are as follows (Timoshenko 1921):

Figure 2. Timoshenko beam model of tunnel on Winkler foundation.

where M= bending moment of the tunnel, Q= shear force of the tunnel, q= transverse load acted on tunnel, w= deflection of the neutral axis of the beam, φ= rotation angle of the cross section, kc= curvature of the neutral axis; (EI)eq = equivalent flexural stiff­ ness, in which E= Young's modulus, I= area moment of inertia; (κGA)eq= equivalent shear stiffness, in which G= shear modulus, A= cross sectional area, κ= the Timoshenko shear coefficient, which depends on the geometry. For annular cross sections, κ=1/2. The overall deformation of the tunnel is an accumu­ lation of the local deformation of ring and joint. The local deformation, such as the opening of the joint and the dislocation between rings can be derived based on the overall deformation curve and the geometric rela­ tionships. According to Wu et al. (2015), the max­ imum opening of the joint, Δ, and the dislocation between the segmental rings, δ, can be expressed as

of the tunnel can be derived using closed-form solu­ tion introduced by Yin (2000a, b). According to Eq(7a),Eq(8a, b), Eq(9a, b), the dif­ ferential equation for a tunnel resting on a Winkler foundation subjected to any pressure can be obtained from:

where D=(EI)eq, C=(κGA)eq, w=vertical displace­ ment of tunnel; q=function of loading pressure in longitudinal direction. ke= equivalent coefficient of subgrade reaction, ke=kb, in which, k= modu­ lus of subgrade reaction, b=outer diameter of tunnel. The external pressure q(x) can be transformed into a Fourier series:

where L = calculation length of the tunnel, , The general solution to the settlement function of the tunnel, w, is expressed as eq

where, θj = rotation angle of the joint; r = outer diameter of the ring; ψ =location of neutral axis angle, derivation of which is described in Liao et al. (2008); γ = shear angle of the calculation unit. 2.2

in which, c1, c2, c3, and c4 are constants; α, β, and dn are expressed as:

Soil-tunnel model and close-form solution

With the longitudinal structural model proposed by Wu et al. (2015), the soil-tunnel interaction can be analyzed combining Timoshenko beam theory with the elastic foundation theory. Figure 2 presents the proposed soil-tunnel interaction model, which is an equivalent Timoshenko beam resting on a Winkler foundation. When a tunnel is subjected to any given pressure, the internal force and overall deformation

698

If w, and φ are known, the bending moment, M, and shear force, Q, can be derived based on Eq. (7) to Eq. (9).Once the bending moment M and shear force Q is determined, the deformation of the joint can be determined based on Eq. (10), Eq. (11). The above equation holds under the condition , then β in the equation should be ; if substituted by The rotation angle function of the tunnel φ can be obtained based on Eq. (7)-Eq. (9), and Eq. (14), and is expressed as follows.

where c5, c6, c7, and c8 are constants that can be cal­ culated from c1, c2, c3, and c4.

Þ

3 DEFORMATION OF SHIELD TUNNELS INDUCED BY OVERLYING EXCAVATION: A CASE STUDY The project is an urban underground expressway tunnel that located in Guimiao Road in Shenzhen, China. The underground expressway tunnel was con­ structed by cut-and-cover method. The excavation width generally ranged from 39.7 m to 46.0 m, with 52.0 m in some local section. 3.09 km of the expressway tunnel was collinear with Shenzhen metro line 11. The tunnel of Shenzhen metro line 11 was constructed by shield-driven method. The exter­ nal diameter of the twin parallel tunnels was 6.7 m, and the thickness and width of concrete segment were 0.35 m and 1.5 m, respectively. The minimum distance between the bottom of the excavation and the top of the tunnel was 6.2 m. The ground condi­ tion in site is as follows. The top are backfills with thickness of 4 to 8 m. The following is a sand layer with a thickness of 0~4 m. Under that is gravelly clay layer with a thickness of 3 to 12 m, and it is residual soil of weathered granite. The next is the fully weathered granite and the strongly weathered granite. The groundwater level is 2 to 3 m under the ground surface. The expressway tunnel was constructed section by section. On April 7, 2018, the tunnel excavation progress is shown in Figure 3. The tunnel uplift reached 20.6 mm, which exceeded the deformation control limits (20mm). Groundwater leakage was observed at more than one section, for the first time. Moreover, compared with longitudinal joints, cir­ cumferential joints are apt to leak, which could be proved through field observations. Severe leakage occurred in K2+833~K2+852. Furthermore, more attention should be paid to K2+847, the worst sec­ tion, where local intrados concrete of segmental ring flaked away. The soil-tunnel interaction model presented above was adopted to analyze the tunnel deformation caused by overlying excavation in Guimiao Road Project. The calculated equivalent flexural stiffness

The constants c1, c2, c3, and c4 in the settle­ ment function w and the rotation angle function φ can be determined based on the following dis­ placement boundary conditions of the tunnel: Figure 3. Excavation above the shield tunnel of metro line 11.

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(EI)eq and equivalent shear stiffness (κGA)eq of Shenzhen metro line 11 are 1.83×108 kN·m2, 1.97×106 kN/m, respectively. Figure 4 gives a comparison of tunnel deformation between field measurement and the proposed method. As shown in Figure 4, the tunnel generally uplift after overlying excavation. The tunnel uplift calculated by proposed method is generally in good agreement with the results of field measurement. This indicates that the model of TBSM resting on Winkler foundation adopted in this paper is reasonable for soil-tunnel interaction analysis. Moreover, the uplift of the tunnel induced by long-distance excavation follows the superposition principle. Figure 5 marked the location of the shield tunnels, with comparison between the joint deformation, and the uplift of the tunnel. It can be seen that the ground water leakage didn’t fall in the maximum uplift of the tunnel, generally occurred between the location with the maximum opening of joint and the location with the maximum shearing dislocation. This further indicated the longitudinal deformation of the shield

tunnel was accumulated by both bending opening and shearing dislocation in the joints, rather than pure opening of joints. 4 CONCLUSIONS A new soil-tunnel interaction model is derived to investigate the longitudinal structural behavior of the segmental tunnel based on the proposed Timoshenko beam simplified model and the Winkler foundation theory. The governing differential equation and the close-form solution for a segmental tunnel resting on a soft foundation subjected to any given pressure are presented. The model of TBSM resting on Winkler foundation is reasonable for soil-tunnel interaction analysis based on the case study in Shen Zhen, China. The uplift of the tunnel induced by longdistance excavation follows the superposition prin­ ciple. The groundwater leakage generally occurred between the location with the maximum opening of joint and the location with the maximum shearing dislocation. This further indicated the longitudinal deformation of the shield tunnel was accumulated by both bending opening and shearing dislocation in the joints, rather than pure opening of joints that com­ monly used in the previous studies and the water­ proofing design.

ACKNOWLEDGEMENTS This paper is supported by National Natural Science Foundation of China (No. 51878267; 51938005) and Natural Science Foundation of Hunan Proince, China (Grant No. 2019JJ30006).

REFERENCES Figure 4. Comparison of tunnel deformation between field measurement and the proposed method.

Figure 5. Comparison between the ground leakage and the tunnel deformation.

Bogaards, P.J., Bakker, K.J.: Longitudinal bending moments in the tube of a bored tunnel. Numerical Models in Geomechanics. pp 317–321. Numog VII, Graz, Austria(1999) Boussinesq, M.J.: Application des potentiels a l'etude de l'equilibre et du mouvement des solides elas­ tiques, principalement au calcul des deformations et des pressions que produisent, dans ces solides, des efforts quelconques exerces sur une petite partie de leur surface ou de leur interieur: Memoire suivi de notes etendues sur divers points de physique mathe­ matique et d'analyse., pp 722. Gauthier Villars, Paris (1885) Koizumi, J., Murakami, H., Saino, K.: Modelling of longi­ tudinal structure of shield tunnel. Journal of Japanese Society of Civil Engineers, 79–88 (1988) (in Japanese). Li, P., Du, S.J., Shen, S.L., Wang, Y.H., Zhao, H.H.: Timoshenko beam solution for the response of existing tunnels because of tunneling underneath. International Journal for Numerical and Analytical Methods in Geomechanics 40(5), 766–784 (2016) Liao, S.M., Peng, F.L., Shen, S.L.: Analysis of shearing effect on tunnel induced by load transfer along

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longitudinal direction. Tunnelling and Underground Space Technology 23(4), 421–430 (2008) Mair, R.J.: Tunnelling and geotechnics: New horizons. Géotechnique 58(9), 695–736. (2008) SCTC (Shanghai Construction and Traffic Committee) Code for investigation of geotechnical engineering. DGJ08-37-2012. Shanghai (2012) Shen, S.L., Wu, H.N., Cui, Y.J., Yin, Z.Y.: Long-term settlement behaviour of metro tunnels in the soft deposits of Shanghai. Tunnelling and Underground Space Technology 40(1), 309–323 (2014) Shiba, Y., Kawashima, K., Ohikata, M., Kana, M.: Evalu­ ation of longitudinal structural stiffness of shield tunnel under earthquake loading. Journal of Japanese Society of Civil Engineers, 385–394 (in Japanese) (1988) Shin, J.H., Addenbrooke, T.I., Potts, D.M.: A numerical study of the effect of groundwater movement on long-term tunnel behavior. Géotechnique 52(6), 391–403 (2002) Shin, J.H., Kim, S.H., Shin, Y.S.: Long-term mechanical and hydraulic interaction and leakage evaluation of segmented tunnels. Soils and Foundations 52(1), 38–48 (2012) Talmon, A.M., Bezuijen, A.: Calculation of longitudinal bending moment and shear force for Shanghai Yangtze river tunnel: application of lessons from Dutch research.

Tunnelling and Underground Space Technology 35, 161–171 (2013) Timoshenko, S.P.: On the correction for shear of the differ­ ential equation for transverse vibration of prismatic bars. Philo. Mag 41, 744 (1921) Wang, R.L.: Analysis on the longitudinal deformation of shield tunnel of Shanghai metro. Underground Engin­ eering and Tunnels 4, 1–6 (2009) (in Chinese). Wang, R.L., Jia, J., Liao, S.M.: Shanghai subway moni­ toring practice. Tongji University Press, Shanghai (2013) Winkler, E.: Die Lehre von der Elastizitat and Feistigkeit. H. Dominicus, Prague, Czechoslovakia. (1867) Wu, H.N., Shen, S.L., Liao, S.M., Yin, Z.Y.: Longitudinal structural modelling of shield tunnels considering shear­ ing dislocation between segmental rings. Tunnelling and Underground Space Technology 50, 317–323 (2015) Yin, J.H.: Comparative modeling study of reinforced beam on elastic foundation. Journal of Geotechnical and Geoenvironmental Engineering 126(3), 265–271 (2000a) Yin, J.H.: Close-form solution for reinforced Timoshenko beam on elastic foundation. Journal of Engineering Mechanics 126(8), 868–874 (2000b)

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Simulating the water-soil leakage induced deformation around the shield tunnel with material point method X.C. Xie, D.M. Zhang, M.L. Zhou & S.J. Feng Department of Geotechnical Engineering, Tongji University, Shanghai, China

D.M. Zhang & S.J. Feng Key Laboratory of Geotechnical and Underground Engineering of Minister of Education, Tongji University, Shanghai, China

ABSTRACT: Water-soil leakage can induce large soil deformation around the urban metro shield tunnel, and potentially threaten the safety of the metro system operation. It is time consuming and operationally diffi­ cult to conduct experiment to test this leakage process in lab condition. Numerical methods were adopted in many previous studies to investigate this complex process: finite element method (FEM), discrete element method (DEM), and CFD-DEM coupled method. However, none of these numerical methods can simulate the tunnel scale leakage-induced soil failure around the shield tunnel. Therefore, the material point method (MPM) is adopted in this study, which has been widely used to deal with large deformation geotechnical prob­ lems. A case study of the incident happened in Shanghai Metro Line 18 is studied to simulate the large quan­ tities of water and soil inrush into the tunnel due to the construction errors in the freezing process. The performance of surrounding soil caused by leakage process, such as the ground settlement, the mass and the velocity of soil leakage, the soil displacement field, the soil stress filed and the pore pressure field, are then carefully calculated by conducting a reasonable numerical simulation. The evolution law of water-soil leakage process sensitivity is analyzed to suggest possible countermeasures and guidance for preventing such tunnel failure in the future.

1 INTRODUCTION With the development of urban economy, Chinese urban rail transit system is booming. For instance, there will be over 20 metro lines in Shanghai by 2020, with total mileage over 870 kilometres. Because shield tunnelling could reduce the influence of construction on surrounding soil and structures, approximately 67% of Shanghai metro lines are con­ structed by shield machine. However, due to the characteristics of the shield tunnel, water seepage or soil erosion have been the major threat to tunnel safety (Asakura & Kojima 2003). Even serious engineering accidents happen from time to time. In 1995, the water and soil leakage in St. Petersburg Metro Line 1 caused the maximal ground settlement to reach 0.9 meters (Wallis 2002). In 2003, during the construction of a passageway of Shanghai Metro Line 4, a large amount of water and sand rushed into the tunnel due to the failure of the freezing method (STEC 2008). Many disasters that occur in shield tunnels are caused directly or indirectly by water seepage. According to the degree of disaster, it can be divided

into three types of water seepage, seepage erosion and water-soil leakage. For water seepage, only water permeate through tunnel defects. Fine soil par­ ticles can be washed into the tunnels through joint openings while coarse particle form a skeleton around tunnel during seepage erosion. As for watersoil leakage, large quantities of water and soil inrush into the tunnel, which may cause the collapse of tunnels. Many scholars have studied water seepage in shield tunnels from many aspects, including waterresistance of tunnel joints (Shalabi et al. 2009), ground surface settlement caused by long-term seep­ age (Wongsaroj et al. 2007, Mair 2008), and the response of tunnel structures to water seepage (Arjnoi et al. 2009, Zhang et al. 2017). In addition, some safety evaluation methods (Shin et al. 2012) considering seepage in the tunnel have been pro­ posed to provide guidance for similar projects. For seepage erosion, previous studies (Liu & Hou 1991, McDonald & Zhao 2001) indicated that seepage erosion around the tunnel can induce soil void, result­ ing in soil stress redistribution. Moreover, soil voids around the tunnel would reduce the support to the

DOI: 10.1201/9780429321559-92

702

tunnel and cause damage to the tunnel structure. Cur­ rent researches on seepage erosion are mainly focused on dam (Crosta & Prisco 1999, Mattsson et al. 2008, Midgley et al. 2013, Wilson et al. 2013, Ouyang et al. 2015, Fern et al. 2017) and pipeline erosion (Zheng, 2007, Ke et al. 2015, Indiketiya et al. 2017, Tao & Tao 2017). For the problems of soil erosion around the tunnel, most studies (Meguid & Dang 2009, Leung & Meguid 2011) focus on the effect of soil-erosion induced voids around the tunnel on the tunnel structure response without condering the process of erosion process. For water-soil leakage, it is definitely the most ser­ ious disasters among the three types of water leakage. Researchers have conducted a series of tests to study the particle motion under gravity to analyze soil leak­ age (Medina et al. 1998, Slominski et al. 2007), and the development of the stagnant zone boundary were predicted using kinematic models (Nedderman 1995). Numerical simulation methods, such as finite element method (FEM) (Wu et al. 2007), discrete element method (DEM) (Balevičius et al. 2011, Xiang et al. 2018), and CFD-DEM coupled method (Zhang et al. 2019), are also widely used. Due to the constraint of mesh distortion limitation, it is difficult to accurately simulate large soil deform­ ation problems caused by water-soil leakage via the traditional FEM. Although DEM is suitable for simu­ lating the particle motion, however, it’s quitetime­ consuming for the large-scale models . So far, a large number of point-based computational methods are developed to simulate large deformation problems, such as material point method (MPM) (Sulsky et al. 1995), smoothed particle hydrodynamics (SPH) (Lucy 1977, Stefanova et al. 2012), element-free Galerkin method (EFG) (Belytschko et al. 1994), etc. Among these methods, the MPM has been widely used to deal with geotechnical problems (Bandara & Soga 2015, Yerro et al. 2015, Martinelli et al. 2017). In addition, it is convenient to implement suitable constitutive models into MPM for further analysis. Therefore, the MPM is adopted to simulate the water-soil leakage induced ground and tunnel response in this paper.

the tunnel crown is about 23 m. The distance between the centerline of twin tunnels is 12 m. The profile of twin tunnels is shown in Figure 1. During the construction of the passageway, the freezing method was used to reinforce the surround­ ing soil. The freezing area is illustrated in Figure 1. After the freezing holes were drilled and freezing pipes were arranged around the twin tunnels, it was supposed to conduct the freezing process. Unfortu­ nately, water and soil leakage were then found at one of the freezing holes on the tunnel lining. The initial leakage points were located at the invert of the twin tunnels, marked as red circular dots in Figure 1. Since the tunnel is embedded in the sandy silt layer with confined aquifer of high water pressure, the inci­ dent was then happened when the leakage occurred through the freezing hole. Consequently, large tunnel deformation happened, which resulted in more ser­ ious tunnel leakage in turn. The ground surface settle­ ment was measured and the monitoring points are marked in Figure 1. Then many reinforcement meas­ ures were carried out, such as the installation of tem­ porary steel support to tunnel segments, loading of cement bag at the tunnel invert, and grouting Polyur­ ethane into the soil to stop the water-soil inflow. 2.2

Soil properties

Soil layers at the tunnel site are mainly composed of saturated clay and sandy soil. According to the field exploration, the profile of soil layers where the tunnel is located are as follows from the ground surface: plain fill (denoted as layer I by local standard), silty clay (layer II), muddy silty clay (layer III), muddy clay (layer IV), clay (layer V), silty clay (layer VI), and sandy silt (layer VII). Generally, it can be divided into the upper layer of clay and the lower layer of sandy

2 PROJECT OVERVIEW In January 2019, a water-soil leakage incident occurred during the construction of Shanghai Metro Line 18. A large amount of water and soil rushed into the tunnel, causing large tunnel deformation and segment damage. A significant settlement was observed on the surface above the tunnel. 2.1

Background of project

The metro line 18 consists of twin tunnels con­ structed by the shield method. The external and internal diameters of the shield tunnels are 6.6 m and 5.9 m, respectively, and the thickness of tunnel lin­ ings is 350 mm. The buried depth of the tunnel from

Figure 1. Profile of the ground and cross-section of twin tunnels.

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Table 1.

Parameters for soil layers.

Soil layer

γ kN/m3

k m/s

c’ kPa

ϕ’ °

Es MPa

K0 ­

Clay Sandy silt

17.8 18.6

4.1×10-9 4.5×10-7

5 0

31.6 35.0

4.49 7.65

0.48

0.39

silt. The clay has low permeability, low strength, and high compressibility. While the sandy silt is noted as high permeability and high water content. The sandy silt layer contains the confined aquifer with a water head of 3–12 m. The tunnel is located at the interface of the clay and the sandy silt. Therefore, the soil layers in the following simulations can be simplified into two layers, and the parameters are listed in Table 1. 3 MPM SIMULATION The tunnel leage is simplied as a two-dimensional plane strain problem and simulated using the MPM software Anura3D. The numerical simulation model dimensions are shown in Figure 2, including soil and tunnels. The soil is divided into two layers, including the clay layer and sandy silt layer. The Mohr-Colomb model was used for the soils. The pore pressure of the confined aquifer is considered for the sandy silt layer.

Figure 2. Typical model configuration with initial material points.

The two-phase single point material type was applied to the soil in the simulation, hence each material point includes both water phase and solid soil phase calcula­ tion. Since the model is simulated with an explicit time integration scheme, this simulation takes the value of the water bulk modulus as 1% of the physical one in order to reduce the computational time. and will not have a significant influence on the results (Martinelli et al. 2017). The tunnel is modelled as one phase dry material with a thickness of 0.35 m. The Young’s modulus of the tunnel and the Poisson’s ratio are taken as 100000 kPa and 0.2, respectively. The tunnel- soil interface needs to be hydraulically con­ strained to simulate the impermeability of the tunnel structure. Thus, to simulate the hydraulically constrain, the tunnel lining is simplified as a rectangle. The grid is composed of triangular cells with 3 material points per cell. Preliminary simulations were conducted to find the optimum cell size. Thus, a characteristic cell size of 0.75 m for the soil domain is adopted in this study for computational efficiency and accuracy. There is a pre-set leakage opening at the tunnel invert as shown in Figure 2. According to the inci­ dent situation, a large-scale soil leakage has occurred through the freezing hole. Therefore, an opening with an inner diameter of 0.2 m is pre-set to induce a sudden leakage. Solid-liquid displacement is restrained at the leakage point in the initial state. The bottom boundary is restrained from both horizontal and vertical movements, and the left and right-side boundaries are only restrained horizontally. To min­ imize the boundary effect on the soil deformation, the size of the model is set as 50 m × 50 m. The procedure of the MPM simulation is mainly divided into two stages: 1) The initialization of stres­ ses is firstly performed, and five calculation steps are set to calculate the equilibrium state of the soil before the leakage. Then, the initial effective stress and the pore water pressure of the soil can be obtained. 2) Water-soil leakage is then simulated by the removal of leakage fixities. As long as the solid and liquid fixities at the opening have been removed, the saturated soil could move into the tunnel under seepage force. For a practical incident, the sustainable tunnel deformation would occur during the leakage, result­ ing in the new leakage channel on the tunnel lining. In addition, the tunnel lining would be damaged and the capacity of the tunnel lining would be decreased. Thus, the leakage would not end until the countermeasures and remedial measures are adopted. This study focuses on the leakage process and aims to reveal the evolution mechanism of soil leakage around the tunnel. Therefore, the leakageinduced tunnel deformation and the adopted meas­ ures during the described incident case are out of the scope of this work, and shall be investigated in the future studies.

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4 SIMULATION RESULTS 4.1

Ground surface settlement trough

Figure 3 shows the ground surface settlement caused by leakage at the final calculation step. Due to the complicated leakage process, this study simulates water-soil leakage in a single tunnel. The surface settlement is almost zero when exceeds 40 m from the tunnel axis. Therefore the boundary effect on the settlement is minimum as the model width is 50 m. The maximal settlement reaches 0.08 m directly above the tunnel centre. Assuming that the leakage process of the twin tunnels is the same, then the ground surface settlement caused by twin tunnels can be obtained by superimposing two settlement profiles. The measured settlements are plotted in Figure 3, with a maximal settlement of 0.137 m. It can be obtained from the curves that the maximal calculated settlement value induced by twin tunnels is about 0.126 m. The calculated value is only 10% smaller than the meas­ ured value. Furthermore, the calculated ground settle­ ment trough is similar to that of the measured one. Therefore, it can be considered that the MPM can rea­ sonably simulate the large soil deformation problem caused by water-soil leakage in the tunnel. It is worth noting that the accuracy of prediction is affected by the assumptions that the interaction of the twin tunnels is ignored. 4.2

Water and soil leakage

Figure 4 shows the cumulative mass of soil inflowed into the tunnel due to the water-soil leakage at the tunnel invert. The mass of the soil is the total mass of soil inflowed into the tunnel of each time. Under the pressure of the confined water head, the mass of inflowed soil increases rapidly at the initial stage of leakage. When the leakage time reaches 10 s, it has accumulated to 73% of the total mass of the inflowed soil. The weight of the inflowed soil works as a loading on the opening. When the water pressure at

Figure 3. Ground surface settlement trough.

Figure 4. Development of soil leakage.

the leakage opening is lower than the weight of the inflowed soil, then the inflowed soil keeps stable and the leakage stops. It can be inferred that the initial stage of the leakage at the tunnel invert is the most critical stage. Once the leakage occurs without timely treatment, a large amount of water and soil will flow into the tunnel shortly, and then result in the serious incident. Thus, the key point to preventing leakage accidents is early detection and early treatment. The trial simulations shows that the permeability only influences the duration of leakage process rather than the leakage form and the final state. Therefore, the permeability in this study was increased considering the computational efficiency and power. Thus the soil inflowing stops after 40 s in the simulation. 4.3

Soil displacement field

Figure 5 shows the development of the soil vertical displacement field caused by soil leakage. In the ini­ tial stage, the soil within 5 m below the tunnel open­ ing heaves significantly, and the soil on the lateral side of the tunnel and above the tunnel settles obvi­ ously. The settlement area extends to the surface. The vertical soil displacement of the underlying soil layer is almost zero, which can be considered as unaffected area. When the leakage time gets to 50 s, the settlement area and the maximal settlement keep increasing. While the upheaval area below the tunnel is slightly increased. To investigate the evolution process of the soil leakage, the velocity field of soil at the first stage is plotted in Figure 6. It can be seen that the flow soil around the tunnel mainly flows from the lateral side of the tunnel to the opening. The affected area of soil in the lateral side of the tunnel is much larger than that below the tunnel. The velocity of the underlying soil is less than 0.1 m/s. Therefore, while controlling the leakage at the tunnel invert, it is also crucial to take reasonable measures, such as grouting cement paste into the soil, to control the soil displacement in the lateral side of the tunnel.

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Figure 6. Velocity contour plot.

Figure 5. Development of vertical displacement field.

4.4

Soil stress field and pore pressure field

Previous studies (Yerro et al. 2015, Fern 2019) show that the MPM software Anura3D can correctly handle effective stresses and pore pressures with numerical stability. Based on the development of the

soil stress field and the pore pressure field, the evolu­ tion law of soil leakage at the tunnel invert can be analyzed, and the development mechanism of the soil displacement field can be revealed. Figure 7 shows the magnitude of the effective soil stress during the leakage. When the leakage at the tunnel invert starts, the effective stress within 5m below the tunnel is almost zero, which indicates that the soil particles are basic­ ally in suspension under the water seepage. The effective stress of the soil on the lateral side of the tunnel is also significantly reduced. This reflects the development law of the leakage path. Leakage process first takes away the completely destabilized soil below the tunnel, and then the soil with reduced stability on the lateral side of the tunnel continuously moves toward to the tunnel opening, forming a leakage path. It is consistent with the distribution of the soil dis­ placement field. Therefore, to control the leakage at the tunnel invert, it is important to improve the stability and strength of the soil not only below the tunnel open­ ing but also on the lateral side of the tunnel. Figure 7 indicates a quick development of the effective stress field during the early stage of the leakage process. Therefore, early measures are vital to prevent poten­ tial serious accidents. Figure 8 shows the development of the pore pres­ sure during the leakage process. It can be seen that the primary cause of the leakage at the tunnel invert is the

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Figure 7. Development of effective stress field.

Figure 8. Development of pore pressure.

larger difference of water pressure inside and outside the tunnel. With the evolution of leakage, the pore pressure in the deep soil layer is basically unchanged because of the confined aquifer. It can be deduced that even if the subsequent leakage process is gradually stopped under gravity, the high water pressure outside the tunnel still poses a huge risk of leakage at the tunnel invert. Thoroughly blocking the tunnel opening is necessary for stopping leakage. Grouting cementsilicate into the opening and fixing steel plate inside the tunnel can be used as possible remidation measures on site. 5 CONCLUSION Based on a case study of Shanghai Metro Line 18, the mechanism and evolution process of water-soil

leakage was investigated using MPM. The following conclusions are obtained: (1) The calculated maximal ground settlement caused by the soil leakage is in reasonable agreement with the measured value. (2) The leakage at the tunnel invert occurs under high water pressure and the initial stage is the most critical stage. The key point to preventing the leakage incidents at the tunnel invert is early detection and early treatment. (3) The developments of the soil displacement and the soil stress reveal that the evolution law of the leakage process: The soil around the tunnel mainly flows from the lateral side of the tunnel to the tunnel opening during leakage. Therefore, while controlling the leakage at the tunnel invert, it is also necessary to take reasonable

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measures, such as grouting cement paste into the soil, to control the soil displacement on the lateral side of the tunnel. (4) The main cause of the leakage at the tunnel invert is the large difference water pressure inside and outside the tunnel. It is necessary to thoroughly block the tunnel opening, such as grouting cement-silicate into the opening and fixing steel plate inside the tunnel.

ACKNOWLEDGEMENT This study is financially supported by National Nat­ ural Science Foundation of China (Grants No. 41772295, 51978517), Innovation Program of Shanghai Municipal Education Commission (Grant No. 2019-01-07-00-07-456 E00051) and key innov­ ation team program of innovation talents promotion plan by MOST of China (No. 2016RA4059).

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Influencing factors and protection technologies of underground diaphragm wall and deep foundation pit construction on metro station H.F. Xing, L.L. Liu & H. Zhang Department of Geotechnical Engineering, Tongji University, Shanghai, China

ABSTRACT: The construction technology of metro station is particularly critical, especially in loose soft stratum with complex construction environment. Based on the construction of Liyang road interchange station of Shanghai metro Line 10, this paper studies the influence of diaphragm wall construction and deep foundation pit construction on metro interchange station and protection tech­ nology. Firstly, this paper discusses the mechanism of groove wall instability and the relationship between stability and slurry specific gravity in grooving construction of mud retaining wall from both global and local aspects. Then the key technology and protection measures for grooving of dia­ phragm wall are put forward. In addition, the constitutive model of station soil, reflecting the unload­ ing rebound effect of foundation pit excavation, is used to simulate and analyze the dynamic process of foundation pit excavation, which is to obtain the quantitative relationship and regularity of influ­ ence of foundation pit excavation on metro interchange station.

1 INTRODUCTION As an important mean of transportation in modern cities, the subway has developed rapidly in recent years. At present, there is a metro line network extending in all directions underground in many big cities, and some even have com­ mercial buildings and entertainment places underground, forming an underground city together with the metro (Xing et al. 2016; Li & Wang 2019; Wu et al. 2019). Modern metro sta­ tion is no longer a simple underground traffic building, but a large comprehensive multifunctional building which integrates metro sta­ tion, underground shopping mall, pedestrian crossing street and municipal engineering. The construction technology of large-scale compre­ hensive metro station has been paid more and more attention. With the development and appli­ cation of subway underground engineering, the technical problems of subway engineering con­ struction under different geological conditions and engineering conditions are constantly emer­ ging, especially the construction of underground diaphragm wall and the construction of deep foundation pit (Mohamed et al. 2018; Lei et al. 2017; Jia et al. 2019; Li et al. 2019). In view of these technical problems, the engineer­ ing and academic circles have done relevant research from different aspects. These research results pro­ mote the progress of related theories, and also

provide theoretical guidance for the development of subway construction (Sun et al. 2018; Luo et al. 2019). However, with the development of metro con­ struction and the increasing density of metro rail transit lines, the development scale of metro stations presents a three-dimensional intersection, and there are more and more interchange stations with two or more floors. With the increase of the number of interchange stations, the forms of transferring are becoming more and more complicated, and there are higher requirements for environmental protection. A single construction method cannot meet the con­ struction requirements of the hub station, so it is necessary to use a variety of construction methods to build the station. At the same time, in the process of construction, we will encounter the crossing of oper­ ating stations, the reservation for future construction of stations, or the reconstruction of operating sta­ tions. This requires a large number of technical research on the construction of complex interchange stations, and constantly accumulate technical experi­ ence to improve the level of construction technology (Xing et al. 2011; Li et al. 2016). There are many pipelines around the inter­ change station, Liyang road of Shanghai metro Line 10. Some of them are in conflict with the structure of the station. Especially, the station is a typical oblique “cross” interchange hub station with complex geological and hydrological condi­ tions. There is no ready-made experience for reference, and there are great risks in the

DOI: 10.1201/9780429321559-93

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about 27 m long. The standard section of the station is 20.78 m wide, and the port well is 5.18 m wide. The south end of the station is a shield receiving well, and the north end is a shield starting well. The main body and the auxiliary structure of station are constructed by open excavation. The main body enclosure of station is underground diaphragm wall, which is used as the waterstop and retaining structure during the excavation of foundation pit and also as part of the permanent underground station external wall structure. The underground dia­ phragm wall is 800 mm thick, and the excava­ tion depth of standard section is about 16.3 m, and the groove depth is 30.5 m; the excavation depth of end section is about 18.3 m, and the groove depth is 33.5 m. According to the data of the site inspection, the site formation is mainly composed of silt clay, sandy silt, and fine sand and appears as complex strata with partial stratum loss, such as ⑤2, ⑤3, ⑤4, ⑥ and ⑦1 layers. The distribution of soil layers and main phys­ ical and mechanical indicators are shown in Table 1 and Figure 2. The groundwater level of the site is

construction process. Through the research of this project, we can master the influence and control technology of diaphragm wall construc­ tion and prevention in saturated sandy silt layer, foundation pit construction on surrounding envir­ onment and especially the interchange station of oblique operation Line 4. The results can not only ensure the smooth progress of the project construction, the safe operation of metro lines, save construction costs and shorten the construc­ tion period, but also has good theoretical guid­ ance and practical reference value for similar projects. 2 PROJECT OVERVIEW The Liyang road station on Shanghai metro Line 10 is located on Siping road and arranged in a north-south direction, which is oblique to the Hailun road station on Line 4, as shown in Figure 1. The main body of Liyang road station is a two-story rectangular frame structure with a total length of 192 m, of which the intersec­ tion part of Hailun road station on Line 4 is

Figure 1. Layout of the metro interchange station.

Table 1.

Main physical and mechanical indexes of soils.

Layer number

Layer name

Water Thickness content (m) (%)

①1 ②3 ④ ⑤1 ⑤2 ⑤3 ⑤4 ⑥ ⑦1 ⑦2 ⑧1

Fill Sandy silt Clay silt Clay Sandy silt Silty clay Silty clay Clay Sandy silt Fine sand Clay

1.6–3.0 9.6–13.7 1.0–4.9 0.7–13.8 0.0–21.5 0.0–11.2 0.0–2.5 0.0–4.8 0.0–8.7 6.8–11.7 —

— 32.2 48.3 36.7 34.5 34.1 24.7 23.6 28.5 27.6 36.9

Figure 2. Geological profile of the site.

Void ratio

Bulk density Es C (kN/m3) (MPa) (kPa) φ(°) a

KH (10-5) (cm/s)

KV (10-5) (cm/s)

— 0.90 1.36 1.07 1.00 0.98 0.77 0.70 0.80 0.82 1.07

— 18.3 16.9 17.7 17.8 18.2 19.0 19.5 18.8 18.6 17.8

— 11.800 0.037 0.023 25.000 0.427 — — 41.600 42.000 —

— 21.700 0.078 0.036 29.000 0.702 — — 45.700 47.600 —

— 9.85 2.17 3.96 8.27 3.56 5.49 6.28 11.90 12.75 4.77

— 6.0 14.0 17.0 5.0 17.0 17.0 36.0 4.0 3.0 —

— 28.5 11.0 16.0 27.0 16.5 20.5 16.0 32.0 33.0 —

— — 0.42 0.56 0.47 0.38 0.45 — — — —

Note: Es = compression modulus; C = cohesion; φ = angle of internal friction; a = static lateral pressure coefficient; KH = horizontal permeability coefficient; KV = vertical permeability coefficient.

711

relatively high; the burial depth is 0.3~1.5 m, and the average groundwater level is 0.5~0.7 m. 3 DIAPHRAGM WALL CONSTRUCTION 3.1 Theoretical analysis of stability of diaphragm wall The construction of underground diaphragm wall is generally carried out by means of forming grooves, placing steel cages and pouring concrete, and the stability of groove wall during the process of forming the diaphragm wall is the key to ensure the grooving quality of diaphragm wall. At present, the construction techniques that are con­ ducive to the stability of groove wall mainly include ground dewatering, slurry retaining wall and retaining wall structure. Because of the adverse effects of sedimentation on the surround­ ing structures caused by ground dewatering, it must be used carefully. The construction cost of retaining wall structure is high and it is not often adopted. So mud retaining wall becomes the pre­ ferred method for the construction of underground diaphragm wall. However, during the construction process of underground diaphragm wall, the groove wall may be unstable, which can be div­ ided into the following two categories: global instability and local instability. Although the slotting depth of underground diaphragm wall is more than 20 m, the instabil­ ity of groove wall often occurs in the topsoil or in the shallow soil layer about 5-15 m deep. A phenomenon of bulging can be observed in the soil directly below the guide wall. The unstable failure surface will spread along the entire length of the groove on the surface of the wall, which is basically elliptical or rectangular. The instability of shallow wall is the main form of the global instability of the groove wall in the construction of mud retaining wall. When the grooved construction is carried out in the foundation with shallow soft soil or sandy heavy interlayer, or the mud level in groove section fluctuates excessively and the liquid level eleva­ tion decreases sharply, it will lead to overexcavation. This will result in an increase in the filling factor of concrete or impermeable mater­ ial which is subsequently poured, thereby increasing the amount of construction materials and the difficulty of subsequent construction. At present, the theoretical basis for the stabil­ ity analysis of groove wall in underground dia­ phragm wall construction is earth pressure theory. Figure 3 is the schematic diagram of global stability analysis of groove wall. The main influencing factors are the natural proper­ ties of soil (including natural gravity γ, buoyant density γ’, cohesion c, and internal friction angle φ), groundwater depth hw, slurry gravity γs,

Figure 3. The schematic diagram of global stability ana­ lysis of groove wall.

slurry level depth hs, and surface overload q. According to Rankine earth pressure theory, the stability safety factor at any depth of groove wall can be calculated as:

Where Fs = global instability safety factor of groove wall (design value generally requires no less than 1.1 or 1.2); p, ps, pw = mud pressure, horizontal earth pressure and water pressure of the unit soil on groove wall at a certain depth, respectively. Figure 4 is the schematic view showing the global stability analysis of groove wall from threedimensional angle. According to the balance of forces, the calculation formula of safety factor for the global stability analysis can be obtained:

Figure 4. The schematic view showing the global stability analysis of groove wall from three-dimensional angle.

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where ΔP, P, Pf, Ps = the combined force of mud pressure and groundwater pressure, ground load, side cohesion and shear resistance in the depth of groove, respectively; W = sliding body weight; α = the angle between sliding surface and horizontal plane. This project is a primary foundation pit with a safety factor of 1.2. According to the thickness of the site and its physical and mechanical parameters (Table 1), the physical and mechan­ ical indexes of the stratum soil are obtained by using the formulas 3 to 5, and the diaphragm wall size and formula 1~2 are used to satisfy the relationship between slurry gravity and dewatering depth required for the stability of diaphragm wall, as shown in Table 2. Figure 5. Ultrasonic testing results of groove wall in test section before (a) and after (b) improved construction technology.

where γi = the natural gravity of each soil layer above the bottom of diaphragm wall; φi = the internal friction angle of each soil layer above the bottom of diaphragm wall; ci = the cohesion of each soil layer above the bottom of diaphragm wall; hi = The thickness of each soil layer above the bottom of diaphragm wall; H = The total thickness of the soil layers above the bottom of diaphragm wall. As can be seen from Table 2, the ideal treatment measure is that the shallow groundwater dewatering depth around diaphragm wall is 6.0 m and the slurry gravity is 11.0 kN/m3. 3.2

Groove wall detection

After the groove is completed, the integrity of the groove is detected by ultrasonic detection, and the detection result is shown in Figure 5a.

Table 2. depth. Depth (m)

Relationship of slurry gravity with dewatering 0.0

2.0

4.0

6.0

8.0

10.0 12.0

12.5 12 11.4 11 10.7 10.4 10.2 Bulk density (kN/m3) 1.25 1.2 1.14 1.1 1.07 1.04 1.02 Mud proportion

It can be seen from Figure 5a that the cement slurry retaining wall with a specific gravity of 1.10, the ground dewatering and the conventional underground diaphragm wall grooving process are inferior, and there are obvious collapses of groove wall. Combined with the stratigraphic distribution the test section: the upper 1~2 m area is artificial fill, and the collapse area is larger; the stratum are mainly sandy silt and argilla­ ceous soil in the depth range of 5~15 m, and the col­ lapse is also obvious and it presents an irregular collapse trend. It can be seen that the theoretically obtained slurry gravity for wall protection, ground­ water dewatering scheme and conventional under­ ground diaphragm wall construction process cannot effectively prevent the collapse of the groove wall. However, it is mainly clay in the range of 15 m deep, and the integrity of the formation is better. To a certain extent, the reasons for the local col­ lapse of the upper formation are as follows: 1) the theoretical calculation of soil parameters is based on the weighted average of soil layers in the grooved area of underground diaphragm wall. Due to the limitation of the value, the theoretical analysis pro­ cess inevitably strengthens the cohesive force of sandy silt, thus, exaggerating its resistance to col­ lapse and causing instability. 2) The 1~2 m soil layer in the upper part of the test section is artificial fill, and the physical and mechanical properties of the soil are poor. 3) The theoretical analysis does not consider the scouring force of the slurry on groove wall caused by the lifting and lowering of the bucket, and construction disturbance to the groove wall. These aggravates the collapse of the groove wall, especially the upper sandy silt layer of the site. The collapse of the groove wall not only increases the difficulty and workload of the construction, but also greatly increases the filling factor of concrete when the underground diaphragm wall is poured,

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thereby increasing the engineering cost. In order to ensure the quality of the underground diaphragm wall of the Liyang Road station and to save the engineer­ ing cost, the construction technology must be improved and perfected according to the characteris­ tics of the artificial fill and sandy soil in the upper part of the site. 3.3 Improvement of groove wall excavation scheme According to the problem in the test section, the fol­ lowing improvement measures are proposed: (1) use reasonable mud parameters, and the specific gravity of the mud is controlled between 1.1 and 1.25. Increase or decrease the specific gravity and viscos­ ity of fresh slurry according to different depths and soil properties, and control slurry performance during construction; (2) Improve the performance of the retaining wall slurry, and add additives such as tackifier and dispersant to the slurry to ensure the integrity of groove wall in the sandy silt. Table 3 is the slurry preparation table. (3) Adopt the “][“ type integral reinforced concrete structure guide wall to ensure the stability of the artificial fill layer at the top. At the same time, it can better ensure the verti­ cality of the underground diaphragm wall; (4) In the process of grooving, the bucket should be kept stable to ensure the relative stability of the slurry level; (5) Shorten the exposure time of the groove wall, and make full use of reverse suction force caused by negative excess pore water pressure because of instantaneous lateral unloading, so as to increase the stability of the side wall and ensure the integrity of the groove wall. Through the above improvement measures, the construction of the standard section is carried out. After the completion of the groove, the ultra­ sonic inspection method is used to carry out inspection of groove integrity. The result is shown in Figure 5b. After the above improvement measures, the qual­ ity of the groove formation in the complex stratum of the Liyang Road station has been greatly improved, and the integrity of the groove wall has been significantly improved. The concrete filling factor has been reduced from 1.25 of the test point to 1.05 of the standard section, which not only lays a technical foundation for ensuring the construction of underground diaphragm wall in Liyang road sta­ tion, but also greatly saves the engineering cost.

Table 3. Preparation of fresh slurry for underground dia­ phragm wall. Slurry Bentonite material (kg/m3)

Soda ash (Tackifier) (%)

CMC (Dispersant) (%)

Water (kg/m3)

Volume 120

0.4~0.5

0.03~0.05

960

4 FOUNDATION PIT CONSTRUCTION 4.1 Influences of foundation pit construction on operating station The impact of foundation pit construction on operat­ ing station and the surrounding environment is mainly reflected in two aspects: foundation pit exca­ vation and dewatering. The influence range of deep foundation pit excavation depends on the plane shape of excavation, excavation depth and soil con­ dition. Because of excavation unloading, the dis­ placement of pit bottom soil is mainly upward. At the same time, the excavation of foundation pit also causes the horizontal displacement of retaining wall and the displacement of soil outside the wall under the action of pressure difference between the two sides. It can be considered that the main causes of surrounding stratum movement caused by excava­ tion of foundation pit are the uplift of pit bottom soil and the displacement of enclosure structure. In order to ensure the safety and normal operation of foundation pit works, foundation pit dewatering is usually required before the construction. However, during the construction period, dewatering measures are taken inside and outside the foundation pit, and the soil will be consolidated due to the water level drop, which will adversely affect the surrounding environment of the foundation pit. The influence of dewatering on the soil layer around the foundation pit is currently considered in the following two aspects: 1. If the dewatering takes away a lot of soil particles, the water content in the soil layer will be reduced, the floating force will be reduced, and the effective weight of the soil will be increased, so that the soil is consolidated and compressed. 2. If there is no large amount of fine particles taken away by the groundwater during the dewatering, ground settle­ ment can be estimated by the stratification sum method. It is generally believed that during the dewatering period, the soil layer below the dewater­ ing surface does not produce obvious consolidation settlement. Because of the drainage, the soil layer between the original groundwater surface and the dewatering surface will cause settlement under the condition of increasing effective self-weight stress, which is usually the main part of the land subsidence caused by dewatering. 4.2 Theoretical analysis of deformation control measures for foundation pit In this study, the PLAXIS program is used to evaluate the deformation of foundation pit during excavation. The finite element simulation in geo­ technical engineering can adopt a variety of soil constitutive models. In the simulation analysis of foundation pit excavation, considering the unload­ ing expansion and load hardening of the soil, the Hardening-soil model is more suitable for soil excavation.

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The Hardening-Soil model uses three different stiffnesses which are axial load stiffness, axial unloading stiffness and isotropic consolidation stiff­ ness to describe the soil. In the simulation of excava­ tion process, it can accurately reflect the mechanical characteristics of soil under loading and unloading, and reflect the dependence of soil stiffness on stress in soil. In the Hardening-Soil model, the relationship between the vertical strain ε1 and the deviating stress q of soil satisfies hyperbola.

where qa = asymptotic value of shear strength when vertical deformation; E50 = stiffness modulus related to confining pressure under main loading. The yield func­ tion can be expressed as:

where Eur = unloading stiffness; εvp = plastic volumetric strain, using orthogonal flow rule. The foundation pit of Liyang Road is constructed in a certain order of operation. When performing finite element analysis on this structure, nonlinear analysis method should be adopted. The incremental method can not only reflect the stress and deform­ ation at a certain stage of the construction process, but also simulate the construction process more closely, so that the simulation results are closer to the actual situation.

Table 4.

Material parameters of structure members.

Type Diaphragm wall Reinforced concrete struts

ρ(kg/ m3)

A E I(10−4 (m2) (GPa) m4)

2,400

—

31.5

833.3

0.2

2,400

0.8

31.5

426.7

0.2

ν

Note: ρ = density; A = cross area; E = Young’s modulus; I ­ = second moment of area; ν = Poisson’s ratio.

192 m. The length of the interchange section with Hailun Road is about 27 m, the net width of the standard section is 20.78 m, the net width of the end well is 25.18 m, and the thickness of the roof is about 2.5 m. In order to study the influence of foundation pit excavation on the operating station and surrounding environment, the excavation of longitudinal section is selected to simulate the excavation process of the founda­ tion pit. The calculated model width and depth dimensions are 120 m and 90 m, respectively. The finite element analysis model is shown in Figure 6. The soil is analyzed by the constitutive relationship of the hardened soil. According to actual construction conditions, the order of the pit construction is shown in Table 5 (Xing et al. 2016). The finite element simulation results of the scheme of without protective piles on the diaphragm walls on both sides of the operating station are shown in Figure 7. It can be seen from Figure 7 that because the length of the foundation pit on both sides of the operating station is quite different, the length of the south end pit is shorter than that of the north pit. Therefore, the operating station is greatly influenced by the earth pressure of the south

4.3 Numerical simulation of deformation control measures for foundation pit In order to obtain the influence of foundation pit excavation on the No. 4 line and the surrounding environment, the finite element method is used to simulate and analyze the excavation and unloading process. The model analysis is divided into two schemes. The first scheme carries out finite element simulation on the model of the foundation pits on both sides of the station without protective piles. The second scheme carries out finite element simula­ tion on the model of 70.0 m protective piles for the foundation pits on both sides of the station. The soil parameters for the hardened soil model in the simu­ lation analysis can refer to Xing et al. (2016) and the material parameters of the excavated structural mem­ bers are shown in Table 4. The main body of the station is a two-story rectangular frame structure with a total length of

Figure 6. Model diagram of foundation pit excavation with protective piles.

