Environmental Challenges in Civil Engineering II 3031268784, 9783031268786

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Lecture Notes in Civil Engineering

Zbigniew Zembaty · Zbigniew Perkowski · Damian Beben · Maria Rossella Massimino · Oren Lavan   Editors

Environmental Challenges in Civil Engineering II

Lecture Notes in Civil Engineering

322

Series Editors Marco di Prisco, Politecnico di Milano, Milano, Italy Sheng-Hong Chen, School of Water Resources and Hydropower Engineering, Wuhan University, Wuhan, China Ioannis Vayas, Institute of Steel Structures, National Technical University of Athens, Athens, Greece Sanjay Kumar Shukla, School of Engineering, Edith Cowan University, Joondalup, WA, Australia Anuj Sharma, Iowa State University, Ames, IA, USA Nagesh Kumar, Department of Civil Engineering, Indian Institute of Science Bangalore, Bengaluru, Karnataka, India Chien Ming Wang, School of Civil Engineering, The University of Queensland, Brisbane, QLD, Australia

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Zbigniew Zembaty · Zbigniew Perkowski · Damian Beben · Maria Rossella Massimino · Oren Lavan Editors

Environmental Challenges in Civil Engineering II

Editors Zbigniew Zembaty Faculty of Civil Engineering and Architecture Opole University of Technology Opole, Poland

Zbigniew Perkowski Faculty of Civil Engineering and Architecture Opole University of Technology Opole, Poland

Damian Beben Faculty of Civil Engineering and Architecture Opole University of Technology Opole, Poland

Maria Rossella Massimino Faculty of Civil Engineering and Architecture University of Catania Catania, Italy

Oren Lavan Faculty of Civil and Environmental Engineering Technion – Israel Institute of Technology Haifa, Israel

ISSN 2366-2557 ISSN 2366-2565 (electronic) Lecture Notes in Civil Engineering ISBN 978-3-031-26878-6 ISBN 978-3-031-26879-3 (eBook) https://doi.org/10.1007/978-3-031-26879-3 © The Editor(s) (if applicable) and The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 This work is subject to copyright. All rights are solely and exclusively licensed by the Publisher, whether the whole or part of the material is concerned, specifically the rights of translation, reprinting, reuse of illustrations, recitation, broadcasting, reproduction on microfilms or in any other physical way, and transmission or information storage and retrieval, electronic adaptation, computer software, or by similar or dissimilar methodology now known or hereafter developed. The use of general descriptive names, registered names, trademarks, service marks, etc. in this publication does not imply, even in the absence of a specific statement, that such names are exempt from the relevant protective laws and regulations and therefore free for general use. The publisher, the authors, and the editors are safe to assume that the advice and information in this book are believed to be true and accurate at the date of publication. Neither the publisher nor the authors or the editors give a warranty, expressed or implied, with respect to the material contained herein or for any errors or omissions that may have been made. The publisher remains neutral with regard to jurisdictional claims in published maps and institutional affiliations. This Springer imprint is published by the registered company Springer Nature Switzerland AG The registered company address is: Gewerbestrasse 11, 6330 Cham, Switzerland

Preface

Dear Readers, We present to you a monograph entitled Environmental Challenges in Civil Engineering II, which is a collection of selected original papers presented at the 5th International Scientific Conference on Environmental Challenges in Civil Engineering (ECCE 2022). The event took place in Opole, Poland, on September 26–28, 2022. The conferences in this series are organized by the Opole Branch of the Polish Association of Civil Engineers and Technicians (PACET), the Faculty of Civil Engineering and Architecture of the Opole University of Technology, the Opole District Chamber of Civil Engineers, and the Commission of Civil Engineering of the Katowice Branch of the Polish Academy of Sciences. The ECCE 2022 was held under honorary patronage of: Prof. Maria Kaszy´nska, the Chairwoman of PACET, Prof. Marcin Lorenc, the Rector of the Opole University of Technology, Mr. Andrzej Buła, the Marshal of the Opole Voivodeship, and Mrs. Aleksandra Drescher, the President of the Provincial Fund for Environmental Protection and Water Management in Opole. The ECCE conferences have been held cyclically every 2 years since 2014, and starting from the 3rd one, i.e., the ECCE 2018, they have an international dimension. The papers from the 3rd ECCE 2018 were published by Matec Web of Conferences (Vol. 174, 2018), and the contributions from the 4th ECCE 2020 were published by Springer (Environmental Challenges in Civil Engineering, Lecture Notes in Civil Engineering, Vol. 122, 2021). In line with the intention of the organizers and the tradition of the ECCE conferences, they focus on the results of interdisciplinary scientific research in the field of sustainable civil engineering, especially in designing, durability, diagnostics, and maintenance of building structures and materials, including also environmental impact assessment and some architectural issues. Hence, the basic purpose of the ECEE meetings follows. Thanks to such a specific “integration” of the above-mentioned topics in one place and from an international perspective, the organizers have always expected and will expect also in the future for the effect of synergy, which allows for an effective exchange of interesting ideas, technical knowledge, and experience between the scientists and practitioners of different platforms, contributing to the deepening of international cooperation between research centers present at the Conference in view of today’s broad, modern, sustainable, and ecological challenges for civil engineering. At the invitation of the organizers of the ECCE 2022, four special plenary lectures were prepared and delivered, which, one might say, determined the main framework of this event in 2022. These were the following ones: • Zbigniew Giergiczny from Silesian University of Technology, Poland: Cement and Concrete in a Low-Carbon Economy. • Oren Lavan and Ohad Idels from Technion – Israel Institute of Technology, Israel: Seismic Retrofitting and Seismic Design Using Advanced Technologies: An Environmental Point-of-View.