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Table 5.

Pit construction order for the metro station.

Step

Clarification of the status

1 2 3 4 5 6 7 8 9 10 11 12

Underground diaphragm wall construction Pit dewatering and then excavating 0.9 m in depth Installing the first horizontal pit bracing Pit dewatering and then excavating 4.9 m in depth Installing the second horizontal pit bracing Pit dewatering and then excavating 7.9 m in depth Installing the third horizontal pit bracing Pit dewatering and then excavating 11.1 m in depth Installing the fourth horizontal pit bracing Pit dewatering and then excavating 13.9 m in depth Installing the fifth horizontal pit bracing Pit dewatering and then excavating 16.3 m in depth

foundation pit. However, because the south end pit is protected by diaphragm wall and six supports, the effect of the south side earth pressure of the south end pit on the operation station is largely resisted. The average horizontal displacement of the operating station to the north is 11 mm, which cannot meet the requirements for the use of the original structure of the operating station. The ver­ tical deformation caused by the foundation pit excavation on both sides of the operating station is 6.6 cm. The station is between two foundation pits. Compared with the size of the foundation pits on both sides, the station is relatively small. Due to the foundation pit excavation on both sides, the soil is unloaded, and the bottom of the pit and the soil below the bottom will rebound in different degrees. This results in a certain amount of upward vertical displacement of the station in the middle of the foundation pits on both sides. The above scheme can reduce the impact of foundation pit construction on the operating sta­ tion to a certain extent. But the metro station is a building with very high safety requirements, and the scheme cannot meet the requirements for safe operation of the station. The maximum horizontal displacement of the station is 11 mm, which has exceeded the control requirements for the horizontal displacement of the station. Therefore, from the perspective of safety, the above scheme is not recommended. The follow­ ing is a finite element simulation of the 70 m protective pile scheme on both sides of the station. The results are shown in Figure 8. From the above finite element analysis results, the maximum horizontal displacement is 7 mm, and the vertical displacement of the station is 8 mm, and the difference of vertical displace­ ment per meter is 0.17 mm. Compared with the previous scheme, the horizontal displacement, vertical displacement or differential deformation of the station is significantly reduced, and the maximum horizontal displacement is reduced by

Figure 7. The simulation results without protection piles (a) vertical displacement nephogram of model (b) horizon­ tal displacement nephogram of model (c) model shear stress nephogram.

36%. The installation of protective piles in front of the foundation pits on both sides plays an important role in the protection of the operating station and the control of the deformation. On the one hand, the protective piles increase the rigidity of the connected walls on both sides. And it forms a pit protection system with dia­ phragm walls and supporting structures, which enhances the ability of diaphragm walls on both sides to resist lateral deformation. On the other hand, the designed length of the protective pile is 70.0 m, and a very rigid barrier is formed below the soil of the foundation pit on both

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5 CONCLUSION The major conclusions drawn from this evalu­ ation of the effect of underground diaphragm wall and deep foundation pit construction on an existing subway station are as follows:

Figure 8. The simulation results with protection piles (a) vertical displacement nephogram of model (b) horizontal displacement nephogram of model (c) model shear stress nephogram.

sides. From the horizontal displacement cloud diagram of the model, the setting of the protect­ ive pile causes a significant change in the hori­ zontal displacement of pit bottom soil. The protection pile is largely resistant to the increase of earth pressure caused by foundation pit exca­ vation of south end pit, which further reduces the lateral displacement of the operating station.

1. The higher groundwater level of the project site is not conducive to the formation of mud cake, and it is more prone to shallow overall instability. Decreasing the groundwater level by 6 m increases the overall stability of the groove wall and the infiltration capacity of the mud to the upper sandy silt layer, which is more likely to form a muddy effect, while the recharge measures reduce the deformation of the surrounding environment. 2. Using special slurry, adding tackifier and dis­ persant and increasing specific gravity of slurry in sandy silt pertinently can effectively improve the integrity of the groove wall. The “][“ type integral reinforced concrete struc­ ture guide wall is used to ensure the integrity of the artificial fill layer at the top. 3. The impact of foundation pit construction on operating station and the surrounding envir­ onment is mainly reflected in two aspects: foundation pit excavation and dewatering. The Hardening-Soil constitutive model can be used to simulate the impact of different con­ struction schemes on interchange station, the foundation pit uplift, the lateral displacement of pit wall and the surrounding environment settlement deformation. It is found that effective reinforcement measures are taken for the pit bottom soil, which can improve the strength and modulus of the pit bottom soil, effectively control the plastic deform­ ation of the saturated soft soil around the foundation pit, and effectively control the ground settlement outside the pit and the horizontal displacement inside the pit. 4. By laying protective piles, the rigidity of the con­ nected walls on both sides of the station can be increased. The protective piles, diaphragm walls and supporting structure form a foundation pit protection system, which enhances the ability of diaphragm walls on both sides to resist lateral deformation, strengthens the anti-pull ability of the station and reduces the influence of soil rebound on station operations. Therefore, the pro­ tective pile plays an important role in the deform­ ation control of the station.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

A centrifuge modelling study on the effect of foundation configuration on tunnel-frame interaction J. Xu & A.M. Marshall Department of Civil Engineering, University of Nottingham, Nottingham, UK

A. Franza ETSI Caminos, Canales y Puertos, Universidad Politécnica de Madrid, Madrid, Spain

ABSTRACT: Engineers need to assess the effects that underground excavations can have on existing structures. Experimental and field data on the tunnel-building interaction problem that accurately consider building character­ istics are rarely reported. This paper aims to address this shortcoming by providing results from geotechnical centri­ fuge tests where tunnel volume loss was simulated beneath model frame buildings with either raft or separate footing foundations. The paper investigates the influence of the foundation configuration on the soil-structure inter­ actions by comparing the displacements and deformation mechanisms of the geotechnical and structural domains (including the soil, the foundation, and the superstructure). A plane-strain experimental package was used, compris­ ing a 2-storey aluminium framed building model and a flexible membrane model tunnel buried within dense sand. During tunnelling, displacements were measured at the front wall of the centrifuge container using digital image analysis. Three centrifuge tests are reported including a greenfield case and two scenarios of tunnelling centrally beneath a framed building; one with a raft foundation and another with separate footings. Insights into the effects of the structure on the soil movements and deformation mechanisms along with a description of the building dis­ placement and distortions are provided to illustrate how the foundation type alters the soil-structure interaction.

1 INTRODUCTION When considering the construction of a new tunnel beneath existing buildings, geotechnical engineers are tasked with the problem of evaluating the poten­ tial for damage of the existing buildings. For this, it is important to reveal the fundamental mechanisms of the interaction between buildings and tunnelling­ induced ground movements. An increasing number of researchers have realised the importance of considering the realistic building char­ acteristics in the tunnel-structure interaction problem (Boone 1996, Laefer et al. 2009, Goh & Mair 2014, Ritter 2017, Boldini et al. 2018, Fu et al. 2018, Haji et al. 2018, Xu et al. 2020). In particular, the response of framed buildings on shallow foundations to excava­ tion-induced ground movements has been studied both experimentally (Laefer et al. 2009, Xu et al. 2020), and numerically (Son 2015, Goh & Mair 2014, Haji et al. 2018, Fu et al. 2018, Boldini et al. 2018). Recent studies have suggested that differential horizontal displacements in a building with a continuous foundation are negli­ gible (Franzius et al. 2006, Dimmock & Mair 2008), whereas a structure on separate footings can experience large horizontal strain at the ground level (Goh & Mair 2014, Franza & Dejong 2019, Laefer et al. 2009). On the other hand, surface structures will also modify

tunnelling-induced ground movements compared with the greenfield case (Potts & Addenbrooke 1997, Ritter et al. 2017). In the tunnel-structure interaction problem, the foundation type inevitably plays an important role. Unfortunately, there are very few experimental studies available that relate to the influence of foundation type on both structure response and excavation-induced ground movements (Laefer et al. 2009). This paper aims at investigating the influence of foundation configuration on tunnel-soil-structure interactions using data from three geotechnical cen­ trifuge tests, all in dry sand: one greenfield tunnelling and two tests where tunnelling takes place beneath a framed building, either on a raft founda­ tion or separate footings. In particular, the effects of foundation configuration on foundation displace­ ments, structural distortions, and tunnelling-induced ground deformations are presented and discussed. 2 BACKGROUND 2.1

Building deformation parameters

Underground excavation-induced ground movements can cause building shear and bending distortions. Sev­ eral approaches are available to assess these deform­ ation parameters (Cook 1994, Mair et al. 1996, Boone

DOI: 10.1201/9780429321559-94

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1996, Finno et al. 2005, Elkayam & Klar 2018). In particular, Cook (1994) proposed an approach to iso­ late tilt (w), bending, and shear displacements of a structure affected by excavation induced ground movements. Based on this approach, Ritter et al. (2020) derived Equation 1 for the calculation of shear strain γ using top and bottom corner displacements of the bay of interest (see Figure 1(a)).

Table 1. Critical tensile strain and categories of damage (Boscardin & Cording 1989). Category

Level

Limiting tensile

of damage 0 1 2 3 to 4 4 to 5

of damage Negligible Very slight Slight Moderate to severe Severe to very severe

strain (%) 0-0.05 0.05-0.075 0.075-0.15 0.15-0.3 >0.3

2.2 where for Ui;j, i ¼ x; z is the displacement direc­ tion, and j ¼ A; B; C; D is the location of the bay corner; as shown in Figure 1(a), C and D are the two lower corners of the base whereas A and B are the upper corners. As reported by Ritter et al. (2020), the shear strain γ inferred from the displacements of the bay/pannel corner points is equivalent to the angular distortion β as defined by Son & Cording (2007). For a framed building with infill walls, Boone (1996) suggested that the shear strain γ can be esti­ mated based on the slope of the slab with respect to the bay tilt (w) by considering the slab to column/wall connections. In the hinged end beam shown in Figure 1(a), the infill walls experience a uniform shear mode, whereas for fixed end beams (Figure 1(b)), a varying profile of shear strain is obtained along the slab, with a maximum value γ1 . Furthermore, the shear strains (γ0 ) at the foundation level may differ from those within the superstructure due to the impact of ground pressure. Note that shear distortion (γ) and tensile strain (εt;max ) are related by εt;max = γ=2 for no hori­ zontal strains (Son & Cording 2005). Then, the level of building deformation due to shear can be evaluated using the building damage criteria proposed by Bos­ cardin & Cording (1989), as shown in Table 1.

Experimental package and building models

Centrifuge tests were performed on the University of Nottingham Centre for Geomechanics (NCG) 4 m diameter geotechnical centrifuge with an elevated gravity level of 68 g. The plane-strain experimental package developed by Zhou et al. (2014) was used, including a strongbox, a transparent acrylic front wall to allow digital images of the subsurface soil to be taken, an aluminium back wall, a flexible membrane model tunnel filled with water, and a tunnel volume loss control system. A fine dry silica sand, known as Leighton Buzzard Fraction E, was used for the soil. In the experiments, the construction of a tunnel beneath a framed building with either a raft foundation or separate footings was simulated. A model tunnel with a diameter Dt of 90 mm and a cover, C, of 117 mm (C=D ¼ 1:3) was used. Figure 2 shows the tunnelling-related and building parameters in model scale. In particular, the width (bfoot ) and length (equal to the height of each storey, hstorey ) of the footing in the building model are 12 mm and 38.1 mm, respect­ ively, giving a ratio, hstorey =bfoot ¼ 3:2. To achieve the plane-strain condition, all the building elements extended the full width (258 mm) of the centrifuge strongbox in the tunnel longitudinal direction, with a 1 mm gap between the building models and front/ back walls of the 260 mm wide strongbox.

Figure 1. (a) and (b) bay distortion modes; (c) parameter definition (Xu et al. 2020).

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Figure 2. Tunnelling-related and building parameters.

The framed building models were manufactured by machining and welding 3.2 mm thick aluminium plates and angles, as shown in Figure 3. For the frame with raft foundation, the welding process can be summar­ ised as follows (referring to Figure 3): 1) with the foun­ dation plate and the right-side wall being placed perpendicular to each other, their junction was welded together to obtain an L-shaped section; 2) the first angle was welded to the L-shaped section and founda­ tion plate; 3) the second angle was then welded to the first angle and the foundation plate; 4) the rest of the angles were welded in a similar sequence. The walls/ slabs were connected by welding 60% of the length along the longitudinal direction (see the top of the frame model in Figure 3a). To create the frame with separate footings, the raft foundation frame was modi­ fied by cutting out sections of the foundation plate between columns and adding material (epoxy adhe­ sive) to the external footings to achieve the designed footing width (12 mm). To replicate a rough soil-foundation interface, a thin layer of sand was bonded to the underside of the building foundations. Digital image analysis using GeoPIV (White et al. 2003) was used to measure both soil and structure displacements during spin-up and simulated tunnel volume loss. Structure displace­ ments were obtained by tracking white dots painted on the front face of the building models (which were

Figure 4. Setup of a centrifuge model.

first painted matt black). A Dalsa Genie Nano-M4020 monochrome camera and single wavelength lightemitting diodes were used (Xu et al. 2020). 2.3 Experimental procedure Three centrifuge tests including 1 greenfield test and 2 tunnel-frame interaction tests were performed. The experimental process is summarised as follows. 1) The sand was poured into the strongbox in line with the model tunnel (consistent with previous studies (Marshall et al. 2012, Farrell et al. 2014, Franza et al. 2019)), achieving a soil relative density of Id ¼ 90%; the model was then rotated to its upright position. 2) The framed building model was carefully placed on the soil surface at 1 g (not necessary for the greenfield test), and the model was mounted on the centrifuge (see the setup of a centrifuge model in Figure 4). 3) The model was spun to 68 g and 2 sta­ bilisation cycles were performed. 4) Tunnel volume loss Vl;t was simulated by extracting water from the model tunnel (by moving the piston connected to the model tunnel) in increments up to 10%, and digital images were taken at each increment. 3 RESULTS 3.1 Building stiffness and pressure beneath the foundation

Figure 3. Framed building models: (a) raft foundation; (b) separate footings.

In the tunnel-building interaction problem, structure stiffness and self-weight play an important role. For the considered frame models, the foundation impacts both of these aspects. Table 2 summarises, the building

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3.3

Table 2 . Building stiffness, self-weight, and pressure beneath the foundation (prototype scale in plane strain condition). Foundation

EIEB;eq

Self-weight

Pressure beneath

type Raft Footings

(Nm2 /m) 2.1E+11 1.2E+11

(kN/m) 4.9E+04 4.0E+04

foundation (kPa) 23 103

equivalent bending stiffness EIEB;eq (obtained using Euler-Bernoulli beam theory), the total self-weight, as well as the average pressure beneath the foundation. Three-point loading tests were performed at 1 g to evaluate the stiffness from the relationship between load and deflection. The stiffness of the building on a raft foundation was found to be 70% greater than for separate footings, which is a marginal difference in tunnel-structure interaction problems. The former is also slightly heavier because of the cut plates. How­ ever, the pressure beneath the separate footings is much larger than for the raft due to the smaller founda­ tion area of the footings. 3.2

Ground reaction

The ground reaction curves, which describe the change of tunnel pressure at tunnel axis level against volume loss, are presented in Figure 5. It is shown that there is a steep drop of tunnel pressure over a range of low tunnel volume loss (51%) for all tunnelling cases. This is due to soil strength mobilisation combined with soil arching that occurs for shallow tunnelling in dense sand (Iglesia et al. 2014, Franza et al. 2019). The pressure reached the lowest point at Vl;t ¼ 3 - 4%, and increased steadily over higher values of Vl;t . Interest­ ingly, the foundation type primarily affected the ini­ tial pressure, and had a minor effect on the ground reaction.

Tunnel and soil volume loss

Figure 6 displays the relationship between tunnel volume loss Vl;t (the ground loss at the tunnel bound­ ary) and soil volume loss Vl;s (given by the integra­ tion of surface soil settlements). The impact of the building on the surface soil settlement trough volume is minor, with a slight increase of soil volume loss for the raft. For both greenfield and tunnel-building interaction tests, the soil transitioned from overall contractive to dilative behaviour at a Vl;t of approximately 1%, consistent with results from Marshall et al. (2012) for dense sand. This out­ come could be useful for engineers tasked with designing excavations to minimise Vl;s (Franza et al. 2020). 3.4

Ground deformation

Figure 7 Shows the ground movements (horizontal Ux and vertical Uz displacements), and strains (engineering shear γ and volumetric εv strains) for the three centrifuge tests. Note that positive horizon­ tal and vertical displacements are oriented towards the right and down, respectively, and contractive volumetric strains are positive in Figure 7. Further­ more, in these plots, spatial coordinates are normal­ ised by tunnel depth zt . Data greater or lower than the contour thresholds were set equal to the closer limit values, and in the regions that are very close to the model tunnel, data was not available due to the presence of a recess in the acrylic wall. From Figure 7(a), the greenfield ground settle­ ment shows a chimney-like pattern due to the low C=D value (Marshall et al. 2012, Franza et al. 2019), and large shear strains are observed at the tunnel shoulders. Soil directly above the tunnel crown experienced high levels of dilation, whereas the soil experienced intermediate levels of contraction within bands spanning from the tunnel springline to the surface. As discussed by Marshall et al. (2012), the zones of dilation and contraction correspond approximately to the areas of high and low shear strains, respectively.

Figure 5. Tunnel pressure against tunnel volume loss.

Figure 6. Soil volume loss against tunnel volume loss.

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Figure 7. Soil displacements (prototype scale), engineering shear (γ), and volumetric (εv ) strains contours at Vl;t ¼ 2%.

Comparing the plots in Figure 7(b) to those of (a) gives an indication of the effects of the framed build­ ings with raft foundations relative to the greenfield case. The soil horizontal displacement near the surface was largely restricted by the foundation roughness and building pressure towards the external part of the build­ ing, whereas the subsurface movements were margin­ ally affected. This is consistent with the mechanism described by Ritter et al. (2017), that the foundation friction restricted horizontal ground movements. Inter­ estingly, the foundation action (both its pressure and frictional shear stresses) altered the soil volumetric strain distribution near the surface and contributed to the development of a thin shear band at the soilfoundation interface, with a decreased maximum con­ traction level and some areas of dilation now present. On the other hand, soil settlements increased through­ out the ground, accompanying a slight increase of shear strain at the tunnel shoulders.

The effect of the separate footings on tunnelling induced ground deformation is now considered. As displayed in Figure 7(c), the footings restricted hori­ zontal soil displacements with a distinct change in magnitude at the location of the footings, which dif­ fers from the more uniform distribution for the raft foundation. The vertical displacement of the soil dir­ ectly above the tunnel decreased compared with the greenfield and raft foundation cases, accompanied by a decrease of the shear strain. However, localised zones of shear strain are noted at the footing posi­ tions, with dilative response directly beneath all foot­ ings except the central footing, and contraction within the soil between footings. 3.5

Foundation and superstructure displacements

Figure 8 shows the displacements of the structure foundations and underlying soil (prototype scale) at

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Figure 8. Displacements of the foundations of the framed buildings and underlying soil (prototype scale).

low (Vl;t = 0.5%) and high (Vl;t = 2%) tunnel volume loss. As shown in Figure 8(a) and (b), because of the relatively high pressure acting on the soil beneath the separate footings (see Table 2), the separate foot­ ings settled more than the adjacent soil and the greenfield settlement at corresponding locations. The separate footings also result in a higher average settlement than the raft foundation due to a smaller foundation area. Furthermore, the relative deflection (given by the difference in settlement between cen­ tral and external foundation edges) of the frame with separate footings is larger than that of the raft foundation. In relation to horizontal displacements, Figure 8 (c) shows that the raft foundation responded rigidly to tunnelling whereas the separate footings experi­ enced large differential horizontal displacements (δUx ). At the high volume loss Vl;t ¼ 2%, the frame on separate footings moved horizontally towards the right (see the average displacement of the footings in Figure 8(d)), indicating that the frame did not respond symmetrically (as ideally it should have). In particular, the central footing (Foot-4 in Figure 3b) has the largest horizontal displacement among all the footings at Vl;t ¼ 2%. This is due to the fact that the framed building was manufactured by welding angles and plates, as described earlier, not achieving a perfect fixed-fixed connection on both sides of the slab. In fact, the central footing has a completely rigid connection with the right side of the elevated slab whereas the footing was only partly welded

with the left side of the elevated slab (see the frame deformed shape in the next section). 3.6

Frame deformed shape and level of damage

In this paper, shear strains are used to estimate the distortion of framed buildings rather than deflection ratio, similar to Boone (1996). Frame deformed shapes and ground movements for the tests at Vl;t ¼ 0:5 and 2% are shown in Figure 9, using different scale factors. To quantify the deform­ ation levels of the bays and panels across the struc­ tures, indicators are provided in Figure 9, which are associated with a range of shear strain, estimated using methods from Boone (1996) (computed from slab slopes with respect to the building tilt) and Cook (1994) (calculated using Eq. 1), and a category of damage inferred from Table 1. Indicators plotted within bays are indicative of γ values that are calcu­ lated based on the displacements at bay corner points (see Figure 1), whereas indicators of γ0 (raft distortion) and γ1 (elevated slab distortion) are placed on the cor­ responding slabs (foundations or elevated slabs). Con­ sistent with Xu et al. (2020), categories of 0-1, 2 and 3 + are defined as low, medium and high levels of deformation using a colour scheme. Figures 9(a) and (b) show that, for Vl;t ¼ 0:5%, the frame on both a raft foundation and separate footings experienced low levels of deformation based on all approaches. For the frame on separate footings, large horizontal displacements (towards the tunnel) are observed for footings 1-3 and 5-7.

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4 CONCLUSIONS This paper presented data from three centrifuge tests aimed at studying the influence of foundation config­ uration on tunnelling-induced soil-framed building interaction in terms of foundation displacements, structural distortion and ground deformation. The following conclusions can be drawn.

Figure 9. Framed building deformed shapes and level of damage.

In Figures 9(c) and (d), deformed shapes are reported for Vl;t ¼ 2%. For both buildings, bays 2 and 5 experienced medium to high levels of deformation based on Cook (1994) and Boone (1996). It is worth noting the distortion response of slabs at different heights from the ground. Using the approach from Boone (1996), the external (bays 1 and 6) and internal (bays 3 and 4) bays of the frame on separate footings underwent medium levels of distortion throughout the building height, which contrasts with the results from the frame on a raft. In Figure 9(d), the first storey slab tends to have larger deformations than the second storey slab for separate footings, possibly because of the bending moments transmitted by the columns. Finally, the results for the frame on separate foot­ ings in Figure 9 show that the footings embedded into the soil to a greater extent than the raft, and that the ground level columns underwent bending deflec­ tions. Furthermore, the presence of a gap beneath the raft at Vl;t ¼ 2% is clearly distinguishable, whereas vertical separation between the soil and the footings did not occur.

• Tunnelling-induced settlements can induce shear deformations of framed building in-fill walls, both for raft and separate footings. Settlements of the foundation are affected by both the soilstructure interaction and soil volumetric behav­ iour. However, centrifuge data at low volume losses clearly indicated a slight increase in the structure settlement level with respect to green­ field conditions. Interestingly, the frame on separ­ ate footings settled more than the building on the raft foundation. • Centrifuge testing confirmed that a frame with separate footings exhibits differential horizon­ tal displacements, whereas minimal horizontal strains occur for the raft. With respect to soil displacements beneath the foundation, both foundation types (raft and separate footings) partially restricted the soil horizontal displace­ ments. On the other hand, contrary to the raft in which a gap formed at medium volume losses, no vertical separation occurred between footings and the underlying soil. • For both foundation configurations, the engineer­ ing shear and volumetric strains of the soil near the surface were altered by the foundation actions, particularly at the foundation-soil interface where zones of localised shear strains and decreased compression/increased dilation were observed (compared to the greenfield case). The existence of the structure affected subsurface soil deform­ ations, with different mechanism occurring depending on the foundation type. • The shear deformation level of the frame was only slightly affected by the foundation type for the considered scenarios. However, the founda­ tion scheme should be considered in the case of serviceability criteria based on settlements.

ACKNOWLEDGEMENTS This project has received funding from the European Union’s Horizon 2020 research and innovation programme under the Marie Sklodowska-Curie grant agreement No 793715. The first author also recog­ nises the financial support provided by the China Scholarship Council (CSC) and the University of Nottingham, UK.

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Iglesia, G. R., H. H. Einstein, & R. V. Whitman (2014). Investigation of soil arching with centrifuge tests. Jour­ nal of Geotechnical and Geoenvironmental Engineering 140(2), 04013005. Laefer, D. F., S. Ceribasi, J. H. Long, & E. J. Cording (2009). Predicting RC frame response to excavation-induced settlement. Journal of geotechnical and geoenvironmental engineering 135(11), 1605–1619. Mair, R. J., R. N. Taylor, & J. B. Burland (1996). Predic­ tion of ground movements and assessment of risk of building damage due to bored tunnelling. In R. J. Mair & R. N. Taylor (Eds.), Proceedings of the International Symposium on Geotechnical Aspects of Underground Construction in Soft Ground, London, United Kingdom, pp. 713–718. Rotterdam: Balkema. Marshall, A. M., R. Farrell, A. Klar, & R. Mair (2012). Tunnels in sands: the effect of size, depth and volume loss on greenfield displacements. Géotechnique 62(5), 385–399. Potts, D. M. & T. I. Addenbrooke (1997). A structure’s influence on tunnelling-induced ground movements. Proceedings of the ICE - Geotechnical Engineering 125 (2), 109–125. Ritter, S. (2017). Experiments in tunnel-soil-structure interaction. Ph.D. thesis , Cambridge University. Ritter, S., G. Giardina, M. J. DeJong, & R. J. Mair (2017). Influence of building characteristics on tunnelling-induced ground movements. Géotechnique 67(10), 1–12. Ritter, S., G. Giardina, A. Franza, M. J. Dejong, & A. M. Asce (2020). Building Deformation Caused by Tunneling: Centrifuge Modeling. Journal of Geotech­ nical and Geoenvironmental Engineering 146(5), 1–17. Son, M. (2015). Response analysis of nearby structures to tunneling-induced ground movements in sandy soils. Tunnelling and Underground Space Technology 48, 156–169. Son, M. & E. J. Cording (2005). Estimation of building damage due to excavation-induced ground movements. Journal of Geotechnical and Geoenvironmental Engin­ eering 131(2), 162–177. Son, M. & E. J. Cording (2007). Evaluation of Building Stiffness for Building Response Analysis to Excavation-Induced Ground Movements. Journal of Geotechnical and Geoenvironmental Engineering 133 (8), 995–1002. White, D., W. Take, & M. Bolton (2003). Soil deformation measurement using particle image velocimetry (PIV) and photogrammetry. Géotechnique 53(7), 619–631. Xu, J., A. Franza, & M. A. Marshall (2020). Response of framed buildings on raft foundations to tunnelling. Jour­ nal of Geotechnical and Geoenvironmental Engineering 146(11), 1–16. Zhou, B., A. M. Marshall, & H.-S. Yu (2014). Effect of relative density on settlements above tunnels in sands. In W. Ding & X. Li (Eds.), 2014 GeoShanghai Inter­ national Congress: Tunneling and Underground Con­ struction, Volume 242 GSP, Shanghai, China, pp. 96–105. American Society of Civil Engineers.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Modelling ground response to TBM tunnelling with active face support T. Xu & W.H. Zhou State Key Laboratory of Internet of Things for Smart City & Department of Civil and Environmental Engineering, University of Macau, Macao, China

A. Bezuijen Department of Civil Engineering, Ghent University, Ghent, Belgium Deltares, Delft, The Netherlands

ABSTRACT: This article presents a simple numerical investigation of ground responses to tunnelboring machine (TBM) tunnelling with active face support. Unlike other numerical modelling in which a flow inward the excavation chamber was assumed, in this modelling a flow inward the ground surround­ ing the face was assumed. The later flow is caused by the penetration of slurry (for slurry shield) or foam (for EPB). The effect of soil layering (semi-confined and unconfined aquifers) was also taken into account. The numerical results show that the ground response to TBM tunnelling with active face support can be significantly influenced by soil layering. In a semi-confined aquifer, at a distance larger than approximately three times of tunnel diameter from the face, the excess piezometric heads are the same at any depths. This finding is helpful to improve the solution for the excess pore pressure caused by TBM tunnelling in a semi-confined aquifer proposed by Xu et al. (2019). The existing analytical solution of excess pore water pressure distribution for the case of semi-confined aquifer matches well with the numer­ ical result. It appears that for an unconfined aquifer, contour of the excess pore water pressures is like a radial distribution. The steady state model of piezometric head for the homogeneous soil results in excellent matches with the values derived from the numerical modelling. For a semi-confined aquifer, the ground surface heave was observed. Though the face collapse potential was found for the unconfined aquifer, the displacement around the face was very small.

1 INTRODUCTION Improper face support probably leads to face instability during tunnel-boring machine (TBM) tunnelling, especially in aquifers. Insufficient support may lead to face collapse and too high may lead to face blow-out. Active face support with pressurised slurry or foam can successfully stabilise the face (Stack, 1992) and hence it has been widely used. With active face support, since the pressure in the excavation chamber is higher than that in the soil, slurry or foam or water from the foam will penetrate into the surrounding soil and thus generates excess pore water pressures. In this condition, the effective face support pressure is reduced. Many analyses have been conducted to investigate the face stability. Some authors assumed that there would be an ideally impermeable layer formed at the face and thus set an uniform (Kasper & Meschke, 2004; Lambrughi et al., 2012; Ukritchon et al., 2017) or a linear face pressure in their analyses (Kim & Tonon, 2010; Zhao et al.2015). Some others considered that the effective face support pressure decreases with the

distance of slurry or foam penetration into the soil in front of the face and proposed some formulas to describe the reduced effective support pressure (Anagnostou & Kovári, 1994; Broere, 2001). How­ ever, only for standstill, there can be a linear pressure over the thickness of the penetrated zone, going from the pressure at the tunnel face to the hydrostatic pressure in the soil. For drilling the penetration of slurry or foam will generate excess pore water pressure in the soil mass. In reality, the cutter head will cut off the soil and carry it into the gap between the TBM and the ground and thus soil-slurry-mix or soil-foam mix rather than ‘clean’ slurry or foam presents at the face. For pure slurry or foam there will be a low permeable layer and the pressure drop will be at the boundary between the slurry or foam and the soil. In this condi­ tion, there will be no low permeable layer formed at the face, but a continuous penetration (Xu & Bezui­ jen, 2019). Since the penetration velocity decreases with the penetration distance, after some time the penetration is very slow and hardly influences the pore pressures in the soil. This has significant

DOI: 10.1201/9780429321559-95

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influence in the pressure transfer at the tunnel face. For this situation the traditional calculation method is valid. The question is then: how is the ground response to TBM tunnelling under such a penetration. This article therefore aims to convince readers that the groundwater flow has to be taken into account and that concepts for the face pressure described in previous research (references) have to be adapted. A simple numerical modelling using a finite element method (FEM) was carried out. The influence of flow generated by the penetration at the face and soil layering are taken into account. Figure 2. Model built-up.

2 NUMERICAL ANALYSIS AND DISCUSSION 2.1

Model built-up

Table 1.

A sketch of TBM tunnelling with active face support through a semi-confined aquifer is shown as Figure 1. An aquifer (sand layer) is overlain by a semi­ permeable layer (peat layer). An impermeable layer is beneath the aquifer. The tunnel is built in the centre line of the aquifer. For an unconfined aquifer, it is assumed that the soil is homogeneous. A two-dimensional (2-D) model was set up in the finite element code PLAXIS 2D, see Figure 2. The model dimensions 480 m × 60 m (length × depth) were selected to balance the boundary effect and the computational efficiency. A 14.5 m diameter tunnel centre is situated at NAP -27.2 m. The inputs for the model are listed in Table 1. Mohr-Coulomb model was selected for the different soil layers. To bring the complex boundary problem to a simple solution, some simplifications and assump­ tions were made. The lining is assumed to be fin­ ished and thus the gap is fully filled with grout and the volume loss behind the shield is neglected. Hence, the tunnel and shield are not modelled. No vertical displacements of tunnel crown and invert are assumed, see Figure 2.

Inputs for the model.

Semi-permeable Parameter layer

Impermeable Aquifer layer

γ’ (kN/ m3) k (m/s)

16

10.5 10-6

4.0×10­ 4

K0 (-) E (kPa)

0.66 1.5×103

v (-) φ (°) ψ (°) c (kPa)

0.15 20° 0° 3

0.47 5.0 × 104 0.3 36.3° 1° 1

0.43 7.5×105 0.3 32° 2 1

The support pressure can be assumed consisting of slurry pressure and excess pressure. The slurry pressure due to slurry self-weight is set as a linear pressure along the depth and the density of slurry is 1300 kg/m3, which was a medium value that moni­ tored in Second Heinenoord Tunnel (COB, 2000). In case of penetration velocity larger than the excava­ tion rate, a discharge is expected at the face. The dis­ charge is assumed as constant. Therefore, a hydraulic boundary of constant discharge at the face is defined. In this condition, according to Broere (2001), the volume of the water discharged by the penetrating slurry is roughly equal to the total pore volume of the excavated material. The dis­ charge per unit area of tunnel was set as, accord­ ing to: q ¼ -n · v

Figure 1. Sketch of TBM tunnelling in a semi-confined aquifer.

16

ð1Þ

with n the porosity of the soil and v the excavation rate of the TBM. In the GHT, the excavation rate is about 40 mm/min and the porosity of the soil is about 0.37. The discharge at the face is estimated as 0.000247 m/s according to Equation (1). According to

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Broere (2001), the starting value of the head at x = 0 is 4.4 m. Hence, Equation (1) is only valid for excavation face pressures higher than 4.4 m height of water and the infiltration rate is higher than the TBM velocity. If not, there should be a head boundary condition. A transient fluid-solid coupled analysis was per­ formed. Undrained condition was assumed to semi­ permeable layer and aquifer, the impermeable layer was set as non-porous. The analysis step lasted 60 mins, corresponding to the normal excavation period of 40 to 60 mins for one ring excavation. Figure 5. Contour of the excess pore water pressures around the face in a semi-confined aquifer (negative value indicates pressure in PLAXIS 2D).

2.2 Numerical result of excess pore water pressures The soil layering may affect the pressures at the face and thus the face stability. Two common types of soil layering are shown in Figure 3. For an uncon­ fined aquifer, contour of the excess pore pressures is a radial distribution, see Figure 4. It is clear from Figure 5 that the magnitudes of excess pore water pressure caused by TBM excavation in a semi-confined aquifer are the same at any depths, when the flow reaches the place where is at least three times of tunnel diameter from the face. This may be helpful for improving the analytical solution for the

excess pore pressure caused by TBM tunnelling in a semi-confined aquifer proposed by Xu et al. (2019). Bezuijen (2001) proposed a steady state model to describe the piezometric head in the case of homoge­ neous soil:

Where ϕ(x) the piezometric head (m) at a distance x (m) in front of the tunnel face, ϕ0 is the piezometric head at the tunnel face (m), and R (m) the radius of the tunnel, assuming a piezometric head of zero far from the tunnel in the pore water. For a semi-confined aquifer, the steady state piezometric head at the face can be derived from the one-dimensional solution of pore pressure generated by TBM tunnelling Broere (2001):

with λ the leakage length of the aquifer (m).

Figure 3. Sketch groundwater flow caused by shield tun­ nelling. (After COB, 1998, drawings by Bezuijen jr, used with permission).

Figure 4. Contour of the excess pore water pressures around the face in an unconfined aquifer (negative value indicates pressure in PLAXIS 2D).

Figure 6. Steady state head change with distance from tunnel face in an unconfined aquifer compared at the axis of the tunnel.

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Figure 6 shows that, for an unconfined aquifer the model of Bezuijen (2001) excellent match the numer­ ical results. For the tunnel in a confined aquifer the agreement is not so good (see Figure 7). The reason is that Eq. (3) is a purely 1D solution and the flow close to the tunnel face has 3D aspects, see Bezuijen & Xu (2018). 2.3

Numerical result of displacements

Vector contour of the ground displacements induced by TBM tunnelling in a semi-confined aquifer can be seen in Figure 8. Neither face collapse nor blow­ out potential was observed. The ground surface heave was clear. The maximum displacement was 226.5 mm. Figure 9 shows vector contour of the

Figure 7. Steady state head change with distance from tunnel face in a semi-confined aquifer compared at the axis of the tunnel.

ground displacements induced by TBM tunnelling in an unconfined aquifer. There was a small blow-out potential at the face, but the maximum displacement around the face was only about 2.5 mm. 3 CONCLUSIONS A simple FEM modelling was carried out. Though some simplifications and assumptions were made, some new findings are found. The steady state model proposed by Bezuijen (2001) has been used to describe the measurements from sites in semi-confined aquifers, but no data from unconfined aquifer was used. In this study, the model was validated by the numerical results. The magnitudes of excess pore water pressure caused by TBM excavation in a semi-confined aqui­ fer are the same at any distance, when the flow reaches the place where is at least three times of tunnel diameter from the face. This is helpful to improve the analytical solution for the excess pore water pressure caused by TBM tunnelling in a semiconfined aquifer proposed by Xu et al. (2019). In that model, the groundwater flow is assumed to be a cylindrically symmetric flow in the horizontal plane, when the flow reaches the place where is tunnel diameter from the face. The soil displacement around the face may sug­ gest that penetration at the face is positive for the face stability. With proper face support, both col­ lapse and blow-out can be avoided. This also depends on the soil layering condition. Further work on the effects of such as diameter, grout pressure, excavation process in the ground response to TBM tunnelling with active face support can be made based on but is not the focus of this study.

ACKNOWLEDGEMENT The research is funded supported by Science and Technology Development Fund, Macao Special Administrative Region of China (File numbers: FDCT/0035/2019/A1 and FDCT/193/2017/A3). Figure 8. Vector contour of the ground displacement for a semi-confined aquifer.

REFERENCES

Figure 9. Vector contour of the ground displacement for an unconfined aquifer.

Anagnostou, G. & Kovári, K. 1994. The face stability of slurry-shield-driven tunnels. Tunnelling and Under­ ground Space Technology, 9 (2): 165–174. Bezuijen, A. & Xu, T. 2018. Excess pore water pressures in front of a tunnel face when drilling in a semiconfined aquifer. ITA-AITES World Tunnel Congress, Dubai, 21­ 25 April 2018. Broere, W. 2001. Tunnel Face Stability & New CPT Appli­ cations. Ph.D. thesis, Delft University of Technology, Delft, The Netherlands. Broere, W. 2015. On the face support of microtunnelling TBMs. Tunnelling and Underground Space Technology 46, 12–17.

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COB (Centre for Underground Construction). 2000. Second Heinenoord tunnel evaluation report, COB report K100­ 06. Gouda, The Netherlands: COB. Kasper, T. & Meschke, G. 2004. A 3D finite element simu­ lation model for TBM tunnelling in soft ground. Inter­ national Journal of Numerical and Analytical Methods in Geomechanics 28, 1441–1460. Kim, S. H. & Tonon, F. 2010. Face stability and required support pressure for TBM driven tunnels with ideal face membrane – Drained case. Tunnelling and Underground Space Technology 25, 526–542. Lambrughi, A., Rodríguez, L. M. & Castellanza, R. 2012. Development and validation of a 3D numerical model for TBM–EPB mechanised excavations. Computers and Geotechnics 40, 97–113. Perazzelli, P., Leone, T., Anagnostou, G. 2014. Tunnel face stability under seepage flow conditions. Tunnelling and Underground Space Technology 43: 459–469.

Stack, B. 1982. Handbook of Mining and Tunnelling Machinery. Chichester, U.K.: John Wiley & Sons. Ukritchon, B., Yingchaloenkitkhajorn, K. & Keawsawasvong, S. 2017. Three-dimensional undrained tunnel face stability in clay with a linearly increasing shear strength with depth. Computers and Geotechnics 88, 146–151. Xu, T. & Bezuijen, A. 2019. Bentonite slurry infiltration in sand, filter cake formation under various conditions. Géotechnique 69 (12), 1905–1106. Xu, T., Bezuijen, A. & Broere, W. 2021. Analytical solu­ tions for groundwater flow caused by TBM tunnelling in a semi-confined aquifer. In preparation. Zhao, C. Y., Lavasan, A. A., Barciaga, T., Zarev, V., Datcheva, M & Schanz T. 2015. Model validation and calibration via back analysis for mechanized tunnel simulations – The Western Scheldt tunnel case. Com­ puters and Geotechnics 69, 601–614.

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Tunnel face stability considering drainage and surface settlements

C. Yi, S. Senent & R. Jimenez ETSI Caminos, Canales y Puertos, Universidad Politécnica de Madrid, Madrid, Spain

ABSTRACT: Tunnels excavated under the water table may need a face support pressure to ensure its stabil­ ity. When such support pressure is too high to be provided by traditional support measures, advance drainage is required. But drainage lowers the water table, hence increasing the surface settlements and their effects on nearby structures. This work studies the surface settlements induced by tunnel face advance drainage bore­ holes, and their relationship with the tunnel face stability, using a numerical model of a shallow tunnel with different values of the applied support pressure. Results show that advance drainage produces a “more stable” situation, as less face support pressure is required for stability, while increasing surface settlements and, con­ sequently, the risk to nearby structures. The critical pressures computed with a rotational Limit Analysis fail­ ure mechanism are also analyzed, preliminarily suggesting that they correspond to the asymptotic values of the stand-up time.

1 INTRODUCTION The face stability of tunnels under the water table is challenging since seepage forces are adverse to face stability (Anagnostou, 2014). Analytical approaches, like wedge-prism mechanisms in the framework of Limit Equilibrium (Anagnostou and Kóvari, 1994 and 1996) or rotational Limit Analysis mechanisms (Pan et al., 2016) can be used to assess the influence of water on the face stability or the need for face sup­ port. However, large support pressure can be unattain­ able in conventional tunneling, hence drainage of the ground ahead of the tunnel face is sometimes required. Zingg (2017) studied the effect of different types of advance drainage (e.g., drainage borehole, pilot tunnel or twin tunnels) under steady state condi­ tions using the Limit Equilibrium wedge-prism mech­ anism of Anagnostou and Kóvari (1994 and 1996). Similarly, Yi et al. (2019), used a Limit Analysis rota­ tional mechanism to discuss the effect of the water table on the collapse mechanism and on the critical pressure of a tunnel face with borehole drains. Both works demonstrate the effectiveness of advance drain­ age to reduce the required face support pressure. Another effect of tunneling under the water table is that they produce surface settlements induced by the tunnel excavation, and by consolidation, which can affect nearby structures. A classic example of this problem is the Holmenkollbanen tunnel in Oslo (Karls­ rud et al., 1978). Anagnostou et al. (2016) studied the surface settlement under transient conditions in soils with different permeabilities, showing that the condi­ tions around the advancing face significantly modify

the settlement profiles. Similarly, Tang et al. (2017) analyzed the surface settlement induced by shallow tunneling in permeable ground, indicating that, for the case history analyzed by them, the effect of drainage boreholes on surface settlements is limited. Several researches (see e.g., Kasper et al., 2006, Yang et al., 2012, Su et al., 2014 and Kim et al., 2017) have reported a clear relationship between face sup­ port pressure and surface settlements for tunnel exca­ vated in dry ground. However, for tunnels under the water table, the relation between the applied face sup­ port pressure and the induced settlement is not yet clear. In this work, we employ a numerical model to analyze the stability of a tunnel face under the water table (with and without advance drainage, and under different support pressures); we also analyze the settle­ ments, assessing how advance drainage affects them. 2 MODEL DESCRIPTION A circular tunnel with diameter D=10m and overbur­ den C=10m, excavated under the water, is considered. The water table is kept at the ground surface (i.e., a sufficient recharge from the far field is assumed) so the initial hydraulic head at the tunnel axis is hw=15m (see Figure 1a). Two drainage boreholes are con­ sidered; they are assumed to be symmetrical in relation to the vertical plane of symmetry of the tunnel, being located in the upper part of the tunnel face with an angle of 45� measured from the crown and with a distance to the center of the tunnel face of r=3.8m (See Figure 1b).

DOI: 10.1201/9780429321559-96

732

Figure 2. Numerical model with advance drainage bore­ holes considered in the analysis.

Figure 1. Schematic of the analyzed problem: a) in the ver­ tical plane of symmetry of the tunnel; b) in the tunnel face cross-section.