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Preface

• Maria Rossella Massimino, Glenda Abate, Angela Fiamingo from University of Catania, Italy, and Dimitris Pitilakis from Aristotle University of Thessaloniki, Greece: Seismic Risk and Environmentally Friendly Solutions: The Geotechnical Point of View. • Mariusz Pustelnik from Pracownia Projektowa MOSTOPOL sp. z o.o., Poland: Technical Aspects of the Designed Railway Bridge over the Odra River in Opole. We are pleased to inform you that two of these interesting lectures, after extensive elaboration by their authors, have been also published in the form of chapters in this monograph. The entire monograph consists of twenty-four chapters, based on the selected papers presented and, importantly, often widely discussed by the participants of the ECCE 2022 Conference. Due to their thematic scope, they have been divided into five parts: • • • • •

Seismic Engineering and Geotechnics – seven chapters Bridge Engineering – four chapters Concrete Engineering – four chapters Architecture and Urban Planning – four chapters Other Issues: Wind Engineering, Innovative Solutions, Reconstruction, and Organization of Building Site – five chapters

At this point, we would like to thank all the authors of submitted papers and chapter proposals, given the difficult time in which we had to meet in 2022 due to the challenges posed by the pandemic, armed conflict, and economic uncertainties. In addition to the authors, we also would like to especially thank the Scientific and Organizing Committees (listed on the next two pages), the Honorary Patrons, and hosting institutions of the ECCE 2022. Without the invaluable contributions of all of you, this book simply would not have been possible. Inviting you to read this monograph. Zbigniew Perkowski Damian B˛eben Zbigniew Zembaty Maria Rossella Massimino Oren Lavan

The Scientific Committee of the ECCE 2022 Conference

Chairman Zbigniew Perkowski

Opole University of Technology, Poland

Members Glenda Abate Monika Adamska Damian B˛eben Dariusz Bajno Mirosław Bogdan Giovanni Bosco Jacek Domski Zbigniew Giergiczny Stefania Grzeszczyk Bo˙zena Hoła El˙zbieta Janowska-Renkas Zbynˇek Keršner Stanislav Kmet’ Seweryn Kokot Peter Koteš Tomasz Krykowski Oren Lavan Maciej Major Andrzej Marynowicz Maria Rossella Massimino Jaroslav Odrobiˇnák Miguel José Oliveira Shehata E. Abdel Raheem Adam Rak Andrea Segalini Halil Sezen Aleksandr Shymanowskiy Elisa da Silva Iveta Skotnicová

University of Catania, Italy Opole University of Technology, Poland Opole University of Technology, Poland Bydgoszcz University of Technology, Poland Opole University of Technology, Poland University of L’Aquila, Italy Koszalin University of Technology, Poland Silesian University of Technology, Poland Opole University of Technology, Poland Wrocław University of Technology, Poland Opole University of Technology, Poland Brno University of Technology, Czech Republic Technical University of Košice, Slovakia Opole University of Technology, Poland University of Žilina, Slovakia Silesian University of Technology, Poland Technion – Israel Institute of Technology, Israel Cz˛estochowa University of Technology, Poland Opole University of Technology, Poland University of Catania, Italy University of Žilina, Slovakia University of Algarve, Portugal Assiut University, Egypt Opole University of Technology, Poland University of Parma, Italy Ohio State University, USA V.N. Shimanovsky Ukrainian Institute of Steel Construction, Ukraine University of Algarve, Portugal Technical University of Ostrava, Czech Republic

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The Scientific Committee of the ECCE 2022 Conference

Pranshoo Solanki Luca Trabattoni Jan Vaslestad Amer Wadi Robert Wójcik Jerzy Wyrwał Zbigniew Zembaty