3 NUMERICAL SIMULATION Figure 2 shows the numerical model built in FLAC3D (Itasca Consulting Group, 2009) and employed in this work. Due to the symmetry of the problem, and to improve the computational effi­ ciency, only half of the model is considered, with dimensions 35m x 55m x 50m in the x, y and z direction, respectively (136 850 elements). The mechanical boundary conditions are defined to avoid displacements at the lateral and lower boundary. Similarly, and for simplicity, the tunnel perimeter is fixed to avoid its movements, hence representing a rigid lining. The fluid boundary condition at the tunnel face is defined using a mixed boundary condition, as described by Anagnostou et al. (2016), due to the consideration of negative pore water pressures gen­ erated by the ‘‘instantaneous’’ excavation. Accord­ ingly, the tunnel face is set to be impermeable when the pore water pressure at the tunnel face is negative; when the pore pressure is positive, an atmospheric pressure is imposed, therefore allowing the water to flow out from the ground. When advance drainage is considered, the permeability of the elements forming the drain is increased by one order of magnitude

(i.e., 10 times higher than the permeability of the ground), without removing them from the model to avoid any mechanical effect on settlement because of such removal. The soil is modelled as elasticperfectly plastic, with the Mohr-Coulomb failure cri­ terion and a non-associated flow rule (with a dilatancy angle l=0). (Note that more advance material models, with strain-hardening behavior, should be used to accurately predict surface settle­ ments, although a comparison with field observa­ tions is not part of this paper). The strength parameters are a cohesionc0 ¼ 20kPa, and a friction angle ’0 ¼ 25� (we use these values, so that the tunnel face is stable in the short term but collapses in the long term.) The deformability of the ground is defined using a Young’s modulus E=20MPa and a Poisson’s ratio v=0.3. Finally, a permeability k ¼ 10-7 m/s and a water modulus Kf ¼ 108 Pa are assigned. To simulate the support pressure, a uniform hori­ zontal total stress is applied at the tunnel face; although this corresponds to an excavated face sup­ ported with air pressure, it is also used to simulate a face reinforced with, for example, bolts (Anagnos­ tou et al., 2016). Three stages are implemented in the numerical modelling: a) stress and pore water pressure initial­ ization; b) excavation and undrained mechanical equilibrium: in this stage, fluid flow is not allowed so that mechanical equilibrium can be achieved under the negative pore pressures generated ahead of the tunnel face (see explanations below); c) fluidsolid coupled process: in this stage, fluid flow and mechanical process are allowed with the mixed

733

boundary condition explained above, so that the evo­ lution of pore water pressures and settlements can be monitored. Two cases are computed and compared: a) without advance drainage and b) with advance drainage (modelling the drainage boreholes shown in Figure 1b, and keeping all the other parameters constant). The most relevant results related to pore water pressure, tunnel face stabil­ ity, and settlement, are shown and disccussed in the following section. 4 BEHAVIOR OF THE TUNNEL FACE 4.1

Pore water pressures

In the second modelling stage (excavation and undrained mechanical equilibrium), the tunnel face extrudes due to the excavation. Consequently, nega­ tive excess pore pressures develop, having a stabilizing effect on the tunnel face. They dissipate over time due to seepage through the ground ahead of the tunnel face. Therefore, the unsupported tunnel face is stable in the short term, but it may fail depending on the ground strength in the long term due to the dissipation of the negative pore pressures. Figure 3 presents the evolution of pore pressures ahead of tunnel face (along the tunnel axis) for dif­ ferent times. The pore pressure distribution in steady state condition, calculated with a flow-only numer­ ical calculation, is also incorporated as a reference. In the case without drainage (Figure 3a), for t=0h (undrained mechanical equilibrium) there is no seep­ age, producing a maximum negative pore pressure at the tunnel face (y=0m). Far from the tunnel face, negative pore pressures reduce until they become positive at around y=5m. The influence of the tunnel excavation on the pore water distribution becomes negligible, with pore pressure similar to the initial hydraulic head, beyond approximately y=10m. This illustrates that the negative excess pore pressures, and their stabilizing effect, develop in a limited region ahead of the tunnel face and decrease with the distance to the tunnel face. (Note that the collapse mechanism usually develops in this narrow region, see e.g., Pan et al., 2016, Lü et al., 2018). Over time, negative excess pore pressures dissipate due to seep­ age and the tunnel face may become instable (see Section 4.2). Similarly, Figure 3b shows the evolution of pore pressures along the tunnel axis, for different times, when the advance drainage boreholes are considered. Dissipation of negative excess pore pressures and dra-inage seepage flow proceed simultaneously, although the effect of drainage at the tunnel axis (ahead of the face) is initially almost negligible. The later effect of advance drainage on the pore pressure distribution can be observed along the whole length

Figure 3. Pore pressures ahead of the tunnel face along the tunnel axis for different times: a) without advance drainage; b) with advance drainage.

of the drains (y � 10m) and even beyond (Zingg, 2017). 4.2

Tunnel face stability

Since the negative pore pressures dissipate with time, collapse occurs after some time, so a stand-up time can be defined as the time that the face is stable before its collapse. Therefore, support pressure needs to be applied on the tunnel face to prevent fail­ ure. Figure 4a and 4b show the evolution of settle­ ments in the monitoring point at the surface (see Figure 1a) for dif­ ferent support pressures in the two cases con­ sidered: without and with advance drainage. This figure clearly shows that applying a higher support

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Figure 5. Stand-up time vs support pressure. (Limit Ana­ lysis results have been computed using the pore pressure distributions for the steady state condition).

Figure 4. Evolution of the surface settlement at the moni­ toring point for different support pressures: a) without advance drainage; b) with advance drainage.

pressure (i) delays the acceleration of settlements associated to failure and (ii) can even avoid the col­ lapse. Note that other criteria can be used to define the failure of the tunnel face, such as the extension of the plastic zone or the rate of volumetric strains (see Anagnostou et al., 2016). For the case without drainage the face collapses in less than about 3 days for support pressures lower than 50kPa. For 50kPa, the rate of settle­ ments within the time interval considered is small and almost constant. Correspondingly, for the case with advance drainage, the face is stable during the first 3 days even for a support pressure of 37.5kPa. Consequently, drainage is clearly helping to stabilize the face, as less support pressure is required. These results are expected and coherent

with previous works (e.g., Zingg, 2017, and Yi et al., 2019). Figure 5 shows the variation of the stand-up time as a function of the support pressure, for the two drainage cases considered. As previously explained, (i) increasing the support pressure rises the stand-up time; and (ii) for the same support pressure the stand-up time is longer with advance drainage. Moreover, this difference enlarges as the support pressure increases. The reader can note that the stand-up time lines tend towards to an asymptotic support pressure, for which the stand-up time is infinite (i.e., the tunnel face is completely stable). Critical pressures com­ puted with the methodology provided by Yi et al. (2019), employing an upper bound solution in the framework of Limit Analysis, are incorporated into Figure 5 as two dotted vertical lines. (These results have been obtained with the pore pressure distribu­ tion corresponding to the steady state condition.) Although it seems that the analytical results corres­ pond to the asymptotic values, more research (and longer computation) are needed to identify the reasons for the difference. (As an illustration, the time required to calculate the stand-up time in the case of advance drainage and σT ¼ 25kPa on an Inter Code CPU i7-5930k 3.50-GHz PC is larger than 4 days). One possible explanation is that the Limit Analysis mechanism assumes an associated flow rule, whereas, in the numerical model a null dilatancy angle is employed as suggested by Ana­ gnostou et al. (2016). Note, however, that other works with dry soil show a limited effect of

735

Figure 6. Longitudinal surface trough for different support pressures and drainage configurations.

dilatancy on computed values of critical pressure (see Senent et al., 2013).

that the tunnel face become unstable after some time. Increasing the support pressure applied on the tunnel face delays such eventual failure, and can even completely avoid it. Similarly, drainage increases the stability of the face, as less support pressure is required. However, for a given support pressure, advance drainage leads to larger surface settlements due to the effect of (i) the higher seepage flow and (ii) of the reduced support pressure required for stability (i.e., of the increased face stability). Finally, a rotational failure mechanism in the frame­ work of Limit Analysis is used to compute the crit­ ical pressure in both drainage cases, employing the steady state pore water pressure distribution. Although more research is needed, results show that the analytical solution can serve to preliminarily esti­ mate the support pressures required for long term face stability.

REFERENCES 4.3

Settlements

Figure 4 shows that settlements at the monitoring point develop faster for the smaller support pres­ sures, and that settlement magnitude do not seem to be heavily influenced by drainage. To analyze this problem, Figure 6 shows the surface settlement troughs (along the tunnel axis) for the minimum sup­ port pressures required to avoid failure (50kPa and 37.5kPa for the case without and with drainage respectively; see Figure 4). (Note that these values are approximated as we do not carry out a detailed analysis following, for example, the bisection method proposed by Mollon et al., 2009). As it can be observed, settlements with advance drainage and for the minimum support required for stability, are slightly larger than without drainage (with a maximum value of 16.1mm and 12.3 mm respect­ ively). Similarly, settlements for the same support pressure (σT =50kPa) are also larger for the case with drainage. Consequently, it is clear that drainage increases surface settlements and that the larger settlements associated to advance drainage develop due to the combined effect of both the effect of the higher seepage flow and of the less required support pressure. 5 CONCLUSIONS We study the face stability of tunnels under the water table considering the presence of advance drainage and its effect on surface settlements. Employing a numerical model of a shallow tunnel, we analyze the evolution of negative excess pore pressures ahead of the tunnel face, and how they affect the stability, showing that the negative excess pore pressures dissipate gradually due to seepage so

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a case history of Shenzhen Metro. Journal of Zhejiang University-SCIENCE A, 18(10), 757–775. Yang, Y. Y., & Li, H. A. 2012. Failure mechanism of large-diameter shield tunnels and its effects on ground surface settlements. Journal of Central South University, 19(10), 2958–2965. Yi, C., Senent, S., & Jimenez, R. 2019. Effect of advance drainage on tunnel face stability using Limit Analysis and numerical simulations. Tunnelling and Underground Space Technology, 93, 103105. Zingg, S. 2017. Static effects and aspects of feasibility and design of drainages in tunnelling. PhD Thesis. ETH Zurich. (Dissertation Nr. 23729).

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Impact of subsoil spatial variability on deformations of immersed tunnel X. Zhang, X. Wu & W. Broere* Department of Geoscience & Engineering, Delft University of Technology, Delft, The Netherlands

ABSTRACT: In this paper, a probabilistic analysis is implemented to determine the settlement of immersed tunnel elements and the effects of subsoil stiffness variability. A soil-structure interaction model is used to study the effects of subsoil stiffness on the shear behavior of tunnel joints. Site inves­ tigation data is obtained from the Hongkong-Zhuhai-Macau Bridge (HZMB) Tunnel project in China. Two probabilistic methods, the Point Estimate Method and Monte Carlo simulation, are compared when determining the tunnel settlements. The first is computationally more efficient and has sufficient accuracy, while the latter is extremely accurate with higher computational costs. Based on the settle­ ment results, spatial variability of the soil stiffness is quantitatively assessed. The soil-structure inter­ action analysis and derivation of shear forces in tunnel joints is performed by coupling FE analysis to a Monte-Carlo model. The results show that tunnel structure behavior is significantly influenced by the soil parameters uncertainty.

1

INTRODUCTION

Immersed tunnels are mainly built for river, canal, or sea straight crossings where the subsoil is soft. Due to increased traffic demands, there is a tendency to construct larger immersed tunnels with larger size tunnel elements, both in transversal and longitudinal direction, resulting in more critical design issues for the tunnel joints. Examples of this trend are the 6.0km long Hongkong-Zhuhai-Macao Bridge (HZMB) Tunnel and the 18km Fehmarnbelt fixed road and rail link. This not only requires a more detailed analysis of the joints themselves, but also of the total behavior of the tunnel and its interaction with the subsoil. Conventional deterministic design­ ing procedures are based on average soil characteris­ tics with sensitivity analyses including upper and lower boundaries of soil stiffness. The trend of larger elements requires a more sophisticated prob­ abilistic approach with better understanding of impact of uncertainties in site investigations and insight in the probability of exceeding serviceability limit state requirements. Natural soils are often inherently anisotropic due to the manner in which they are deposited. It has been observed that the performance of foundations is considerably affected by the inherent spatial variabil­ ity of the soil properties. According to (Kulhawy 1992) uncertainty in soil behavior can be attributed to three different sources:

DOI: 10.1201/9780429321559-97

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- Variability of the soil parameters - Measurement errors - Transformation errors and uncertainty. The most likely issue affecting immersed elem­ ents is differential settlements, when zones of weaker subsoil will settle more than stiff zones (Grantz 2001a,b). For most immersed tunnels, the tunnel length is much larger than its width, leading to more uncertainties in behavior in longitudinal dir­ ection, whereby differential settlements and rotations may lead to shear failure in the tunnel joints. However, these kilometers long immersed tun­ nels are designed based on a limited offshore site investigation, so it is hard to determine the exact settlements and spatial distribution of soil stiffness along the tunnel. The site investigation is often limited due to economic reasons, so assumptions have to be made based on interpol­ ation from a limited number of borings and on geological models. In this paper, a probabilistic analysis is used to determine the potential tunnel element settlement and the effects of subsoil stiffness variability. Two probabilistic methods, the Point Estimate Method and a Monte Carlo simulation are compared when determining tunnel settlements. Furthermore, spatial variability of the soil stiffness is quantitatively assessed. The results provide insight into the uncer­ tainties in immersed tunnel behavior.

2 PROJECT INFORMATION 2.1 Introduction to The Hong Kong-Zhuhai-Macao Bridge (HZMB) Tunnel The Hong Kong-Zhuhai-Macao Bridge (HZMB) is located at the entrance of the Pearl River, crossing the Lingding Bay in the South China Sea. HZMB consists of three navigable bridges, two artificial islands, and one immersed tunnel. The HZMB Tunnel is composed by a 5664 m-long underwater immersed tunnel and two 326 m cut-and-cover approaches. The underwater tunnel consists of 33 rectangular concrete elements, with a standard length of 180 m each. Each element (shown in Figure 1) is 37.95 m wide and 11.40 m high. 2.2 Geotechnical conditions According to the project geological investigation report (CCCC 2009), Holocene deposits between 10 and 25 m thick are found below the seabed surface. These soils consist of muck or mucky clay mixed with sand, and can be classified as very soft, highly compressible and normally consolidated soils. Below the Holocene deposits, Late Pleistocene deposits are found, with a thickness that varies between 37m and 102m (locally). The Pleistocene deposits appear to be over-consolidated and mainly consist of clay, silty sand, sand and gravel. The sand and gravel generally underlay the cohesive soils. Underneath the Pleistocene deposits, the bedrock is encountered, which is mainly composed of mixed granites. Generally, the site geological conditions are very soft and there is a strong need for ground improvement.

ground improvement, including replacement of soft soils by sandy gravels or gravel, use of Settlement Reduction Piles in soft cohesive layers, cement deep mixing piles in soft cohesive layers and sand com­ paction piles; and (2) a pile foundation, using con­ crete piles, which is employed close to the artificial islands. Tunnel elements on both ends (elements 1 to 9 and 25 to 33) are placed on zones with extensive artificial ground improvements, and hence a lower degree of soil uncertainties. The middle parts of the tunnel (elements 10 to 24) are placed on natural untreated soil paved with a gravel layer on the sur­ face. The natural soil conditions (in the middle part) without deep soil improvement, lead to a higher degree of soil uncertainties, and therefore this study mainly focuses on settlement analysis of element 10 to 24. The effective load on the foundation is estimated considering the tunnel elements themselves, as well as backfilling and siltation based on the design docu­ ment. Details on the loading conditions can be found in Wu 2017. 2.4 Deterministic settlement estimation methods The settlements of the clay layers are calculated using one-dimensional consolidation theory, assum­ ing a constant load. Due to the dredging and tunnel construction, most layers will initially be unloaded and the unloading-reloading stiffness and stress history (foundation soil is over-consolidated) are considered. The final settlement are calculated using (Eq.1):

1Þ z0

2.3 Immersed tunnel foundation

zc

The HZMB Tunnel is placed mainly on muck and silty clay at both ends and on silty clay or sand in the middle. As a consequence, differential settlements along the tunnel will definitely occur without ground improvement measures. In this project ground treat­ ment is required over a considerable part of the tunnel alignment. Longitudinally, two types of foundation were adopted: (1) a composite foundation formed by

Where, H0 is the thickness of the soil layer; Cc is the compression index; e0 is the 0 initial void ratio; Cr is the recompression index; σzc is0 the maximum effective stress before unloading; σz0 is the current 0 vertical effective stress; σfin is the final vertical effect­ ive stress. Settlements of the sand layers are assumed to occur almost immediately when the foundation load is imposed and are determined using (Eq.2):

Figure 1. Cross section of HZMB immersed tunnel (unit: cm, based on Zhang et al. 2016).

where C1 is a depth correction factor, C2 is a creep factor; Δp the net load increase at foundation level, B the width of load area, Iz the vertical strain influ­ ence factor, ES the secant Young’s modulus and ΔZ the thickness of the soil layer, following Schmert­ mann 1970.

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is associated with the natural variability of soil prop­ erties. This inherent variability of soil can be attrib­ uted to the deposition processes that lead to the formation of the soil, and ongoing geological pro­ cesses that continue to alter the soil body. It is often modelled with random variables and can be quanti­ fied by soil investigation measurements, statistical approximations and engineering experience. 3.2 Figure 2. Deterministic settlement analysis results.

2.5

Deterministic settlement results

Soils in the considered central section can be divided into two main units, clayey soil and sandy soil. These soils can be further divided into subunits using CPT results from the site investigation. For the sandy layers, the soil stiffness is calculated based on the relationship between cone resistance and Young’s modulus (Schmertmann et al. 1978). From the site investigation, around 8 to 10 CPTs are avail­ able in the vicinity of each tunnel element, although only a summary of the means, maximum and min­ imum CPT values for each element is reported. For the clay layers, parameters are obtained from labora­ tory tests reported by HZMB, with a summary reported by Wu 2017. The deterministic settlement analysis results are shown in Figure 2, based on mean parameters for each element, which do not consider the soil uncer­ tainties. These show significant settlement differ­ ences between neighboring elements, which would lead to significant shear reinforcement in the joints. These results will be compared with the probabilistic method results below. 3 PROBABILISTIC METHOD FOR SETTLEMENT ANALYSIS 3.1

Uncertainties in Geotechnical engineering

Geotechnical designs are influenced by different kinds of uncertainties, amongst which the spatial variability of soil. Since variations into the thickness of soil layers, local transitions between layers and variations in properties within the area of an element are not taken into account in the traditional determin­ istic design, a proper probabilistic analysis is needed. This provides the possibility of including uncertainties and making a better assessment of the reliability of the structure. In general, soil uncertainties can be categorized as aleatory and epistemic uncertainties, which coincide in most geotechnical practical applications (Baecher et al. 2005). Epistemic uncertainties are due to a lack of data or information about events, or lack of under­ standing of physical laws. The aleatory uncertainty

Stochastic description of foundation soil

In this study, the inherent soil variability is con­ sidered, and uncertain variables are introduced as stochastic variables, described by their mean m, and standard deviation σ. The coefficient of variation (COV) is a non-dimensional statistical parameter that can describe the dispersion of a probabilistic dis­ tribution relative to the mean, and is defined as the ratio of standard deviation over the mean of a parameter (COV ¼ σ=μ). COV is considered the most straightforward and widely used parameter to describe the uncertainty of soil properties. Based on the local site investigation report, the mean values of COV for all input variables are cal­ culated and listed in Table 1. In the probabilistic analysis, it is important to choose a probability distribution properly based on the values of COV for this variable. According to Kamp 2016, for small coefficients of variation (COV), a normal distribution is suitable to describe the variables, and this is used in this analysis. 3.2.1 Scale of fluctuation Even within homogenous layers, the soil proper­ ties can vary considerably from one location to another. This variability is associated with geol­ ogy and the conditions during soil deposition. Considering the variability of soil, the probabilis­ tic modeling of soil profiles was put forward by Vanmarcke 1977. Reliability of a structure is determined as a function of both local statistics (characterized by COV of the individual model parameters) and a spatial correlation property or autocorrelation function. The autocorrelation function is usually expressed in terms of an exponential decaying function, termed the scale of fluctuation θ, which is defined as the distance beyond which the correlation between soil prop­ erties becomes negligible.

Table 1.

Coefficient of variation of different variables.

Soil property

Symbol Averaged COV Unit

Thickness of sand layer Thickness of clay layer Void ratio Recompression index Reloading Modulus

hs hc e Cr Eur

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0.16 0.15 0.11 0.28 0.19

[m] [m] [–] [–] [MPa]

variables involved. Steps to generate normally dis­ tributed correlated variables can be found in Wu 2017.

Many methodologies have been developed to determine the scale of fluctuation. In this research, the space average method (see Vanmarcke 1977) is used to determine the scale of fluctuation of soil stiffness. 3.3

3.5

Point Estimate Method

The Point Estimate Method (PEM) is a relatively simple method to evaluate the reliability of a structure. This method is a computationally straightforward approach to explicitly account for uncertainty of input parameters. The general idea is to simplify the entire distribution of a variable by a discrete equivalent distribution. This is done by assigning the same three first statistical moments from the complete original distribution to the new equivalent distribution. Before performing calculations with PEM, evalu­ ations points need to be defined. Generally, two evaluations points are defined, located at one stand­ ard deviation on either side of the mean value. This is done for each stochastic input parameter. Next the performance function is calculated for every possible combination of the evaluation points. This results in 2n calculations, where n is the number of included stochastic variables. For more details, see Rosenble­ uth 1975. 3.4

Parameter determination of randomized soils

Based on the relationship between CPT values and the lab test results, the soil properties used in the settlement calculation are determined. There are seven different subunits defined for the central area, which are labeled as Sand 1-5 and Clay 1-2. For brevity, only the means and standard deviations for the first three elements are listed in Table 2 and Table 3. The full set of parameters can be found in Wu 2017. 3.6

Settlement results based on PEM

The settlement along tunnel line are firstly calcu­ lated using the Point Estimate Method, simplify­ ing to a case where elements are directly installed on the bottom of trench and no add­ itional foundation layer is used. First the values for each input variable are determined. For each

Table 2. Parameters used for stochastic settlement calculations-a.

Number

Sub-layer

Layer thickness (m)

10

Clay1 Sand1 Sand4 Sand3 Sand5 Sand4 Sand3 Sand4 Sand3 Sand5

0.47 0.44 19.57 5.63 3.00 16.90 14.24 15.78 5.21 4.9

Monte Carlo simulation

As reliability related issues are becoming more crit­ ical in engineering design and analysis, proper assessment of stochastic behavior of an engineering system is essential. However, due to the complexity of physical systems and mathematical functions, der­ ivation of the exact solution for the random charac­ teristics of the system response is difficult. In such cases, a Monte Carlo simulation is a viable tool to provide numerical estimations of the stochastic fea­ tures of the system response. A Monte Carlo simulation uses multiple realiza­ tions of the system response of interest under various parameter sets generated from the known or assumed probabilistic distribution. It offers a practical approach to reliability analysis because the stochas­ tic behavior of the system response can be probabil­ istically duplicated. The frequency of each outcome can be plotted by means of a histogram. When suffi­ cient simulations have been performed a probability density function can be fitted on the output histo­ gram (Kamp 2016). In many practical engineering analyses, random variables are often statistically and physically dependent. To properly replicate such systems, Monte Carlo simulations should be able to preserve the correlation relationship among the stochastic parameters and their distributions. As a practical alternative, this section describes procedure to gener­ ate multivariate random variables that preserves the marginal distributions and correlation of the random

11 12

Std (m)

Unit weight (KN/m3)

0.12 0.1 3 1.2 0.2 3.5 3 3.2 1.2 1

18.7 18.8 19.9 19.6 20.9 19.9 19.6 19.9 19.6 20.1

Table 3. Parameters used for stochastic settlement calculations-b. Sublayer 10 Clay1 Sand1 Sand4 Sand3 Sand5 11 Sand4 Sand3 12 Sand4 Sand3 Sand5

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Void ratio

Std (-)

0.958 0.826 0.470 0.650 0.460 0.470 0.650 0.470 0.650 0.460

0.10 0.034 0.01

Cr

Std (-)

Eur/ MPa

Std/ MPa

179.6 310.0 292.4 251.2

15.7 14.7 28.1 27.8

193.2 147.6 252.5 173.4 215.2

14.7 13.8 24.3 15.1 16.8

Figure 3. Settlement results based on PEM.

variable, two additional evaluation points need to be determined next to the mean, as one standard deviation below or above the mean value. Sec­ ondly, for each tunnel element, settlement calcu­ lations are made for each possible combination of parameter values, which leads to 2n combinations, with n is the number of variables. As illustrated in Figure 3, the largest settlements occur at element 15 and element 24, and it can be derived that the subsoil close to these two zones is relatively weak and more compressible. The averaged coefficient of variation of the calculated settlements is 0.27. 3.7 Settlement results based on Monte Carlo Simulation Next, a Monte Carlo Simulation is performed based on the stochastic soil properties. Input variables are div­ ided into two main groups, consisting of clay layers and sand layers, which are further characterized by their sublayers. Following the analytical settlement cal­ culation methods, in clay layers the considered vari­ ables are the thickness of the layer hc, recompression modulus Cr and void ratio e. In sand layers the vari­ ables are thickness hs and unloading-reloading modu­ lus Eur . The mutual correlations between the different input variables are assumed constant for each soil layer, with the correlations estimated based on expert judgment (Rebonato et al. 2011). The detailed correlation between parameters is shown in the cor­ relation matrix in Figure 4.

Figure 4. Correlation among parameters.

The Monte Carlo analysis of settlements along the tunnel are performed using Matlab. For each element, 1000 settlement calculation are performed, and the his­ tograms of the magnitudes of settlements for each element are obtained, from which the stochastic char­ acteristics (including max value, mean value, mean value, standard deviation and coefficient of variation) are collected. For design purposes, the mean values are com­ monly of most interest. As seen in Figure 4, the mean settlements of elements 15 and 24 are the lar­ gest among the total fifteen elements, while the first three elements, from elements 10 to 12, settle the least, with the settlements all below 20 mm. That indicates that the subsoils below elements 10 to 12 are much stiffer than those in other locations of this area. In Figure 5, the upper blue line and the lower green line represent the maximum and minimum probable settlement respectively. Comparing the geological pro­ file and the Monte Carlo realizations, it is straightfor­ ward to see that settlements fluctuate along the tunnel alignment, and the maximum settlement occurs at element 24, as here the thickest soft soil layers can be found. 3.8 Comparison of deterministic and probabilistic results A comparison of the settlement results between deterministic method and two probabilistic methods (PEM and MCS) is made in this section. In Figure 6, the mean values of both PEM and MCS are plotted together with the deterministic results. This shows the Monte-Carlo results per­ fectly match the deterministic method while small differences can be found in the PEM results. This confirms that for the MCS, a 1000 realizations are sufficient to get converging results with a sufficient reliability for the determination of the variation in settlements. Although small differences are observed with the PEM results, overall, it still provides a high similar­ ity to the MCS results and can be considered accur­ ate enough for this application.

Figure 5. Settlement results based on MCS.

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Table 4. PEM.

Advantages and disadvantages for MCS and

Methods Advantages MCS

PEM

Figure 6. Mean expected settlements comparison.

High accuracy; Provide complete output distribution; Require little know­ ledge about probabilis­ tic theory Considerable accuracy Lower computational cost; Requite little knowledge about probabilistic theory

Disadvantages High computational cost

Less accurate than Monte Carlo; Does not provide complete output distribution; Not accurate in the “tails”, which are out of the range (m ± σ)

4 JOINT SHEAR FORCE CALCULATION BASED ON MONTE CARLO SIMULATIONS 4.1 Stiffness back-calculated from settlement distribution

Figure 7. Settlement results comparison.

3.9 Comparison of Monte Carlo Simulation and Point Estimate Method The settlement results from both methods (MCS and PEM) including the minimum and maximum expected values are plotted together in Figure 7. As shown above, the results from the both methods show highly similar mean values of settlement. On the other hand, the fluctuation of settlements obtained with PEM is slightly wider than from MCS, which means, the maximum and minimum values of PEM are slightly higher and lower than those (μ±σ) from MCS. This can be caused by the limitation of extreme input variable combinations and the discontinuous variable selection. The input values for each variable in PEM are selected only at two extreme values (μ±σ). Therefore, the occurrence of the values outside of that range (μ±σ) is not considered. Thus, the maximum and minimum values from PEM are not particularly significant, as that they cannot represent the entire distribution of output. This problem can be easily solved in Monte Carlo simulations where nearly all the possibil­ ities (μ±3σ) of input variables can be generated, given sufficient calculation times. Table 4 shows the advantages and disadvantages for those methods, and the method that is best suited for similar projects depends on the engineering requirements, including the required accuracy level, computational costs and the expectations from the results.

In order to analyze the shear force in the tunnel joints, a beam-spring structural model is built, where each “spring” is simulated by a laterally uniform soil column below each element. The stiffness of each soil column is calculated based on the settlement results from the previous ana­ lysis and is considered as the input variable for the beam-spring model. The equivalent stiffness distribution for each soil column is calculated from the settlement distribution and thickness of entire column as (Eq.3):

where E is the equivalent stiffness for soil column, P is the acting load, H is the equivalent thickness of soil column and S is the displacement of the soil column. To easily model this using PLAXIS, the equiva­ lent layer thickness H is fixed as 30 meters for all elements, and variations in layer thickness are effect­ ively captured by adapting the layer stiffness. Stiff­ ness values and its stochastic characteristics are listed in Table 5. 4.2 Model for joint shear force analysis In this model, the stiffness of neighboring soil units is assumed to be independent from each other. In order to achieve best efficiency and save manual input and calculation costs, a Python script is written

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Table 5.

Spring stiffness & stochastic characteristics.

Table 6.

Stochastic results of joint shear forces.

Min Max value value E-No. (MPa) (MPa)

Mean value (MPa)

Std. (MPa) COV

Max value Joint (kN)

10 11 12 13 14 15 16 17 18 19 20 21 22 23 24

334.30 372.47 338.64 120.49 102.32 69.30 102.18 95.97 111.88 168.69 151.15 205.52 103.08 122.01 79.21

54.06 47.93 43.55 26.69 26.88 14.18 22.76 27.89 21.81 24.69 30.97 37.28 23.31 27.80 23.49

1 2 3 4 5 6 7 8 9 10 11 12 13 14

601.29 560.49 473.02 290.03 348.36 145.12 201.47 236.61 259.28 276.29 414.68 443.52 268.24 394.30 221.36

185.83 232.90 171.05 74.30 54.14 38.93 59.55 38.52 67.12 94.66 88.38 120.77 58.11 71.23 39.64

0.161 0.128 0.128 0.221 0.262 0.204 0.222 0.290 0.194 0.146 0.204 0.181 0.226 0.227 0.296

3779.86 2517.58 15798.27 15117.17 9211.34 10187.34 6607.49 7322.12 9202.83 8736.71 7942.75 13251.32 12411.89 18871.59

Min value (kN)

Mean (kN)

Std (kN)

COV

0.26 0.33 398.82 1.67 2.27 3.28 4.65 4.24 1.75 2.21 0.70 28.52 0.66 14.99

569.51 532.84 3486.37 1436.87 2857.89 2646.04 1703.95 1595.17 2090.90 1467.39 2053.81 4120.25 2317.20 2740.93

467.98 412.74 1293.72 1249.49 1720.19 1738.06 1284.96 1271.86 1366.11 1173.66 1337.19 1894.06 1881.24 2032.59

0.82 0.77 0.37 0.87 0.60 0.66 0.75 0.80 0.65 0.80 0.65 0.46 0.81 0.74

* E-No. means element number;

to direct PLAXIS to perform Monte Carlo Simula­ tions with 1000 realizations, and automatically save the outputs from PLAXIS. The model is set up in Plaxis2D, as shown in Figure 8, with the tunnel elements simulated by plates elements placed on the top of the soil vol­ umes. The circles in between plate elements are modelled as joints. For the plate elements, the normal stiffness EA is set at 5.24E9 kN and the flex­ ural rigidity EI at 99.5E9 KN*m2 simulating the stiffness of the tunnel elements. The soil domain is divided into 15 soil volumes, one below each element, with each volume 30 meters high and 180 meters wide. The soil is assigned normally distributed stiffness values with different means (ui ; i ¼ 1; 2; . . . 15) and standard deviations ðσi ; i ¼ 1; 2; . . . 15Þ, which are independ­ ent from each other. Shear forces are calculated for each element joints for each realization. 4.3

Shear forces in tunnel element joints

The stochastic values after 1000 realizations for each joint are listed in Table 6. In the table, “Joint No.1” means the first joint, between the first two tunnel elements (elements 10 and 11), etc. In general it can be observed that, as shown in Table 6, the maximum possible shear forces in each joint are much larger than mean value, and the high COV indicates that a high degree of variation in joint

Figure 8. Soil-structure interaction model in PLAXIS.

shear forces can occur. Although this doesn’t neces­ sarily indicate that in reality very large shear forces occur in the joints, it shows a necessity for a structural reliability analysis. Considering the uncertainties in the soil, the shear forces calculated show a large variation. It should be noted that in this analysis each element is modelled as a single plate element, and a reduction in stiffness local to the seg­ ment joint is not considered. Still, the uncertainty in soil parameters does potentially affect the joint deformation, which shows the necessity of probabil­ istic analysis in complex designs. 5 CONCLUSIONS In this paper, a probabilistic analysis is implemented to determine the settlements of immersed tunnel elements and the effects of subsoil stiffness variabil­ ity. Site investigation data from the Hongkong­ Zhuhai-Macau Bridge (HZMB) Tunnel project in China is used in this case study. Two probabilistic methods, the Point Estimate Method and Monte Carlo Simulation, are compared for the determin­ ation of tunnel settlements. This shows that PEM is computationally more efficient and has sufficient accuracy, while the latter is more accurate with higher computational costs. A soil-structure interaction model is built to study the effects of variations in sub-soil stiffness on the shear behavior of tunnel joints. Based on the settle­ ment results, the spatial variability of the soil stiffness is quantitatively assessed. The soil-structure interaction analysis and derivation of shear forces in tunnel joints is performed by coupling FE analysis to a MonteCarlo simulation. The results show that tunnel struc­ ture behavior and attainable shear forces in the joints are significantly influenced by the uncertainty in soil

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parameters. As uncertainty in soil parameters affects the settlements and joint deformation, this calls for the use of probabilistic analysis in future project designs.

REFERENCES Baecher, G.B. & Christian, J.T. 2005. Reliability and statis­ tics in geotechnical engineering. Chichester: John Wiley & Sons. CCCC Second Flight Engineering Survey & Design Insti­ tute Co., Ltd. (CCCC SFES&DI). 2009. Geological investigation report on immersed tunnel of Hong Kong-Zhuhai-Macao Bridge in construction documents design phase. Guangzhou (in Chinese). Grantz, W. C. 2001a. Immersed tunnel settlements. Part 1: nature of settlements. Tunnelling and Underground Space Technology 16(3): 195–201. Grantz, W. C. 2001b. Immersed tunnel settlements: Part 2: case histories. Tunnelling and Underground Space Tech­ nology 16(3): 203–210. Kulhawy, F.H. 1993. On the evaluation of static soil properties. In Stability and performance of slopes and embankments 2: 95–115.

Kamp, S.P. 2016. Reliability-based ultimate limit state design in finite element methods. Delft: TU Delft. Rosenblueth, E. 1975. Point estimates for probability moments. Proceedings of the National Academy of Sciences 72(10):3812–3814. Rebonato, R. & Jäckel, P. 2011. The most general method­ ology to create a valid correlation matrix for risk man­ agement and option pricing purposes. Available at SSRN 1969689. Schmertmann. 1970. Static cone to compute static settle­ ment over sand. ASCE Journal of Soil Mechnicas and Foundation Division 96: 1101–1043. Schmertmann, J.H., Hartman, J.P. Brown, P.R. 1978. Improved strain influence factor diagrams. Journal of Geotechnical and Geoenvironmental Engineering 104: 1131–1135. Vanmarcke, E.H. 1977. Probabilistic modeling of soil profiles. Journal of the geotechnical engineering division 103(11):1227–1246. Wu, X. 2017. Impact of spatial variability of subsoil stiffness on immersed tunnels. Master thesis. Delft: TU Delft. Zhang, Z., Lin, W., Ji H. & Liu, X. 2016. Layout and Design Techniques of Cross Section for the Large Immersed Tunnel. Procedia Engineering 166:37–44.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Coupled elastoplastic analysis of the soil-pile foundation interaction induced by deep excavations C. Zheng, A. Franza & R. Jimenez ETSI Caminos Canales y Puertos, Universidad Politécnica de Madrid, Madrid, Spain

ABSTRACT: Deep excavations can detrimentally affect existing structures by causing a ground displace­ ment field of vertical and horizontal movements. Consequently, in the case of deep foundations, excavationinduced ground movements result in complex soil-pile and soil-pile groups interaction mechanisms that result in foundation displacements and additional distress. Using an elastoplastic two-stage continuum-based ana­ lysis, the response of single pile and pile groups behaviour close to an ongoing and finalised excavation is analysed. Both the excavation-induced displacements and internal forces along the pile axis and at their heads are discussed. Results show that the excavation depth with respect to the pile tip, and the presence of a cap (for pile groups) have a significant influence on the bending and axial deformations of piles; and that the induced pile bending can vary during the excavation sequence. In addition, comparing the results obtained from single pile analyses with results from capped pile groups, preliminary insights are given into the import­ ance of pile-to-pile effects when a rigid superstructure is present.

1 INTRODUCTION In urban areas, deep excavations inevitably cause vertical and lateral ground movements, which may affect existing foundations. When deep excavations are carried out close to the existing piles, they may lead to foundation distress or failure (Finno et al. 1991, Goh et al. 2003). Centrifuge tests were con­ ducted to investigate the response of a single pile (Leung et al. 2000, Ong et al. 2006, Ng et al. 2017) and of pile groups (Leung et al. 2003, Ong et al. 2009), confirming that the impact of excavation to adjacent piles can be significant. Hence, developing a reliable and straightforward method to estimate the behaviour of single pile and pile groups next to deep excavations is important for geotechnical engineers. When performing preliminary risk assessments, current methods to analyse the response of the pile foundations adjacent to a deep excavation can be classified into two types. 1) Advanced numerical analyses (Liyanapathirana & Nishanthan 2016, Soomro et al. 2019) modelling the entire geotech­ nical domain can provide refined results by consider­ ing, among other aspects, the detailed excavation sequence, an advanced soil and pile-soil interface models. However, they are computationally expen­ sive and require a careful calibration of suitable model parameters, which may not be easily achieved during preliminary stages. 2) Simplified two-stage analysis methods, for which the greenfield soil movements are first computed either from advanced

models (Poulos & Chen 1997) or from analytical solutions (Sagaseta 1987, Xu & Poulos 2000). Sub­ sequently, the response of pile foundations embed­ ded into elastic half-space or Winkler/Pasternak’s soil models is obtained by solving the soil-structure interaction. Although the two-stage analysis method has been extensively used to evaluate the effects of tunnelling on adjacent structures with shallow and deep founda­ tions (Chen et al. 1999, Kitiyodom et al. 2005, Huang & Mu 2012, Basile 2014, Franza et al. 2017, Franza & DeJong 2019, Elkayam & Klar 2019), its previous applications to deep-excavation problems is limited. For instance, Poulos & Chen (1997) esti­ mated the influence of a deep excavation on an exist­ ing single pile inputting greenfield movements obtained from a finite element analysis into a boundary element method (BEM) model. Subse­ quently, Xu & Poulos (2000) suggested the use of a source-sink imaging technique and the response of a single pile was solved by using an elastic BEM model, named GEPAN. Both works considered linear elastic soil behaviour. Recently, Zhang et al. (2011) and Korff et al. (2016) proposed, for a single pile, Winkler-based two-stage methods that consider nonlinear load transfer mechanisms between the pile and the soil. Despite previous research, the response of pile groups (affected by the presence of a cap or a superstructure) to deep-excavations has not been yet fully analyzed with simple nonlinear soilstructure interaction models.

DOI: 10.1201/9780429321559-98

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Figure 1. Studied cases.

As a first attempt to deal with pile group and piled structure response to deep-excavations, this study uses an elastic-perfectly plastic two-stage model to simulate the progress of an excavation, and to inves­ tigate the effects of two different excavation depths with respect to the pile tip on (i) a single pile and (ii) on pile rows (see Figure 1). Excavation-induced dis­ placements and internal forces along the pile axis, and at their heads, are discussed. The impact of pile group head condition (i.e., free head or rigid cap) is analyzed, comparing the results from single pile ana­ lyses (used to simplify the problem) with the results from pile groups analyses. These results provide insights into the importance of pile-to-pile effects resulting from rigid superstructures. 2 MODEL A Finite Element Method (FEM) model (continuum­ based, two-stage, elastoplastic soil behaviour) is adopted to investigate the effects of a deepexcavation next to a single pile or to pile groups con­ sisting of six vertical piles.

of the diaphragm wall (i.e., its later displacement) is associated with a distributed ground loss. By assuming that each infinitesimal wall depth, and the corresponding wall displacement, is associ­ ated an infinitesimal ground loss, the horizontal ux and vertical uz greenfield soil movements are obtained by integrating the infinitesimal volume losses along the wall depth, which are linked to the undrained soil deformation by the solution from Sagaseta (1987).

2 where ;h is the integration variable; Hd is the diaphragm wall depth; x is the horizontal distance to the diaphragm wall; z is the depth to the ground surface; and f ðhÞ is the deflection of the diaphragm wall.

2.2 2.1

Estimation of greenfield soil movements

To estimate the deep-excavation movements with a simple closed-form solution, the superposition of singularities method (Xu & Poulos 2000, Zhang et al. 2011) is used. In this approach, the deflection

Analysis of soil-pile interaction

To solve the soil-structure interaction due to the greenfield movements, the soil-pile group system is subjected to an equivalent set of forces associated with the greenfield input. The equilibrium equations are solved with the Finite Element Method (FEM)

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approach, as fully detailed in the companion paper (Franza et al. 2020). This FEM model has been val­ idated in Franza et al. (2019b) and Franza et al. (2019a) for tunnelling-induced ground movements against Boundary Element Method models (Basile 2014, Loganathan, Poulos, & Xu 2001) that were tested using field and centrifuge measurements. Spe­ cific validations for deep-excavation cases are not conducted herein, since they will be studied in future works, subsequently to an improved characterisation of the greenfield displacement field caused by deepexcavations that is currently underway. The piles are modelled as Euler-Bernoulli beams connected to fully coupled vertical and horizontal springs obtained from the integration of Mindlin’s solution (Mindlin 1936), describing the half-space continuum behaviour for loading. This allows one to overcome the limitations of Winkler soil models and interaction factors (Huang et al. 2009). Both linear elastic (EL) and elastoplastic (EP) analyses are per­ formed. For the EL analyses, perfect compatibility is imposed between the pile and the continuum. In the EP analyses, the soil plasticity (i.e., yielding) at the pile-soil interface is considered using vertical per­ fectly plastic sliders at the interface; such sliders limit the shaft and base capacities to ultimate values, and allow slippage to occur at the shaft and a gap beneath the base. On the other hand, a linear elastic response was considered in the horizontal direction for both the EL and EP solutions. 3 RESULTS 3.1

Investigated scenarios

Figure 1 summarises the simulation cases con­ sidered, comprising two possible excavation depths of a wall close to either a single pile (labelled pile 1) or to a row of six piles (labelled pile 1-6), in which the depth of the wall (undergoing lateral displace­ ments due to the excavation) is set to twice the pile. Scenarios with piles shorter than the excavations are known to be critical in terms of pile settlements (Korff et al. 2016). Uniform head loads prior to the excavations are selected for individual piles, while analyses of the pile rows are carried out considering both a free-head condition and a rigid cap (to cover a wide range of boundary conditions that includes most field applications). The soil elastic modulus Es is 24MPa and its Pois­ son’s ratio ls is 0.5. Piles have a diameter dp of 0:5m, a length Lp of 15m, and a Young’s modulus of 30GPa. For the interface, it is assumed that the ultimate base and shaft stresses are qb;f = 540 kPa and τf = 48 kPa, respectively. Note that a service load P0 ¼ 618kN is considered, giving an initial safety factor SF0 ¼ Qtot =P0 ¼ 2, where Qtot is the ultimate pile capacity given by the resistance of the shaft Qs and base Qb , Qtot ¼ Qb þ Qs .

Two excavation depths are considered for the dia­ phragm wall: H d1 ¼ 15m and H d2 ¼ 30m; they simulate different stages of a given deep excavation. Importantly, Equation (1))requires to select the func­ tion f (i.e., the profile of deflection of the diaphragm wall). For the sake of simplicity, a second order par­ abola is assumed for its deformation mode with an average lateral displacement of 0.1% Hd , following the field measurements of Wang et al. (2010). How­ ever, further research is currently ongoing to explore the impact of other ground loss distributions and wall deflections on the greenfield displacement field. 3.2

Single pile

Considering that complex scenarios are often simpli­ fied to a single pile, the results obtained from consider­ ing the front pile 1 as isolated are considered first. Figure 2 shows both the axial and lateral response of such single pile for two different excavation depths (hence modelling, in a simplified manner, an inter­ mediate and the final excavation stages). In this work, dark and light colours in plots are used, respectively, for the intermediate and final excavation depths. Importantly, results compare EL and EP outcomes, while plotting greenfield movements. Note that the pile displacements (uz =dp or ux =dp ) are normalised by its diameter. Results in Figures 2(a) and (b) show that the mag­ nitude of vertical and horizontal movements increases with the excavation depth Hd =Lp , as expected. However, it is interesting that the green­ field input shape changes for the intermediate and final conditions. While the impact of this on the axial force distribu­ tions in Figure 2(c) is limited, bending moments in Figure 2(d) are affected, with a shift in the location of maximum excavation-induced bending distress. Regarding the pile axial response, the limiting skin friction is mobilised along the pile in the EP solutions for both excavation depths (Hd1 and Hd2 ). Considering pile plasticity reduces the maximum compression axial force, while slightly increasing the pile settlement. For instance, for the EP analyses the maximum settlement of the pile are large: 15.4 mm (3.08% dp ) and 48 mm (9.6% dp ) for Hd1 and Hd2 , respectively. Furthermore, comparing the EP and EL analyses for the final normalised depth there is a decrease in the maximum axial force of 69.5% and an increase in settlement of 17%. Such reduction of excavation-induced axial forces could be considered when estimating the risk for bending failure of piles that occur due to horizontal ground movements, as discussed next. Next, the horizontal response is addressed. Pile deflections in Figure 2(b) are very similar to the soil horizontal displacement, except at the pile head, where the curvature of the greenfield movements is significant. Interestingly, for the intermediate excava­ tion depth the largest bending moments

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Figure 2. Response of a single pile (i.e., pile #1) to excavation induced ground movements considering EL and EP behav­ iour in case of SF0 = 2.

concentrate around pile mid-depth. For the final depth (Hd =Lp ¼ 2) there is an increase of the bend­ ing moment (opposite to the wall) close to the pile head and decreased moments around mid-depth. This qualitative and quantitative variation in the bending moment distribution is due to the variation in the shape of the horizontal movements: greenfield ux are about parabolic for the intermediate excavation depth (Hd =Lp ¼ 1), while they are almost linearly increas­ ing with depth for the final stage (Hd =Lp ¼ 2).

3.3

Pile groups

The response of the pile row (piles 1-6) — displace­ ments and internal forces at the pile heads (i.e., the surface level), for piles with free and capped heads — is illustrated in Figure 3. Firstly, the results obtained for free-head conditions (triangular markers) are analysed. Interestingly, in Figure 3(a), pile heads tend to settle equal to, or slightly less, than the surface; this was expected, due to deep-excavation movements mostly decreasing with

Figure 3. EP Response of pile groups to excavation induced ground movements for pile #1 � #6 (SF0 ¼ 2).

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depth korff 2016. The relationship between the pile head and greenfield surface settlements for pile 1 is affected by Hd =Lp, which changed the magnitude and shape of the greenfield settlement shape (see Figure 2). Similarly, in the absence of a cap, the horizontal pile heads displacements in Figure 3(b) are similar to the surface greenfield profile. Clearly, because of the freehead condition, the excavation-induced forces at the pile head were nearly zero. Subsequently, the impact of the rigid cap is dis­ cussed (see starred markers). The constraint intro­ duced by the rigid cap imposed a linear variation of the pile head settlement with the horizontal offset, and a constant lateral pile head displacement in Fig­ ures 3(a) and (b), respectively. Consequently, the cap must react against the pile to satisfy this kinematic constraint, by uplifting external piles 1 and 6 and embedding the central piles 2-4; these results, respectively, in tensile and compressive variations of axial forces at the heads. Similarly, bending move­ ments are caused by the combined cap effect of restraining relative pile horizontal displacement while imposing an identical rotation to each pile cap. The magnitude of both axial and bending moments at the pile heads in the presence of the cap (see Fig­ ures 3(c) and (d)) are significant, when compared with the results for a single pile (see Figures 2(c) and (d)). Finally, the effects of the excavation sequence are commented. In Figure 3, the tilt and settlement of the capped pile groups increases as the depth of excavation (Hd =Lp ) increases. On the other hand, the lateral displacement of different pile heads within

the pile group are equal due to the constraint of the cap, and the cap horizontal displacement also increases with the excavation depth. Maximum com­ pression and tensile axial forces of pile groups also increase as the excavation depth increases. 3.4 Comparison between single pile and pile group results Figure 4 compares results obtained for the front pile 1 for its analyses as a single pile and as a pile within a capped pile group. It therefore illustrates the impact of the cap action along the entire pile axis. Figure 4(a) confirms that the uplift due to the cap (mobilising the axial reaction of the remaining piles 2-6) is significant; for instance, the settlement for pile 1 within the capped pile group is 36.4% smaller than for single pile. To induce this relative pile uplift, the cap induced a tensile variation of force at the pile head, with decreases the pile settlement and the nega­ tive friction caused by the greenfield movements (see Figure 4(c)). On the other hand, Figure 4(b) shows that the variation of horizontal pile displacements is signifi­ cant only for the final excavation depth (Hd =Lp ¼ 2). Despite this, large bending moments at the top of the pile, but with opposite signs, are induced by the cap for both excavation depths (Hd =Lp ¼ 1; 2) (see Figure 4(d)). Finally, results obtained along the entire pile for single piles and for pile groups with free heads (ana­ lyses not reported) displayed limited differences. This is likely to be due to the significant transverse spa­ cing of 5m between piles (sp =dp ¼ 10) causing the

Figure 4. Comparison of pile #1 in the EP analyses of the pile groups (SF0 ¼ 2).

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settlements, lateral movements and tilt of the cap gradually increased.

pile-to-pile interaction (also referred to as shielding effect in such excavation problems) to be negligible.

Future works will aim at estimating the efficiency of this model against centrifuge and field data.

4 CONCLUSIONS This study analyzed and compared the responses of (i) a single pile and of (ii) pile-groups (with free heads or with a rigid cap) to displacements produced by deep excavations. A coupled continuum-based two-stage model is employed, accounting for limit shaft and base forces along the pile. The effect of the excavation depth was investigated by simulating excavation levels at the pile tip and below it. The following preliminary conclusions may be made from this work, initiating a line of research on the response of pile groups and piled structures to deepexcavations. • It is recommended to use soil-structure interaction models that consider the complex excavationinduced displacement field, consisting of horizon­ tal and vertical movements. For a realistic estima­ tion of the axial pile response, the soil plasticity (i.e., yielding, or limit shaft friction and base cap­ acity) should also be accounted for. Therefore, the proposed elastoplastic two-stage analysis method (EP) is considered more appropriate than the linear elastic models (EL). The reason is that EL analyses could lead to overestimated compressive forces (thus, with potentially unconservative assessments of risk for pile damage). • For a single pile close to the excavation, the settlement and axial force were affected by the excavation depth. As the excavation progresses, a larger excavation depth (associated with greater ground losses) produces larger settlements but slightly lower axial forces along the single pile, while keeping their profile with depth nearly unchanged. On the other, lateral movements and bending profiles of the pile were altered, both qualitatively and quantitatively, by changes of the excavation depth. • For pile groups, the fixing conditions at the head had significant effects on their response. Firstly, rigidly capped groups underwent large excava­ tion-induced bending moments and large vari­ ations of axial forces at the pile heads, as compared with results of the free-head conditions. For the external piles of the cap, tensile excava­ tion-induced force increments were a significant portion of the uplift pile capacity. Furthermore, these additional distresses propagated with depth down to the pile mid-depth. This aspect can be crucial for risk assessments and is not currently captured by free-head single pile analyses. Sec­ ondly, the variation in the pile head movements due to the presence of the cap was marginal for the case considered. Finally, as expected, with the increase of the excavation depth, the maximum

ACKNOWLEDGEMENTS This project has received funding from the European Union’s Horizon 2020 research and innovation programme under the Marie Sklodowska-Curie grant agreement No 793715. The first author is supported by China Scholarship Council (CSC). The support of these institutions is deeply acknowledged.

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Tunnelling and Underground Space Technology 24(4), 410–422. Kitiyodom, P., T. Matsumoto, & K. Kawaguchi (2005). A simplified analysis method for piled raft foundations subjected to ground movements induced by tunnelling. International Journal for Numerical and Analytical Methods in Geomechanics 29(15), 1485–1507. Korff, M., R. J. Mair, & F. A. F. Van Tol (2016, aug). PileSoil Interaction and Settlement Effects Induced by Deep Excavations. Journal of Geotechnical and Geoenviron­ mental Engineering 142 (8),04016034. Leung, C. F., Y. K. Chow, & R. F. Shen (2000). Behavior of Pile Subject to Excavation-Induced Soil Movement. Journal of Geotechnical and Geoenvironmental Engin­ eering 126(11), 947–954. Leung, C. F., J. K. Lim, R. F. Shen, & Y. K. Chow (2003). Behavior of Pile Groups Subject to Excavation-Induced Soil Movement. Journal of Geotechnical and Geoenvir­ onmental Engineering 129(1), 58–65. Liyanapathirana, D. S. & R. Nishanthan (2016). Influence of deep excavation induced ground movements on adja­ cent piles. Tunnelling and Underground Space Technol­ ogy 52, 168–181. Loganathan, N., H. G. Poulos, & K. J. Xu (2001). Ground and pile-group responses due to tunnelling. Soils and Foundations 41(1), 57–67. Mindlin, R. D. (1936). Force at a point in the interior of a semi-infinite solid. Journal of Applied Physics 7(5), 195–202. Ng, C. W. W., J. Wei, H. Poulos, & H. Liu (2017). Effects of Multipropped Excavation on an Adjacent Floating Pile. Journal of Geotechnical and Geoenvironmental Engineering 143(7), 04017021.

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Design and application of ground improvement for underground construction

Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Microcapsule-based self-healing cement stabilised clay

B. Cao, C. Litina, L. Souza & A. Al-Tabbaa Engineering Department, Cambridge University, Cambridge, UK

ABSTRACT: The in-situ mixing of soft clays with cementitious binders is a common soil stabilisation method for a wide range of underground construction applications, varying from the improvement of foundation ground to the construction of earth retaining walls within the framework of excavation works. However, it is shown that the mechanical and transport properties of cement stabilised clay change considerably when subjected to aggressive environments. Therefore, there is a need to improve the resilience of cement stabilised clay against mechanical and environmental stresses. The development of self-healing cement stabilised clay could provide more reliable materials that are likely to reduce the maintenance and repair costs. Microcapsule-based self-healing approaches have been investigated recently. The mechanism of self-healing microcapsule system is that when cracks propagate in the cementitious matrix, they rupture the microcapsules, leading to the release of healing agents into the crack volume. The objective of this paper is to investigate the self-healing performance of cement stabil­ ised clay incorporating microcapsules. In this study, microencapsulated sodium silicate was incorporated in cement stabilised kaolin clay and the healing agent reacts with the calcium hydroxide in the cementi­ tious matrix to form the calcium-silicate-hydrate gel to heal the cracks. Properties investigated were the distribution and release of microencapsulated sodium silicate, the permeability, and microstructure of the stabilised clay samples. CT scan and scanning electron microscopy analysis revealed the rupture of microcapsules and the generation of healing products around the microcapsules. The effective healing of cracks was verified by the recovered permeability properties of microcapsule-containing stabilised clay samples.

1 INTRODUCTION The in-situ mixing of soft clays with cementitious binders has become a common soil stabilisation method for a wide range of underground construc­ tion applications, including the improvement of foundation ground and the construction of earth retaining walls. The main objective of mixing a binder with soft clays is mechanical and transport properties improvement, i.e., strength and permeabil­ ity. Soft clays are stabilised with binders, including cement, lime, fly ash and ground granulated blast furnace slag, which usually possess cementitious properties. The cracking of stabilised clays in geotechnical engineering applications can occur due to mechan­ ical factors and environmental stresses, adversely impacting the serviceability of the soil-cement sys­ tems (Osman, 2007; Indraratna et al., 2009; Wang et al., 2015). However, the formation of damage is not problematic as long as counteracted by selfhealing processes. The self-healing concept is par­ ticularly valuable for underground construction applications, many of which are inaccessible, installed in aggressive environments rendering them

prohibitively expensive to maintain. For cementi­ tious materials, the mechanisms for achieving selfhealing are classified into two categories: autogenous and autonomic self-healing. The self-healing process is termed autogenous when the recovery process uses components that could otherwise be present when not specifically designed for self-healing (Figure 1). Conversely, the engineered addition of materials or components to promote self-healing in cementitious material characterises autonomic selfhealing (Figure 2). Amongst the latter, microcapsule-based selfhealing shows great promise and has been exten­ sively investigated at Cambridge University through EPSRC funding of a programme grant: Resilient Materials for Life (2017-2022, £4.9M). This approach relies on sequestering the healing agent in discrete capsules until damage triggers rupture, lead­ ing to the release of healing agents into the crack volume (Giannaros et al., 2016; Souza & Al-Tabbaa, 2018; Al-Tabbaa et al., 2019). Mineral core materials including sodium sili­ cate are considered as excellent healing agents for self-healing of cementitious materials because of their good compatibility with matrix materials.

DOI: 10.1201/9780429321559-99

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2 MATERIALS AND METHODS 2.1

Binder and clay materials

This binder used in this study for clay stabilisation was the CEM I 52.5 N cement with a particle density of 2.7-3.2 g/cm3 and a specific surface area of 0.30­ 0.40 m2/g, supplied by Hanson UK. Polywhite E China Clay, which is a high-quality medium par­ ticle size kaolin consisting of 50 wt% silicon dioxide and 35 wt% aluminium oxide, was utilised as the soft clay. Figure 1. Schematic representation of the mechanisms of autogenic self-healing (De Rooij et al., 2013).

Figure 2. Schematic representation of the mechanisms of autonomic self-healing (Souza, 2017).

Sodium silicate reacts with calcium hydroxide in the presence of water to form a calcium silicate hydrate (CSH) gel which can heal cracks. Mortar and concrete incorporating microencapsulated sodium silicate have shown enhanced self-healing performance in terms of decreased crack width and depth, recovered water toughness and regained strength (Kanellopoulos et al., 2016; Kanellopoulos et al., 2017; Al-Tabbaa et al., 2019). However, the above work has all focused on structural applications of concrete, with no attention paid on geotechnical underground con­ struction applications (Al-Tabbaa & Harbottle, 2015). Although there is great potential of using the developed microcapsules for stabilised clays, validation is needed as the microcapsule-based self-healing process is highly dependent on the composition and properties of the matrix. This study investigates the applicability of self-healing microencapsulated sodium silicate to stabilised clay. CT-scanning was used to examine the survival, distribution and rupture of the microcapsules. The release of the healing agent and the generation of secondary hydration prod­ ucts was investigated using SEM-EDX. Perme­ ability tests were conducted to assess the selfhealing efficiency in terms of the transport prop­ erties of the stabilised clay.

2.2

Microcapsules

For self-healing applications, microcapsules contain­ ing crystalline sodium silicate as a cargo material were adopted. The microcapsules were manufactured by Thies Technology Inc using interfacial polymer­ isation with polyurea as the shell material. The total sodium silicate content is calculated by analysing the element composition of the core materials. The weight fraction of sodium silicate is ~40% based on the EDX analysis, and the rest of the core materials is a mixture of the reactive monomers that were used for the synthesis of the shell. The microcapsules were observed under an SEM (Figure 3) and an aver­ age diameter of 210 μm was obtained using particle size analysis software. The buckling of the microcapsule is attributed to the loss of liquid during the drying process and thus they can be stored in powder form. Residual debris resulting from the microencap­ sulation process could also be observed among the microcapsules. 2.3

Sample preparation

The microcapsules were added in cement grout at a mass fraction of 6% with respect to the grout

Figure 3. SEM image of Thies microcapsules.

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weight, which was then mixed with kaolin clay for the preparation of cement stabilised clay samples. The grout with a water-to-cement ratio of 1:1 was prepared in a mixer of a rotational speed of 190 rpm. A soil-to-grout ratio of 3:1 was selected based on previous studies. The kaolin with a water content of 50% was utilised to produce a medium stiff clay. The density of the kaolin clay is 1700kg/m3 and it has a shear strength cu of 6.5kPa. The stabilised clay samples were placed in cylin­ drical moulds (50 mm in diameter and 100 mm in height) for 28-day curing at 20˚C and 100% relative humidity in curing tanks. The relevant tests were con­ ducted in due course and then the samples were returned in curing tanks for the 28-day self-healing process. 2.4

Permeability test

The vertical permeability of the stabilised clay speci­ mens was determined by a constant flow rate test using triaxial cell permeameters. Each sample was positioned on the pedestal of the base of the permea­ meter, with porous stones and filter paper at both ends of the specimen. Two rubber membranes secured by O-rings were placed around the samples to prevent the surrounding water in the cell from penetrating the sample. To prevent the rubber mem­ branes from separating from the sample and to improve saturation, the cell water pressure was raised to 50kPa and maintained throughout the test. A steady flow rate was applied at the bottom of the sample using a peristaltic pump which created a vertical upward flow through the sample. The water that permeated the samples in all the perme­ ability tests was potable water. A pressure transducer positioned at the inflow position measured the water pressure generated by the flow. This pressure was recorded every second, and the hydraulic conductiv­ ity k of the sample was calculated using Darcy’s law when a constant pressure was reached. 2.5

SEM-EDX analysis

can be obtained with the help of the image recon­ struction process after X-ray micro-CT test. A stabilised clay specimen with the size of 10×10×10 mm was cut from a cylinder sample after curing and it was cracked by unconfined compres­ sion. The sample was then scanned by the X-ray micro-CT setup (Nikon XT H 225 ST). The X-ray micro-CT system consists of a microfocus X-ray source, a rotation stage which allows for 360˚ imaging, and an image intensifier detector (Figure 4). A radiography of the test object is obtained from a comprehensive viewing angle. The radiography is an image of the attenuation of the primary X-ray beam based on the differences of chemical properties and physical properties (e.g., density and thickness) of testing materials. A flat panel detector with 2000 × 2000 pixel is used. Four parameters of X-ray micro-CT test were set as follow: X-ray energy is set as 70 kV and 85 mA, which are parameters determined by the sample geometry and the material composition; the geomet­ ric magnification is 39.5; the reconstructed image matrix had a volume of 2000 × 2000 × 1000 pixels; the effective pixel size is 5.05 μm. The X-ray absorption in each voxel, represented by the material specific X-ray absorption coefficient, is normalized to 16-bit gray values. 3 RESULTS 3.1

Recovery of permeability

The change in permeability of stabilised clay was used to assess the self-healing performance of con­ trol and microcapsule-containing samples. Stabilised clay samples were damaged using an unconfined compressive strength load frame until the peak strength was reached, after which the permeabilities of the cracked samples were determined. After 28 days of self-healing, the permeability tests were con­ ducted again to obtain the recovered permeabilities. The decrease in permeabilities shown in Figure 5

The scanning electron microscope (SEM) instrument used to obtain images of microcapsules embedded in soil mixed wall materials was a Phenom ProX equipped with an EDX (energy-dispersive X-ray spectroscopy) detector. Small chipped pieces were extracted from the crack faces of the specimens that were damaged by unconfined compression. Samples were coated with gold and examined under a 10kV accelerating voltage for imaging and 15kV for EDX analysis. 2.6

CT analysis

The X-ray micro computerised tomography (microCT) is applied to in-situ monitor the rupture process of the microcapsules in stabilised clay upon crack­ ing. Herein, the internal slice images of the specimen

Figure 4. Nikon XT H 225 ST X-ray micro-CT setup.

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Figure 5. The permeability recovery of control and microcapsule-containing samples.

indicates the sealing of cracks and the consequent closure of interconnected flow paths in both control and microcapsule-containing samples. Autogenous self-healing contributed around 50% reduction in per­ meability (from 6×10-8 to 3×10-8 m/s) after 28 days, while the microcapsule-containing samples showed a significant decrease from 1.2×10-7 to 1×10-8 m/s due to the combined effect of autogenous and autonomic self-healing. This suggests that the addition of microcapsules at a dosage of 6% by grout weight can heal the cracks and recover the permeability effectively, improving the self-healing performance of stabilised clay samples noticeably. 3.2

SEM-EDX analysis

SEM-EDX analysis was conducted to investigate the microstructure of the ruptured microcapsules and heal­ ing agents in the stabilised clay samples. The microcapsules were seen to disperse well throughout samples that were extracted from crack surfaces and showed clear breakage along with remnants of shell material (Figure 6a). Copious healing products in the cementitious matrix generated by the healing agent deposited on and around the microcapsules can also be observed. Sodium is a unique element that only exists in the sodium silicate core materials, so the traces of sodium found in the hydration products around the ruptured microcapsule by EDX analysis confirm that the sodium silicate healing agent was successfully released onto the crack’s surface and reacted with the surrounding matrix (Figure 6b). SEM images were also used to measure the shell thickness of microcapsules which was seen to vary between 2-3μm. Hydration products that appear to be calcium silicate hydrate (C-S-H) and other car­ bonation products, including clear ettringite needles and hexagonal portlandite crystals, were also observed deposited around the microcapsule shell (Figure 7). Despite the gaps between the microcap­ sule shell and the matrix, the microcapsules clearly

Figure 6. (a) SEM images of a ruptured microcapsule along with (b) EDX analysis of a point around the microcapsule.

showed breakage instead of debonding that would occur in the case of weak shell-matrix interfacial bonding, suggesting that fairly good bonding was maintained between the microcapsule shell and the stabilised soil matrix.

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Figure 7. The SEM image of the bonding between the cementitious matrix and the microcapsule shell.

3.3

CT scan analysis

For deeper examination, micro-CT scanning was conducted to observe the microcapsules embed­ ment in the matrix. The rupture of microcapsules triggered by cracking is revealed by the recon­ struction of CT scan results. After compression, several cracks can be observed on the cross sec­ tion in Figure 8a. Each crack passed through sev­ eral microcapsules and triggered the shell rupture. The microcapsules were extracted from the stabil­ ised clay matrix, and Figure 8b shows the good dispersion of microcapsules in the specimen, which increases the chance of propagating cracks passing through microcapsules. The roughness of the ruptured shell material can be seen by zoom­ ing in on the microcapsule and such surface texture can enhance adhesion to the matrix (Figure 8c). 4 CONCLUSIONS The work presented here is the first attempt to intro­ duce the concept of self-healing to cement stabilised clays by incorporating microencapsulated sodium silicate. The uniform dispersion and crack-triggered rupture of microcapsules were observed under CT, and the release of healing agents as well as the gen­ eration of healing products were confirmed by SEM­ EDX analysis. The effective healing of cracks was also verified by recovered permeability. The results demonstrated the significant potential of microcap­ sules as a self-healing approach for the development of more resilient and reliable cement stabilised clay materials.

Figure 8. CT scan 3D image reconstruction results of (a) CB matrix incorporating microcapsules; (b) extraction of embedded microcapsules; (c) crack-triggered rupture of microcapsules.

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ACKNOWLEDGEMENTS This research is financially supported by the EPSRC-funded project Resilient Materials for Life (EP/P02081X/1). The first author would also like to acknowledge the support from China Scholarship Council and Cambridge Trust.

REFERENCES Al-Tabbaa, A. & Harbottle, M.J. 2015. Self-healing mater­ ials and structures for geotechnical and geo-environmental applications. In: Proceedings of the XVI ECSMGE., pp. 589–594. Al-Tabbaa, A., Litina, C., Giannaros, P., Kanellopoulos, A., & Souza, L. 2019. First UK field application and per­ formance of microcapsule-based self-healing concrete. Construction and Building Materials, 208, 669–685. De Rooij MR, Tittelboom K Van, De Belie N, & Schlangen E. 2013. Self-Healing Phenomena in Cement-Based Materials: State-of-the-Art Report of RILEM Technical Committee 221-SHC: Self-Healing Phenomena in Cement-Based Materials. Springer Giannaros, P., Kanellopoulos, A., & Al-Tabbaa, A. 2016. Sealing of cracks in cement using microencapsulated sodium silicate. Smart Materials and Structures, 25(8), 084005.

Indraratna, B., Muttuvel, T., & Khabbaz, H. 2009. Modelling the erosion rate of chemically stabilized soil incorp­ orating tensile force–deformation characteristics. Canadian Geotechnical Journal, 46(1), 57–68. Kanellopoulos, A., Giannaros, P., & Al-Tabbaa, A. 2016. The effect of varying volume fraction of microcapsules on fresh, mechanical and self-healing properties of mortars. Construction and Building Materials, 122, 577–593. Kanellopoulos, A., Giannaros, P., Palmer, D., Kerr, A., & Al-Tabbaa, A. 2017. Polymeric microcapsules with switchable mechanical properties for self-healing con­ crete: synthesis, characterisation and proof of concept. Smart Materials and Structures, 26(4), 045025. Osman, A. A. M. 2007. Durability and mechanical proper­ ties of deep-mixed clays (Doctoral dissertation, Univer­ sity of Cambridge). Souza, L., & Al-Tabbaa, A. 2018. Microfluidic fabrication of microcapsules tailored for self-healing in cementi­ tious materials. Construction and Building Materials, 184, 713–722. Souza, L.R. 2017. Design and synthesis of microcapsules using microfluidics for autonomic self-healing in cementitious materials (Doctoral dissertation, University of Cambridge). Wang, F., Wang, H., & Al-Tabbaa, A. 2015. Timedependent performance of soil mix technology stabil­ ized/solidified contaminated site soils. Journal of haz­ ardous materials, 286, 503–508.

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Application of biopolymer hydrogel for ground hydraulic conductivity control under pressurized conditions I. Chang School of Engineering and IT (SEIT), University of New South Wales (UNSW), Australia

G-C. Cho Department of Civil and Envinronmental Engineering, Korea Advanced Institute of Science and Technology

A.T.P Tran Department of Hydrogeological and Geological Engineering, Hue University of Sciences

ABSTRACT: The use of hydraulic barriers to control the water leakage and flow into underground struc­ tures such as tunnels has become a common method in geotechnical engineering practice. There are different types of materials used for hydraulic conductivity control purposes such as soil-bentonite and cement­ bentonite mixtures, and chemical admixtures including asphalt, sodium silicate and so on. Underground hydraulic barrier materials are required to well-resist against the in-situ earth pressure and hydrostatic water pressure. In this study, a gel–type polysaccharide biopolymer gellan gum–has been used to reduce the hydraulic conductivity of sand. It results from the fact that gellan gum biopolymer forms high viscous hydrogel once thermo-gelated, which is expected to have a high structural resistance even at high external pressure conditions. Therefore, the hydraulic conductivity of gellan gum-treated sand is assessed with vertical confine­ ment and water pressure increase. It is interesting to observe that the hydraulic conductivity of sand exponen­ tially decreases with gellan gum content.

1 INTRODUCTION Recently, the possibility and feasibility of using gellan gum-treated soils in geotechnical engineer­ ing practices have been postulated and verified (Chang and Cho 2019; Chang et al. 2016). As gellan gum reduces the hydraulic conductivity of soils significantly via pore clogging, gellan gumtreated soil is suggested to be applied as a hydraulic barrier or a grouting material to control the ground­ water around underground structures (Chang et al. 2016). To investigate the pore-clogging effect of hydrogels and biopolymer treatment to soils, current studies mostly accept 2 different approaches. The first type is flowing pure (deionized) water through hydrogel (or biopolyemer) mixed soils (Al-Darby 1996; Andry et al. 2009; Bouazza et al. 2009; Chang et al. 2016; Narjary et al. 2012), while the other type is using hydrogel solutions as the main fluid to flow through clean (untreated) soils (Ete­ madi et al., 2003; Khachatoorian et al. 2003). This study introduces a new apparatus which can simu­ late and apply external pressures, i.e., vertical

confinement and hydrostatic water pressure to clean and saturate the biopolymer-treated sands. As pure water is used to be the main fluid for flowing, the test method suggested by this study can be regarded as an improved method for the first type approach mentioned above. Gellan gum biopolymer–sand mixtures with dif­ ferent biopolymer to soil contents in mass (0%, 0.5%, 1.0%, and 2.0%) were pre-consolidated under different confinement pressures before commencing hydraulic conductivity tests. Once pre-consolidation is completed, excess hydraulic pressure has been applied with restricted volumetric strain to represent an in-situ pressurized condition at a K0 geometry, then, the hydraulic conductivity of gellan gumtreated sand was measured. 2 MATERIALS AND METHOD 2.1

Materials

Juminjin sand. Jumunjin sand is a standard sand material in Korea classified as SP (USCS), having an

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average particle size of 0.46 mm, specific gravity (Gs) of 2.65. The coefficient of uniformity (Cu) and the coefficient of gradation (Cc) are found to be 1.39 and 0.76, respectively (Chang and Cho 2019). Gellan gum biopolymer. Gellan gum is a hydrogel-typepolysaccharide biopolymer produced by Pseudomonas elodea bacterium, which has been investigated in the fields of pharmaceutical technol­ ogy, biomedical applications (Osmałek et al. 2014), food industry (Morris et al. 2012). In geotechnical engineering perspective, gellan gum has been attempted to enhance shear strength parameters, especially at saturated conditions due to its high sta­ bility in water induced by thermogelation (Chang and Cho 2019; Chang et al. 2017; Chang et al. 2015; Im et al. 2017). In this study, low acyl gellan gum biopolymer (Sigma Aldrich; CAS No.71010-52-1) has been used. Gellan gum-treated sand. Biopolymer hydrogels were mixed with soils at target biopolymer to soil contents in mass as 0.5%, 1.0%, and 2%. To allow thorough mixing, the initial water content has been set at 33%. Gellan gum powder was first dissolved and hydrated into heated (100°C) deionized water to obtain uniform gellan gum solution. Then, dry sand and hot gellan gum solution were uniformly mixed and cooled to ensure adequate thermogelation (Chang et al. 2015). 2.2

Hydraulic conductivity test

The hydraulic conductivity of gellan gum-treated sands was determined by using a pressurized hydraulic conductivity apparatus (Figure 1). The gellan gum hydrogel-sand mixtures were placed into a cylindrical cell which is 9.3 cm in height and 8.0 cm in diameter. At the top and bottom of the specimen, filter paper and porous plate were placed thus water can evenly distribute within specimen during the experiment. After the specimen was fully set up and cooled down, confining pressure was loaded to the soil under drained condition using a pneumatic air compressor to apply effective

Figure 1. Diagram of hydraulic conductivity setup.

stresses (100 and 200 kPa) to soil samples. The consolidation process lasted for 24 hours at where the vertical strain of soil shows no more changes. An initial water pressure of 70 kPa was applied to the specimen using a high-pressure precision syr­ inge pump. Once the inlet flow rate of water reached zero, it has been regarded to be the point where specimen is fully saturated and then the per­ meability test was conducted. The permeability of soil was observed at various water pressure condi­ tion beginning from 70 kPa and reaching maximum value which is identical to the external confinement (i.e effective stress) level. The saturated hydraulic conductivity is calculated based on Darcy’s law

where V is collected volume of water, L is the height of soil specimen, A is the cross-section area of soil specimen, h is the head difference, and t is the time required to collect the outlet V. 3 EXPERIMENTAL RESULTS Figure 2 shows the changes in the hydraulic con­ ductivity of sands with water pressure increase. For untreated sand, the hydraulic conductivity values were approximately 1 × 10-6 m/s, regardless of the confinement pressure and water pressure conditions (Figure 2a). Meanwhile, the change in hydraulic conductivity of gellan gum-treated sand with water pressure increase shows a recognizable difference affected by the confinement pressure (Figures 2b, c and d), where higher vertical confinement renders lower hydraulic conductivity due to the denser packing between sand grains and gellan gum hydrogels. Furthermore, hydrophilic water absorption and water holding capacity of gellan gum plays an important role for the water flow behavior through sands. For pure sand, pressurized water easily and quickly flew through inter-granular pore spaces due to the lack of hydrophilic interactions. However, gellan gum-treated condition shows significant water flow retention with higher gellan gum to soil con­ tents (%) as shows in Figures 2(b, c & d). The relationship between average hydraulic con­ ductivity and gellan gum content is expressed in Figure 3 using measurement data from Figure 2. The hydraulic conductivity rapidly drecreases for gellan gum contents up to 1%, and then converges to a steady value, which is also observed by Chang et al. (2016). Thus, the 1% biopolymer to soil con­

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Figure 3. Reduction in hydraulic conductivity of treated sand with gellan gum content.

4 CONCLUSION In this study, in order to see the effects of gellan gum hydrogel on the hydraulic conductivity reduction of sand, the hydraulic conductivity tests were conducted under pressurized conditions using pressurized hydraulic conductivity apparatus. The hydraulic test­ ing results show that gellan gum can reduce the hydraulic conductivity of sand via water absorbing, and holding processes. The gellan gum treated sand can withstand higher water pressure. The confinement pressure affects the hydraulic conductivity via con­ trolling pore size within the treated-soil system. According to the effective soil hydraulic conductivity control behavior of gellan gum under in-situ stress conditions, gellan gum can be considered to be used for underground grouting/injection practice. The effect of hydraulic reducing with gellan gum content is shown to decrease nonlinearly and level off at con­ tent higher than 1% of gellan gum. It can be sug­ gested that the most economical and efficient content of gellan gum for grouting should be 1%.

ACKNOWLEDGEMENT This research was supported by a grant (19AWMP­ B114119-04) from the Water Management Research Program funded by the Ministry of Land, Infrastructure, and Transport (MOLIT) of the Korean Government; a National Research Founda­ tion of Korea (NRF) grant funded by the Korean Government (MSIP) (No. 2017R1A2B4008635); a grant (19SCIP-B105148-05) from the Construc­ tion Technology Research Program funded by the MOLIT of the Korean Government.

Figure 2. The change in hydraulic conductivity with vary­ ing water pressure.

tent in mass can be postulated to be the condition where gellan gum hydrogels fully clogs intergranular pore spaces of the the sand used in this study.

REFERENCES Al-Darby, A. 1996. The hydraulic properties of a sandy soil treated with gel-forming soil conditioner. Soil technol­ ogy, 9(1-2), 15–28.

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Andry, H., Yamamoto, T., Irie, T., Moritani, S., Inoue, M., and Fujiyama, H. 2009. Water retention, hydraulic con­ ductivity of hydrophilic polymers in sandy soil as affected by temperature and water quality. Journal of Hydrology, 373(1-2), 177–183. Bouazza, A., Gates, W. P., and Ranjith, P. G. 2009. Hydraulic conductivity of biopolymer-treated silty sand. Géotechnique, 59(1), 71–72. Chang, I., and Cho, G.-C. 2019. “Shear strength behavior and parameters of microbial gellan gum-treated soils: from sand to clay.” Acta Geotechnica, 1–15. Chang, I., Im, J., and Cho, G.-C. 2016. Geotechnical engineering behaviors of gellan gum biopolymer trea­ ted sand. Canadian Geotechnical Journal, 53(10), 1658–1670. Chang, I., Im, J., Lee, S.-W., and Cho, G.-C. 2017. Strength durability of gellan gum biopolymer-treated Korean sand with cyclic wetting and drying. Construction and Building Materials, 143, 210–221. Chang, I., Prasidhi, A. K., Im, J., and Cho, G.-C. 2015. Soil strengthening using thermo-gelation biopolymers. Con­ struction and Building Materials, 77, 430–438.

Etemadi, O., Petrisor, I. G., Kim, D., Wan, M.-W., and Yen, T. F. 2003. Stabilization of metals in subsurface by biopolymers: laboratory drainage flow studies.” Soil and Sediment Contamination, 12(5), 647–661. Im, J., Tran, A. T., Chang, I., and Cho, G.-C. 2017. “Dynamic properties of gel-type biopolymer-treated sands evaluated by Resonant Column (RC) tests.” Geomechanics and Engineering, 12(5), 815–830. Khachatoorian, R., Petrisor, I. G., Kwan, C.-C., and Yen, T. F. 2003. “Biopolymer plugging effect: laboratory-pressurized pumping flow studies.” Journal of Petroleum Science and Engineering, 38(1-2), 13–21. Morris, E. R., Nishinari, K., and Rinaudo, M. 2012. “Gel­ ation of gellan–a review.” Food Hydrocolloids, 28(2), 373–411. Narjary, B., Aggarwal, P., Singh, A., Chakraborty, D., and Singh, R. 2012. “Water availability in different soils in relation to hydrogel application.” Geoderma, 187, 94–101. Osmałek, T., Froelich, A., and Tasarek, S. 2014. “Applica­ tion of gellan gum in pharmacy and medicine.” Inter­ national journal of pharmaceutics, 466(1-2), 328–340.

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Soil enhancement via microbially induced calcite precipitation C. Konstantinou & G. Biscontin Department of Engineering, University of Cambridge, Cambridge, UK

ABSTRACT: Poor performance of soils due to low strength has been always a key challenge for the con­ struction industry, especially for underground structures. While several methods have been proposed in litera­ ture to strengthen granular soils, these often lead to high disturbance. Microbially induced calcite precipitation (MICP) is a bio-cementation technique that could be applied in granular materials to enhance strength and stiffness. This paper measures the tangent modulus and the shear wave velocity of bio-cemented sands at varying cementation levels to assess the strength and stiffness enhancement. SEM images were taken and microCT scans were performed on those specimens to identify the changes in porosity due to the intro­ duction of the cementation within the matrix.

1 INTRODUCTION 1.1

Background

In recent years, the use of biological processes in civil and environmental engineering has emerged as a new area with very promising results. An example is a bio-cementation technique called microbially induced calcite precipitation (MICP), in which granular material is bound with a carbonate based agent (Whiffin, 2004). Compared to other chemical methods, MICP is considered to be an environmen­ tally friendly technique, since it is a naturally occur­ ring process. MICP has been successfully applied in cases where low amount of cementation is needed and the soil’s permeability needs to be retained, such as soil stabilization in liquefiable grounds or internal or external erosion control in dams (Montoya & DeJong, 2015; Jiang et al., 2017). The MICP technique is also appropriate when heavy cementation is required in order to target strength enhancement. Under careful design and implementation of the method, the cementation level and the binding material’s characteristics could be easily controlled (Konstantinou, 2020). This bio-chemical process is described by two chemical equations. The first one describes the hydrolysis of urea where the molecule splits into two parts (carbonate and ammonia). The reaction is pHdependent, rather slow in aqueous solution, and irre­ versible. The urease activity is influenced by a variety of environmental conditions (Whiffin, van Paassen and Harkes, 2007; van Paassen, 2009; Mar­ tinez et al., 2013). The carbonate ions then react with the available calcium ions in the environment

to precipitate at particle contacts (second chemical equation). The source for the reactants could be both Na2CO3 and CaCl2 (Kawano et al., 2002). This paper investigates the potential of applying bio-cementation on sandy soil to increase strength and stiffness, and assesses the reduction of porosity. 1.2

Method optimization

The specimens being produced should be uniformly and isotropically cemented to allow for meaningful mechanical testing. The testing protocol should also aim to produce repeatable specimens at the targeted cementation levels. The parameters of the protocol should be carefully selected in order to accommo­ date the needs of this study. Literature contains a vast amount of information regarding MICP and the parameters that control the bio-cementation pro­ cess (Whiffin, 2004; Rebata-Landa, 2007; van Paas­ sen, 2009; Al Qabany et al., 2012; Al Qabany & Soga, 2013; Cheng et al., 2013; Martinez et al., 2013; Montoya et al., 2013; Zhao et al., 2014; Cheng et al., 2017). The amount and population of bacteria (optical density), the injection intervals (time between two subsequent injections), the chem­ ical solution concentration, the total duration of the injection program, and the type of injection (via gravity, flow rates etc.) are proven to be of para­ mount importance to achieve the above targets. Several bacterial optical densities have been pro­ posed in literature (OD600 0.3- 2.0) (DeJong et al., 2006; Mortensen et al., 2011; Al Qabany et al., 2012; Martinez et al., 2013; Zhao et al., 2014; Feng & Montoya, 2015; Montoya & DeJong, 2015). Although there is no conclusion with regards to the

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amount of bacteria that should be injected, the gen­ eral attitude is that very high optical densities (large bacterial population) could lead to clogging of the pores. The bacteria population profile versus time is an exponential curve, indicating that growth is quite fast. Higher bacteria concentration at the injection phase, though, results in clogging of the pores near the source. Therefore, it is suggested that a bacteria solution with lower optical density should be injected. The reaction rates seem to depend on the urease activity (the rate at which bacteria hydrolise urea) and the amount of chemicals being injected. It is also recognized that the distribution of the microbes within the granular matrix controls the uni­ formity of the specimens, which in turn depends on the injection intervals (or retention periods) and the initial concentration of the microbes (Martinez et al., 2013). Larger retention times allow for all reactions to take place even in the most isolated areas within the grain network (van Paassen et al., 2010). For this reason, stopped flow (in the form of injections) is preferred over a continuous augmentation scheme, whilst the flow direction or flow reversal has very little effect on the overall efficiency of the process (Martinez et al., 2013). Flow via gravity is con­ sidered one of the best injection schemes because high flow velocities are achieved without dislocating the particles (Mujah et al., 2017). A low calcium chloride concentration is generally considered the optimum way of injecting to avoid any potential clogging of the pores (Al Qabany & Soga, 2013). However, although very low calcium chloride concentrations can lead to uniformly cemented samples, they are not proven to be efficient in bonding the grains and, hence, the stiffness of the samples may not increase as much as expected. Research has shown that calcium chloride is the limiting factor over urea in the MICP reactions with the optimal ratio (urea/CaCl2) of more than 1 and sometimes even above 7 (Whiffin, 2004; RebataLanda, 2007; van Paassen, 2009; Al Qabany et al., 2012; Al Qabany & Soga, 2013; Cheng et al., 2013; Martinez et al., 2013; Montoya et al, 2013; Zhao et al., 2014; Cheng et al., 2017). The relationship between the chemical solution concentrations and bacterial density was also studied and showed that the cementation solution consump­ tion depends on the microbes population (Zhao et al., 2014). Even when higher concentrations of chemicals are used, they will not be fully utilized if the bacterial density is low. The mechanical properties of bio-treated sands specimens are affected by the experimental protocol used. Extensive studies have been conducted on all experimental parameters to observe their effects on the physical and engineering properties of the bio­ enhanced sands (Al Qabany & Soga, 2013; Feng & Montoya, 2015; Montoya & DeJong, 2015). For example, low chemical concentrations lead to better strength enhancement, and very low concentrations

have the opposite effect (Al Qabany & Soga, 2013; Lin et al., 2015). Cheng et al. (2013) demonstrated that lower levels of saturation with chemical solution and bac­ terial suspension lead to higher strength. This is attributed to a better distribution of the calcite crys­ tals within the matrix at contact-to-contact locations. The same study also proved the ability of the method to retain permeability. (Al Qabany & Soga, 2013) showed that the use of a higher chemical concentra­ tion solution results in less homogeneous precipita­ tion, and therefore greater reduction of permeability. The resulting mechanical properties are also affected by the characteristics of the grains. For example, relative density is also a key parameter as it controls the number of contacts between the par­ ticles and thus strength (Al Qabany & Soga, 2013). 1.3

Scope of this study

This paper presents a parametric study to assess the stiffness parameters of bio-cemented sands utilizing a fixed protocol (bacterial population, chemical solu­ tion concentration and type of injection) in order to create cemented samples with similar microstructural characteristics (calcite crystals size, shape and type). All samples were deposited at the same relative dens­ ity and were bio-treated in the same way. The intro­ duction of the chemical solution was achieved via several injections. A full range of cementation levels was secured by varying the number of injections, and thus varying the chemical solution volume. Speci­ mens with similar initial pore networks, but filled by different amounts of calcite, were obtained. 2 EXPERIMENTAL WORK 2.1

Materials

Sporosarcina pasteurii were used as their urease­ synthesis behaviour is well-defined and their ureoly­ tic activity has been demonstrated to be higher than many other alternative species (Whiffin, 2004). Batch experiments were conducted under aerobic conditions and in a sterile environment for the bac­ terium strain. The medium is NH4-YE, consisting of 20 g/L yeast extract, 10 g/L ammonium sulphate, 20 g/L agar, and 0.13 M Tris buffer (base). After 24 h of incubation at 30oC, the culture was harvested and stored at 4oC. Before MICP treatment, bacteria colonies extracted from the NH4-YE medium were introduced into in a urea-rich NH4-YE solution medium (without agar), which was then placed in a shaking incubator for 24 hours. Silica sands supplied by David Ball Specialist Sand were used to prepare the soil specimens for the experiments of this study. Sand fraction D was used for these experiments with mean particle diameter of about 180 μm. The particle size distribution curve is shown in Figure 1. The grains’ shape is subrounded

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Figure 2. The experimental setup: injection via gravity; the removal of the previous solution and introduction of new solution was achieved by removing the placing the tube at a lower level.

Figure 1. Particle size distribution curve of the sand used in this study.

trimmed to reach a height of 150-160 mm to achieve as uniform precipitation profiles as possible. Once the injection phase was completed the speci­ mens were removed from the moulds and placed in an oven to dry at 105 oC.

with medium sphericity. Silica (SiO2) constitutes more than 98% of the sand composition. 2.2

Identification of protocol parameters

A set of preliminary experiments was conducted to establish the experimental protocol. Syringes of internal diameter 35 mm were used as moulds for sample preparation and oriented with the injection point at the top. The samples had a height of 9 cm and the injection method was a constant rate of 10 ml/min by gravity. Once the sample was placed in the mould and densified to a targeted porosity of 0.38-0.42, it was sat­ urated with water. Bacterial suspensions with optical density of 1.5-2.0 were injected first. Subsequent injec­ tions with cementation solution had a retention time of 24 hours. A range of calcium chloride concentrations was used (0.25 M, 0.33 M, 0.5 M, 0.75 M, 1 M). The ratio of urea to calcium chloride was also fixed at 3:2. The results indicated that, as the chemical concen­ tration of calcium chloride increased, the chemical efficiency decreased and a less uniform calcium car­ bonate precipitation profile was observed. Therefore, a low chemical concentration was chosen for cal­ cium chloride (0.25 M) and urea (0.375 M). In order to address the uniformity issues, the injection rate of the bacteria had to be higher. 2.3

2.4

Measurement of tangent modulus

Unconfined compressive tests were performed to examine the correlation between the degree of cementation and the modulus of the samples. The specimens were prepared and tested according to ASTM (2004) and ASTM (2013). The loading of the samples was strain-controlled with a rate of 1.14 mm/min. During testing, the specimens were placed in plastic bags to preserve all the fragments after failure. The tangent modulus was calculated as the slope of the stress-strain curve when the stress reached 50% of the ultimate strength, as shown in Figure 3.

Large scale experiments

Larger samples of 70 mm diameter were prepared once the protocol was shaped with the aid of the small scale experiments. As shown in Figure 2, the sand was first placed in the mould and densified to reach a targeted density. The injections were from top to bottom via gravity. The retention time was 24 hours for both bacteria and cementation injections. Before each injection the previous solution was removed. In all phases the sample was saturated. The samples were initially 220 mm and the top and bottom parts were

Figure 3. An example of a stress-strain curve obtained through the unconfined compressive strength tests and the calculation of the tangent modulus.

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2.5

Shear wave velocity measurements

A convenient way to measure the small strain shear stiffness of soil is using shear waves measurement with bender elements. The oven dried specimens were placed in a triaxial system, on pedestals equipped with bender elements (see Figure 4). A slot was created for the bender elements on the two ends of the specimens and filled with soft clay to achieve coupling between the rock and the ceramic material. The frequency of the shear waves was varied from 2kHz to 5kHz. The peak-to-peak method was selected to define the travel time from the signal. The shear wave travel time is defined as the time interval between the positive peak of the transmitted signal and the first major peak on the received signal, and the shear wave velocity is represented by the velocity calculated by this travel time. This method relies on the quality of the signals (Leong, E. C., Yeo, S. H. & Rahardjo, 2005). The distance between the tips is defined as: Ltt ¼ Hspecimen - hbottom tip - htop tip

ð1Þ

In the current setup, the top bender element’s length (h) was 3.7 mm and the bottom’s was 3 mm. Shear wave velocity is and the shear stiffness is calculated as follows Gmax ¼ ρVs2 . tcorr is the correction for the time delay due to the equipment (in this case it was 0.0336 ms). Figure 5 shows an example of a transmitted and received signal and the travel time identification. A negative peak was observed in all received

Figure 5. An example of a transmitted and a received signal. The peak-to-peak method was chosen for the travel time calculation.

signals; this was attributed to mean-field effects that can mask the arrival of the shear wave for 0:255λ=L54 where λ is the wavelength and L the distance between source and receiver (Viggiani & Atkinson, 1995). Post-test calculations showed that this ratio is around 1, which falls in the range defined above. However, the purpose of this test was to iden­ tify the relative change of the shear wave velocity with respect to the cementation level change which showed consistent results. 2.6

Calcite content

A calcium carbonate measurement chamber (calci­ meter) as described by ASTM (2014) was used. The method relies on quantifying the gas released by the treatment of the sample with hydrochloric acid (HCl). The reaction CaCO3(s) + 2HCl(aq) → CaCl2(aq) +CO2(g) + H2O(l) leads to the dissolution of calcite and the release of carbon dioxide (ASTM, 2014). 30 gr of sample and 30 mL of hydrochloric acid were used to measure the calcium content. The sam­ ples were first dried, grounded, and weighed before they were placed in the chamber. To correlate the pres­ sure reading and the amount of calcium carbonate, a calibration curve was constructed: calcium carbonate (Fisher Scientific Calcium carbonate 98+ %) was left under varying concentrations to react with hydrochlor­ ide acid. The cement content (Cw) calculation in this study is defined as the weight of the calcite over the total weight of the specimen (particles and calcite). 2.7

Figure 4. The experimental setup for the measurement of the shear wave velocity.

SEM imaging and MicroCT analysis

Scanning electron microscope (SEM) images of bio­ treated samples were taken to investigate the morph­ ology and distribution of calcite crystals and their

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bonding network. The microscopy investigation was carried out with a Hitachi S3400 scanning electron microscope. All bio-treated samples were dried at 100oC for 24h before conducting the microscopy analysis. Scans of the bio-treated samples were performed with the X-ray μ-CT high energy micro-tomography scanner (X-Tec Systems). The voxel size of the images was around 25-30 μm. An example of a cubic reconstruction is shown in Figure 6, with its associated gray values histogram depicted in Figure 7. The histogram was analysed to identify the threshold of the gray value between the solid and gas phases. As shown in Figure 7, the two peaks identify the two different phases and the boundary is defined in the middle of the two peaks.

Since the densities of calcite and silica are very similar, it was hard to identify the threshold between them at this resolution; however, a qualitative assess­ ment of this threshold was enough for the purposes of this paper. The interest focused on the pore net­ work, rather than the calcite bonds’ location. 3 RESULTS & DISCUSSION 3.1

Stiffness assessment

The shear stiffness is significantly enhanced when cementation is added to the granular matrix. The tan­ gent modulus of the bio-treated sands with respect to the cementation level is presented in Figure 8. The cemented sands ’stiffness is enhanced by the add­ ition of cementation from 60 MPa at 5% cementation to 400 MPa at 10% cementation. The exponential fit has an R-squared value of 0.879 (shown in the graph) while the power fit has an R-squared value of 0.873 in the following forms:

The sharp peak that is immediately followed by a dramatic decrease of the strength indicates brittle failure which occurred at less than 2% strain as shown in Figure 3. At lower cementation levels the samples show a characteristic axial splitting mode, while at higher degrees of cementation the failure mode transi­ tions to a shear-like mode. The threshold at which the fracture mode changes was identified at around 8% cementation level. Failure planes are shown in Figure 9 for low and high cementation levels. The shear wave velocity increases as a linear function of the cement content, as shown in Figure

Figure 6. An example of a reconstructed volume of a bio-treated sample.

Figure 7. An example of a histogram of the gray values obtained from the x-ray analysis. The first peak corresponds to the gas phase (air) and the second peak corresponds to the solid phase (silica and calcite).

Figure 8. The tangent modulus with respect to the cemen­ tation level of the bio-treated specimens.

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Both measurements of strength and stiffness showed a significant enhancement of the properties of the sands. The bio-cementation method could indeed be applied in the field to increase the soil’s shear strength and to lower the deformations and compressibility of the soil. 3.2

Figure 9. The two failure modes being observed in the unconfined compressive strength tests. Axial splitting was observed at low cementation levels (left image) while shear failure was observed at high cementation levels (right image).

10. The shear wave velocity is 250 m/s at about 4% cementation, increasing up to 500 m/s for the highest cementation level studied. The rate of increase is about 35 m/s per 1% by weight of cementation. Tangent modulus measures the modulus for larger strains whilst shear wave velocity gives an estimate of the small strain shear modulus. Although there is a difference in the scale, the two measures are expected to be related. A linear fit in shear wave vel­ ocity would produce a quadratic expression for small strain G with respect to the cementation level. The tangent modulus data points in this study could be fitted with an R-squared value of 0.84 providing a comparable relation between the two metrics. However, the rate of shear modulus degradation may also be nonlinear with respect to cementation, in which case the tangent modulus and the small strain modulus would not necessarily follow the same functional dependency on cementation.

Figure 10. Shear wave velocity measurements with respect to the cementation level of the bio-treated specimens.

Microstructure

The microstructure was investigated through SEM imaging and microCT analysis. The SEM images in Figure 11 show that the cal­ cite crystals land on spots against the grain surfaces and their size is small compared to the void size. At lower cementation levels these crystals are not enough in number to fill the pores. At the highest possible cementation levels the calcite crystals seemed to clog the pores in such way that the poros­ ity and permeability reduced dramatically. The CT-scanning uncovered more information regarding the pore network. In Figure 12 the pore net­ work is in black, calcite crystals in a light gray and grains in red. The calcite crystals are formed on pore throats (or at the contact points between particles) and the voids remain intact. The final measured porosity

Figure 11. SEM images of the bio-treated sands. The cal­ cite crystals are significantly smaller compared to the grain size.

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Figure 12. MicroCT images of the bio-treated sands. The black colour is the pore network, the red colour shows par­ ticles and the gray is calcite.

was calculated from the microCT images and was found to fall in the range between 0.3-0.4, reducing by 0.1 when heavy cementation was introduced into the host material. This gradual reduction of porosity can be easily controlled under a fixed MICP protocol at a targeted cementation level. 4 CONCLUSION The stiffness of bio-treated sands were investigated to assess the applicability of microbially induced calcite precipitation (MICP) as a way to enhance the stiffness of the soil without causing much disturbance on the other properties of the soil. The method is successful in increasing the strength and stiffness of the granular materials as the tangent Young’s modulus shows an exponential increase as the cementation level increases whilst the shear wave velocity follows a linear path. MicroCT scanning and SEM images demonstrate that the MICP technique can reduce permeability dramatic­ ally when heavy cementation is introduced as all the pores are filled with calcite.

AKNOWLEDGEMENTS This work has been carried out at the Department of Engineering at the University of Cambridge. We thank Simon Marshall and Graham Smith for their help in microCT imaging. The authors would like to acknow­ ledge the funding and technical support from BP through the BP International Centre for Advanced Materials (BP-ICAM) which made this research possible.

REFERENCES ASTM, 2004. Compressive Strength and Elastic Moduli of Intact Rock Core Specimens under Varying States of Stress and Temperatures. Stress: The International Jour­ nal on the Biology of Stress, 4–11.

ASTM, 2014. Standard Test Method for Rapid Determin­ ation of Carbonate Content of Soils. ASTM Inter­ national, 1–5. Cheng, L., Cord-Ruwisch, R., and Shahin, M. a., 2013. Cementation of sand soil by microbially induced calcite precipitation at various degrees of saturation. Canadian Geotechnical Journal, 50 (1), 81–90. Cheng, L., Shahin, M.A., and Mujah, D., 2017. Influence of Key Environmental Conditions on Microbially Induced Cementation for Soil Stabilization. Journal of Geotechnical and Geoenvironmental Engineering, 143 (1), 04016083. DeJong, J.T., Fritzges, M.B., and Nüsslein, K., 2006. Microbially induced cementation to control sand response to undrained shear. Journal of Geotechnical and Geoenvironmental Engineering, 132 (11), 1381–1392. Feng, K. and Montoya, B.M., 2015. Influence of con­ finement and cementation level on the behavior of microbial-induced calcite precipitated Sands under monotonic drained loading. Journal of Geotechnical and Geoenvironmental Engineering, 142 (1), 04015057. Jiang, N.-J., Soga, K., and Kuo, M., 2017. Microbially Induced Carbonate Precipitation for Seepage-Induced Internal Erosion Control in Sand–Clay Mixtures. Jour­ nal of Geotechnical and Geoenvironmental Engineering, 143 (3), 04016100. Kawano, J., Shimobayashi, N., Kitamura, M., Shinoda, K., and Aikawa, N., 2002. Formation process of calcium carbonate from highly supersaturated solution. Journal of Crystal Growth, 237–239 (1–4), 419–423. Konstantinou, C., 2020 Hydraulic fracturing of artificially generated soft sandstones, University of Cambridge. PhD thesis. Leong, E. C., Yeo, S. H. & Rahardjo, H., 2005. Measuring shear wave velocity using bender element. Geotechnical Testing Journal, 28 (5), 488–498. Lin, H., Suleiman, M.T., Brown, D.G., and Kavazanjian, E., 2015. Mechanical Behavior of Sands Treated by Microbially Induced Carbonate Precipitation. Journal of Geotechnical and Geoenvironmental Engin­ eering, 142 (2), 04015066. Martinez, B.C., DeJong, J.T., Ginn, T.R., Montoya, B.M., Barkouki, T.H., Hunt, C., Tanyu, B., and Major, D., 2013. Experimental Optimization of Microbial-Induced Carbonate Precipitation for Soil Improvement. Journal of Geotechnical and Geoenvironmental Engineering, 139 (4), 587–598. Montoya, B.M. and DeJong, J.T., 2015. Stress-Strain Behavior of Sands Cemented by Microbially Induced Calcite Precipitation. Journal of Geotech­ nical and Geoenvironmental Engineering, 141 (6), 1–10. Montoya, B.M., DeJong, J.T., and Boulanger, R.W., 2013. Dynamic response of liquefiable sand improved by microbial-induced calcite precipitation. Géotechnique, (4), 302–312. Mortensen, B.M., Haber, M.J., Dejong, J.T., Caslake, L.F., and Nelson, D.C., 2011. Effects of environmental fac­ tors on microbial induced calcium carbonate precipitation. Journal of Applied Microbiology, 111 (2), 338–349. Mujah, D., Shahin, M.A., and Cheng, L., 2017. State-of-the -Art Review of Biocementation by Microbially Induced Calcite Precipitation (MICP) for Soil Stabilization. Geomicrobiology Journal, 34 (6), 524–537.

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van Paassen, L., 2009. Biogrout: Ground Improvement by Microbially Induced Carbonate Precipitation. Technol­ ogy. Delft University of Technology. van Paassen, L.A., Ghose, R., van der Linden, T.J.M., van der Star, W.R.L., and van Loosdrecht, M.C.M., 2010. Quantifying Biomediated Ground Improvement by Ureolysis: Large-Scale Biogrout Experiment. Journal of Geotechnical and Geoenvironmental Engineering, 136 (12), 1721–1728. Al Qabany, A. and Soga, K., 2013. Effect of chemical treat­ ment used in MICP on engineering properties of cemented soils. Geotechnique, 63 (4), 331–339. Al Qabany, A., Soga, K., and Santamarina, C., 2012. Fac­ tors Affecting Efficiency of Microbially Induced Calcite Precipitation. Journal of Geotechnical and Geoenviron­ mental Engineering, 138 (8), 992–1001.

Rebata-Landa, V., 2007. Microbial Activity in Sediments: Effects on Soil Behavior. Georgia Institute of Technol­ ogy Atlanta. Viggiani, G. and Atkinson, J.H., 1995. Interpretation of bender element tests. Geotechnique, 45 (1), 149–154. Whiffin, V.S., 2004. Microbial CaCO3 Precipitation for the Production of Biocement. Murdor University Reposi­ tory. Murdor University. Whiffin, V.S., van Paassen, L.A., and Harkes, M.P., 2007. Microbial carbonate precipitation as a soil improvement technique. Geomicrobiology Journal, 24 (5), 417–423. Zhao, Q., Li, L., Li, C., Li, M., Amini, F., and Zhang, H., 2014. Factors Affecting Improvement of Engineering Properties of MICP-Treated Soil Catalyzed by Bacteria and Urease. Journal of Materials in Civil Engineering, 26 (2004), 1–10.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Base grouting against uplifting water for a deep excavation in Taipei basin Hung-Jiun Liao & Shao-Jie Weng Department of Civil and Construction Engineering, National Taiwan University of Science and Technology, Taipei, Taiwan

Shih-Hao Cheng Taiwan Building Technology Center, National Taiwan University of Science and Technology, Taipei, Taiwan

Ricky K.N. Wong Taipei Branch, SanShin Construction Corporation Ltd., Taipei, Taiwan

ABSTRACT: An deep excavation up to 33 m in depth and retained by 53 m deep diaphragm wall was carried out in the silty sand and gravel layers of Taipei basin with high groundwater level. A base grouting work was conducted to cut off the uplifting groundwater coming from gravel layers. Tube-a-manchette (TAM) grouting method was adopted in this project. Based on the recorded pressures response and the spatial coordinates of each grout port on the Manchette tube (M-tube), 2D grouting pressure response contours at given depths can be established using commercial software and then used to pin-point the areas of low grouting pressure. If neces­ sary, TAM grouting might be repeated in the low pressure spots/areas. On-site permeability tests were run on the base grouting zone. The measured coefficients of permeability were all below the required value of 5 ×10-5 cm/sec. In addition, the in-situ transmissivity (T) and storage coefficient (S) of the conglomerate gravel layer were low. Therefore, only six 90 m-deep wells were needed to lower the groundwater level of gravel layer to keep a good base stability (FS > 1.2) against uplifting water pressure for a 160 m x 26 m excavation site with a depth of 33 m.

1 INTRODUCTION Deep excavation in the inter-layered silty sand and silty clay formation of Taipei basin under high groundwater pressure is always a challenge to the geotechnical engineers. Over the past three decades, the construction of Taipei MRT system was not with­ out geotechnical failures (Moh et al. 1997). To deal with the conditions of high groundwater pressure and the notorious Taipei silty sand layer, water cut­ off grouting (e.g., base grouting) was commonly chosen to reduce the risk of piping and/or ground­ water uplifting in deep excavations. After years of practice, base grouting had become a mature grout­ ing practice in Taipei basin. Similar water cut-off grouting works had also been practiced in other South East Asian countries (Cheng et al. 2015). This paper will report a base grouting job for the deep excavation of a MRT station in New Taipei city. The excavation was 29 - 33 m in depth and was retained by 53 m deep diaphragm wall (D-wall). The ground­ water level of silty soil was at around GL -6 m; while the water head of the gravel layer was at GL -10 m. The bottom of the D-wall was expected to be

embedded in the highly permeable gravel layer; and base grouting was to be carried out above and below the bottom of D-wall. It was to cut-off the ground­ water seepage from the underlying gravel layer and to reduce the risk of groundwater uplift during exca­ vation. The tube-a-manchette (TAM) grouting method was adopted in this project. Based on the grouting pressure records gathered from the grouting process, 2D grouting pressure response contours will be used to show the results of TAM grouting. After completion of grouting, on-site permeability tests were run to check the water tightness of base grout­ ing zone. During the excavation, the groundwater level inside the D-wall enclosed area was lowered in synchronization with the staged excavation to keep the working area dry. Meanwhile, the water head in the underlying gravel layer was also lowered by pumping. When the excavation reached the bottom (GL -33 m), only 6 (out from 22) deep pumping wells were in operation to keep the factor of safety against base uplift at the final depth (GL -33 m) of excavation be larger than 1.2. This paper will report the grouting and water pumping works carried out in this underground station construction.

DOI: 10.1201/9780429321559-102

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2 SITE CONDITIONS The soil profile of the excavation site is shown in Figure 1. Generally, the excavation (to a final depth of 33m) was carried out in the interlayered formation of silty clay and silty sand layers. Below the bottom of excavation, it was mainly silty sand layer underlain by gravel layer (Figure 1). The depth of interface which separates the silty sand and the gravel layers varies from GL-46 m to GL -54 m. The base grouting zone was located in silty sand (SM 1) layer with the fines content (FC, passing #200 sieve) varies from 10 ~ 25% (Figure 2). The coefficient of permeability (k) of the silty sand layer before grouting was about equal silty sand (fines content = 10 ~ 40%) obtained from other sites in Taipei basin. Based on the pumping test carried out on two 90 m-wells (PW-01 and PW-02, Figure 3) and using other wells as observation wells, it had been found that the influence radii of the pumping wells were equal to 160 m and 231 m respectively. The transmis­ sivity (T) of the gravel layer ranged from 0.181 to 0.315 m2/min with an average of 0.251 m2/min; the

storage coefficient (S) ranged from 7.13 × 10-5 to 8.49 × 10-2 with an average of 5.35×10-3. The coeffi­ cient of permeability or hydraulic conductivity (k) of the conglomerate gravel layer was estimated using this equation: T = k × thickness of aquifer. It was about equal to 1.2×10-2 cm/sec, assumed the thickness of aquifer equal to the embedded length of well in the gravel layer (~ 35 m). The dimensions of the excavation site were 160 m in length, 26 - 34 m in width and 33 m/29 m in depth as shown in Figure 3. The bottom-up construc­ tion method was adopted for this underground station project. Since the length of site was 160 m long, a large lateral deformation of the retaining diaphragm wall in to 1×10-3 ~ 6 × 10-4 cm/sec as measured from the in-situ constant head permeability test. It was in line with the k values (= 1×10-3 ~ 5×10-5 cm/sec) of the movement, the ground settlement around the excavation site could be controlled. Meanwhile, these cross D-walls could also sub-divide the excavation site into 6 smaller zones to reduce the risk of total base failure due to piping or uplifting if the base grouting were not unsuccessful. This is similar in principle to the compartment design of a ship. Generally, the thickness of base grouting zone was 5 m (from GL-49 to -54 m). But the thickness was increased to 8 m (from GL-46 to -54 m) along the per­ ipheral area adjacent to the D-walls of each zone (both cross wall and retaining wall, Figure 3). The past experience had shown that the interface between D-wall and soil could be a potential passage for groundwater to seep through the base grouted zone. So, it was decided to increase the thickness of grouted zone adjacent to the wall-soil interface to reduce the risk of groundwater seep-in. 3 BASE GROUTING FOR THE EXCAVATION SITE

Figure 1. Soil profile of the site.

Figure 2. Gradation curves of the silty sand (SM layer) in the base grouting zone.

TAM grouting method was adopted for the base grouting job of this project (Figure 4). The grout holes were drilled with duplex drilling system. The outer casing of the drilling system had an outside diameter of 118 mm and a cutting bit diameter of 125 mm. The inner drill rod had an outside diameter of 73 mm. When the drilling reached the design depth, the string of inner drill rod was withdrawn and CB grout was injected to the cased drill hole. Then the M-tube was inserted to the drilled hole. After M-tube was in place, the outer casing was withdrawn and additional CB grout was added to fill up the drill hole. Wait until the CB grout hardened, the M-tube was ready for grouting. On the M-tube, sleeved grout ports were placed at an axial spacing of 33 cm. Upon grouting, an inner grouting monitor with expandable air packers above and below the injection holes was inserted to the M-tube. Grout injection was carried out at constant flow rate (10±2 liter/min), the grouting pressure response at each

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Figure 3. Wells layout and pore water pressure sensors of the underground station excavation.

injection port on the Manchette tube (M-tube) was monitored. Two grouting stages were adopted in this base grouting work: the suspension type cement and ben­ tonite grout (CB grout) was used in the first-stage; the solution type sodium silicate grout (SL grout, with MK reagent) was used in the second-stage. The porosity of silty sand (fines content = 10 ~ 25%) was about equal to 45%. The goal of the base grouting was to fill 90% of the soil porosity. So the grout injection volume was set to 40% (ffi 45% × 0.9) of soil volume in this project. The grout volume injected was further divided to 10 % by soil volume for CB grouting and 30 % for SL grouting. Using 40% of soil volume as the injection volume had been

Figure 4. Schematic of TAM grouting method.

a common practice for base grouting in silty sand layer in Taipei MRT underground station excavation over the past years. When doing the suspension type CB grouting, the injection rate was kept at 8 ~ 15 liter/min/port and the injection pressure was kept < 50 bars to prevent leak­ ing from the packers of the inner grouting pipe; when doing the solution type SL grouting, the injection rate was kept at 8 ~ 12 liter/min/port and the injection pres­ sure was kept < 50 bars. The injected grout volumes per port of CB grout were set at 40 liters for grout holes spacing = 1 m; and 140 liters for grout holes spa­ cing = 2 m respectively. In comparison, the injected volumes per port of SL grout were set at 120 liters for grout holes spacing = 1 m; and 420 liters for grout holes spacing = 2 m respectively. For each grout hole, grouting was started from the bottom and then pro­ ceeded upward step-by-step (0.33 m per step). Once the injection volume of a port has reached the refusal criteria set in Table 1, grouting was terminated at that port. The applied grouting pressure to meet the refusal criteria was recorded. The TAM grouting was con­ ducted in a multiple-port injection manner. Normally, 5 - 10 grout holes were injected as a group following the either sequences: inside-out from the center or outsidein from the periphery. Grout holes spacing was reduced to 1 m for the peripheral grouting. In the central part of each grout zone, the thickness of grouting zone was kept at 5 m and the grout hole spacing was 2 m. From GL -46 to -49 m, grouting was only carried out along the peripheral side of each zone; from GL -49 to -54 m, grouting was carried out for the entire excavation site. Table 1 lists the mixing proportions of CB and

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Table 1. Mixing proportions and grouting parameters of the grouts used in this project. Cement-bentonite grout (CB grout) Cement (kg)

150 – 250

Silicate grout (SL grout)

Sodium silicate (liter) Bentonite (kg) 50 – 80 MK reagent (kg) Water (liter) The rest Water (liter) ∑ = (liter) 1000 ∑ = (liter) Gel time (min) Gel time (grout itself)(min) Refusal criter­ 40 (spa­ Refusal Criter­ ion per port cing =1m) ion per port (0.33m per 140 (spa­ (0.33m per step)(liter) cing =2 m) step)(liter) Injection rate 8 ~ 15 Injection rate (liter/min /port) (liter/min /port)

250 40 – 60 690 – 710 1000 60 120 (spa­ cing =1m) 420 (spa­ cing =2 m) 8 ~ 12

Figure 5. Pressure distributions of SL grouting for the entire excavation site (plane section).

SL grouts used in this project. SL grout was mixed with the MK reagent first on a mixer before injected to the ground. The gel time of grout itself was set at 60 minutes. But the actual gel time might be shorter when injected to the ground. 4 GROUTING PRESSURE DISTRIBUTION The advantage of TAM grouting is that each injec­ tion port has its own spatial coordinate and each port can be repeatedly injected. So it can keep a spatial record of injection pressure and grout take inside the grouting zone for each injection stage. By knowing the grout take and grouting pressure at each injection port of a grout hole, grouter can have some idea on the ground response (either the pressure or the grout take) in the grouted zone. But to get the whole pic­ ture of the entire grouting zone, the TAM records per grout hole need to be interpreted in a more sys­ tematic way, with the help of 3D image processing programs. This paper uses GRAPHER to present the 2D grouting pressure contours at given depths inside the grouting zone. Such 2D pressure distribution contours are very helpful information to evaluate the outcome of base grouting. Figures 5 & 6 show the cross section and plane section of the grouting pressure distribu­ tion contours for the SL grout injection (2nd stage grouting) in the base grouted zone. It can be seen that zone 4 showed a higher grouting pressure (up to 32 bars), while zone 1 yielded a lower grouting pressure (down to 10 bars). In average, the grouting pressure of the site was around 20 bars, which was relatively low com­ pared to that of CB grouting. Since the purpose of grouting was for waterproofing, it was the SL grout which played the key role of base grout­ ing. So only the grouting pressure response of

Figure 6. Pressure distributions of SL grouting for the entire excavation site (cross section).

SL grout will be discussed here. As shown in Figures 5 & 6, the grouting pressure of SL grout was quite evenly distributed and most of the grouting pressure was around 10 bars and up. So a good water proofing effect of base grouted zone was expected, except in certain low pressure areas/zones. Having the information of grouting pressure distribution in hand, engin­ eer can be more confident to select the proper locations to do the supplementary grouting (if needed). As a result, the overall good water tightness of the base grouted zone can be better assured. 5 ON-SITE PERMEABILITY TEST RESULTS To verify the effectiveness of this base grouting work, on-site pumping test and rising head test were carried out to measure the permeability of each grout zone (Zone 1 ~ 6) in the excavation site. The pumping test was carried out by pumping out

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where h1: water level in observation well at time t1; h2: water level in well at time t2; A: area of the test site; D: thickness of base grout­ ing zone; Δh: drop of groundwater level in the observation well; a = ΔQ/(h2–h1) and ΔQ = (Δhrising/Δhlowering)×Q. Table 2 shows the klowering and krising of different zones after the completion of base grouting. In gen­ eral, the k-values of base grouting zone were in the order of magnitude of 10-5 ~10-6 cm/sec. It met the requirement set by the owner that k-value of the base grouting zone must be < 5×10-5 cm/sec, which is about two orders of magnitude lower than the original permeability (= 1×10-3 ~ 6×10-4 cm/sec) of the in-situ sandy soil and conglomerate gravel layer before grouting. By cutting off the uplifting groundwater at the bottom of base grouting zone, it could gain more weight (i.e., the bulk weight of soil between base of excavation and bottom of base grouting zone) against uplifting groundwater. 6 GROUNDWATER PRESSURE MONITORING RESULTS

Figure 7. On-site pumping test and rising head test carried out to measure the permeability of base grouting zone.

an adequate amount of water to lower the phreatic surface of groundwater table in the test area (i.e., each zone) and kept the water head in the observa­ tion wells at a constant level (equivalent to constant head test) (Figure 7). The coefficient of permeabil­ ity (klowering) of the test site was calculated using Equation 1.

where Q: pumping rate; A: area of the test site; D: thickness of base grouting zone; Δh: drop of ground­ water level in the observation well. After pumping test, turned off the pumping well. Let the water level in the observation wells rise up and measure the water level in wells at certain elapsed time. The coefficient of permeability (krising) of the test site was calculated using Equation 2.

Table 2.

To secure the stability of this excavation site against uplifting groundwater pressure, 22 wells of 90 m deep with 120 HP pumps (i.e., No. PW-01~ 22, 90 m-well) were installed through the wall body of D-wall to lower the groundwater level in conglomerate gravel layer. 8 wells of 40 m deep with 2 HP pump (i.e., No. SPW-1 ~ 8, 40 m-well) were installed within the D-wall enclosed area to lower the groundwater level in synchronization with the staged excavation and kept the working area dry. The riser pipe of the 90 m-well was welded to the rebar cage of D-wall before it was lowered to the D-wall trench. The pipe went all the way from top to bottom (53 m deep) of D-wall. To go beyond the bottom of D-wall, the well was drilled fur­ ther down to 90 m. The drilled diameter of well was 71.1 cm (28 inch) but the diameter of the riser pipe was 50.8 cm (20 inch). These 90 m-wells were installed to control the uplifting groundwater pressure coming from the underlying gravel layer. The number of wells in operation depended on how much water head drop in the conglomerate gravel layer was required.

Coefficients of permeability measured at different zones after completion of base grouting.

Pumping test kowering(cm/sec) Rising test kising(cm/sec)

Zone 1

Zone 2

Zone 3

Zone 4

Zone 5

Zone 6

1.57×10-5

1.15×10-6

1.52×10-6

3.86×10-6

1.99×10-6

3.8×10-6

1.17×10-5

1.16×10-5

9.69×10-6

4.32×10-6

4.28×10-6

4.62×10-6

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Figure 8. Monitored groundwater pressure changes above (40 m deep) and below (55 m deep) the base grouting zone at zones 1, 3 and 5.

After the pumping of 90 m-deep wells started, the water pressure monitored by the 55 m deep electrical piezometers (installed inside the gravel layer) dropped from the initial level (~ 45 t/m2 which was equivalent to water head at GL -10 m) gradually down to about 30 t/m2 (water head at GL -25 m) when the excava­ tion reached to the bottom (GL -33m) and during the period of mat foundation construction (Figure 8). In summary, when the excavation reached to the final depth, the GWL inside the excavation area was at GL -34 ~ -38 m. In the meantime, the GWL in the gravel layer was at GL -25 m. So, the water head dif­ ference was around 9 ~ 13 m (= -25 m - (-38 ~ -34 m)) above and below the base grouting zone. The total overburden pressure of soil above the base grout- ing zone was equal to (55 m – 38 m) × 1.8 t/m3 + (38 m – 33 m) × 1.6 t/m3 = 38.6 t/m2, which was enough to keep a factor of safety (= 1.29 = 38.6/30) against uplift­ ing groundwater pressure of 30 t/m2 (= 55 m – 25 m) > 1.2 (the required factor of safety). At this moment, only 6 pumping wells (PW-03, 05, 09, 18, 19 and 22) were in operation at a pumping rate of 1.0 ~ 1.2 m3/min each and the total volume of groundwater discharging was around 6 ~ 7 m3/min. The layout of operating wells is shown in Figure 9. Although the grouting pres­ sure response in zone 1 was relatively low (see Figures

Figure 9. Number of pumping wells in operation when excavation reached the final depth.

5 & 6), no extra effort was needed to lower the uplift­ ing groundwater pressure at zone 1 by pumping more water out. In comparison, the 40 m-deep wells were mainly used to keep the groundwater level within the D-wall enclosed area below the excavation base at each excavation stage. The riser pipe of the 40 m well had only 5 cm in diameter and the bored hole had 12.7 cm (5 inch) in diameter. It took 10 excavation stages to get down to the final stage of excavation (33 m deep). Since the bottom of D-wall enclosed area was base grouted, not much ground­ water could be pumped out from the 40 m-well. After the pumping of 40 m-deep wells started, the water pressure inside the excavation site dropped from the initial level (~ 30 t/m2, i.e., the water head was 10 m below ground level) gradually down to about 2 - 6 t/m2 (i.e., the water head was 34 ­ 38 m below the ground level) when the excavation reached to the bottom (33 m deep); and remained at about 6 - 10 t/m2 (i.e., the water head was 30 ­ 34 m below the ground level) during the construction of mat foundation slab (Figure 8). As mentioned earlier, based on the site pumping test carried out on two 90 m-wells (PW-01 and PW­ 02) before the excavation started, the average transmis­ sivity (T) of the gravel layer was 0.251 m2/min and the storage coefficient (S) had an average of 5.35×10-3. T value of 0.251 is more than one order of magnitude smaller than the T values of the conglomerate gravel layer obtained from other areas in Taipei basin. Geo­ logically, this site is located in the downstream of the river deposits in the Taipei basin. The particles of the deposit are finer. The fines content (FC) of the in-situ silty sand layer in the base grouting zone (GL -40 m ~ -54 m) varies from 10 ~ 25% (Figure 2). The D10 ranges between 0.01 ~ 0.1 mm. Use the Hazen’s equa­ tion (1930),

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The coefficient of permeability (k) of this silty sand layer is estimated between 1x10-2 and 1x10-4 cm/sec, which is in the range of fine sand and silty clay (Das et al. 2014). It explained why the influence radii of both pumping wells (PW-01 and PW-02) were limited to a small distance of 160 m and 231 m, respectively; and the water head difference between pumping well (say, PW-20) and nearby (within 30 m range) observation wells or standby wells could be as much as 20 m (due to small T). Therefore, only six 90 m-deep wells were needed to lower the groundwater level of the gravel layer to keep a good base stability against uplifting water pressure (FS > 1.2) for a 160 m x 26 m excavation site with a depth of 33 m.

(2) The coefficients of permeability (k) measured after the completion of base grouting using field pumping test were in the order of magnitude of 10-5 ~ 10-6 cm/sec. It met the requirement (i.e., < 5×10-5 cm/sec) set by the owner and was about two orders of magnitude lower than that of insitu silty sand layer (=1×10-3 ~ 6 × 10-4 cm/sec) before grouting. (3) When the excavation reached the bottom (GL -33 m), only 6 deep pumping wells (90 m deep) were in operation to maintain the groundwater level of gravel at around GL -25 m and kept the factor of safety against base uplift > 1.2. It indi­ cated that the base grouting zone had success­ fully controlled the uplifting water coming from the underlying gravel layer. In addition, the number of deep wells in operation was less than expected could also be resulted by the low trans­ missivity (T) and storage coefficient (S) of the conglomerate gravel layer.

7 CONCLUSIONS The following conclusions can be drawn from the findings of the base grouting work carried out for this MRT underground station: (1) Each injection port of TAM grouting has its own spatial coordinates, so the grouting information at a given depth inside the grout zone can be pre­ sented in 2D image. The 2D mapping of grouting pressure distribution is a very helpful informa­ tion to evaluate the TAM grouting results; and if needed it can be used as the supporting informa­ tion for the supplementary grouting to further enhance the water tightness of the base grouting zone.

REFERENCES Cheng, S. H., Liao, H. J., Hatakeyama, K., Wong, Ricky K. N., & Iwakubo, T. 2015. TAM Grouting to Reduce Artesian Water Pressure Acting on the Base of an Excavation, Proc. of International Conference on Soft Ground Engineering (ICSGE 2015), Singapore. Das, Braja M. and Sobhan, Khaled 2014. Principles of Geotechnical Engineering, 8th Edition, SI, Cengage Learning. Hazen, A. 1930. Water Supply, American Civil Engineers Handbook, Wiley, New York. Moh, Z. C., Ju, D. H. and Hwang, R. N. 1997. A small hole can be-come really big, Momentous Lecture, Proc. of 14th International Conference on Soil Mechanics and Foundation Engineering, Hamburg, Germany.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

The largest tunnels in freshly consolidated soft soil: Tuen Mun Chek Lap Kok Link Subsea tunnels, Hong-Kong T. Lockhart Bouygues Travaux Publics

ABSTRACT: The Northern Landfall, part of the Tuen Mun Chek Lap Kok Link tunnel project, is a large reclamation on very soft Marine Deposits and Alluvium, inside which two large-diameter tunnels have been bored. The realization of this challenging, world-record-holding project implies the use of consolidation under surcharge, to give the compressible soils enough strength to withstand the excavation of the tunnels and ensure its durability. This paper describes the design and monitoring of this consolidation scheme, and out­ lines its close relationship with the design of the tunnel linings.

1 INTRODUCTION 1.1

The project

The Tuen Mun-Chek Lap Kok Link (TMCLKL), in Hong-Kong, is a new highway connection pro­ posed to meet the anticipated traffic demand between Northwest New Territories and Lantau Island. It serves to connect the Hong Kong Boundary Crossing Facilities (HKBCF) situated to the northeast of Hong Kong International Air­ port (HKIA), the future Tuen Mun Western Bypass (TMWB) and the existing Lung Mun Road in Tuen Mun (Figure 1). The Contract HY/2012/08 comprises the design and construction of the Northern Landfall, a 16.5-hectare reclamation on the Tuen Mun side, two tunnels cross­ ing the maritime navigation channel, 56 cross-passages for emergency egress and ancillary structures such as Ventilation buildings and control rooms. 1.2

The Northern landfall

The Northern Landfall is reclaimed on the sea, with the double purpose of serving as a storage yard for containers in the future, and housing the tunnel portal and access ramp to the subsea portion of the project (Figure 2). The geological configuration of this area (Figure 3) exhibits features commonly encountered in the Hong-Kong region. At the seabed level, a 2m to 7m-thick layer of very soft unconsolidated Marine deposits, overlaying 5m to 10m of Alluvium deposits of various nature (either clayey or sandy), with sand and gravel lenses. Under the Alluvium, weathered Granite of various degrees of alteration

is found, its thickness varies from 5m to over 20m. The bedrock is made of Granite, crossed by several faults and shearing zones. The associated geotech­ nical parameters are summarized in Table 2 below. The extensive dredging of the very weak Marine deposits is prohibited due to environmental reasons, which leaves no other option than achieving an improvement of their strength parameters to enable the construction of the reclamation and the subse­ quent tunneling activities. 1.3

The Northern ramp TBM tunnels

The Northern landfall houses the access ramp to the subsea tunnels, which is a combination of a cut-and-cover tunnel and two bored tunnels. The descending lane tunnel hosts a regular 2-lane carriageway, the other tunnel hosts a 3-lane car­ riageway, with an additional lane dedicated to slow vehicles when the road gradient becomes steeper than 5%. Table 1 summarizes the main parameters of both TBM tunnels: The Northbound tunnel, with an internal diameter of 15.60m and an excavation diameter of 17.63m, holds to-date the world record of size for a TBM tunnel. Figure 4 shows a typical cross-section with the Northern reclamation, on which the following elem­ ents can be noticed: – The spacing between the tunnels is around one tunnel radius (8m) – The TBMs will start in the backfill, then cross the compressible layers as they go down, which will have an impact on the choice of materials used for the consolidation.

DOI: 10.1201/9780429321559-103

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Figure 4. Cross-section in the Northern Reclamation.

– The seawalls are close to the tunnels, and their stability needs to be assessed under the surcharge placed over the middle. The design and construction of these tunnels in the recently consolidated ground must therefore take these elements in consideration. Figure 1. Location map of the TMCLKL project.

2 OVERALL DESIGN METHODOLOGY 2.1

Figure 2. From May-2014 [left] to Dec-2015 [right].

Available data

The design of the Northern ramp tunnels was initi­ ated on data available from a previous ground inves­ tigation campaign, prior to the start of the backfilling of the reclamation. Table 2 provides typical geotech­ nical values: Both the Marine deposits and the clayey Alluvium layers are fine-grained soils, which will consolidate under the additional vertical load brought by the backfilling of the reclamation. The parameters con­ sidered in all consolidation studies are provided in Table 3. However, this in-situ state will be perturbed by the backfilling works of the Northern reclamation, and then by the tunneling activity. In other words, the design of the reclamation must take into account the pre-existing state, its own requirements, but also prepare the ground for the subsequent boring of the tunnels. Table 2.

Unit weight Soil layer kN/m3

Young modulus Friction MPa angle °

Undrained cohesion kPa

Public fill1 Sand fill1 Marine deposits2 Alluvium clay2 Alluvium sand2 CDG3 Granite

19

20

30

0

19

40

35

0

14.5

2

0

5

19

8

0

35

19

12

34

0

20 23

70 >500

36 0 UCS 30 to 150 MPa

Figure 3. Geological profile of the Northern reclamation.

Table 1.

Northern landfall TBM tunnels.

Tunnel

ML03

ML02

Direction Traffic lanes Internal diameter Lining thickness Segments per ring

Northbound 3 15.60 m 700 mm 12

Southbound 2 14 m 550 mm 9

Ground cover*

From 3 m (Launch shaft) to 26 m (Reclamation tip)

* Distance between the tunnel crown and the surface

Typical geotechnical parameters.

1

Material placed during the backfilling works In-situ parameters, prior to reclamation works 3 Completely Decomposed Granite 2

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Table 3.

line with the predictions – have been finally inte­ grated in the calculations.

Consolidation parameters1.

Soil layer

k m/s

Cc -

Cs -

Cα -

Marine deposits Alluvium clay

10-8 10-7

0.8 0.11

0.07 0.03

1.6 0.5

3 ANALYSIS OF THE CONSOLIDATION 3.1

1

k: permeability, Cc: compression index, Cs: Recompression index, Cα: creep index

Furthermore, the design of the tunnels must be based, not on these initial conditions, but on the state of the ground at the end of the reclamation works, which leads to an interlinked process described in the next section. 2.2

Design development process

A preliminary calculation of the tunneling activ­ ity, both in temporary stage (operation of the TBM) and in permanent stage (final concrete lining) showed without ambiguity that the strength parameters of the existing in-situ soil layers, summarized in Table 1, were significantly too low to allow for a safe excavation and a long-term performance of the final lining. Therefore, it is necessary to incorporate in the design of the reclamation the requirements of the tunneling activity, as shown on Figure 5. In the next section, we will describe how the per­ formance criteria given by the design of the tunnel are incorporated in the design and construction of the Northern landfall, including the consolidation period. Then we will highlight several aspects of the design of the TBM tunnels, including how the actual results of the consolidation phase – not always in

Assessment of the initial conditions

In addition to the geotechnical data available from earlier investigation campaigns, a thorough survey has been conducted along the tunnel alignment to map the geological conditions under the reclamation. To get the initial state of the strength parameters of the soft, compressible layers, a total of 157 Cone Penetration Tests (CPTs) have been performed along three lines: between the tunnels and between each tunnel and the edge of the reclamation (Figure 6). The choice of the CPT as a ground investigation tech­ nique has been governed by the following considerations: – It is a reliable tool in soft soils, and the correlation between the results and the actual mechanical prop­ erties of the ground are well backed by experience; – It can be efficiently implemented using simple and robust maritime equipment, from a barge, without delay. Figure 7 gives an example of CPT result, on which the distinction between the soil layers becomes clearly visible. To derive the mechanical properties of the soil layers, the following correlation have been used:

Figure 6. Pre-consolidation CPT campaign.

Figure 5. Design development process.

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Table 4.

Strength of the compressible layers. Pre-consolidation

Soil layer Marine deposits Alluvium clay 1

Post-consolidation1

cu kPa

qc MPa

cu kPa

qc MPa

5 – 10

0.1 – 0.2 65 –85

1.3 – 1.7

35 – 45

0.7 – 0.9 65 –85

1.3 – 1.7

Post-consolidation also means ‘prior to tunneling’

a significant amount of data exists about the properties of the Marine Deposits and the Alluvium, which allows us to use correlation formulae backed by experi­ ence. After full consolidation, the increase in undrained cohesion Δcu is linked to the increase in vertical effect­ ive stress Δσ’v by

Figure 7. A typical pre-consolidation CPT result.

and

where cu and Eu are the undrained cohesion and Young modulus, respectively, and qc is the CPTmeasured cone resistance, all numbers being expressed in the same unit (kPa or MPa). The following table summarizes the ranges of values measured on site, as well as the target values required by the tunnel design to ensure the safe oper­ ation of the TBMs and the long-term stability of the segmental concrete lining. 3.2

Consolidation design

In order to bring strength parameters of the com­ pressible soils – namely, the Marine Deposits and the clayey Alluvium – to the minima given by the design of the tunnels, a classical scheme of consoli­ dation under surcharge has been chosen. Due to the surcharge imposed on the compressible soils, in which drains of high permeability have been placed, the water is squeezed out of the soil matrix and the grains are packed more tightly against each other, leading to an increase in the shear strength of the material. It is therefore necessary to establish a link between the surcharge (i.e. the increase in effective vertical stress, in geotechnical terms) and the resulting increase in strength of the soils. 3.2.1 Determination of the surcharge height Since the development of Hong-Kong has been pos­ sible thanks to the extensive use of reclaimed land,

with the parameter a ranging typically from 0.20 to 0.24, for both the Marine clays and the Alluvium. Therefore, to achieve the increase in strength defined by the tunnel design and given in Table 4, the height of the surcharge can be calculated by the following equation:

where Hs is the height of the required surcharge measured from the top of the compressible layer (i.e. the seabed in our case), F% is the efficiency of the consolidation (see below) and γfill is the unit weight of the fill. The efficiency of consolidation F%, ranging from 0 to 100%, is used to integrate the fact that the final equilibrium will not be reached in practice, since it would require an infinite amount of time. Formula (3) is based on test results at full consolidation, therefore it is necessary to add this F% coefficient to account for the fact that only a fraction of the increase in strength will be gained due to the limited (finite) time of the works. The design of the Northern landfall is based on a consolidation efficiency of 90%, which gives the Table 5.

Minimum Maximum

Surcharge heights. cu kPa

σ’v MPa

Hs m

Level1 mPD

65 85

295 385

22 26.5

+10

+14.5

1 during the works. The final platform level is +6.0m.

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platform elevations summarized in Table 5, and then allows for a choice of the drain pattern. When using the unit weight of the fill, attention must be paid to the fact that part of the reclamation is backfilled under the sea level, and the submerged unit weight (usually noted γ’) shall then be used; for the backfill placed above the sea level, the total unit weight γ applies. Another subtle important point in such case of significant thickness of very compress­ ible soils is that the backfilling will induce a vertical settlement of the whole platform, of the order of mag­ nitude of the meter or more; therefore, part of the ini­ tially dry backfill will become submerged, thus bringing less contribution to the increase in effective vertical stress that is required to improve the cohe­ sion. This has been solved in practice by an iterative calculation to estimate the settlement induced by the surcharge, then adding this value to the surcharge height target (Figure 8). The levels calculated to bring the necessary increase in vertical effective stress, +10mPD and +14.5mPD, respectively became +12mPD and +15.5mPD when including the provision for future settlements. 3.2.2 Determination of the drain pattern The schedule of works allowed for a consolidation period of 4 to 6 months, this period being defined as the duration of the settlement under the full surcharge of the ground as calculated previously. The pattern and spacing of the band drains has therefore been defined using the radial consolidation theory (Barron, 1948), to reach the final efficiency of 90% discussed above. Barron’s chart, used for upper and lower bounds of the duration of consolidation, give a triangular drain pattern of 1.5m and 1.2m respectively. 3.3

– They have to be installed quickly, to fit within the tight construction schedule; – They need to be ultimately ‘chewed’ by the TBMs (see Figure 4), without causing any damage to the muck-out system. Plastic band drains, installed by a derrick lighter using a hydraulic hammer (Figure 9), have been chosen. The band drains are placed inside a steel tube, which is then hammered into the ground down to refusal. When comparing the refusal levels to the corresponding CPTs, it has been noticed that the toe level of the band drains reached roughly the qc value of 2.5 to 3MPa, being enough to ensure that the com­ pressible layers would be drained (refer to Table 4). Once the installation of the vertical band drains is complete, the reclamation (Figure 10) is backfilled up to the required elevation. Monitoring of the consolidation-induced settlements then starts. 3.4

Consolidation monitoring

The monitoring of the consolidation is of paramount importance: what is at stake is to guarantee that the target degree of consolidation of 90% has been attained, which in turn is a sine qua non condition to the success of the tunneling works. A commonly used graphical method has been developed by Asaoka, based on the identification of the differential equation of consolidation and its

Overview of the works

The main constraints for the realization of the drains are: – They have to be installed from the sea, prior to the backfilling of the landfall;

Figure 9. Installation of plastic band drains from a barge.

Figure 8. Theoretical platform (top); levels given to earth­ works survey (bottom) to account for settlements.

Figure 10. Reclamation and surcharge in progress.

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finite difference (first order) approximation. A complete theoretical demonstration can be found in the original article (Asaoka, 1978). To follow the consolidation, it is proposed to monitor the settlement at a constant time interval Δt, at the same location, and plot the settlement meas­ ured at a time (T + Δt) with respect to the settlement measured at time T. The logarithmic nature of the evolution in time of the settlements implies that the obtained plotted points will form a straight line (for constant external parameters: loads etc…), which intersects at some point the y = x line. At this point, the settlements at (T + Δt) are equal to the ones at T, which means mathematically that the recurrence has converged, and practically that the consolidation has finished. An array of deep and surface settlement plates has been installed (Figure 11) and monitored daily (Δt = 1 day) during the consolidation period. The evolution in time of the surcharge (linked to the progress of the earthworks on site), as well as the pres­ ence or absence of secondary consolidation (‘creep­ ing’) is also visible on the Asaoka diagram (Figure 12). However, the reality of the consolidation in-situ is much more complicated that the idealized cases of the vertical consolidation (Terzaghi), radial (Barron) or even combined (Carillo). In addition, in the case of the Northern landfall, two different materials need to be consolidated: the Marine Deposits and the clayey Alluvium, each with its own set of defining parameters (permeability, consolidation characteris­ tics, and layer thickness); in addition, local features such as the presence of draining gravel lenses, can remain unnoticed during the ground investigation campaign, yet have a significant impact on the con­ solidation process.

Therefore, the Authors have adapted the Asaoka method using the equivalent consolidation coefficient idea (Magnan & al., 1980). In practice, the real consolidation coefficients (commonly written cv or cr) of the pure soil, as determined in a lab by an oedometer test, are not necessarily representative of the full scale situation; it is pro­ posed to use an equivalent time factor, Tv,eq, in which are contained all the uncertainties, and dir­ ectly linked to the slope β of the Asaoka curve by the following equation:

with Tmax the duration of the consolidation at the date of interest. It shall be noted that this expression is independ­ ent from the thickness of each layer, thus reducing the uncertainty linked to the geological profile; it then becomes possible to calculate the degree of con­ solidation U% by:

The consolidation process can be considered advanced enough when U% reaches the target value of 90%; however, this does not guarantee that the material has achieved the required shear strength as expected following Formula (3), and a post-consolidation verification is necessary. 3.5

Figure 11. Array of surface and deep settlement markers.

Assessment of the final state

To check the efficiency of the consolidation on the strength parameters of the ground, a postconsolidation CPT campaign has been conducted, with the values summarized in Table 4 as targets. Figure 13 shows a typical example of the results the post-consolidation CPT campaign. The ovalshape region contains the Marine Deposits, which

Figure 13. Comparison CPTs before/after consolidation.

Figure 12. Asaoka curve for consolidation monitoring.

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have both settled by approximately 1m and reached the dashed line of the target qc = 1.7 MPa. The underlying clayey Alluvium show a similar gain in strength. After the consolidation has reached 90%, as moni­ tored by the modified Asaoka method, and the postconsolidation CPTs have validated the increase in strength necessary for the excavation of the tunnels, the consolidation is judged successful and the sur­ charge is removed. Where one of these conditions was not met, a local ground improvement by jet grouting has been performed. The reason for this lack of performance has been investigated, and was often correlated to difficulties during the installation of the band drains. After the end of the consolidation and surcharge period, the twin tunnels under the newly constructed Northern landfall can be excavated. 4 DESIGN OF THE SEGMENTAL LINING The purpose of this section is not to detail the entire design process of the TBM tunnels, but rather highlight the specificities brought by the fact that the Northern reclamation was still to be built in parallel with the development of this design. This is a radical difference to the normal case where geological conditions preexist and can be evaluated before the underground structure is conceived. 4.1 The explicit Convergence-Confinement approach The design of the TBM tunnels relies on the Convergence – Confinement method (Panet, 2001), in its explicit formulation (Aristaghes & al., 2003) using a finite-element soil-structure interaction software. The action of the slurry-shield TBM onto the ground is modeled with a pressure, corresponding to a weighted average between the pressure of the slurry and the pressure of the tail void mortar injected around the concrete rings. This is of great importance in geological configurations such as the Northern Landfall, where the heterogeneity of the ground layers make the use of other methods (for instance, the ones based on the ‘volume loss’ approach) far less physically obvious to use. All construction stages are modeled in sequence:

Figure 14. Virtual tunnels in non-consolidated ground.

Figure 14 shows the outcome of the first pre­ liminary round of calculations, performed on the pre-consolidation geotechnical parameters. The red dots represent the plastic points, and it is obvious that the ground surrounding the tunnels is not strong enough to withstand the additional vertical stress brought by the excava­ tion of the tunnels, let alone the influence of their proximity. This calculation has been iterated by increas­ ing the shear strength of the soil materials until the plastic deformation with the ground had been brought back to an acceptable level (typic­ ally, 5% of the total strain). The results are the target parameters in Table 4. 4.2 Integration of the ground treatments After the consolidation works had been carried out, it turned out that some areas did not achieve the expected strength, and as mentioned above, add­ itional treatment by jet grouting was necessary. The explicit Convergence – Confinement method make the incorporation of such treat­ ments in the finite-element model straightfor­ ward: appropriate clusters of soil with the corresponding properties of the jet grouting have been added (Figure 15). The design of the TBM tunnels has been closely linked to the reclamation works, and has incorpor­ ated the in-situ conditions resulting from the consoli­ dation process. Figure 16 shows the completely backfilled reclamation with the two tunnels visible inside the Launch Shaft.

– Initial stage, – Construction of the reclamation, – Excavation of the first tunnel, supported temporar­ ily by the slurry pressure, – Erection of the concrete lining of the first tunnel, – Excavation of the second tunnel, – Erection of the second concrete lining, – Long term evolution of the ground parameters (residual settlements).

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Figure 15. FE Model with ground treatments.

ACKNOWLEDGEMENTS The Authors would like to express their gratitude to François BERBET, Senior Geologist at Bouygues Travaux Publics, for his help and support during the development of the Consolidation studies and throughout the follow-up of the works on site.

REFERENCES

Figure 16. The Northern reclamation and its twin tunnels.

5 CONCLUSION The design of the consolidation scheme of the North­ ern Landfall, performed in close relationship with the design of the tunnels, has proven to be a challenging part of this project. The careful monitoring of the execution of the works, to ensure that this consolidation had met its purpose, has enabled the successful construction of the two twin tunnels.

Aristaghes, P., Autuori, P., 2003. Confinement Efficiency Concept in Soft Ground Bored Tunnels. Amsterdam Asaoka, A., 1978. Observational procedure of settlement prediction. Soils and Foundations, Vol. 18, No. 4, pp. 87–101 Barron, R.A., 1948. Consolidation of fine-grained soils by drain wells. Transactions of ASCE, Vol. 113, pp. 718–724 Biot, M.A., 1941. General theory of three-dimensional consolidation. Journal of Applied Physics, Vol. 12, No. 2, pp. 155–164 Bjerrum, L., 1967. Engineering geology of Norwegian normally-consolidated marine clays as related to settle­ ments of buildings. Milestones in Soil Mechanics, pp. 173–209 Magnan, J.-P., Deroy, J.-M., 1980. Analyse graphique des tassements observés sous les ouvrages. Bull. liaison Labo. P. et Ch., Vol. 109, pp. 45–52 Panet, M. 2001. AFTES Recommendations on the Conver­ gence-Confinement method, AFTES document GT7R6A .

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Soil conditioning for EPB tunnelling in coarse grained soils based on laboratory model tests A.S. Merritt Pells Sullivan Meynink, Sydney, Australia

S.A. Jefferis Environmental Geotechnics Ltd, Banbury, UK

R.B. Storry Bouygues Construction, Melbourne, Australia

ABSTRACT: Earth Pressure Balance (EPB) tunnel boring machines (TBMs) are commonly used for tunnel construction in soft ground. Their application range has been extended through the use of soil conditioning to improve tunnel face stability and control of spoil flow through the TBM and screw conveyor. Effective condi­ tioning is challenging for coarse grained soils, particularly with significant groundwater pressures. This paper presents results from laboratory tests undertaken with a model EPB screw conveyor on coarse grained soil samples with a range of gradings, conditioned with foam, polymers and fines addition. The tests modelled the EPB excavation process to assess the performance of different conditioning treatments. The results provide experimental support for published guidance on EPB TBM application ranges and conditioning requirements, and demonstrate the potential to expand the application range through the use of specialised water absorbent polymer conditioning treatments.

1 INTRODUCTION Earth pressure balance (EPB) tunnel boring machines (TBMs) are used to construct tunnels in a wide range of soils and weak rocks. These TBMs operate in a pressurised mode with the support pressure trans­ mitted to the tunnel face through the excavated spoil contained within the TBM head chamber. The face pressure is dissipated along the screw conveyor as the spoil is extracted from the head chamber during the TBM advance. Control of the spoil flow through the TBM to maintain the face support pressure and pre­ vent groundwater inflows are key factors which influ­ ence the face stability and ground movements during tunnelling. Soil conditioning is used to transform the exca­ vated soil into a soft plastic paste, to improve control of the excavation process and spoil flow through the TBM by forming a plug along the screw conveyor. Effective conditioning of different soil types requires different treatments, which have extended the appli­ cation range of EPB TBMs. However, in coarse grained sands and gravels it can be challenging to effectively condition the soil to form a paste. This paper presents results from a series of physical model tests undertaken with an advanced

laboratory scale model EPB screw conveyor. Tests were performed with a range of coarse grained sand and gravel samples conditioned with different treatments using foam, polymers and a thick mortar for fines addition. The test results provide experimental support for published guidance on EPB TBM application ranges and conditioning treatments for coarse grained soils with ground­ water pressures. The results also demonstrate the potential to expand the range of EPB TBM oper­ ations in these ground conditions through the use of specialised water absorbent polymers for soil conditioning. 2 EPB TBM APPLICATION RANGES Ideal ground conditions for EPB TBMs consist of well graded soils with at least 30% fines content (silt and clay fraction below 0.06 mm) and permeability below 10-6 m/s, such as clayey or silty sands con­ taining a mixture of clay, silt and fine to medium sands (Maidl et al, 1996; Milligan, 2000; BTS, 2005). These soils can readily form a homogeneous plastic paste of soft consistency with a low perme­ ability when excavated.

DOI: 10.1201/9780429321559-104

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Most natural soils do not have these ideal character­ istics and soil conditioning is used to improve the spoil properties for EPB operations to allow application in a wider range of ground conditions. Additives such as water, foam, polymers and fines (high density slurry or thick mortar) are injected from the TBM to modify the properties and consistency of the excavated soil. The conditioning treatments required for different soil types and groundwater conditions vary widely, influenced by geotechnical factors including the particle size distribu­ tion, permeability, water content and groundwater pres­ sure. In granular soils foam is used to fluidify the spoil and improve workability, and polymers are used to bind and viscosify the soil and groundwater. In coarse sands and gravels, fines can be added to improve the soil grading to aid paste formation. The objective of conditioning for coarse sands and gravels is to form a soft, homogenous plastic paste from the fines and sand to surround and support the coarser size fractions. The paste needs to be sufficiently workable to flow through the TBM and form a low permeability plug along the screw conveyor to control groundwater inflows and pressure dissipation. If the paste is too stiff it may exacerbate clogging of the machine, whereas if it is too fluid the screw conveyor cannot confine the face support pressure and uncon­ trolled groundwater inflows and face loss can occur. Various guidance based on field experience and laboratory testing has been published on the applica­ tion of EPB TBMs and soil conditioning requirements in different ground conditions. DAUB (2010) states that the ideal ground conditions described above repre­ sent the ‘main field’ of EPB TBM applications, and that operations are possible in soils with fines contents as low as 5%, although with additional technical meas­ ures or ground modification required and reduced TBM performance. Thewes (2007) presents four zones with particle size distribution envelopes for EPB TBM applica­ tions with typical soil conditioning requirements (Figure 1). Coarse grained soils in Zones 2, 3 and 4 require progressively more extensive conditioning,

with reducing groundwater pressures that can be suc­ cessfully managed. Foam, polymer and fines add­ ition are recommended for soils in Zone 4 with more than 55% gravel and less than 10% fines, with no groundwater pressure. Ground conditions beyond the Zone 4 envelope are outside the typical application range for EPB TBMs, and are more suited for slurry TBMs. Similar guidance on EPB TBM applications and soil conditioning requirements based on particle size distributions have been published by others, e.g. Maidl et al (1996), EFNARC (2005). 3 SOIL CONDITIONING LABORATORY TESTS FOR COARSE GRAINED SOILS 3.1

Port of Miami Tunnel project

The 12.9 m diameter twin bore Port of Miami Tunnel (POMT) in Florida, USA, was con­ structed through layered carbonate sedimentary formations comprising a variable sequence of limestone, cemented sand/shell and sand layers, with a highly porous variably cemented coralline limestone layer also present. The soil and weak rock layers had permeability of 10 -2 to 10-5 m/s, with water pressures up to 3.5 bars encountered along the tunnel beneath the harbour, with saline water at the tunnel horizon. The ground condi­ tions for the project were towards the limits of experience for EPB TBM applications. During planning of the tunnel construction, a laboratory test program was undertaken to assess the feasibility of EPB tunnelling and evaluate the most effective soil conditioning treatments for the expected ground conditions. The testing included a series of advanced laboratory model EPB tests using the ‘OBYONE’ apparatus at Bouygues’ Pôle Ingénierie Matériaux laboratory. Further details of the POMT project ground con­ ditions, soil conditioning tests and application during tunnelling are given in Merritt et al (2015). 3.2

Laboratory model EPB tests

The OBYONE model EPB system (Figure 2) com­ prises a pressurised sample tank with mixing rotors, representing the head chamber of an EPB machine, connected to a 1.2 m long, 100 mm diameter screw conveyor. Test samples of approximately 60 L volume are prepared by mixing soil with foam and polymer conditioning agents injected into the pressurised tank. The screw conveyor extracts the sample from the tank under pressure to model the EPB excavation process. The system is instru­ mented to monitor the total pressures in the mixing tank and at the start (inlet), middle and end (outlet) of the screw conveyor. The laboratory model tests allowed assessment of the performance of different soil conditioning treat­ ments under conditions representative of the EPB

Figure 1. EPB TBM application ranges (after Thewes, 2007).

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Table 1.

Figure 2. OBYONE model EPB screw conveyor system.

excavation process, based on confinement of the tank pressure and development of a stable pressure gradient along the screw conveyor during the extrac­ tion tests. Characterisation tests including water con­ tent, grading and slump tests were performed on the extracted material to assess the sample properties and homogeneity of the conditioned soil samples. The tests were performed on samples prepared from mixtures of the soil and weak rock materials obtained from excavations at the POMT project site. The samples were prepared with varying gradings with a range of gravel, sand and fines sized particles. Due to the reduced size of the laboratory model screw conveyor, the maximum particle size in the samples was limited to 10 mm. The samples were prepared with a range of water contents using seawater from the project site to represent the in situ ground conditions. Water con­ tents of 8% to 25% were used for the samples, based on the estimated in situ porosity of the materials at the project site. The samples were conditioned with different treatments using varying combinations and quan­ tities of foam, two polymers (in liquid form) and thick mortar for the addition of fines. One standard polymer (denoted P.1) was a soil structuring agent, designed for use in saturated, coarse grained soils with low fines content, to bind the soil into a plastic paste with a water soaking and swelling effect. The second more specialised polymer (denoted P.2) contained a super water-absorbent/viscosifying poly­ mer. This polymer acts by absorbing substantial vol­ umes of water, so reducing the free water content and increasing the viscosity/cohesion of the conditioned soil mixture. The P.2 polymer was developed specific­ ally to enhance performance in saline waters, and in addition to the super-water absorbent polymer, it appeared to also include a PHPA (partially hydrolysed polyacrylamide) type polymer. Solutions of high molecular weight PHPAs tend to form strings when poured. In soils, these PHPAs can bind the soil and water to produce a soft, cohesive paste. The thick mortar was prepared as a mixture of silt and fine sand with bentonite and water, with a density of approximately 1.7 t/m3 and water con­ tent of 40%. This conditioning agent was used as

Model screw conveyor test samples.

Test No.

Sample grading (% G/S/F)[1]

Water content (%)

T-1

39/47/14

10

T-2 T-3

30/57/13 40/47/13

11 11

T-4

46/41/13

8

T-5 T-6

75/18/7 75/19/6

20 20

T-7

65/24/11

12

T-8

58/30/12

10

T-9 65/24/11 T-10 75/19/6 T-11 64/18/18

12 20 20

T-12 67/18/15

20

T-13 59/32/9 T-14 72/21/7 T-15 69/22/9

20 20 25

Conditioning treatment[2]

Grading zone[3]

30% FIR 0.1% P.1 20%FIR 20% FIR 0.3% P.1 30% FIR 0.2% P.1 1.5% P.1 10% FIR 1.5% P.1 15% FIR 1% P.1 21% T.M. 20% FIR 0.1% P.1 1%P.2 1.5% P.2 5% T.M. 1.25% P.2 10% T.M. 1.25% P.1 1.5% P.2 1.75% P.2 2% P.2 15% T.M.

2/3 2 2/3 3 >4 >4 4

3/4 4 >4 4 4 4 5% T.M. >4

Percentage gravel (G), sand (S) and fines (F) in sample. FIR – Foam Injection Ratio; P.1, P.2 = Polymer type and injection ratio; T.M. = Thick Mortar injection ratio. [3] Zone of sample grading from Thewes (2007); see Figure 1. [1] [2]

a fine filler to increase the fine sand and silt content of the samples. The test samples and conditioning treatments are summarised in Table 1. 4 WATER ABSORBENT POLYMER CHARACTERISATION Super water-absorbent polymers can be used in soil conditioning to reduce the amount of free water present in spoil and so promote paste formation. These poly­ mers are based on high molecular weight cross-linked anionic PHPAs, and function through the polymer grains swelling to absorb large quantities of water with­ out dissolving. For EPB soil conditioning, these poly­ mers are generally used by injection into the TBM head chamber or screw conveyor to manage high groundwater inflows or high water content soils, to restore plasticity of the spoil. The polymers are usually injected in very small quantities, typically at about 0.1% to 0.2% of the excavated soil volume. Higher dosage rates can be used in difficult ground conditions, but there are risks with continuous use during

790

tunnelling as over-dosing the polymer can cause exces­ sive stiffening of the paste and unstable paste proper­ ties. For water absorption, these polymers in liquid form must be injected directly without prior dilution in water, as this would negate their water absorption effect. Water absorbent polymers are sensitive to saline water which can significantly reduce their perform­ ance relative to fresh water. In saline water the poly­ mer structure coils up which reduces the swelling and water absorption capacity. The P.2 polymer properties used on the POMT project and laboratory tests were characterised in fresh water and sea water (35 to 40 grams salt per litre) by: • Water absorption tests to measure the weight of water absorbed per gram of polymer after varying swelling times, determined from the mass of material retained on a 600 micron screen; • Flow funnel tests to provide an index measure­ ment of the relative viscosity of the polymerwater mixtures at a high shear rate. Mixtures of polymer and water were prepared at varying con­ centrations with a 30 minute swelling period, and the time for 1.0 L of the liquid to flow through the funnel with a 10 mm orifice was measured.

Figure 4. Flow funnel measurements for P.2 polymer in fresh and saline water. Data labels indicate percentage of water absorbed by polymer after 30 minutes swelling time.

Results from the P.2 polymer water absorption tests after various swelling times are shown in Figure 3. In fresh water approximately 80 to 100 grams of water is absorbed per gram of polymer after 30 to 60 minutes, compared to approximately 15 to 20 grams of sea water absorbed in the same period. Results from the flow funnel tests are shown in Figure 4, as flow time measurements for different polymer concentrations in fresh and sea waters. The reference flow time for water (with no polymer) is approximately 6 seconds. For each test, Figure 4 also shows the percentage of water absorbed by the polymer, based on the absorption measurements after 30 minutes (from Figure 3). The flow test results show a rapid increase in relative viscosity when more than 50% of the water is absorbed. The polymer mixtures in fresh water and sea water had

similar flow times (relative viscosities) for similar proportions of water absorbed, but significantly higher polymer concentrations were required in the sea water. The increase in concentration required for a similar viscosity was approximately proportional to the ratio of water absorption in fresh water and sea water (i.e. a factor of about 4.5) from the meas­ urements in Figure 3. The test measurements show that the P.2 super water-absorbent polymer properties are significantly affected by the water salinity. This influences the amount of polymer and swelling time required to effectively condition a soil, relative to that for fresh water. When used at sufficiently high concentration and mixed for sufficient time, these polymers can improve the properties and plasticity of coarsegrained, high water content soils in both fresh water and saline water. 5 MODEL SCREW CONVEYOR TESTS Results from the model screw conveyor tests were assessed based on the observed confinement of the 3.5 bar pressure in the tank during extraction of the sample. When the conditioned soil formed a stable homogenous paste, the tank pressure was confined with a stable pressure gradient along the screw con­ veyor. When a paste was not formed with the condi­ tioning treatment the tank pressure was not confined; the spoil blew out through the screw conveyor. In these tests the flow was uncontrolled and the mater­ ial segregated during extraction with the fines, sand and water first flowing through the conveyor, fol­ lowed by the gravel fraction. 5.1

Figure 3. Water absorption measurements for P.2 polymer.

Standard conditioning treatments

The samples for tests T-1 to T-8 in Table 1 were con­ ditioned with different combinations and quantities

791

Figure 5. Model screw conveyor test pressure measurements: Test T-4 with pressure confinement and controlled flow.

of the standard foam and P.1 polymer agents, and with thick mortar for one sample. Figure 5 shows example screw conveyor test results from test T-4, with the tank pressure remaining stable at 3.5 bars throughout the extraction test and a stable pressure gradient along the screw. The reduc­ tion in pressure towards the end of the test occurs as the sample tank and screw conveyor empty. This test sample contained 46% gravel and 13% fines, with 8% water content. The conditioning treatment was 30% foam and 0.2% P.1 polymer, which formed a homogeneous plastic paste with a slump in the range 20 to 26 cm. The slump was higher than the 10 to 20 cm range typically recommended as ‘suitable’ for EPB operations (e.g. Peila et al, 2009), but the relatively fluid paste in this test allowed confinement of the tank pressure with controlled extraction. As shown in Figure 6, the test T-4 sample grading was within the Zone 3 envelope from Thewes (2007). The conditioning treatment with standard foam and polymer agents was consistent with typical treatments for Zone 3 soils. However, the 3.5 bar pressure confinement sustained was higher than the recommended maximum of 2 bars.

Figure 6 shows grading curves for the test T-1 to T-8 samples with standard conditioning treatments. Confinement of the 3.5 bar pressure was achieved for the T-1 to T-4 samples with gradings in Zones 2 and 3 (solid lines), confirming the recommended condition­ ing treatments and showing that pressures greater than 2 bars can be managed with effective soil conditioning. Slump values measured for these samples ranged from 5 to 25 cm, indicating that controlled spoil flow with pressure confinement is possible for plastic soil pastes with consistency stiffer and more fluid than the typical ‘suitable’ range for EPB operations. For the T-5 to T-8 samples in Zone 4 and coarser with approximately 55% to 75% gravel and 6% to 12% fines (dashed lines), confinement of the 3.5 bar test pressure was not achieved. The conditioning treat­ ments for these very coarse grained samples did not form a homogeneous paste, even with high injection ratios of the standard P.1 polymer for the samples with higher water contents. Segregation of the samples occurred during extraction, with uncontrolled flow through the screw conveyor. These results are consist­ ent with the Thewes (2007) EPB application enve­ lopes, as the conditioning treatments were not effective for such coarse grained soils and the test pressure exceeded the no water pressure recommendation. 5.2 Water-absorbent polymer conditioning treatments

Figure 6. Grading curves of model screw conveyor test samples with standard conditioning treatments.

Tests T-9 to T-15 were performed on samples with grading curves in the Zone 4 envelope or coarser, with approximately 60% to 75% gravel and 6% to 18% fines, and water contents in the range 20% to 25%. These samples modelled soils with gradings at or beyond the typical range for EPB applications with groundwater pressures. The samples were con­ ditioned with ‘special’ treatments using the super water-absorbent P.2 polymer at various injection ratios, in combination with thick mortar for some samples. The polymer injections were mixed with the soil samples for at least 30 minutes before

792

Figure 7. Model screw conveyor test pressure measurements: Test T-9 with no pressure confinement and uncontrolled flow.

extraction, to allow time for water absorption to occur. In tests T-9 to T-12 the pressure was not confined during the screw conveyor extraction. The samples segregated with a rapid drop of pressure at the start of the test and uncontrolled flow with no pressure gradient along the conveyor. Figure 7 shows example results of pressure measurements from test T-9. The test T-9 to T-12 samples were conditioned with the P.2 polymer at 1.0% to 1.5% injection ratio, with 5% and 10% thick mortar for samples T-11 and T-12. Sample grading curves are shown in Figure 8, includ­ ing the additional sand and fines content from the thick mortar. Based on the amounts of polymer injected and the water absorption results shown in Figure 3, the P.2 polymer absorbed approximately 40% to 55% of the saline water in the samples during the mixing stage. From Figure 4, this gave very little increase in the vis­ cosity of the fluid alone. The conditioning treatments were not sufficient to form a homogeneous paste from the coarse grained soils with high water contents, lead­ ing to the segregation and non-confinement of the pres­ sure during extraction.

Figure 8. Grading curves of model screw conveyor test samples with water absorbent polymer conditioning; no pressure confinement.

For tests T-13 to T-15 the pressure was confined during the extraction tests, as shown in Figure 9 for test T-15. These samples were conditioned with larger amounts of the P.2 polymer at injection ratios of 1.5% to 2.0%, in combination with 5% and 15% thick mortar for tests T-14 and T-15. The grading curves in Figure 10 show the increase in sand and fines content from the thick mortar addition. With the higher injection ratios, the P.2 polymer absorbed approximately 80% to 90% of the saline water in the samples during the mixing stage. From Figure 4, this also caused a significant increase in the viscosity of the polymer/water fluid within the samples. The conditioning treatments formed a homogeneous paste from the coarse grained soils with low fines con­ tent and high water content, allowing confinement of the pressure and controlled material flow during the extraction tests. These samples had slump values in the range 12 to 25 cm; that is, generally within the ‘suit­ able’ range for EPB operations. These results show that the special conditioning treatments using sufficient quantities of super waterabsorbent polymer allow effective conditioning of coarse grained soils with high water contents, to con­ fine relatively high pressures. This is achieved through the action of the polymer absorbing the water and binding the soil, so increasing the viscosity/cohesion of the soil matrix and aiding formation of a stable plas­ tic paste to support the coarse grained soil fractions. The sample gradings in Figure 10 contain approxi­ mately 10% more gravel that the Zone 4 envelope by Thewes (2007), and with the 3.5 bar pressure repre­ sent ground conditions beyond the typical range for EPB applications. As shown in Figure 10, these tests illustrate the potential to successfully use EPB TBMs in a wider range of coarse ground conditions with high water contents and groundwater pressures, through the use of specialised polymers to effectively condition the soils. The effects of the polymer in coarse grained soils are enhanced by the addition of fines as part of the conditioning treatment to improve the soil grading.

793

Figure 9. Model screw conveyor test pressure measurements: Test T-15 with pressure confinement and controlled flow.

polymer dose required for effective soil conditioning can be designed respecting these controls, by deter­ mining the water absorption capacity, rate and relative viscosity of the polymer based on the test procedures set out in this paper. The laboratory model tests presented demonstrate the potential to extend the application range of EPB TBMs into saturated coarse grained soils and weak rocks with less than 10% fines and up to approxi­ mately 80% gravel with high permeability and groundwater pressures, through the use of appropri­ ately designed, specialised polymer conditioning treatments. The results presented are specific to the polymer, water and soils used, but appropriate site specific tests can be undertaken to design treatments for different ground conditions. For application in practice, the required polymer swelling time also needs to be considered in relation to the injection point on the TBM and the advance rate, to provide sufficient mixing time for effective conditioning of the spoil. In difficult ground condi­ tions these conditioning treatments can also be mixed with spoil inside the chamber and screw con­ veyor while the TBM is stationary, to restore spoil plasticity prior to continuing advance.

Figure 10. Grading curves of model screw conveyor test samples with water absorbent polymer conditioning; with pressure confinement.

6 CONCLUSIONS The ground conditions for application of EPB TBMs have extended into coarse grained soils through the use of soil conditioning. The laboratory model tests presented here provide experimental support for published guidance on typical conditioning treatments for EPB applications in a range of sand and gravel soils under pressurised conditions. With standard foam and polymer conditioning treat­ ments, formation of a homogenous soil paste for con­ trolled EPB operations is challenging for saturated soils with more than 50% gravel and less than 10% fines, particularly with groundwater pressures. Super water-absorbent polymers can be used to absorb large amounts of water and increase the vis­ cosity/cohesion of the soil. The amount of water absorbed by these polymers and hence their perform­ ance is sensitive to the salinity of the groundwater. They also require sufficient swelling time to develop their properties when mixed with the spoil. The

ACKNOWLEDGEMENTS The views expressed are those of the authors and not necessarily those of the Florida Department of Transport.

REFERENCES BTS 2005. Closed-face tunnelling machines and ground sta­ bility, A guideline for best practice. London: Thomas Telford. DAUB 2010. Recommendation for selection of tunnel boring machines. Koln: German Tunnelling Committee.

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EFNARC 2005. Specification and guidelines for the use of specialist products for mechanised tun­ nelling in soft ground and hard rock. Farnham: EFNARC. Maidl, B., Herrenknecht, M. & Anheuser, L. 1996. Mechanised shield tunnelling. Berlin: Ernst and Sohn. Merritt, A., Storry, R. & Brais, L. 2015. Soil condi­ tioning testing and monitoring for the Port Miami Tunnel. In Proc. ITA World Tunnel Congress, Dubrovnik, 2015.

Milligan, G. 2000. Lubrication and soil conditioning in tunnelling, pipe jacking and microtunnelling: A state-of­ the-art review. London: The Pipe Jacking and Tunnelling Research Group. Peila, D. Oggeri, C., & Borio, L. 2009. Using the Slump Test to Assess Behaviour of Conditioned Soil for EPB Tunnelling. Environmental and Engineering Geoscience, Vol. XV (3): 167–174. Thewes, M. 2007. TBM Tunnelling Challenges – redefin­ ing the state of the art. In. Proc. ITA World Tunnel Congress, Prague, 2007.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Influence of mineral or polymeric modification on bentonite-based tunnel face support P. Mianji, W. Baille & T. Wichtmann Department of Civil and Environmental Engineering, Ruhr-Universität Bochum, Bochum, Germany

R. Verst & M. Pulsfort School of Architecture and Civil Engineering, University of Wuppertal, Wuppertal, Germany

ABSTRACT: The time-dependent penetration of bentonite-based support fluids into non-cohesive soil is a current subject of investigation, especially in the field of shield tunnelling with fluid face support due to the cyclic excavation process. Research has so far focused on linking the time-dependent penetration of clean bentonite suspensions with the penetration rate of the cutting wheel. However, experience shows that benton­ ite slurries in-situ often contain suspended fines, which can change the penetration behaviour of the slurry significantly. Moreover, in recent years, it has become common practice to improve stability or rheological behaviour of bentonite suspensions by adding water-soluble polymers. Understanding the effect of such slurry modifications on the time-dependent penetration behaviour is an essential basis for realistic face support models. The present study analyses the influence of a mineral or polymeric modification of a typical bentonite slurry by means of penetration tests with sand columns at over-pressure. The sand type was chosen so that a pure unmodified bentonite slurry does not create an external filter cake, but penetrates into the pore space of the soil to be supported. Pore water pressure sensors along the height and measurements of inflow and outflow using electronic balances enable a time-dependent determination of penetration depth and pressure transfer. For selected tests, a special focus lies on the evaluation of the effect of filter cake formed and penetration on the soil permeability. From the results, it can be concluded that a bentonite modification with fines results in a significant reduction in the penetration depth with the formation of an external filter cake. The addition of specific polymers reduces the amount of bentonite needed to create a stable suspension with defined rheology. Special care must be taken regarding the choice of fines and polymers.

1 INTRODUCTION Mechanized tunnelling with a bentonite-based sup­ port system is a widely accepted method, especially when the excavation is performed in a saturated ground. In this method, the slurry chamber is par­ tially filled with slurry and pressurized towards the tunnel face to counteract the ground and hydrostatic pressures. To ensure a safe movement of the tunnel boring machine (TBM) as well as to minimize the ground-surface subsidence, a steady support pressure has to be guaranteed at the working face during the excavation process (Anagnostou & Kov´ari 1994, Kirsch 2010, Thewes et al. 2016). The effective slurry pressure, i.e. the pressure dif­ ference between chamber and hydrostatic water pres­ sures, leads to the penetration or/and infiltration of the slurry into the soil in front of the TBM. This pro­ cess is influenced by various factors, mainly the grain size of native soil and slurry, the rheological properties of the slurry and the effective slurry

pressure. Three main mechanisms are found in litera­ ture: (I) external filter cake formation, (II) pure slurry penetration, no filter cak, (III) combined pene­ tration and filter cake formation, internal and exter­ nal cake (Min et al. 2013, Arwanitaki 2009, Thienert 2011, Haugwitz & Pulsfort 2018). They are schemat­ ically depicted in Figure 1. In mechanized tunnelling, the time-dependency of infiltration and/or penetration of the slurry plays a significant role due to the cyclic excavation process (Zizka et al. 2017). The time-dependent penetration and infiltration behaviour of ‘clean’ bentonite slur­ ries has been extensively studied (Broere & Van Tol 2000, Bezuijen et al. 2005, Talmon et al. 2013, Yin et al. 2016, Xu & Bezuijen 2018, Zizka et al. 2017,. Haug-witz & Pulsfort 2018). It is known that a penetration of the slurry reduces the efficacy of the support pressure as only a part of the support pres­ sure counters the excess pore water pressure. The support pressure being only partly available on the working face has to be taken into account for face

DOI: 10.1201/9780429321559-105

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Figure 1. Infiltration and transport mechanisms of bentonite slurry.

stability calculations. Therefore, the formation of an almost impermeable external filter cake is the ideal scenario, in which the support pressure is entirely transferred to the soil structure (Anagnostou & Kov ´ari 1994). In coarse-grained soil, where the perme­ ability is high, the penetration of pure bentonite slurry is inevitable. According to Kanayasu et al. (1995) and Krause (1987) schildvortrieb, a pure bentonite slurry is inapplicable when the permeability of the ground exceeds the range of 1 · 10-4 to 5 · 10-3 m/s. Equally, the approach of German standard to calculate a minimum static yield strength τF of the support fluid based on empirical correlations with the d10 value of the ground would result in τF values unattainable with regular bentonite slurries for these cases (DIN 4126 2013, Haugwitz & Pulsfort 2018). In recent years, it has been found by several authors that bentonite slurries can be applied in coarser soils, if certain fines are suspended in the slurry which significantly diminish the penetration by clogging of the larger soil pores and subse­ quent ceasing or deceleration of the infiltration process. Xu & Bezuijen (2019) investigated the infiltration characteristics of slurry containing sand. They found that the existence of a small amount of sand in the slurry reduces the infiltra­ tion rate significantly. They reported that the per­ meability of the infiltrated sample decreased drastically with increasing sand content in the slurry. Arwanitaki (2009) found a significant influence of the fines content in the slurry on the filter cake thickness and composition of slurrysupported diaphragm walls (resulting in an increase of the wall friction angle (¢) with the diameter of fines in the filter cake after concret­ ing). Thienert (2011) found a significant influence of sand particles suspended in the bentonite slurry on the penetration depth as well as the pressure gradient along the penetration length. He developed an empirical formula to predict the maximum penetration (or rather infiltration) depth of bentonite suspensions loaded with specific sizes of fines similar to the filter criteria of Ter­ zaghi & Peck (1948). The effect of suspended fines on the behaviour of polymer solutions as support fluids has also been studied experimen­ tally by Verst & Pulsfort (2019) showing the same tendency as the aforementioned authors. The choice of suitable fines was thereby drawn

based on the criteria for filtration according to Terzaghi & Peck (1948) and the support fluid was chosen according to Stoke’s law (Stokes 1850) in order to ensure a viscosity at low shear rates sufficient to carry the fines. Moreover, ‘blended’ bentonites were developed by adding water-soluble polymers to improve the rheological and physical or chemical properties of bentonite slurries for an improved filtration resist­ ance and stability. One of the first applications of polymers as additives in slurry shield tunnelling is the Grauholz tunnel (Jancsecz & Steiner 1994), in which polymer was added alongside sawdust and sand to the slurry. It was revealed that saw­ dust is not appropriate under regular advancement as it had clogged not only the soil pores but also the sieves in the separation plant. Subsequently, Fritz & Tandler (1999) and Fritz et al. (2002) have performed extensive laboratory investiga­ tions to find appropriate additives. They suggested a combination of polymer and vermiculite as an additive for a tunnel project in Switzerland, which had the high capacity to form a tight mem­ brane in these specific site conditions. According to the literature, there is no doubt that polymer and fines can be used to improve clean ben­ tonite slurries. As a result, the application of benton­ ite slurries can be widened to more unfavourable ground conditions, i.e. ground with higher perme­ ability and instability conditions, while keeping the bentonite-slurry-specific performance. However, some uncertainties remain regarding the appropriate types and concentrations of fines and of polymer for bentonite slurry modification, espe­ cially concerning their influence on the timedependent penetration behaviour of the mixture. These comprise the penetration depth as well as the pressure transfer and the hydraulic properties of the infiltrated soil. These aspects shall be addressed experimentally in the present paper. Firstly, the influ­ ence of the polymer type on the stability and proper­ ties of the slurry are discussed based on visual and bulk rheological evaluation. Accordingly, appropri­ ate combinations of additives are chosen for slurries with different bentonite concentrations in order to perform penetration tests in a vertical large-scale column with a reference ground material. Finally, a detailed discussion on the influence of various combinations of slurry additives on the time­

797

3 METHODS

dependent penetration behaviour and pore water pressure generation as well as on the hydraulic prop­ erties of the infiltrated soil is provided.

3.1

2 MATERIALS The properties of reference ground material (‘host sand’) and fines used for the experimental tests are given in Table 1. The host sand is chosen with poor grading and high permeability (2 · 10-3 m/s) to create full permeation of a commonly used pure bentonite slurry. Concerning the fines, three ‘non-reactive’ options were selected as slurry additives: sand equal to the natural ground, a fine sand matching the filter cri­ teria of Terzaghi & Peck (1948) to ensure filtration at the soil surface and a finer material (quartz powder). A common Na-activated bentonite powder with high swelling capacity, stability and plasticity prop­ erties is used for the preparation of the slurry. The used bentonite has a liquid limit of wL ¼ 381:8% and a plasticity of IP ¼ 345:9%. Concerning the choice of suitable polymers, four polymers having different properties and genesis, either synthetic-based or cellulose-based, were used to modify the pure bentonite slurry. The specifica­ tions of the utilized polymers (based on manufac­ turer information) are provided in Table 2. Two types of water-soluble polymers, a biopolymer derivative and a synthetic product, were studied with variations in molecular weight (chain length) and charge density. The quantities suggested from manu­ facturers vary according to their application (viscosi­ fier, stabilizer, etc.) with significant deviations from 0.02 g of polymer per 1 g of bentonite to 5 g of poly­ mer per 1000 ml of solution.

Table 1.

Properties of host sand, fine sand, and fines.

Soils

Grain size

Host sand Fine sand Quartz powder

Table 2.

Z ZLV S SHC

d50

d10

Cu

[mm] [mm] [mm] [-] 1-2 1.53 1.2 1.32 0.1-0.5 0.24 0.14 1.86 d99 50:125 mm, d85 50:063 mm

Cc [-] 1.05 5.27

Polymer properties. Type

Molecular weight

Charge

Cellulosic Cellulosic Synthetic Synthetic

High Low Low Low

High DS*

High DS*

Medium DA**

High DA**

* DS = degree of substitution (= anionic functional groups) ** DA = degree of anionicity

Slurry mixing

To prepare the pure bentonite slurry, bentonite powder was added to tap water while mixing at approximately 2000 rpm for 10 minutes using a 4-bladed propeller. Thereafter, the mix was kept in a temperature-controlled environment (to dispel the effects of temperature on the bentonite swelling) for 18 hours to attain its maximum swelling capacity. The slurry was remixed by hand for one minute before each experiment. With regard to possible methods to mix polymer, bentonite and water, polymer manufacturers suggest different approaches depending on the type of poly­ mer and the desired property modification. In this case, different mixing methods with different poly­ mer concentrations were studied with the objectives to both reduce the amount of bentonite needed and to increase the static yield strength or gel strength to keep the chosen fines in suspension. The evaluation criteria were the visual homogeneity (stability) of the mixture directly after mixing and after full hydra­ tion and swelling (18 hrs), bentonite-slurry­ resembling bulk rheological behaviour (see next sec­ tion) and a visual evaluation of the settling behaviour of the chosen fines within the slurry. Several methods (such as powder-to-powder mixing with subsequent addition to water at high shearing rates, the addition of unhydrated bentonite to a polymer solution as well as direct mixing of vis­ cous polymer solution and hydrated bentonite slurry) were ruled out with respect to the shear sensitivity of hydrated polymers as well as the behaviour of vis­ cous polymer solutions to encapsulation rather than dispersion. The best mixing method was found to be the addition of polymer powder to the fully dispersed bentonite-water mix at low shearing rates for about 50 minutes directly after dispersion (10 minutes) and before bentonite hydration. Polymer addition to the slurry after bentonite hydration and swelling resulted in slightly less homogeneous mixes in some cases. Only slurries which were able to keep the fines in suspension were chosen for the penetration tests. The addition of fines into the slurry was performed directly before the test started and differed according to the size of the fines to ensure a maximum disper­ sion of the particles. The air-dried sand material was slowly sprinkled and stirred into the slurry, while the quartz powder was mixed into the slurry with prior dispersion in a small amount of tap water. 3.2

Rheological testing

The standard rheological testing employed to evaluate the slurry properties comprised bulk rheological meas­ urements with rheometer (at 20� C), ball harp and Marsh funnel, the latter two according to the German standard DIN 4127 (2014). Ball harp and rheometer provide device-specific values for the static yield

798

strength τF . Additionally, rheometer flow curves and Marsh funnel times were used as indicators of the grade of viscosity.

conditions. The slurry penetration depth over time was calculated based on balance measurements from inflow and outflow assuming uniform fluid displace­ ment in the pore structure defined by porosity.

3.3 Penetration tests The experimental base of the present study consists of penetration tests performed with water-saturated vertical soil columns at different scales. The devices used are shown in Figure 2. Based on the observations in small-scale tests, minimum concentrations for bentonite and polymer with a medium amount of fines were chosen for four large-scale tests (‘LS-T’) allowing for a detailed evaluation of the time-dependent pore pressure devel­ opment along the penetration depth over a 24 hrs period. All large-scale tests were performed with downwards flow to allow direct contact between host sand and slurry for a more realistic filter cake build­ up. An effective pressure of 50 kPa was applied to the slurry in both cases. For this, the pressures of inflow and outflow chambers were regulated to 90 kPa and 40 kPa, respectively. The applied back pres­ sure prevented the samples from desaturating due to gravity. Pore pressure sensors radially distributed along the height of the cylinder enabled the evalu­ ation of the pressure drop along the penetration depth over time. Water permeability tests were performed after slurry infiltration through the filter cake as well as through the subsoil after removal of the cake. The air-pluviation method was used in both col­ umns for sample preparation to ensure reproducible density and permeability (ρd ¼ 1:67 g/cm3 , kw ¼ 2 · 10-3 m/s) of the host sand before infiltration. Constant-head water flow tests were performed prior to every test for verification of similar flow boundary

4 RESULTS 4.1 Selection of polymer and fines based on SS-T and rheological evaluation 66 small-scale penetration tests (‘SS-T’) were per­ formed for pre-selection to achieve the best possible results for the combined action of all additives. The variations included pure bentonite suspensions at 40­ 60 g/l, polymer addition at concentrations of 0.25-1.5 g/l and all listed types of fines at 5-40 g/l with upwards flow to eliminate possible effects from sedimentation. As expected, only the fines which fit the Terzaghi filter criteria created a thin layer of colmated fines and sub­ sequently reduced the permeability of the tested sand interface to a minimum to create stagnation. Concerning the choice of polymer for bentonite suspensions (at 20-30 g/l), visual and rheological evaluation revealed the cellulosic polymer of high molecular weight (‘Z’) to create the most homoge­ neous slurry at concentrations of around 0.25-1.5 g/l (here: g per litre of slurry). A significant influence was found on the shear stress of the resulting slurry mix at low shear rates. As expected from their low molecular weight, the other polymers did not have any effect on the bulk rheological behaviour at low concentrations. Higher concentrations of SHC entirely flocculated and thus completely destabilised the slurry due to its high charge density. Its medium charge density representa­ tive S created barely visible small micro-flocs (without

Figure 2. Laboratory setup for small-scale (SS-T left) and large-scale (LS-T right) tests.

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the tendency to destabilise, i.e. separate into water and bentonite) at higher dosages of around 0.25-1.5 g/l, which increased the rheometer yield value. Smallscale penetration tests were therefore performed with the addition of polymer (0.25-1.5 g/l) of types Z and S. All tests with polymer type Z and fines added in the range of 5-40 g/l resulted in full stagnation. The tests with polymer type S revealed measurable amounts of residual (clear) filtrate water despite a filter cake formation in some cases, so the stability of the slurry mix was not guaranteed. Therefore, polymer type Z with fine sand (‘F’) was chosen for bentonite slurry modification. A minimum concentration of 0.5 g/l of polymer (‘Z0.5ʹ) was selected as sufficient to carry the fines needed to create a filter-cake formation through col­ mation of fines in the small-scale tests. 4.2 Bulk rheology

Table 3.

B20+Z0.5 B30+Z0.5 B40 B50 B60

Bulk rheological evaluation. Marsh funnel

Ball Harp

Rheometer

1l/1.5l [sec] 40/66 54/94 36/59 42/72 52/100

τF [Pa] 55 16.9 16.8 21.2 31.5

η/τF [Pa*s]/[Pa] 0.0051/0.4 0.0049/2.2 0.0046/3.3 0.0045/7.6 0.0047/13.5

concentrations did not carry the fines. They were therefore subjected to polymeric modification. 4.3 Penetration behaviour and pore pressure development

Figure 3 illustrates the rheometer test results of the slurries used in the large-scale column. To accentu­ ate the polymeric influence, the untreated B20 and B30 (the number behind B denotes the bentonite content in the slurry in g/l) are also presented. Table 3 provides an overview of the bulk rheo­ logical properties of the slurries used in the largescale column tests. Polymer addition to the lowconcentration bentonite slurries clearly enhanced their bulk rheological parameters. For instance, the Marsh funnel times of B20+Z0.5 slurry resembled a pure bentonite slurry at 50 g/l, and the B30+Z0.5 slurry reached a yield point similar to that of B40 slurry, yet, with significantly higher Marsh times. From the rheological investigations, it was revealed that the bentonite slurries at concentrations of 40 and 50 g/l (‘B40ʹ and ‘B50ʹ) could carry the fines without the need of polymer addition while slurries at lower

Figure 4 presents the slurry penetration depth over time for the tested slurries with reference to B60. The penetration curves of the tested modified slurries can be partitioned into a steep part (initial) and a flat part (subsequent until 24 hrs), before and after visual stagnation due to formation of a low-permeable membrane-like filter cake. Table 4 summarises these results for the modified slurries. The reference curve B60 shows immediate slurry penetration of the sand without stagnation, visible as a vertical line at the y-axis. In contrast, the modified slurries showed a distinct border clearly separating ini­ tial (s1 ) penetration and subsequent (Δs24 ) infiltration. Concerning the initial penetration, B20+Z0.5+F10, B40+F10 and B50+F10 are ordered according to their static yield strength. B30+Z0.5+F10, however, despite inferior rheological properties, displays the least initial penetration. Interestingly, the subsequent infiltration through the filter cake formed with fines without

Figure 3. Rheometer flow curve of suspensions.

Figure 4. Penetration depth over time.

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Table 4.

(4) at the end of the test (24 hrs). Figure 6 shows a comparison of penetration depth s and pressure drop Δpinter face between the slurry pressure immedi­ ately above the interface and the first sensor in the sand over the first few minutes of penetration. At the beginning, due to the high permeability of the sand, the applied pressure could readily distribute

Evaluation of penetration and filter cake.

B20+Z0.5+F10 B30+Z0.5+F10 B40+F10 B50+F10

Initial penetration t1 /s1

Subsequent infiltration Δs24h

Filter cake thickness h24h

[sec]/[mm] 16.5/224 4.5/99 14/161 10.5/129

[mm] 30 29 87 69

[mm] 3-4 2-3 5.5-6.5 6-7

polymer addition is higher by a factor of 42 com­ pared to the tests with polymer modification, even though the filter cake measured after 24 hrs is thicker. Here, a faster stagnation (parameter t1 ) is correlated with increasing bentonite concentration and a slightly thicker filter cake. In the case of polymer modification, the opposite can be observed, while the subsequent penetration depth, which represents the permeability of the formed cake, is fairly similar. In practice, the least penetration is desired for minimum fluid consumption, so the best values are obtained with B30+Z0.5+F10. However, all values range between 100 and 250 mm, and could be con­ sidered as acceptable. A quicker filter-cake build-up could be achieved with a higher fines content as the colmation of fines in the filtration zone is the pri­ mary reason for stagnation. The pore pressure development at distinct time steps corresponding to the penetration behaviour is given in Figure 5. The selected time steps comprise: (1) the pressure distribution before stagnation (3-9 sec), (2) at the time of visual stagnation/filter cake formation (5-17 sec), (3) within the first few minutes after filter cake formation, when the pressure is com­ pletely adsorbed by the filter cake (220-232 sec), and

Figure 6. Comparison of fs0 (assumption Δh ¼ 2 cm), respective pressure drop Δp at the fluid-sand interface and penetration depth s over time for different slurry types used.

Figure 5. Pore pressure development along cylinder height in large-scale tests (LS-T).

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along the height. The low viscosity of the slurry resulted in a subsequent high inflow velocity of the slurry. The maximum penetration depth is therefore reached shortly after the start of the penetration. A striking drop in pore fluid pressure (‘PFP’) at the interface of soil and slurry, between the first sensor (at height 260 mm) and the second sensor (at height 220 mm) is observed at approximately the same time (3-9 secs). This shows the link between filter cake formation, rapid decrease in permeability and pres­ sure adsorption at the interface. The resulting dense and low permeable layer further increases the pres­ sure drop at the interface to approximately 100%. This optimum sealing capacity reached immediately after the filter cake formation is maintained until the end of the test. The pressure gradient (Figure 6 bottom) can be calculated according to Equation 1

with Δh ≈ 2 cm (at maximum) being the length over which the pressure is adsorbed, calculated from the filter cake thickness in addition to the depth of the first sensor in the sand. According to DIN 4126 (2013), full fluid support can be assumed with a stagnation gradient 4200 kN/m3 or a reduction 55% of the support pressure at the soil interface. The required support pressure at the tunnel face can therefore be considered as fully mobilised within the first minute of penetration for the given sand and the chosen four slurry mixtures. 4.4

Permeability

Long-term penetration tests were performed to obtain stable filter cakes and the possibility to investigate the influence of different slurries on the hydraulic proper­ ties of the prior penetrated soil as summarised in Table 4 and Table 5. Penetration tests with water after 24 hrs of slurry penetration were used as an indicator of the soil permeability. However, piping effects were observed, especially from penetration through the cakes, which increased the measured flow rates on which the permeability calculations are based. According to the time-dependent penetration results summarised in Table 4, the polymer-modified slurries have similar initial and notably smaller Table 5.

Results of permeability measurements.

B20+Z0.5+F10 B30+Z0.5+F10 B40+F10 B50+F10

kw;start [m/s]

kw;cakeþsoil [m/s]

kw;soil [m/s]

2 · 10-3 2 · 10-3 2 · 10-3 2 · 10-3

8…400 · 10-6 3…50 · 10-6 8…20 · 10-5 3…200 · 10-6

7 · 10-4 2 · 10-4 8 · 10-4 4 · 10-4

subsequent penetration depths, which is due to the creation of a lower permeable filter zone. However, the lower permeability is not governed by the thick­ ness of the filter cake but by its porosity, as the poly­ mer-loaded slurries create a thinner filter cake compared to the slurries without polymer. In contrast, the permeability of the slurry-penetrated zone below the filter cake is not significantly influ­ enced by the type of slurry, e.g. the used polymer. Similar permeability coefficients were measured for all slurries. However, it is observed that the slurry pene­ tration causes a clear reduction in permeability of the subsoil. 5 CONCLUSION From the results of the present experimental investi­ gation, it can be concluded that the presence of appropriate fines in a slurry can decrease the penetra­ tion depth via colmation. The addition of fines is also an effective method to extend the applicability of bentonite slurries in coarse grounds without the need to significantly increase the bentonite content. The small-scale tests’ results with varying sizes of fines confirm that the application of the Terzaghi filter criterion is a qualified method to choose a suitable grain size of fines leading to stagnation of the slurry due to colmation. The bulk rheological investigation and smallscale penetration tests showed that certain types of polymer at specific ranges of concentration signifi­ cantly reduce the required bentonite concentration. A combination of bentonite concentrations of 20 g/l and 30 g/l with small amounts of polymer resulted in slurries with rheological properties almost equal to the slurries with 40 g/l and 50 g/l of concentrations of the bentonite while creating a stable slurry cap­ able of suspending all tested fines. Moreover, the large-scale penetration tests with slurries containing appropriate fines revealed that the penetration curve over time can be clearly divided into an initial penetration and a subsequent infiltra­ tion. The former involves the largest fraction of the penetration depth of the slurry and ends with the filter cake formation and full pressure transfer at the slurrysoil interface within a short period of time. The latter reflects the permeability of the formed filter cake. The polymer-modified slurries exhibited signifi­ cantly lower subsequent infiltration depths after the formation of the colmation zone and a lower filter cake thickness after 24 hrs compared to the slurries without polymer, which evidences of an improve­ ment in slurry properties by polymer addition. The water permeability of the filter cake and the infil­ trated soil below the filter cake did not show any marked influence of the type of slurry. Generally, with regard to the excavation cycle in mechanized tunneling with slurry face support, it can be concluded that a large majority of the pres­ sure difference (480%) is transferred at the slurry­

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soil interface already within the first few seconds during filter cake formation, so within one revolution of the cutting wheel. However, full pressure drop at the interface can only be safely assumed after several minutes, which should be taken into account with regard to the revolution rate. Further investigations are needed for the verifica­ tion of the applicability to a wider range of soils, bentonite and polymer products.

REFERENCES Anagnostou, G. & K. Kova´ri (1994). The face stability of slurry- shield-driven tunnels. Tunnelling and Under­ ground Space Technology 9(2), 165–174. Arwanitaki, A. (2009). U¨ ber das Kontaktverhalten zwischen einer Zweiphasenschlitzwand und nichtbindi­ gen B¨oden. Ph. D. thesis, Ruhr-Universit¨at Bochum, Fakult¨at f¨ur Bau- und Umweltingenieurwissenschaften. Bezuijen, A., J. P. Pruiksma, & H. H. van Meerten (2005). Pore pressures in front of tunnel, measurements, calcula­ tions and consequences for stability of tunnel face. In Tunnelling. A Decade of Progress. GeoDelft 1995–2005, pp. 42–49. CRC Press. Broere, W. & A. Van Tol (2000). Influence of infiltration and groundwater flow on tunnel face stability. Geotech­ nical as- pects of underground construction in soft ground 1, 339–344. DIN 4126 (2013, 09). Nachweis der Standsicherheit von Schlitzwa¨nden. DIN 4127 (2014, 02). Erd- und Grundbau–Pru¨fverfahren fu¨r Stu¨tzflu¨ssigkeiten im Schlitzwandbau und fu¨r deren Aus- gangsstoffe. Fritz, P., R. H. Stengele, & A. Heinz (2002). Modified ben­ tonite slurries for slurry shields in highly permeable soils. In 4th International Symposium Geotechnical Aspects of Un-derground Construction in Soft Ground, Toulouse, France. Fritz, P. & C. Tandler (1999). Hydroschildvortrieb Her­ metschloo–Werdho¨lzli in hochdurchla¨ssigem Schotter. ETH Zu¨rich–Institut fu¨r Geotechnik Weiterbildungs­ kurs 27, 28. Haugwitz, H.-G. & M. Pulsfort (2018). Pfahlwa¨nde, Schlitzwa¨nde, Dichtwa¨nde, pp. 823–907. Wiley Online Library. Jancsecz, S. & W. Steiner (1994). Face support for a large mix- shield in heterogeneous ground conditions. In Tun­ nelling 94, pp. 531–550. Springer. Kanayasu, S., I. Kubota, & N. Shikibu (1995). Stability of face during shield tunneling–a survey of japanese shield

tunnel- ing. Underground construction in soft ground, 337–343. Kirsch, A. (2010). Experimental investigation of the face stabil-ity of shallow tunnels in sand. Acta Geotechnica 5 (1), 43–62. Krause, T. (1987). Schildvortrieb mit flu¨ssigkeits- und erd­ gestu¨tzter Ortsbrust. Mitteilungen des Instituts fu¨r Grundbau und Bodenmechanik der Technischen Univer­ sita¨t Braunschweig (24). Min, F., W. Zhu, & X. Han (2013). Filter cake formation for slurry shield tunneling in highly permeable sand. Tunnelling and Underground Space Technology 38, 423–430. Stokes, G. G. (1850). On the effect of internal friction of uids on the motion of pendulums. Trans. Camb. Phil. Soc. 9, 8106. Talmon, A., D. Mastbergen, & M. Huisman (2013). Inva­ sion of pressurized clay suspensions into granular soil. Journal of Porous Media 16(4). Terzaghi, K. & R. B. Peck (1948). Soil mechanics in engin­ eering practice. John Wiley & Sons. Thewes, M., B. Schoesser, & Z. Zizka (2016). Transient face support in slurry shield tunneling due to different time scales for excavation sequence of cutting tools and penetration time of support fluid. In Proceedings of the ITA world tunnel congress. Thienert, C. (2011). Zementfreie Mo¨rtel fu¨r die Ringspalt­ ver- pressung beim Schildvortrieb mit flu¨ssigkeitsges­ tu¨tzter Orts- brust. Ph. D. thesis, Universita¨t Wuppertal, Fakulta¨t fu¨r Ar- chitektur und Bauingenieurwesen. Verst, R. & M. Pulsfort (2019). Einfluss des Polymertyps auf die Standsicherheit polymerflu¨ssigkeitsgesttzter Erdwa¨nde, pp. 33–52. Shaker Verlag. Xu, T. & A. Bezuijen (2018). Pressure infiltration charac­ teristics of bentonite slurry. Ge´otechnique 69(4), 364–368. Xu, T. & A. Bezuijen (2019). Bentonite slurry infiltration into sand: filter cake formation under various conditions. Ge´otechnique, 1–12. Yin, X.-s., R.-p. Chen, Y.-c. Li, & S. Qi (2016). A column system for modeling bentonite slurry infiltration in sands. Journal of Zhejiang University-SCIENCE A 17 (10), 818–827. Zizka, Z., B. Schoesser, & M. Thewes (2017). Excavation cycle dependent changes of hydraulic properties of granu­ lar soil at the tunnel face during slurry shield excavations. In Geotech- nical Aspects of Underground Construction in Soft Ground: Proceedings of the 9th International Symposium on Geotech-nical Aspects of Underground Construction in Soft Grounds (IS-Sa˜o Paulo 2017), April 4–6, 2017, Sa˜o Paulo, Brazil, pp. 137. CRC Press.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Compensation grouting for conventional tunnelling with low overburden at the Oberau Bypass Tunnel E. Neun EDR GmbH, Munich, Germany

S. Sabew Renesco GmbH, Lörrach, Germany

ABSTRACT: The 3 km long twin-tube Oberau Bypass tunnel is the longest road tunnel in the state of Bav­ aria and crosses two mountain ranges, as well as the Gießenbach valley which is characterised by fluvial deposits. With an overburden of only 10 m and sensitive industrial buildings on top of the tunnel alignment, the project called for over 7,700 m² of compensation grouting as a mitigating measure to counteract the settle­ ments caused by the tunnelling. The paper describes the design concept for tunnelling and compensation (FE analysis), the impact on the surface settlements, the interaction between tunnelling and compensation grouting as well as the experience gained during the project execution, in particular regarding drilling deviations and effectiveness of the compensation grouting. The project has been successfully completed without any damage to the buildings or interruption of the tunnelling works.

1 INTRODUCTION 1.1

Project area and constrains

The Oberau bypass tunnel (diameter ~11.8m) runs under two mountain ranges and crosses the village of Oberau in the Gießenbach valley. As a consequence of the alignment, the tunnel passes under a large industrial building, as shown in (Figure 1). The area influenced by tunnel related settlements under this building had been assessed to be about 7,000 m². 1.2

According to the results of the borehole prob­ ing, the upper and lower gravel predominantly have a medium density to very dense bedding. The upper gravel can then be classified as highly permeable at a permeability value of approx. 1.5 x 10-4 m/s to 2.0 x 10-3 m/s. The groundwater level in the Gießenbach valley can typically be found 1 - 2 m below the tunnel invert, but can rise by several metres in cases of heavy rainfalls. 1.3

Geology

The tunnel drive in the area of the Gießenbach valley traverses alluvially deposits. From ground level down to a depth of approx. -50 m gravel had been encoun­ tered in the geological surveys with underlying glacial deposits of a ground moraine. The gravel can be div­ ided into an upper and a lower gravel layer. The bound­ ary of this layer was largely explored at a depth of approx. 15 m to 22 m below ground level. The tunnel runs mainly in the upper gravel. Only bench/invert excavation touches on the lower gravel in some areas. The geological longitudinal section is shown in (Figure 2). The upper gravel shows a comparatively higher stone and fine grain content and is inter­ spersed with stones and blocks with a diameter of up to 50 cm. The lower gravel consists mainly of sandy, slightly stony gravel.

The challenge

The building above the tunnel had been deter­ mined to be sensitive to differential settlements (tilt) of the individual pillar foundations. Strict criteria for these differential settlements had been defined based upon a detailed analysis of the building. The maximum tilt between two adjacent foundations had to be limited to 1:250 of the span between the foundations (Figure 3), which translates into 2.5 cm for a typical 6.25 m spacing between the columns. In order to avoid reaching or surpassing this maximum limit, warning and alarm limits were also defined (Table 1). Reaching the alarm value called for information of all parties involved by SMS and E-mail, whereas the intervention limit called for a stop in tunnel exca­ vation and intervention by compensation grouting.

DOI: 10.1201/9780429321559-106

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no observable positive impact from the pilot tunnel. Therefore full crown excavation plus bench/invert excavation has been adopted. Given the strict limits for deformation and the prevailing ground conditions, self-drilling spiles and radial anchors were adopted (Figure 4). After considerable settlements in excess of 6 cm were measured when entering the Gießenbach valley, the type of spiles underwent a thorough review: With the spile configuration shown in (Figure 4), there were up to 4 layers of spiles stacked on top of each other. The drilling of each hole causes a certain amount of settlement due to borehole instability in the overcut despite the use of mortar during drilling (Figure 5). The typical overcut of a standard self-drilling spile, especially in case couplers need to be used, measures about 2 cm and is used to extract the material behind the drill-bit, creating a certain flow of material from the top of the drill to the drill base. This movement allows for instabilities in the borehole and in consequence leads to settlements on the surface. With the boreholes in the crown inclined upward the cement slurry used for stabilisation of the annular void around the spile has the tendency to fall out of the drilling base. The key to solving this problem was to switch from traditional self-drilling spiles to tube spiles. Tube spiles are characterised by a very small overcut of about 2 mm only. The drill­ ing spoil is transported through the tube itself rather than the annular void. The surface settlements caused by the drilling (rather than the soil/tunnel deformation) have been reduced to a few millimetres thanks to the use of tube spiles instead of standard self-drilling spiles.

Figure 1. Industrial building and tunnel alignment.

Figure 2. Geological longitudinal section.

Figure 3. Tilt between column foundations.

Table 1.

Values for Alarm, Intervention and Limit.

Type

Alarm

Intervention

Limit

Tilt

1/500

1/350

1/250

Figure 4. Excavation class in the Gießenbach valley.

2 CONCEPT FOR EXCAVATION AND COMPENSATION GROUTING 2.1

Excavation concept

The original concept foresaw a pilot tunnel in the centre of the top heading which would have been widened to the full tunnel width, followed by the bench/invert excavation. Measurements from a section of the tunnel excavated with this concept at the northern portal of the tunnel and extensive 2D and 3D numerical analysis indicated that there was

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Figure 5. Annular void around self-drilling spiles.

2.2 Criteria and concept for compensation grouting There were several criteria that had to be taken into account when developing the concept for the com­ pensation grouting. The most important criteria were: 1) Limit the tilt between pillar foundations 2) Avoid the use of compensation grouting close to the tunnel face as this might have a negative impact on the face stability and may reduce the efficiency of the compensation grouting 3) Avoid impact on construction sequence due to the need to wait for any grouting to compensate for settlements during the tunnel drive.

Figure 6. Location of hose pipe levels.

With these criteria in mind the following concept was adopted: First, the expected surface settlements for each construction stage (top heading, bench/ invert) were predicted with a 2D finite element model. This also showed the impact of the exca­ vation of the second tube on the surface above the first tube, taking into account that the cri­ teria was based on the tilt between individual foundations rather than absolute values for settlements. The result of this analysis proved that the tilt would be in excess of the maximum acceptable tilt if no compensation grouting were applied. If compensation grouting were to be applied as soon as the tilt reaches the interven­ tion limit, this would contradict criteria 2 and 3 (grouting close to the face and having to stop tunnelling). Therefore the decision was made to lift the building by about 50% of the predicted settlement for the top heading prior to starting the excavation in the Gießenbach valley. As cri­ teria 1 (max tilt) also applies to heave it had to be checked that the tilt between all columns of the building was within acceptable range during the heaving process. Following the pre-heave and the settlements caused by the top heading of both tubes, interim-heave grouting was to be executed before excavation of bench and invert and last but not least a final compensation after the finish of all tunnelling activities in the area. As the on-site settlements can differ from the pre­ dictions, the amount of intermediate and final compensation could not be determined in advance and had to be decided upon once the real on-site settlement had been measured. The settlement of every single column had to be monitored. Therefore, each of the 122 col­ umns of the industrial building was equipped with hose pipe levels and real-time measure­ ments (Figure 6). The deformations were recorded automatically every minute and compared with the limits of differ­ ential deformations. All data was available at any time via an online platform.

3 GENERAL MECHANISM OF COMPENSATION GROUTING Compensation grouting can be used to compensate for settlements of buildings through locally limited heav­ ing of the ground and by this of the building. For this purpose, a nearly horizontal screen of sleeve pipes is installed under the building to be lifted. The injection screen consists of several rows of sleeve pipes, each inserted in the cased drilling filled with coat mixture. With the help of a double-packer, a cement containing suspension (grout) is pumped into the ground through the injection valves (rubber sleeves) (Figure 7). This injection process generates fractures and leads to an increase of stress in the ground (hydraulic fracturing). The cracks are filled with grout material which in turn increases the volume in the ground and lifts the build­ ing. The needed heave can be reached very precisely through multiple injections of small quantities of grout (injection stages). The execution of compensation grouting is tech­ nically challenging and expensive. Significant cost factors include the drilling of the injection screen, the necessary grout volume and execution time. It is par­ ticularly difficult to perform compensation grouting in highly permeable types of gravel because here, in contrast to fine-grained soils, the grout penetrates the large pore volume without any heaving taking place.

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Figure 7. Injection technique.

Figure 10. Predicted settlements and pre-heave. Figure 8. Section through shaft, tunnel and building.

To reduce the costs of compensation grouting in the highly permeable types of gravel as encountered in the Gießenbach valley, the following grouting concept has been implemented: The aim is to avoid an uncontrolled and inefficient penetration of the subsoil, by using a grout mix with adopted grain size and high viscosity and yield limit. This also avoids the need for additional injection screens for permeation grouting before the compensa­ tion grouting. The injection screen was arranged in the middle between the tunnel roof and the foundations of the buildings (Figure 8).

The compensation grouting before and after each excavation step had to be determined by an iterative simulation process, checking the tilt between foun­ dations in each step and ensuring that the limit values were not reached or exceeded at any time. (Figure 10) depicts the maximum settlements pre­ dicted for the top heading without any compensation grouting and the planed pre-heave executed prior to tunnelling. In addition a 3D model has been set up to verify the assumptions taken in the 2D model. As the 3D model did not account for the possible ground loss due to the spiles, the max credible settlements were still taken from the lower bound 2D analysis. 5 EXECUTION OF COMPENSATION GROUTING

4 FEM ANALYSIS For the prediction of the settlements, a 2D FEM ana­ lysis has been set up using a model with hardening soil material parameters. The 3D arching effect in the longitudinal direction of the tunnel has been modelled with a pre-relaxation factor of α=0.6 on the finite elements to be excavated in the next step as an upper bound value (min. deformations) as well as a lower bound value of α =0.3 (max. credible settle­ ments). (Figure 9) presents the FE model used for the simulation. The area of the compensation grout­ ing has been modelled as a 1 m thick soil layer with thermal properties to model the volume increase by the grout material by applying heat to individual sections.

(Figure 11) illustrates the location of the injection screen below the industrial building. The injection screens were arranged in a fan shape, with shafts up to 12 m in depth and 6 m in diameter. In total 7,731 m of drillings with lengths of up to 53.2 m were drilled out of 4 shafts with up to 52 boreholes in each shaft. Due to a large number of boreholes below the buildings, the settlements caused by the drilling pro­ cess was to be reduced to a minimum. Therefore, an overburden drilling method was applied. In this

Figure 11. Overview of the injection screen below the industrial building.

Figure 9. FE model.

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method, the outer rods (external piping) and the inner rods were drilled simultaneously. A down-the­ hole hammer with air flushing was used to drill through anticipated obstacles such as blocks and boulders. The debris from the drilling was flushed out between the external piping and the inner rods and extracted by a preventer connected to a dust extraction system. The drillings were performed using a shaft drilling rig specifically designed for this project. The double-head drilling system has a maximum torque of 28,000 Nm, with feed and retraction forces of up to 100 kN. The drilling rig was equipped with a jib crane with a magnetic grip­ per, two external hydraulic units and a dust extraction system (Figure 12). The drilling and installation of the sleeve pipes were performed in a back step method. After reach­ ing the final depth, the inner rods were removed, and the deviation of the drilling was measured. Then the sleeve pipes (2'' steel pipes, sleeves spacing at 0.50 m) were installed. While dismantling the exter­ nal piping, the annular space between the sleeve pipe and subsoil was successively filled with a grout mix of low compressive strength. The injection processes were computercontrolled. The specified flow rate was controlled automatically, and the injection process stopped when the defined grout volume or maximum pres­ sure was reached. During the injection process, the flow rates were kept constant with flow rates between 5 to 10 l/min. The grout volume varied between 30 and 60 l. The properties of the selected grout mix are listed in Table 2.

Table 2. w/z ratio [-] 0,7

Properties of the grout mix.

Density

Marshtime

Flowlimit

28d­ Compressive strength

[kg/l] 1,60

[s] 42

[N/m²] 54

[N/mm²] 2

The grout was easily pumpable and its low com­ pressive strength of approx. 2 N/mm² after 28 days allowed repeated injection of the same sleeve over an extended period. The following quantities resulted for the compen­ sation grouting of an area of approx. 7,700 m²: • 1,650,000 l of grout mix • 36,500 nos. of injections • 6,500 h of injection time The pre-heave grouting was executed between June 2016 and February 2017 and the interimheave grouting between August 2017 and Octo­ ber 2017. The compensation grouting was con­ ducted on a total of 228 days, with a total grout volume of approx. 1,650,000 l. Further compen­ sation grouting after the bench/invert excavation was not required. The compensation grouting was completed in October 2017. The grouting works were executed with two grout­ ing containers each with six plunger pumps and a batching plant consisting of two mixer and agitators and two cement silos. Due to the very cold winter months in the foothills of the Alps, the batching plant, as well as the water and suspension lines, were enclosed. 6 EXPERIENCE FROM EXECUTION OF THE WORKS 6.1

Figure 12. Shaft drilling rig connected to a dust extraction line.

Efficiency of the compensation grouting

Efficiency is defined by the ratio of the heave volume to the quantity of grout volume. The first heaves were achieved once a total quantity of approx. 99 l/m² of grout had been injected. As expected the efficiency for the first grouting stages was quite low with an efficiency value of approx. 1%. As the number of injection stages increased, the injection pressures increased, caus­ ing efficiency to improve substantially. After just the second injection stage, a significantly higher efficiency was achieved in the middle of the injection screen than at the peripheral areas. As the number of injections stages increased, this dif­ ference becomes more and more apparent. Lower quantities of grout were needed for areas in the middle of the screen, as the peripheral areas made it more difficult for the grout in the middle to flow off.

808

Evaluations of the heave efficiency show that overall efficiency of approx. 7 % was achieved in the peripheral areas and approx. 15 % in the middle of the injection screen. While in the last injection stages the efficiency in the middle of the screen was approx. 30% to 50%, the maximum efficiency in the peripheral areas of the screen was only approx. 10% to 20%. (Fillibeck, et al., 2019). 6.2

Drilling deviations

The measurement of the alignment of the borehole with relation to positional and directional accuracy was carried out within the outer casing with a borehole probe, consisting of an inclinometer and a fibre-optic gyroscope. The following factors had an impact on the drilling deviation: a) Deviations at the drilling point b) Deviations due to the alignment of the drill rig c) Subsoil (e.g. changes in geological strata, drilling obstacles, blocks and boulders) d) Deformation of the drill bit e) Type of drilling tools f) Feed force g) Rotation speed of the drill bit h) Flow rate of bore flushing

Figure 13. Evaluation of drilling deviations as a function of drilling length.

Factors (a) and (b) are caused in part by equipment and were reduced to a minimum by precise alignment checks. It was possible to sub­ stantially reduce the deviation of the horizontal drillings, in particular by choice of drilling tools (e) as well as by the adjustment of the feed force (f) and the flow rate of bore flushing (h) in accordance with the encountered subsoil con­ ditions (c). An evaluation of the different drill­ ing systems used to produce the injection shield showed that the overburden drilling with a down-the-hole hammer and a full face bit achieves more consistent results with fewer drill­ ing deviations than with an eccentric drill bit Fillibeck, et al., 2019). (Figure 13) and Table 3 show the evaluation of the drilling deviations as a function of the drill length. The use of arithmetic means for the evalu­ ation of drilling deviations reacts very sensitively to individual outliers. So to obtain a better assessment of the mean variation and distribution, drilling devi­ ations were displayed in the form of a box plot dia­ gram. The median (the horizontal line of dashes inside the box) indicates that 50% of the drilling deviations are less than or equal to this value. The upper and lower limits on the box represent the cen­ tral 50 % of the measured drilling deviations. The elongation of antennas (limitation outside the box) indicates the range of drilling deviations for the majority of the drillings. The evaluation of the drilling deviations is listed in Table 3. The evaluation shows that the majority of boreholes could be drilled with a low drilling deviation.

Table 3.

Evaluation of drilling deviations.

Number of boreholes Total length Median of the deviation Median of the deviation Drill lengths ≤ 30 m Median of the deviation Drill lengths >30 colloidal silica for both clean sand and with 10% CaCO3. Clearly, the micro-cement indicates higher UCS values. UCS tests on a pure micro-cement grout with w/c ratio of 1 and with a superplasticiser added to the cement slurry in a concentration of 1.5% by mass of cement provided:

Table 4. UCS results on clean sand and sand with fines injected for three different binders. Binder

Soil type

Age (day)

UCS (kPa)

Colloidal silica

Sand Sand Sand Sand+CaCO3 Sand+CaCO3 Sand+CaCO3 Sand Sand Sand Sand Sand+CaCO3 Sand+CaCO3 Sand+CaCO3 Sand Sand Sand Sand+CaCO3 Sand+CaCO3 Sand+CaCO3

1 7 28 1 7 28 1 7 14 28 1 7 28 3 7 28 3 7 28

57 66 120 92 313 337 772 10200 10170 12170 1190 9660 14370 1440 1180 1100 1760 1400 1780

Micro-cement

Geopolymer

• • • •

Figure 4. UCS tests on sand with fines permeated with col­ loidal silica at different curing times. Pictures of the sam­ ples at failure.

includes some pictures of the specimens at failure (samples at higher curing times displayed surface detachment). From Figure 4 is observeded a more ductile behaviour for 1 day curing ageing, while a brittle/ fragile behaviour is observed for longer curing ages (7 and 28 days). Porcino et al. (2011) observed interparticle bonds at the grain contacts imparting a small cohesion to

at 1 day 2.0 MPa; at 3 days 9.4 MPa; at 7 days 11.5 MPa and at 28 days 15.2 MPa.

During the hydration of cement, calcium hydrox­ ide is released. The cement hydration product has high strength, which increases as it ages, while cal­ cium hydroxide contributes to the pozzolanic reac­ tion. In general, the effect of the soil deteriorates the UCS with respect with the pure grout, except for the 3 days curing where the treated sand provided slightly higher UCS value (8% higher). The curing mechanisms of the geopolymer are given by the reaction between activator (potassium silicate) and the metakaolin (consisting of reactive disordered silicate and aluminate layers). The alka­ linity of the waterglass (high degree of activation with low modulus water glass as used in this study) activates the system and yields, in the end, an alumo­ silicate (an oxidic 3D network) containing aluminate and silicate tetrahedra. Figure 5 and 6 show respectively the relation UCS vs curing time and UCS vs void ratio for col­ loidal silica and geopolymer treated sand and sand + CaCO3. Figure 5 shows that a straightforward behav­ iour is observed for the soils injected with colloidal silica. After 7 days curing for both types of soils, the maximum strength remains constant. Besides, the addition of fines decreases the void ratio, explaining, therefore, the higher UCS values. Regarding the geopolymer for 3, 7 and 28 days the UCS decreases with increasing curing age. No explanation is provided so far. The addition of 10% CaCO3 decreases the void ratio as well, increasing, therefore, the UCS values and a slight drop in strength is observed for 7 days curing age. Figure 6 shows the relation between UCS and void ratio. It is clear for the colloidal silica that the strength increase for the same type of soil is not driven by the void ratio decrease but more likely by

833

Regarding the geopolymer fluid injected in pure sand, the highest UCS value corresponds to the highest void ratio, i.e. 0.876. For 7 and 28 days the void ratios are respectively 0.847 and 0.854, which are very simi­ lar and might explain, within the range of experimental error, the similar UCS values obtained for these curing ages. For the geopolymer injected in sand+10% CaCO3 the trend is similar for 3 and 7 days, however, at 28 days the strength increases again. While for 3 and 28 curing days the void ratios are quite similar, for 7 days the void ratio is higher, which may explain the drop in UCS. These results clearly show the very different impact of the binders used in this research program. 4 CONCLUSIONS

Figure 5. UCS vs curing time for colloidal silica and geo­ polymer treated sand and sand + CaCO3 (void ratio indi­ cated as labels).

A laboratory program has been carried out to under­ stand the impact of fines on clean sand with nonconventional grouts. Colloidal silica is the material with the lowest viscosity among the fluids used, but delivering the lowest unconfined compressive strength (UCS) values over the investigated ageing periods. Detailed information is still missing to com­ pletely comprehend the curing mechanism and action on the sand with and without fines. However, mechanical properties are strongly improved.

ACKNOWLEDGEMENT The authors wish to thank Master Builders Solutions for the permission granted to publish the results of this research.

REFERENCES

Figure 6. UCS vs void ratio for colloidal silica and geopo­ lymer treated sand and sand + CaCO3 (ageing time indi­ cated as labels).

bonding created during the gelling of the fluid. For the pure sand injected with the colloidal silica there is a slight increase in void ratio which seems not to affect the increase in strength.

Arroyo, M., Ciantia, M., Castellanza, R., Gens, A., & Nova, R. (2012). Simulation of cement-improved clay structures with a bonded elasto-plastic model: A practical approach. Computers and Geotechnics 45: 140–150. ASTM D2166/D2166M-16 2016. Standard test method for unconfined compressive strength of cohesive soil, ASTM International, West Conshohocken, PA. Beyer, W. 1964. Zur Bestimmung der Wasserdurchlässig­ keit von Kiesen und Sanden aus der Kornverteilungs­ kurve. Wasserwirtschaft und Wassertechnik 14(6): 165–168. Bolisetti, T., Reitsma, S. & Balachandar, R. 2009. Experi­ mental investigations of colloidal silica grouting in porous media. Journal of Geotechnical and Geoenviron­ mental Engineering 135(5): 697–700. BS EN 12715 2000. Execution of special geotechnical work: Grouting. London: BSI. Cambefort, H. 1977. The principles and applications of grouting. Quarterly Journal of Engineering Geology and Hydrogeology 10(2): 57–95. Colombo, P. & Colleselli F. 1996. Elementi di Geotecnica. Bologna: Zanichelli.

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Consoli, N. C., Foppa, D., Festugato, L., & Heineck, K. S. (2007). Key parameters for strength control of artifi­ cially cemented soils. Journal of Geotechnical and Geoenvironmental Engineering 133(2): 197–205. Dano C., Hilcher, P.Y. & Tailliez, S. 2004. Engineering properties of grouted sands. Journal of Geotechnical and Geoenvironmental Engineering 130(3): 328–338. De Paoli, B., Bosco, B., Granata, R. & Bruce, D.A. 1992. Fundamental observations on cement based grouts (2): Microfine cements and the CemillR process. In Proceed­ ings of the Conference on Grouting, Soil Improvement and Geosynthetics, New Orleans, 25-28 February, 1992. Geotechnical Special Publication 30. New York: Ameri­ can Society of Civil Engineers. Vol. 1, 486–499. Dickes, G. 2012. Support of tunneling with chemical grout­ ing in New Jersey. In North American Society for Tren­ chless Technology (NASTT), Nashville 11­ 15 March 2012. Federal Highway Administration 1977. Chemical Grouts for Soils. Report no. FHWA-RD-77-51, Washington, USA. Gallagher, P.M. & Mitchell, J.K. 2002. Influence of col­ loidal silica grout on liquefaction potential and cyclic undrained behavior of loose sand. Soil Dynamics and Earthquake Engineering 22: 1017–1026. Ganeshan, V., Chun Nam, O.W., Marotta, M., Yoshimatsu, A.Y.A., Yi Yng, J.E.E. 2008. Grouting and its application in Tunneling. Proc. of International Con­ ference on Deep Excavation, 10-12 November 2008, Singapore. Karol, R.H. 2013. Chemical Grouting and Soil Stabiliza­ tion, Revised and Expanded. Boca Raton: CRC Press. Littlejohn, G.S. 1993. Underpinning by chemical grouting. In S. Thorburn & G.S: Littlejohn (eds), Underpinning and Retention: 242–275. New York: Springer. Mollamahmutoglu, M. & Avci, E. 2015. Ultrafine Portland cement grouting performance with or without additives. KSCE Journal of Civil Engineering 19(7):2041–2050. Naudts, A., & Naudts, W. 2008. Pre-conditioning/pre­ excavation grouting of the soils prior to tunnelling below highways & railways: case histories from the past

decade. North American Tunnelling Conference, San Francisco, June 7-11, 2008. Porcino, D., Marcianó, V. & Granata, R. 2011. Undrained cyclic response of a silicate-grouted sand for liquefac­ tion mitigation purposes. Geomechanics and Geoengi­ neering 6(3): 155–170, DOI: 10.1080/ 17486025.2011.560287. Romero, E. & Simms P.H. 2008. Microstructure investiga­ tion in unsaturated soils: a review with special attention to contribution of mercury intrusion porosimetry and environmental scanning electron microscopy. Geotech­ nical and Geological Engineering 26(6): 705–727. Salvatore, E., Modoni, G., Mascolo, M.C., Grassi, D. & Spagnoli, G. 2020. Experimental Evidence of the Effect­ iveness and Applicability of Colloidal Nanosilica Grouting for Liquefaction Mitigation. Journal of Geotechnical and Geoenvironmental Engineering 146(10): 04020108. Semprich, S. & Stadler, G. 2002. Grouting in geotechnical engineering. In U. Smoltczyk (ed), Geotechnical Engin­ eering Handbook-volume 2: 57–78. Berlin: Ernst & Sohn. Sichardt, W. 1927. Die Grundlagen der Theorie des Grund­ wasserabsenkungsverfahrens und die Entwicklungsaus­ sichten des Verfahrens. In Das Fassungsvermögen von Rohrbrunnen und seine Bedeutung für die Grundwasser­ absenkung, insbesondere für größere Absenkungstiefen. Berlin, Heidelberg: Springer. Spagnoli, G. 2018. A review of soil improvement with non-conventional grouts. International Journal of Geotechnical Engineering, https://doi.org/10.1080/ 19386362.2018.1484603. Stadler, G. & Krenn, H. 2013. Permeation grouting. In K. Kirsch & A. Bell (eds), Ground Improvement: 169–206. Boca Raton: CRC Press. Zebovltz, S., Krizek, R.J. & Atmatzidis, D.K. 1989. Injec­ tion of fine sands with very fine cement grout. Journal of Geotechnical Engineering 115(12): 1717–1733. Zhang, M., Guo, H., El-Korchi, T. Zhang, G. & Tao, M. 2013. Experimental feasibility study of geopolymer as the next-generation soil stabilizer. Construction and Building Materials 47:1468–1478.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Permeability characteristics of coarse-grained soil conditioned with foam for EPB shield tunnelling S. Wang, S. Huang, Q. Hu & Z. Liu School of Civil Engineering, Central South University, Changsha, People’s Republic of China

ABSTRACT: Foam is often used to reduce the permeability coefficient of coarse-grained soil to avoid water spewing during earth pressure balance (EPB) shield tunnelling. A series of permeability tests were carried out on foam-conditioned soil. The test results show that with an increase in d10, the initial permeability coefficient of the conditioned soil increased greatly, however, it only changed slightly with increases in Cc and Cu. In addition, permeability coefficient was much lower than that of the unconditioned soil, even though the slow growth period had been entered for a long time. The results also indicate that d10 greatly affected the perme­ ability characteristics of conditioned soil. By contrast, Cc and Cu had less effects. In addition, it is not reason­ able to consider that the permeability coefficient of the conditioned soil will reach or be close to that of unconditioned soil when most, or all, of foam in the foam-conditioned soil dissipates.

1 INSTRUCTIONS The soil in the chamber and screw conveyor is required to have a low permeability coefficient for EPB shield tunnelling below the water table, otherwise, water spewing can happen easily, thereby inducing instability of the excavation face (Psomas, 2001; Borio and Peila, 2010). An effective measure to solve the problem is soil conditioning, which uses conditioning agents to make the soil pulpy. This pulpy soil has suitable fluid­ ity, low permeability, suitable compressibility, low adhesion and low internal friction angle (Budach and Thewes, 2015; Mori et al., 2017). The foam is widely used for soil conditioning because of its low cost and simple preparation process during EPB shield tunnelling in water-rich sandy ground. Some researchers have studied the permeability characteristics of foam-conditioned sandy soil. Bezui­ jen et al. (1999) showed that the permeability coeffi­ cient of the conditioned soil is closely related to the fact that foam fills in the soil pores. The better the foam fills in the pores, the lower the permeability coef­ ficient of foam-conditioned soil is. Quebaud et al. (1998) found that, as the foam injection ratio (FIR, which is defined as the ratio of volume of foam to that of soil) increased, the permeability coefficient of foamconditioned soil gradually decreased. However, the coefficient remained almost unchanged or even tended to grow when the ratio was too high. Both of too coarse or too fine soils cannot be conditioned well by foam. Only when the particle size of soil is moderate, the permeability coefficient can be strongly reduced after being conditioned with foam. Borio and Peila

(2010) and Peila (2014) pointed out that the flow quan­ tity can better characterize the permeability of foamconditioned soil than the permeability coefficient for EPB shield tunnelling. The results showed that the finer the particle size of the soil is, the lower the per­ meability of the conditioned soil is. Budach and Thewes (2015) carried out permeability tests on nine types of soils and found that the foam clearly reduced the permeability coefficient of sandy soil over those of silt and gravel. Kim et al. (2019) found that the add­ ition of foam can greatly reduce the permeability coef­ ficient of soil, and the reduction extent can be up to approximately 3 orders of magnitude. Huang et al. (2019) revealed the effect of grain gradation on perme­ ability characteristics of coarse-grained soil condi­ tioned with foam. In summary, there are still limited research findings on permeability characteristics of foam-conditioned soil, especially considering the effect of grain grad­ ation. In this study, a series of large-scale permeability tests were carried out, and then the permeability char­ acteristics of foam-conditioned, coarse-grained soil were investigated under the influence of grain gradation. 2 TESTING PROGRAM 2.1

Testing materials

Natural sand was collected and sieved to prepare the test materials. Three groups of grain size distributions were designed with variations in d10, Cc, and Cu in

DOI: 10.1201/9780429321559-110

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Figure 2. Photo of the Self-designed Permeameter.

Figure 1. Grain Size Distributions of the Test Soils: (a) with Constant Cc (=1.5) and Cu (=10) and Various Values of d10, (b) with Constant d10 (=0.23) and Cu (=10) and Various Values of Cc, (c) with Constant d10 (=0.23) and Cc (=1.5) and Various Values of Cu.

Figure 1. An anion and cation compound foam agent with a concentration of 3% was used as the raw mater­ ial of the conditioning agent, and foam was generated using a foaming system that followed the requirements of the EFNARC (2005). The foam expansion ratio (FER) was approximately 10, and the foam had a halflife duration (defined as the time required for the foam to dissipate for half of its mass) of 300 s.

the specimen is under 37.5 mm, the effect of the per­ meameter boundary can be ignored. Thus, the par­ ticle sizes of all of the specimens were designed to meet this requirement. This study does not examine the effect of conditioning parameters on the perme­ ability of coarse-grained soils, so, the FIR and water content (ω) for preparing the soil specimens for all permeability tests were set to 20% and 7.5%, respectively. The permeability coefficient was deter­ mined at a time spacing of 5 min. The water flowing out of the permeameter bottom was collected within the time of 5 min, and the specimen height at differ­ ent time was also recorded using a ruler on the per­ meameter wall. According to the Darcy’s law, the permeability coefficient of the foam-conditioned soil at different time was calculated using the recorded water volume and specimen height. When the per­ meability coefficient stopped significantly changing with time, the permeability test was halted.

2.2 Testing approach The constant head permeability tests were carried out with a self-designed, large-scale permeameter whose diameter and height were 300 mm and 750 mm, respectively, as shown in Figure 2. Follow­ ing the ASTM (2006), soil having maximum particle size less than 1/8 of the diameter of the permeameter (37.5 mm for this permeameter) can be tested using this permeameter. If the maximum particle size of

3 TESTING RESULTS 3.1 Permeability characterisitcs of foam-conditioned soil Figure 3 shows the time-varying curves of perme­ ability coefficients (k) of the foam-conditioned soils under the variation of characteristic gradation

837

3.2

Effect of grain gradation on initial permeability coefficient of the foam-conditioned soil

The minimum value of the permeability coefficient in the initial stable period was defined as the initial permeability coefficient (ki), and the changes in the initial permeability coefficients against grain grad­ ation parameters are shown in Figure 4. (1) Effect of d10 on initial permeability coefficient Figure 4(a) shows the change in the initial perme­ ability coefficients versus d10. As the d10 increased, the ki increased greatly. For example, as the d10 increased from 0.13 mm to 0.50 mm, the ki of the foam-conditioned soil increased from 5.70×10-8 m/s to 1.21×10-6 m/s, increasing by two orders of magni­ tude. The curve in the Figure 4(a) is “step-like”, indicating that the growth rate of ki first increased

Figure 3. Changes in permeability coefficients vs. time for the foam-conditioned, coarse-grained soils: (a) with con­ stant Cc (=1.5) and Cu (=10) and various values of d10, (b) with constant d10 (=0.23) and Cu (=10) and various values of Cc, (c) with constant d10 (=0.23) and Cc (=1.5) and vari­ ous values of Cu.

parameters. It can be seen that all of the timevarying curves for the permeability coefficients generally experienced an initial stable period, a fast growth period and a slow growth period. The foam structure in the soil became relatively stable, and the curve entered the initial stable period, during which the change of permeability coefficient was very small. Moreover, because of different grain gradations, there was a certain dif­ ference in the duration of the initial stable period. After this period, the fast growth period occurred (see Figure 3) due to the dissipation and/or mer­ ging of foam in the soil. Finally, as most of the foam disappeared, the permeation structure of the conditioned soil tended to be stable, and the curve entered into the slow growth period, during which the permeability coefficient of the conditioned soil slowly increased or remained almost unchanged.

Figure 4. Changes in Initial Permeability Coefficients vs. Grain Gradation Parameters for the Foam-conditioned, Coarse-grained Soils: (a) with Constant Cc (=1.5) and Cu (=10) and Various Values of d10, (b) with Constant d10 (=0.23) and Cu (=10) and Various Values of Cc, (c) with Con­ stant d10 (=0.23) and Cc (=1.5) and Various Values of Cu.

838

and then slowed down and later increased again as d10 increased. The reason for this trend is that when the particle size of the soil is small, the soil particles can easily work with the foam to form a good waterblocking structure. In other words, the injection of the foam is equivalent to increasing the content of the fine particles in the soil. Nagy (2011) showed that more fine particles fill the pores between the coarse particles, causing the coarse particles to stay in a “suspended” state in the soil. As a result, the soil structure is denser, and the permeability coeffi­ cient of the soil is reduced. Thus, in this study, as the particles of conditioned soil became coarser, the number of fine particles was relatively reduced, causing the initial permeability coefficient and its growth rate to increase. The soil particles continued to become coarser, and the pores between the par­ ticles became larger. According to the model of Hazen (1892), the formula for the permeability coef­ ficient of unconditioned sandy soil is as follows:

where Ce is a parameter related to the pore char­ acteristics of the soil. From the above formula, it can be seen that when the d10 is moderately large, the permeability coefficient of the soil part itself (not including foam) of the foamconditioned soil is also large. Thus, it can be thought that the soil part itself in the conditioned soil has less effect on water blocking in the overall conditioned soil. Therefore, the water-blocking ability of the conditioned soil mainly depends on foam filling in the pores of the soil. The coarser the soil particles are, the weaker the ability of the soil itself to block water is and the larger role the foam will play in water-blocking. Thus, as the particle diameter increases, the initial permeability coefficient of the whole conditioned soil tends to approach the permeability coefficient of the foam itself. As a result, there occurred a period in which the growth rate of the permeability coefficient was gentle. After that, when the d10 increased further, the coarse particles were in contact with each other to form a skeleton, and the volume of pores among the soil par­ ticles increased so sharply that the foam could not fill the pores completely, meaning that the soil condition­ ing changed from being sufficient to being insufficient. Furthermore, a large amount of water flowed directly over the permeation passages among the soil particles, causing a sharp increase in the initial permeability coefficient. (2) Effect of Cc on initial permeability coefficient Figure 4(b) shows the change in the initial perme­ ability coefficients versus Cc. As the Cc increased, the ki first increased and then decreased; however, the variation was not large. Specifically, the max­ imum and the minimum ki were 6.11×10-7 m/s and 3.12×10-7 m/s, respectively; so, the maximum value was equal to only 1.96 times of the minimum value.

In the prediction of the permeability coefficient in unconditioned soil, Hazen (1892) and Terzaghi et al. (1964) considered d10 to be the most important factor affecting soil permeability coefficient. This study also verifies that d10 is the determinant of the permeability coefficient of foam-conditioned soil under the influence of grain gradation. When the d10 was held constant, a change in Cc caused a small change in the ki. This is because, once d10 is fixed, the pore size of soil does not vary greatly no matter how the Cc changes. The foam can lift the soil skel­ eton allowing the permeability coefficient to be maintained at a low level and only change in a small range. (3) Effect of Cu on initial permeability coefficient Figure 4(c) shows the change in the initial perme­ ability coefficients versus Cu. With an increase in Cu, the ki of the conditioned soil first remained stable and later gradually decreased. However, similar to the change in Cc, the whole change in ki was small ranging from 6.27×10-7 m/s to 4.30×10-7 m/s (the maximum value of ki was equal to 1.46 times of the minimum value of ki). As the Cu increased, the nonuniformity of the soil increased. To a large degree, the pores between coarse-grained skeletons in the soil were filled by the medium and fine particles. When the foam was added to the soil, it could easily fully fill the pores and lifted the coarse particle skeleton. As a result, the permeation passages among the coarse particles were blocked, con­ sequently reducing the initial permeability coefficient of the conditioned soil. 3.3 Effect of foam addition on the stable permeability coefficient of the coarse-grained soil As shown in Figure 3, the permeability coefficient of each conditioned soil eventually entered the slow growth period in permeation, during which the per­ meability coefficient remained virtually constant or increased very slowly. The stable permeability coef­ ficient in the slow growth period (ks) is determined to be the value at the intersection point, as shown in Figure 5. The ratios of the permeability coefficient

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Figure 5. Schematic Diagram of the Method for Determining the Stable Permeability Coefficient of the Foam-conditioned Soil.

Table 1.

Changes in two different permeability coefficients with grain gradation parameters for different specimens.

Cc = 1.5, Cu = 10 d10 (mm) 0.13 0.18 0.23 0.35 0.50

ku (m/s)

d10 = 0.23 mm, Cu = 10 ks (m/s)

-5

3.71×10 6.22×10-5 1.61×10-4 8.92×10-4 1.20×10-3

ku/ks -5

1.30×10 1.53×10-5 1.65×10-5 2.75×10-5 4.45×10-5

2.85 4.07 9.76 32.44 26.97

Cc 0.7 1.5 2.3 3.0 4.0

d10 = 0.23 mm, Cc = 1.5

ks (m/s)

ku (m/s) -5

7.92×10 1.61×10-4 5.81×10-4 7.91×10-4 1.00×10-3

of the unconditioned soil (ku) to the ks for soils with different grain gradation parameters are presented in Table 1 to investigate the effect of foam addition on the long-term soil permeability coefficient. The ku/ks for all test cases are always larger than 1, indicating a certain gap between the ks and the ku. For example, Figure 6 shows the permeability variation of the soil specimen with d10 = 0.23 mm, Cc = 1.5, and Cu = 15. The permeability test was conducted for approximately two months. When the time was approximately 30000 min (approximately 22 days), the permeability coefficient of the foam-conditioned soil was 3.35×10-5 m/s and still far from that of the uncon­ ditioned soil (6.04×10-5 m/s). After experiencing 78323 min (approximately 54 days), the permeability coefficient of the foam-conditioned soil reached 5.83×10-5 m/s and approached that of the uncondi­ tioned soil. From this result, it can be inferred that the foam added to the soil gradually dissipated in the early stage of permeation so that the permeability coefficient of the conditioned soil rapidly rose to a relatively high level and entered into the slow growth period. There­ after, due to the delayed adjustment of soil microstruc­ ture due to foam dissipation and gradual loss of fine particles in the soil under the action of water, the per­ meability coefficient of the conditioned soil changed slowly. However, since the structure and saturation state of soil were changed due to the addition of foam, there was still a large gap between the permeability coefficient of the conditioned soil and that of the

Figure 6. Permeability Variation of the Soil Specimen with d10=0.23 mm, Cc=1.5, Cu=15.

ku/ks -5

1.06×10 1.65×10-5 2.05×10-5 2.90×10-5 3.78×10-5

7.47 9.76 28.34 27.28 26.46

Cu 3 5 10 15 25

ks (m/s)

ku (m/s) -4

3.16×10 1.81×10-4 1.61×10-4 6.04×10-5 2.64×10-5

ku/ks -5

2.83×10 1.98×10-5 1.50×10-5 1.40×10-5 1.10×10-5

11.17 9.14 9.76 4.31 2.40

unconditioned soil, even after a long time of permeability. Thus, it is not reasonable to consider that the per­ meability coefficient of the conditioned soil will reach or be close to that of unconditioned soil when most, or all, of foam in the foam-conditioned soil dissipates. This point has significance for guiding the post-treatment of foam-conditioned soil, such as for the analysis of the waste muck yard and embank­ ment drainage performance and the selection of their corresponding reinforcement measures. 4 CONCLUSIONS The time-dependent permeability coefficients of coarse-grained soil conditioned with foam gener­ ally experienced an initial stable period, a fast growth period and a slow growth period. With an increase in d10 (constant Cc and Cu), the durations of the initial stable period and the fast growth period of the permeability coefficient became gradually shorter, which was clearly due to the poorer filling conditions among the coarser soil particles. Meanwhile, the growth of the permeabil­ ity coefficient was faster in the rapid growth period, and the permeability coefficient in the slow growth period was larger. With an increase in Cc (constant d10 and Cu), the durations of the initial stable period and the fast growth period of the per­ meability coefficient become gradually shorter due to the fewer medium sized particles and weaker water-blocking structure formed by the foam and the fine particles in the soil. Meanwhile, the growth of the permeability coefficient was faster in the fast growth period, and the permeability coefficient in the slow growth period became larger. With an increase in Cu (constant d10 and Cc), the durations of the initial stable period and the fast growth period of the permeability coeffi­ cient gradually increased due to the better inner filling performance and easier sealing of foam in the pores among the soil particles. Meanwhile, the growth of the permeability coefficient was slower in the fast growth period, and the permeability coefficient in the slow growth period was lower.

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By comparing the effects of d10, Cc and Cu on the foam-conditioned soil, it is shown that the d10 had the most significant influence on the initial perme­ ability coefficient and the safety of permeation of the conditioned soil. The permeability of the foamconditioned soil depends on the filling and stability of foam in the soil pores. When the d10 was constant, the changes in Cc and Cu did not cause large changes in the whole particle size scale, so, both of these parameters did not have much influence on the initial permeability coefficient and the safety of permeation of the conditioned soil. Although the change in the permeability coefficient of foam-conditioned soil entered the slow growth period, the permeability coefficient was far below that of the soil without being conditioned by foam. This result has signifi­ cance for guiding the post-treatment of foamconditioned soil in the waste muck yard and road embankment, etc.

ACKNOWLEDGEMENTS The financial support from National Natural Science Foundation of China (No. 51778637) and National Key R&D Program of China (No. 2017YFB120120 4) is acknowledged and appreciated.

REFERENCES ASTM. 2006. Standard test method for permeability of granular soils (constant head), D2434-68, ASTM Inter­ national, West Conshohocken, PA. Bezuijen, A., Schaminee, P. E. L. and Kleinjan, J. A. 1999. Additive testing for earth pressure balance shields. Proc. 12th Eur. Conf. on Soil Mechanics and Geotechnical Eng., Amsterdam, Netherlands: 1991-1996. Borio, L. and Peila, D. 2010. Study of the permeability of foam conditioned soils with laboratory tests. American Journal of Environmental Sciences, 6(4): 365–370.

Budach, C. 2012. Untersuchungen zum erweiterten Einsatz von Erddruckschilden in grobkörnigem Lockergestein [Investigations for extended use of EPB Shields in coarsegrained soils], Ph.D. Thesis, Ruhr-Universität Bochum, North Rhine-Westphalia, Germany (in German). Budach, C. and Thewes, M. 2015. Application ranges of EPB shields in coarse ground based on laboratory research. Tunnelling and Underground Space Technol­ ogy, 50: 296–304. EFNARC 2005. Specifications and guidelines for the use of specialist products for mechanized tunnelling (TBM) in Soft Ground and Hard Rock. Recommendation. Farn­ ham, UK, http://www.efnarc.org/publications.html. Hazen, A. 1892. Some physical properties of sands and gravels, with special reference to their use in filtration. 24th Annual Report, Massachusetts State Board of Health, Pub.Doc., 34: 539–556. Kim, T. H., Kim, B. K., Lee, K. H. and Lee, I. M. 2019. Soil conditioning of weathered granite soil used for EPB shield TBM: a laboratory scale study. KSCE Journal of Civil Engineering, 23(4): 1829–1838. Mori, L., Alavi, E. and Mooney, M. 2017. Apparent density evaluation methods to assess the effectiveness of soil conditioning. Tunnelling and Underground Space Tech­ nology, 67: 175–186. Nagy, L. 2011. Permeability of well graded soils. Periodica Polytechnica Civil Engineering, 55(2): 199–204. Peila, D. 2014. Soil conditioning for EPB shield tunneling. KSCE Journal of Civil Engineering, 18(3): 831–836. Psomas, S. 2001. Properties of foam/sand mixtures for tun­ neling applications, MSc. Thesis, University of Oxford, Oxford, UK. Quebaud, S., Sibai, M. and Henry, J. P. 1998. Use of chem­ ical foam for improvements in drilling by earth-pressure balanced shields in granular soils. Tunnelling and Underground Space Technology, 13(2): 173–180. Huang, S., Wang, S., Xu, C., Shi, Y., Ye, F., 2019. Effect of gradation on the permeability characteristics of coarse-grained soil conditioned with foam for EPB shield tunneling. KSCE Journal of Civil Engineering, 23 (11): 4662–4674. Terzaghi, K., Peck, R.B. and Mesri, G. 1964. Soil mechan­ ics in engineering practice, John Wiley & Sons, New York, USA.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Stabilisation of Singapore soft marine clay using a novel sustainable binder for underground construction H. Yu & Y. Yi School of Civil and Environmental Engineering, Nanyang Technological University, Singapore

R. Liu School of Civil Engineering, Tianjin University, Tianjin, China

N. Jiang Department of Civil and Environmental Engineering, University of Hawai’i at Mānoa, Honolulu, USA

ABSTRACT: There are many underground construction projects, including tunnelling and deep excava­ tions, in Singapore. However, the underground construction is challenging when encountering Singapore soft marine clay due to its poor engineering properties, such as low shear strength and stiffness. Cement stabilisation through deep mixing or jet grouting has been commonly used to treat the soft marine clay for underground construction in Singapore, which can significantly increase the shear strength and elastic modulus, as well as decrease the permeability of soft clay. However, the process of manufacturing trad­ itional Portland cement (PC) leads to negative environmental impacts, e.g. high CO2 emissions and energy consumptions. In addition, the PC is not very effective for stabilisation of marine clay with high water con­ tent. Hence, this study proposes to use a novel sustainable binder consisting of two industry by-products, namely carbide slag (CS) and ground granulated blast furnace slag (GGBS), to stabilise Singapore soft marine clay for underground construction. By replacing PC with industry by-products, the associated envir­ onmental impacts are significantly reduced. Specimens of soft clay stabilised by both CS-GGBS and PC with two binder contents were prepared in laboratory and then tested at different curing ages. The key engineering properties related to underground construction, including unconfined compressive strength, elastic modulus and permeability of the stabilised clays with different binders were compared. The results indicate that the CS-GGBS-stabilised soft clay can achieve much higher unconfined compressive strength (up to ~300%) and elastic modulus (up to ~600%), and lower permeability (one order of magnitude) than the corresponding PC-stabilised soft clay, which are beneficial to the soft stabilisation for underground construction.

1

INTRODUCTION

The Singapore marine clay is a unit of the Kallang for­ mation that is a deposit mainly consisting of soil of marine and covers approximately 25% of Singapore Island (PWD 1976). With rapid development in Singa­ pore in recent decades, many underground construc­ tions, such as deep excavations and underground rail systems, have been in areas of soft marine clay (Shirlaw et al. 2006). Soft marine clays typically have poor engineering properties, such as low shear strength, low stiffness and high compressibility, which can induce low stability and high (vertical and/or horizontal) deformation during underground construction activ­ ities. To improve the geotechnical properties of soft marine clay, stabilisation with Portland cement (PC) is widely used in many underground construction works, including tunnelling and deep excavations (Nakagawa

et al. 1996; Koshima & Guatteri 2006). PC stabilisa­ tion, through the deep mixing or jet grouting, is also commonly used in Singapore to enhance the stability and reduce the movement associated with tunnelling and excavation in clays (Shirlaw 2012). For these applications, the key engineering properties of stabil­ ised clays mainly include unconfined compressive strength, elastic modulus, and permeability (Holt & Griffiths 1992; Tan et al. 2002). However, from the perspective of environmental impact, the major issues associated with PC manu­ facture include high CO2 emissions (~0.95 ton CO2 /ton PC) and energy consumptions (~5000 MJ/ton PC). By comparison, ground granulated blast-furnace slag (GGBS), an industrial by-product, shows signifi­ cant advantages in environmental impacts including low CO2 emissions (~0.07 ton CO2/ton GGBS) and energy consumptions (~1300 MJ/ton PC) (Higgins

DOI: 10.1201/9780429321559-111

842

2007). Due to the low rate of hydration and strength development of GGBS, alkaline activators are used to accelerate the hydration (Shi et al. 2006). Lime was generally used to activate GGBS in soil stabil­ isation (Nidzam & Kinuthia 2010) and, compared to conventional PC, lime-GGBS showed significant advantages in improving the strength of clays (James et al. 2008; Yi et al. 2014; Yi et al. 2015). Carbide slag (CS) is an industrial by-product, majorly com­ posed of Ca(OH)2 (85%–95%) (Cardoso et al. 2009), and it is possible that the CS can replace lime to acti­ vate GGBS for stabilisation of Singapore soft marine clay for underground construction. The major objective of this study is to investigate the engineering properties of CS-GGBS-stabilised Singapore soft marine clay, including water content, unconfined compressive strength, elastic modulus, and permeability, for underground constructions. The effects of CS/GGBS ratio, binder/dry soil ratio, and curing time on the properties of CS-GGBS­ stabilised clays are also studied. 2 MATERIALS AND LABORATORY EXPERIMENTS 2.1

Materials and mixture proportions

A Singapore soft marine clay, with an in-situ water content of 50-69%, plastic limit of 25.2% and liquid limit of 58.9%, was used in this study. The mineral­ ogical composition of Singapore clay mainly includes kaolinite, smectite, and mica (Bo et al. 2015). Refer­ ring to the in-situ water contents, a water content of 60% was used for the clay for the specimen prepar­ ation in the laboratory. GGBS and PC were purchased from the Engro Co. Ltd., Singapore. CS was collected from the WKS Industrial Gas Pte. Ltd., Singapore. The chemical composition of all raw materials was tested by X-ray fluorescence as shown in Table 1. In this study, binders refer to PC and CS-GGBS. Considering the commonly used binder content for underground constructions in Singapore, two binder/ dry clay mass ratios (binder content) of 0.2 and 0.3 were used. For the CS-GGBS binder, four CS/GGBS mass ratios of 0.05/0.95, 0.1/0.9, 0.15/0.85 and 0.2/ 0.8 were used according to Yi et al. (2014, 2015). Since the wet deep mixing is widely used in Singa­ pore, a water/binder mass ratio of 1.0 was used to calculate the additional water for producing the binder slurry.

Table 1.

2.2

Experimental procedure

The air-dried clay was ground into small particles using a mechanical grinder and sieved with a sieve of an aperture size of 2.36 mm. The sieved clay was then oven-dried at a temperature of 110°C for 24 hours to ensure the dry state. The CS was also oven-dried at 110°C for 24 hours and then sieved with a sieve of an aperture size of 1.18 mm. The dry clay and binder, weighted based on the test program, were mixed and homogenised for five minutes in a laboratory mixer with rpm of 60. The weighted tap water was added and mixed for another five minutes to form a slurry-form mixture. The prepared mixture was then filled in cylindrical moulds with 50 mm of diameter and 100 mm of height in three layers. The moulds were filled with the mixture up to one-third of the volume of the mould at a time, and after each addition of the mixture, the mould was struck to the laboratory table 50 times to remove air bubbles and compact the mixture (Kitazume & Terashi 2013). The moulds filled with mixture were placed in air-tight plastic bags each before being stored in a moisture room with the constant relative humidity of 95% ± 3% and the temperature of 23 °C ± 2 °C. The three different curing periods for the stabilised clay specimens were 7, 28 and 56 days. For each curing period and mixture proportion, three repli­ cate specimens were prepared. After reaching each curing period, stabilised clay specimens were demoulded for testing. The specimens were weighed on a laboratory scale and measured for diameters and heights using an electronic caliper. The unconfined compressive strength (UCS) tests were conducted with the con­ stant displacement rate of 1 mm/min until failure. The stress-strain behaviour was obtained during the UCS test for calculating elastic modulus (E50) and axial strain at failure (εf). After finishing UCS tests, the crushed samples were hammered to finer particles and sieved through the sieve of 2.36 mm for water content tests. Considering the hydration of binder in the stabilised clay samples at early age, the sieved clay particles were immersed in ethanol for 24 hours before transfer­ ring to the oven at 60 °C. Samples were initially dried for one week and weighted. Additional measurements (at least two) were conducted for each period of one week to ensure that the sam­ ples reached the constant mass. For samples cured

Chemical composition (by % weight) of raw materials.

Material

CaO

SiO2

Al2O3

Fe2O3

SO3

MgO

K2O

TiO2

MnO

Na2O

LOI*

PC GGBS CS

62.13 43.24 71.83

20.77 28.41 4.88

3.66 13.84 2.48

2.93 0.49 0.14

1.76 4.25 0.18

4.11 7.05 0.22

0.35 0.42 -

0.69 1.40 0.05

0.06 0.34 -

0.37 0.35 -

3.16 0.24 20.18

* Loss on ignition

843

for 56 days, the permeability (or hydraulic con­ ductivity) was tested in duplicate using a flexiblewalled permeameter according to ASTM D5084 (ASTM 2016), with a confining pressure of 250 kPa and constant back pressure of 200 kPa.

comparison, the stabilised clays with CS/GGBS = 0.15/0.85 have the maximum water contents as shown in Figure 1b. 3.2

Unconfined compressive strength

Figure 1 shows the water content of PC- and CS­ GGBS-stabilised clays. The water content decreases with increase of curing time due to the hydration of binder. The slope of water content-curing time curve generally decreases as curing time increases, which can be contributed to the hydration reaction mainly occurred at the early age. The water content of CS-GGBS-stabilised clays is slightly higher (< 7%) than that of PC-stabilised clays at each curing time, indicating a slower hydration rate of CS­ GGBS than PC. For stabilised clays cured for 7 days, the difference in water content is limited for various CS/GGBS ratios. For stabilised clays cured for 28 and 56 days, the stabilised clays with CS/GGBS = 0.1/0.9 have the maximum water contents as shown in Figure 1a. By

Figure 2 shows the relationship between UCS and curing time of CS-GGBS- and PC-stabilised clays. Generally, for stabilised clay with a fixed binder type and binder/dry soil ratio, the UCS increases with increase of curing time. The UCS of CS-GGBS sta­ bilised-clays is higher than that of PC-stabilised clays with the same binder content and curing age. A possible reason is that a relatively higher amount of hydration products (e.g. CSH) is formed to fill pores in the CS-GGBS-stabilised clay, which is bene­ ficial on increasing the strength by filling in pores and decreasing porosity of stabilised clay. Yi et al. (2015) confirmed that CS-GGBS-stabilised clays had lower porosities than PC-stabilised clays. It is found that, compared to other stabilised clays, the UCS of the stabilised clay with CS/GGBS = 0.05/0.95 increases rapidly and reaches the highest values at 56 days. Higher UCS of stabilised clay is desired for may underground constructions to increase stability. As shown in Figure 2a, for stabilised clays cured for 7 days, the UCS of stabilised clay with CS/ GGBS = 0.05/0.95 is close to that of PC stabilised

Figure 1. Water content of CS-GGBS- and PC-stabilised clays with (a) binder/dry clay = 0.2, (b) binder/dry clay = 0.3.

Figure 2. UCS of CS-GGBS- and PC-stabilised clays with (a) binder/dry clay = 0.2, (b) binder/dry clay = 0.3.

3 EXPERIMENTAL RESULTS AND DISCUSSION 3.1

Water content

844

clay. As CS/GGBS ratio increases from 0.1/0.9 to 0.2/0.8, the change in UCS of stabilised clays is limited. As curing time increases to 28 days, the highest value of UCS is achieved by the stabilised clay with CS/GGBS = 0.1/0.9. For stabilised clays cured for 56 days, the UCS of stabilised clays decreases with increase of CS/GGBS ratio. For stabilised clays with binder/dry clay = 0.3, as shown in Figure 2b, the effect of CS/GGBS ratio on UCS of stabilised clays is limited for stabilised clays cured for 7 days. By comparing stabilised clays cured for 28 and 56 days, a similar trend is observed, which is that the UCS decreases with increase of CS/GGBS ratio. 3.3

Elastic modulus (E50)

for 56 days. Higher E50 is beneficial for underground constructions, which can reduce the deformation asso­ ciated with underground constructions. The E50 is usually correlated to UCS in previous studies (Porbaha et al. 2000; Lorenzo & Bergado 2006; Kitazume & Terashi 2013). Therefore, E50 of CS-GGBS- and PC-stabilised clays are also plotted against the corresponding UCS in this study as shown in Figure 4. Generally, the E50 increases with increase of UCS. The upper and lower limits of the E50-UCS data can be bounded by using the follow­ ing linear equation:

where λ is a coefficient. For the stabilised Singapore soft marine clays in this study, λ ranges from 50 to 200 as shown in Figure 4, which is consistent with previous results on PC-stabilised clays (Tsuchida et al. 2001; Kita­ zume & Terashi 2013). The general trend between E50 and UCS of CS-GGBS-stabilised clays is similar to that of PC-stabilised clays.

The E50 is defined as the secant modulus at 50% of failure stress in the stress-strain curve (Kitazume & Terashi 2013). The E50 of CS-GGBS- and PC­ stabilised Singapore soft marine clays is plotted against the curing time as shown in Figure 3. Generally, the E50–curing time cures show a similar trend with that of UCS-curing time. The E50 of CS-GGBS-stabilised clays is higher than that of PC-stabilised clays with the same binder content and curing time. The maximum E50 is obtained by the CS-GGBS-stabilised clay with CS/GGBS = 0.05/0.95 and binder/dry clay = 0.3 cured

3.4

Axial strain at failure

The relationship between εf and UCS was also com­ monly investigated for stabilised clays (Kitazume & Terashi, 2013). In this study, εf of CS-GGBS- and PC­ stabilised clays is plotted against the UCS as shown in Figure 5. A general trend of decreasing εf with increase UCS in stabilised clays is observed. The εf of stabilised clays ranges from 0.4% to 2%. The similar results were also reported in previous studies (e.g. Jegandan et al. 2010; Kitazume & Terashi, 2013). 3.5

Permeability

Table 2 summarizes the permeability results for PCand CS-GGBS-stabilised clays cured for 56 days. As shown in Table 2, the permeability of CS-GGBS-

Figure 3. Elastic modulus (E50) of CS-GGBS- and PC-stabilised clays with (a) binder/dry clay = 0.2, (b) binder/dry clay = 0.3.

Figure 4. Elastic modulus (E50)-UCS of CS-GGBS- and PC-stabilised clays.

845

Figure 5. Axial strain at failure-UCS of CS-GGBS- and PC-stabilised clays.

Table 2. Permeability of stabilised clays cured for 56 days. Binder/dry

clay 0.2

0.3

Binder

Permeability (m/s)

PC CS/GGBS = 0.05/0.95 CS/GGBS = 0.1/0.9 CS/GGBS = 0.15/0.85 CS/GGBS = 0.2/0.8 PC CS/GGBS = 0.05/0.95 CS/GGBS = 0.1/0.9 CS/GGBS = 0.15/0.85 CS/GGBS = 0.2/0.8

2.28 × 10-1° 2.14 × 10-11 2.15 × 10-11 1.93 × 10-11 1.95 × 10-11 1.57 × 10-1° < 1 × 10-12* 1.71 × 10-11 1.84 × 10-11 1.35 × 10-11

when the curing time is more than 28 days. Higher UCS of stabilised clay can increase the stability of underground constructions. 2. The elastic modulus (E50) ranges from 50 to 200 times of UCS (i.e. E50=50-200UCS). The general trend of E50-UCS of CS-GGBS-stabilised clays is similar to that of PC-stabilised clays. Higher E50 can reduce the deformation associated with underground constructions. 3. The permeability of CS-GGBS-stabilised clays was more than one order of magnitude lower than that of PC-stabilised clays. A lower perme­ ability can reduce the seepage-induced risk during underground constructions. 4. This study indicates that replacing PC with CS­ GGBS for stabilisation of Singapore soft marine clay, not only significantly reduces the environ­ mental impacts associated with PC production, but also significantly improves key properties of stabilised clay for underground construction.

ACKNOWLEDGMENT The authors sincerely appreciate the financial sup­ port from Ministry of Education (AcRF Tier 1, RG184/17), Singapore.

REFERENCES

* no water detected during the test

stabilised clay is one order of magnitude lower than that of PC-stabilised clays, which is desired for many underground constructions to reduce the seep­ age-induced risk during underground constructions. As shown in Table 2, the effect of binder/dry soil ratio on permeability is limited for PC- and CS­ GGBS-stabili-sed clays. For example, as binder/dry soil ratio increases from 0.2 to 0.3, the permeability of PC-stabilised clays slightly decreases from 2.28 × 10-1° to 1.57 × 10-1°. Furthermore, no significant change in permeability is observed in stabilised clays due to different CS/GGBS ratios. 4 CONCLUSIONS The following conclusions are drawn from this study: 1. Compared to PC-stabilised soft clay with the same binder content and curing time, the uncon­ fined compressive strength (UCS) of CS-GGBS­ stabilised clays is up to 300% higher, especially

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Experimental study of pore water pressure development in soil when foam infiltrates into saturated sand D. Zheng Department of Civil Engineering, Ghent University, Ghent, Belgium

A. Bezuijen Department of Civil Engineering, Ghent University, Ghent, Belgium & Deltares, Delft, The Netherlands

M. Thewes Institute for Tunnelling and Construction Management, Ruhr-University Bochum, Bochum, Germany

ABSTRACT: Foam is often used as an additive during tunneling in soft ground conditions in an Earth Pressure Balance (EPB) shield. Field measurements show that excess pore pressure builds up in front of the tunnel face during drilling and dissipates during standstill. This excess pore pressure will have con­ sequences over the tunnel face stability. In this study, the pressure transfer mechanism of a slurry driven shield is reviewed as a starting point for the mechanisms to be expected during drilling with an EPB shield. Foam infiltration tests were conducted to study the pore water pressure development during foam infiltration into saturated sand. The main focus is to investigate the change of pore water pressure during foam infiltration that can be expected during drilling of a tunnel. The results indicate that a pressure drop can be realized through the foam infiltrated area, thus the supporting pressure can be applied on the soils in front of a tunnel face. Further pressure development reveals that this foam infil­ trated area functions similarly as an internal cake in a slurry shield. Mechanisms of the pressure change are discussed.

1 INTRODUCTION The use of the Earth Pressure Balance (EPB) shield tunneling under soft ground conditions has gained quite popularity since its first successful application in Japan in the year of 1974 (Herrenknecht et al. 2011). The widened application of an EPB TBM is primarily attributed to the development of the soil conditioning technology. Additives like foam, poly­ mer and bentonite can help to change the properties of the excavated soil with suitable conditions. Foam is just one of the important additives which extended the use of an EPB shield from fine soil conditions to coarse soil conditions. Tunneling under soft ground often requires extra support pressure to prevent face collapse. In an EPB shield, this support pressure is usually achieved by the excavated soil in the mixing chamber. Condition­ ing agents like foam, bentonite and polymer are often utilized to condition the excavated soil into a low permeable, homogenous plastic paste in order to apply the support pressure uniformly to the tunnel face (Anagnostou & Kovári, 1996). Field measurements (Bezuijen, 2012) and model test result (Bezuijen, 2011) have shown that increasing

the porosity of the mixture in the mixing chamber to values higher than the maximum porosity of the sand will help decrease the torque on the cutting wheel. With a porosity higher than the maximum porosity, there will be no effective stress in the mixture and thus the support pressure in the mixing chamber will be fluid pressure (Bezuijen, 2002). However, the mechan­ ism of using foam to provide face support is still not very well understood (Bezuijen & Xu, 2019). Limited field measurements (Hoefsloot, 2001) and experimental study (Bezuijen & Schaminée, 1999, Thewes & Budach, 2012, Peila et al. 2013, Xu, 2018) have shed some light on this research focus. While more investi­ gations should yet to be carried out. 1.1 Pore pressure development in front of a slurry shield TBM In a TBM with a slurry shield, extensive studies have shown that the slurry pressure can be trans­ ferred to the soil skeleton by the formation of a filter cake (Talmon et al. 2013, Xu et al. 2018). An exter­ nal filter cake can form in a slurry shield during standstill and the pore pressure at the front face is expected to be transferred to the soil skeleton. The

DOI: 10.1201/9780429321559-112

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pressure drop over such a thin filter cake can be large, thus transforming the fluid pressure in the pressure chamber successfully onto the soils in the front face, creating an effective support to prevent face instability (Anagnostou & Kovári, 1994). Such a filter cake will not form during drilling, because when it starts to form, it will be ‘eaten’ by the TBM (Broere & van Tol, 2000, Bezuijen et al. 2016). During tunnel drilling, excess pore water pressures are generated in front of the TBM depending on the permeability of the soil and the advance rate of the TBM. A slurry infiltrated area can form in front of the TBM, which is often addressed as the internal cake in front of a slurry shield TBM. Kaalberg (2014) introduced a factor α to describe the influence of such a pressure drop over the internal cake when calculating the face stability. The factor α can be described by:

1.2 Pore pressure development in front of an EPB shield

Xu (2018) carried out an experimental study of foam infiltration into sand. The result was compared with the slurry infiltration characteristics. Well known are the two different infiltration stages during slurry infiltration: mud spurt and filter cake forma­ tion (Talmon et al. 2013). Xu’s finding suggests that a comparable ‘foam spurt’ stage happens at the very beginning of the tests with foam. During this foam spurt, the water discharge from the sample is rela­ tively high, but on a decreasing trend and the foam bubbles and the water between the bubbles infiltrate with the same velocity, thus the process is undrained. After the foam spurt, the foam bubbles are blocked by the sand grains and only the water between the bubbles flows into the soil. The bubbles fill up most of the pores in the soil and this creates a low perme­ able layer. Water discharge will be reduced due to that low permeable layer. During drilling in an EPB shield, formation of the low permeable layer may help to maintain the face stability. How quick this low permeable layer forms depends on the diameter of the tunnel and the permeability of the sand. In this study, it is assumed that the foam infiltration velocity is smaller than the drilling velocity. During drilling, the rotating cutting wheel constantly removes the soils ahead which has been infiltrated by foam. After which the freshly uncut soils will be exposed to be infiltrated by foam. Then the infiltration process can be renewed. Under such condition, foam spurt seems to be always the situation that happens in front of a tunnel face. Qualitative study by Bezuijen and Xu (2019) illustrated the mechanism of permeability reduction through foam infiltration. Their study sug­ gests that the air bubbles will be blocked at the foamsand boundary which creates a low permeability mem­ brane on the boundary between the foam and the ori­ ginal sand, functioning like a filter cake in case of a slurry shield (Figure 1). The low permeability

Different from a slurry shield, model tests revealed that an external cake formation process couldn’t happen in front of an EPB shield when foam is used as an addi­ tive (Bezuijen & Schaminée, 1999). In order to gain a successful support at the front face, the fluid pressure can only be transferred to the soil skeleton by the for­ mation of a low-permeability zone. Maidl (1995) and Quebaud et al. (1998) carried out some experiments of foam infiltration into sand. Their results indicate that depending on the pressure drop and grain size of the sand, some foam infiltration can be reached, resulting in a sand-foam mixture that has a low permeability compared to the original sand. In the light of this con­ clusion, the foam penetrated area has some similar function as the penetration zone that forms an internal filter cake in a slurry shield. This suggests that the excess pore pressure is transferred to the soil skeleton by the foam infiltration zone. A recent study has shown that the reduction in permeability by foam is realized through replacing the pore water with foam, which in turn creates an unsaturated flow in the soil (Bonab et al., 2014).

Figure 1. Sketch of air bubble and pressure distribution at a tunnel face (after Bezuijen & Xu, 2019).

with φ0 the piezometric head in the soils around the tunnel, φmx the piezometric head in the mixing cham­ ber and φ∞ the piezometric head far from the tunnel in the soil. The value of the factor α in Equation 1 depends on the slurry infiltration velocity. When the penetration velocity is lower than the drilling velocity, no effective plastering is present in the front face and thus α will be close to 1. α will be smaller than 1, if the slurry infiltration vel­ ocity is larger than the drilling velocity (Bezuijen et al. 2016).

849

membrane in turn will cause a large pressure drop. This way, the pressure can be applied on the sand and face stability can be achieved. To explore the pressure transfer mechanism during foam infiltration into saturated sand, foam infiltration tests were conducted. Special attention is focused on the pressure development during fast infiltration period when the foam infiltration velocity is relatively high, and mostly happens during tunnel excavation. The results will be analyzed, and the pressure transfer mechanism will be discussed.

schematic view of the testing system and the related dimensions. The set-up is similar to what Xu & Bezuijen (2018) used in their experiments where the hydraulic gradient was comparable to what can be expected in front of a slurry shield TBM. In this set-up, a small cylinder was added at the bottom of the 100 mm diameter cylinder to create an extra flow resistance which makes the equivalent length of the sand column to be equal to:

2 EXPERIMENTAL SET-UP Four pore pressure transducers (PPT: k1, k2, k3 and k4) are installed at the wall of a Perspex cylinder with a vertical distance of 4 cm between them to measure the pressure head in the foam and also in the sand column during the test. Specifically, k1 is placed in the foam and 2 cm above the foam-sand boundary. So k2 is 2 cm below the foam-sand boundary, k3 and k4 are located below k2 and the depths are 6 cm and 10 cm with respect to the foamsand boundary, respectively. Figure 2 shows the

with Ls1 the length of the big cylinder, Ls2 the length of the small cylinder, D1 the diameter of the big cylinder and D2 the diameter of the smaller cylin­ der. A filter cloth is placed beneath the small cylinder which is supported by a wire mesh. The filter cloth allows water flow but blocks the sand grains. With the dimension in this set-up (two layers: 45 mm thick in the small cylinder and 290 mm thick in the large cylinder), the equivalent length calculated by Equation 2 will be about 5 m. Applying an air pressure of 50 kPa will result in a hydraulic gradient of i = Δϕ/Ls = 1. This is comparable to the hydraulic gradient predicted by the groundwater flow model described in Bezuijen (2002) with a shield diameter of 10 m and an extra pore pressure of 50 kPa. 3 MATERIALS 3.1

Foam

Foam was produced using a laboratory foam generator according to Freimann (2012), which was calibrated to get comparable foam to EPB shields. Surfactant used was Condat CLB F5/TM in a concentration of 3.0 Vol %.The foam used in the experiment was first gener­ ated under atmospheric pressure and then filled inside the cylinder. Applying an excess pressure will cause a volume decrease of the foam and thus influences the Foam Expansion Ratio (FER) of the foam. The EFNARC (2005) recommends a FER between 5~30 in engineering application. For this study, the FER was chosen to be 10. Due to the fact that foam is highly compressible, the FER will be influenced under different pressures. Therefore, the FER of the foam represents the value calculated under applied pressure (50 kPa) throughout this study. 3.2

Figure 2. Schematic view of the experimental set-up (unit: mm).

Sand

In this study, a medium sand was utilized for the preparation of the sand column. The type of sand is a commercial product purchased from Euroquarz GmbH (Product type: Siligran). Particle size distri­ bution curve of the sand is shown in Figure 3.

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5 RESULTS 5.1

Volume and discharge

The infiltration depth (x) is calculated assuming all pore water is replaced by foam and that there is also no drained behavior of the foaming liquid during the test. With these assumptions, the infiltration depth can be calculated by:

with x the infiltration depth, Vpw the volume of expelled water, AC the cross-sectional area of the large cylinder and n the porosity of the sand column. The specific discharge (u) is the same to the Darcy’s velocity and is calculated by:

Figure 3. Grain size distribution curve of the used sand.

4 EXPERIMENTAL PROCEDURE For sample preparation, the cylinder was filled with a specific volume of water from bottom to avoid any air bubbles in sand and sensor front. After the filling of water, sand was rained in and densified with a compactor in water environment layer by layer. This procedure assures the non­ existence of any unsaturated condition of the sand column. The final height of both the sand column and the water as well as the weight of the sand used were recorded for the calculation of the sample parameters. The remaining water on top of the sand column was removed by a syringe. The permeability was obtained by the measured pore pressures between two adjacent sensors and the water discharge. The sand column in this study has a porosity of around 0.38. The tested water permeability is about 6×10-4 m/s. A foam layer of 300 mm thickness was then filled on top of the sand column, an air pressure of 50 kPa was then applied and kept constant during the test. According to Boyle’s Law, the foam will be com­ pressed and the height of the foam in the cylinder under 50 kPa (150 kPa of absolute pressure) will be reduced to about 200 mm. This also has consequences for the FER of the foam since its total volume has decreased. In this research, the FER will represent the ratio between the volume of foam and the volume of the foaming liquid under applied pressure in accord­ ance with EFNARC (2005). Under this principle, the FER of the produced foam in atmospheric pressure will be approximately 1.5 times of the FER under pressure. The test started by opening the valve at the bottom. Pressures and discharged water were recorded at a time interval of 1 s. After 1 hours of test, the valve was closed and the pressure was released.

with Q the discharge determined at a given time interval, which was smoothed with a moving average of 15. Calculation results with Equations 3-4 are pre­ sented in Figure 4. Figure 4 shows that at the beginning of the test, the specific discharge is about 5.5×10-4 m/s, a little smaller than the water permeability of the sand. Because the hydraulic gradient of the sand column under the applied pressure of 50 kPa is nearly 1, and no foam has yet infiltrated into the sand, the specific discharge at the start of the test should be almost the same to the water permeability of the sand. The specific discharge quickly decreased by one order of magnitude at about 100 seconds. During the whole test period of 1 hour, the specific discharge continued to drop. Upon the end of the test, it dropped to about 2×10-6 m/s. As is pointed out by Xu (2018), the first two stages come with the following definition. The foam spurt defines the stage when foam bub­ bles are quickly entering the sand pores. The pore water in the sand will be replaced by the foam bub­ bles and the behavior of the foam is undrained. In the second stage, the permeability reduction, where foam bubbles will be blocked and entrapped among the sand pores. The dominating flow in this stage will be the drainage of the foaming liquid. Which is similar in the foam aging process (Wu et al. 2018, Koehler et al. 2004, Magrabi et al. 1999). For this study, however, it seems difficult to deter­ mine the foam spurt period since the specific dis­ charge keeps relatively large all the time. Most likely there is no sharp boundary where the foam spurt ends. More likely, the velocity of the bubbles decreases

851

over time and after some infiltration time, the velocity is significantly lower than the velocity of the water between the bubbles. Although on a decreasing trend, the total discharge remained larger than 10-6 m/s even at the end of the test. As in the foam spurt period, the behavior of foam should be undrained, effective dis­ crimination over the foam spurt period could not be realized with merely the discharged water. Since the main focus is to study the change of pore water pres­ sure during foam infiltration, the foam spurt period will not be discussed any more. 5.2

Pore water pressure

The measured pore water pressures at different loca­ tions were plotted against time. As is shown in Figure 4, the infiltration depth could reach nearly 12 cm, indicating that the foam front has passed all the pressure transducers located in different depths. Therefore, the pressures were also plotted as a function of the infiltration depth. The plots are shown in Figure 5. Figure 5 shows the measured pore water pressures against both time (Figure 5a) and infiltration depth (Figure 5b). Upon the start of the test, the pressures at different locations were more or less the same. After the valve at the bottom was open, the pore pressures below the sand level quickly decreased, while the pore pressure measured by k1 only decreased a little. Below the sand level, the pore pressures decreased fast and almost linearly with time within the first 25 seconds. Which corresponds with the quick decrease of specific discharge shown in Figure 4. The fast dis­ sipation of the pore water pressures in the sand column suggests that the excess pore pressures could be transferred to the soil skeleton and that very soon the permeability of the upper part of the sand column (where the foam infiltrated) is nearly 4 orders of mag­ nitude lower than the permeability of the rest of the column. It should be noted that until 1000 s still excess pore water pressure is present in the soil (measured with k4) due to the foam penetration.

Figure 4. Discharge and infiltration depth over time.

Figure 5. Pore water pressure distribution curve. The pore pressures were plotted against both time (a) and infiltration depth (b).

Shortly after the foam front passed k2, which means the infiltration depth is bigger than 2 cm, the measured pore pressure by k2 started to increase. While k3 and k4 continued to decrease. The contrast in the pressure development at this infiltration depth indicates that the foam infil­ trated area has a strong influence over the pore pressure distribution. As the foam front traveled even deeper, the pore pressure measured by k2 got even higher. But with a slow progress in the pressure rise. Similar to k2, at the infiltration depth of about 6.5 cm, the pore pressure meas­ ured by k3 also started to increase. With more infiltration depth, the pore pressure measured with k4 continued to drop to smaller values. Figure 5 shows that after about 200 seconds, the pore pressure in the original sand (10 cm below the sand level) almost reached steady values. While the pore pressures measured by k2 and k3 continued to increase. Figure 6 shows the measured pore water pressure distribution at different depths. Prior to the start of the test, the pore water pressure increases with depths because they are hydrostatic pressures. Once the test started, flow resistance in the upper part of the sand column increased due to foam infiltration and the pore water pressure dissipated fast. At 10 seconds, the pore water pressures in the sand

852

measured by k2 changes from P2 to P2ʹ during this process. Assume there is a linear distribution of pore water pressure over the foam infiltrated sand, which seems a reasonable assumption, from the measure­ ment results shown in Figure 6. An increase in pres­ sure at location k2 is expected when the foam front passes k2 and infiltrates further into the sand. The increase in pressure measured by k2 and k3 illus­ trates that the foam front has passed 2 cm and 6 cm, respectively, see Figure 5b. 6 CONCLUSION Figure 6. Pore water pressure distribution at different times as a function of location.

column were more or less the same, but with a much smaller value than in foam. After 20 seconds, the pore water pressure at 2 cm below the sand level shows some higher value than at 6 cm and 10 cm, indicating that the foam front has passed k2. This is confirmed that as time went on, the distribution of pore water pressure at the depth of 6 and 10 cm continued to drop, while at 2 cm, the pore water pressure showed some increasing trend. Similar development in pore water pressure happened at the depth of 6 cm. A schematic view over the change in piezo­ metric head of k2 after the foam front has passed is shown in Figure 7. The upper part is pure foam, and the foam has penetrated in the blued dotted area and below that it is still pure sand. In the diagram the foam front settles further to deeper sand (from x to x’). The pore pressure

Figure 7. Change in piezometric head measured by k2 after foam front passed.

The pressure transfer mechanism of a slurry driven shield is reviewed, especially during drilling when the formation of an internal cake plays an important role. Laboratory experiments on foam infiltration into saturated sand have been conducted to investi­ gate the pressure transfer mechanism during foam infiltration which is anticipated during drilling. A few conclusions can be made according to the test results: 1. There is no impermeable layer formed at the foam-sand boundary as in a slurry infiltration test. While the foam infiltrated area could main­ tain a pressure drop of the excess pore pressure. Which is in line with the previous research by Maidl (1995), Quebaud et al. (1998) and Bezui­ jen & Xu (2019). 2. It is shown that the large pressure drop of the excess pore pressure happens at the first few centimeters of soils where the permeability reduces nearly 4 orders of magnitude due to the infiltrated foam. The infiltration depth by foam is fast at the first few centimeters, and the excess pore water pressure dissipates within a minute in the original sand for the conditions tested, indicat­ ing that the support pressure will be effective during standstill but only partly effective during drilling when it can be expected that a layer of soil is cut away every half a minute. The influ­ ence of this on the stability of the tunnel face is a topic of further investigation. 3. Analogous to the conditions expected in front a slurry shield during drilling, the foam infil­ trated area functions similarly as an internal cake in a slurry shield condition. The pressure drop over the foam infiltrated area is realized through the permeability reduction on the sand. This is proven by the measured pore pressure inside the infiltrated area. That the pore water pressures in the foam infiltrated sand are relatively high compared with those in the original sand. Further research will focus more on the influence of foam and sand on foam infiltration. The mech­ anism behind the permeability reduction by foam is also interesting to investigate.

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ACKNOWLEDGEMENT This research was carried out at the Institute for Tun­ nelling and Construction Management of Ruhr Uni­ versity Bochum, the permission to work on this research and the assistance are highly appreciated. The first author would like to acknowledge the s­ cholarship funded by China Scholarship Council and the CWO Mobility Fund of Ghent University.

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Geotechnical Aspects of Underground Construction in Soft Ground – Elshafie, Viggiani & Mair (eds) © 2021 ISSMGE, London, UK, ISBN 978-0-367-33733-9

Author Index

Acikgoz, S. 545, 605

Afshani, A. 357, 493, 590

Akagi, H. 357, 493, 590

Alagha, A.S.N. 415

Alhaddad, M. 3

Alloul, B. 51

Alonso, E. 153

Al-Tabbaa, A. 115, 755

Ambrosi, M. 211

Anh, P.T. 681

Anthony, C. 98

Aradas, R. 300, 309

Ares, J. 98

Arroyo, M. 829

Arson, C. 379

Astle, P.J. 60, 68

Baille, W. 796

Bakker, K.J. 501

Baldi, A.M. 199

Barker, C. 98

Beesley, M.E.W. 217

Belhai, D. 51

Bengtsson, E. 83

Bezuijen, A. 226, 234, 408,

467, 727, 848

Bilfinger, W. 26

Bilotta, E. 605, 613

Bin-Chen, H.B. 283

Biscontin, G. 765

Bobet, A. 395

Boldini, D. 510

Bolton, M.D. 430

Boone, S.J. 242

Bragard, C. 459

Broere, W. 211, 738

Brown, O. 250

Buket, Z. 569

Burd, H.J. 545

Burland, J.B. 191

Busbridge, J. 554

Cafaro, M. 325

Cao, B. 755

Casini, F. 401

Chaiyaput, S. 681

Chan, D.Y.K. 259

Chang, I. 761

Chapman, D. 444

Chen, H.Z. 696

Chen, J. 681

Chen, R.P. 517, 696

Chen, S. 696

Cheng, H.Z. 517

Cheng, S-H. 773

Cheng, W-C. 12, 19

Chien, S.C. 365

Cho, G-C. 761

Crispin, J.J. 523

Dang, T.S. 226

De Gori, V. 266

de Lillis, A. 266

De Rivaz, B. 385

de Wolf, W.J. 529

Derriche, Z. 51

Dewhirst, M. 98

Di Mariano, A. 536

Di Murro, V. 177

Divall, S. 584

Elshafie, M.Z.E.B. 275

Embley, T. 115

Endou, K. 493

Faramarzi, A. 444

Faustin, N.E. 275

Feng, S.J. 702

Ferrari, A. 421

Fillibeck, J. 576, 640

Franza, A. 510, 719,

746

Fujiwara, K. 563

Fukuda, R. 74

Gangrade, R. 35

Gao, Y.B. 43

Gens, A. 153, 536

Geuder, S. 640

Glab, K.B. 211

Goh, S.H. 812

Goisis, M. 115

855

Gómez, R. 829

Goodey, R.J. 584

Grasmick, J. 35

Guida, G. 401

Gulen, D.B. 545

Gutierrez, M. 459

Hai, P.K. 283

Haig, B. 98

Haigh, S.K. 415, 430, 613

Hanssen, R.F. 137

Hashida, K. 554

Hashimoto, T. 554, 563

Hassan, G. 493

He, Y. 108

Heath, I. 821

Hebib, R. 51

Heron, C.M. 673

Ho, C.C. 91

Hof, C. 83

Hsieh, P.G. 365

Hu, Q. 836

Huang, H. 444

Huang, S. 836

Hung, C. 283

Isa, M. 563

Işık, S. 569

Ito, S. 343

Ito, Y. 357

Izquierdo, J. 153

Jefferis, S.A. 788

Jiang, N. 842

Jimenez, R. 732, 746

Jones, B.D. 60, 68

Kahlström, M. 632

Kaneko, S. 357

Karlsrud, K. 83, 161

Kechavarzi, C. 177

Klar, A. 430

Klinger, A. 576

Konishi, S. 74, 343

Konstantinou, C. 765

Korff, M. 137, 529

Kuszyk, R. 169

Kydland Lysdahl, A.O. 632

Labanda, N.A. 300, 309

Lam, L.G. 681

Lande, E.J. 317, 647

Langford, J. 83, 632

Lavasan, A.A. 451

Le, B.T. 584

Leung, C.F. 812

Li, G. 12, 19

Li, W. 357, 590

Liao, H-J. 91, 773

Lile, C. 129

Lim, A. 365

Litina, C. 755

Liu, L.L. 710

Liu, R. 842

Liu, Y. 696

Liu, Z. 836

Lockhart, T. 780

Lopes dos Santos, A. 26

López, A.R. 597

López, J. 829

Losacco, N. 325, 510

Luciano, A. 605

Mace, N. 812

Madabhushi, S.P.G. 259,

613

Maekawa, K. 343

Mair, R. 605

Mair, R.J. 275

Manabe, K. 681

Marazzita, R. 325

Marshall, A.M. 673, 719

Martínez-Bacas, B. 664

Martini, M. 300, 309

McNamara, A.M. 371

Meng, F.Y. 517, 696

Merritt, A.S. 788

Meschke, G. 226

Mianji, P. 796

Miliziano, S. 266, 401, 408

Mimnagh, F. 821

Miraei, S.M. 510

Miranda, G. 613

Mitew-Czajewska, M. 620

Modoni, G. 350

Mooney, M. 35, 459

Moriya, T. 74

Morton, R.F. 177

Murakami, T. 74

Murphy, E. 60, 68

Nadim, F. 632

Nagaya, J. 563

Nair, R. 554

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