College of Applied Science and Technology – Illinois State Univ., USA University of Pavia, Italy Norwegian University of Life Sciences, Norway ViaCon International AB, Sweden University of Warmia and Mazury, Poland Opole University of Technology, Poland Opole University of Technology, Poland

Scientific Secretaries Dominika Bysiec Mariusz Czabak Krystian Jurowski

Opole University of Technology, Poland Opole University of Technology, Poland Opole University of Technology, Poland

The Organizing Committee of the ECCE 2022 Conference

Chairman Wiesław Baran

Opole University of Technology/Polish Association of Civil Engineers and Technicians, Opole Branch

Members Damian B˛eben Dominika Bysiec Zbigniew Perkowski

Adam Rak

Opole University of Technology Opole University of Technology Opole University of Technology/Commission of Civil Engineering, Katowice Branch of Polish Academy of Sciences Opole University of Technology/Opole District Chamber of Civil Engineers

Secretaries Małgorzata Gody´n Krzysztof Irek

Opole District Chamber of Civil Engineers Opole University of Technology

Contents

Seismic Engineering and Geotechnics Seismic Risk and Environmentally Friendly Solutions: The Geotechnical Point of View . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Maria Rossella Massimino, Glenda Abate, Angela Fiamingo, and Dimitris Pitilakis Seismic Design Using Advanced Technologies: An Environmental Point-of-View . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Oren Lavan and Ohad Idels

3

23

Sustainable Building: The Role of the Soil Parameters on Earthquake Safety . . . Glenda Abate, Angela Fiamingo, and Maria Rossella Massimino

32

The Role of DSSI on the Seismic Risk Assessment of a Building . . . . . . . . . . . . . Glenda Abate, Angela Fiamingo, and Maria Rossella Massimino

48

Response of Geodesic Domes on the Seismic Excitation with Time History Analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Dominika Bysiec, Adriana Janda, and Tomasz Maleska

64

Algorithms for the Near-Real Time Identification and Classification of Landslide Events Detected by Automatic Monitoring Tools . . . . . . . . . . . . . . . Alessandro Valletta, Andrea Carri, Roberto Savi, and Andrea Segalini

74

Analysis of Dynamic Compaction Effects – Tavira Case Study . . . . . . . . . . . . . . . Elisa M. J. da Silva and Miguel Oliveira

85

Bridge Engineering Stress State and Several Problems of Estimating the Actual Operation of Effective Bridge Superstructures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Oleksandr Shymanovskyi, Valerij Shalinskyi, Maryna Shymanovska, and Wiesław Baran

99

Durability and Assessment of Early Post-tensioned Bridges . . . . . . . . . . . . . . . . . . 118 ˇ Petra Bujˇnáková, Jakub Kralovanec, Martin Moravˇcík, and František Bahleda

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Contents

Impact of Duration and Intensity of Seismic Records on the Soil-Steel Bridge Behavior . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 127 Tomasz Maleska, Nashwan Al Zubairi, Adriana Janda, Damian B˛eben, and Kareem Embaby Impact of Reinforcement Layer in Soil-Steel Culvert on Laboratory and Numerical Tests . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 139 Tomasz Maleska, Adam Wysokowski, and Damian B˛eben Concrete Engineering Comparative Analysis of the Use of Different Concrete Models in the Evaluation of the Propagation of Damage in the Reinforced Concrete Sample . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 151 Kseniya Yurkova and Tomasz Krykowski Diagnostic and Design of Reconstruction of Building Váhostav . . . . . . . . . . . . . . 165 Peter Koteš, Michal Zahuranec, and Martin Vavruš Rheological Properties of Concrete Based on Waste Materials . . . . . . . . . . . . . . . 175 Mateusz Zakrzewski, Artur Sanok, and Jacek Domski Effect of the Addition of Waste Fibers on Some Properties of Concrete . . . . . . . . 185 Artur Sanok, Mateusz Zakrzewski, Marek Lehmann, and Jacek Domski Architecture and Urban Planning Ecological Aspect of Water Retention Through Algae Based Hydraulic Systems in Interactive Architecture . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 199 Anna Grajper-Dobiesz and Sebastian Dobiesz Summer City: Campsites as a New Ecological Approach to Sustainable Living . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 211 Luca Trabattoni, Carlo Berizzi, Margherita Capotorto, Gaia Nerea Terlicher, and Marta Mazurkiewicz The Use of a Prefabricated Mass Timber Structure in the Design of Single-Family Houses in Terms of Sustainable Development . . . . . . . . . . . . . . 226 Kamila Wilk Rural Building in Opole Silesia During the Period of the Frederician Colonization . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 234 Dariusz Bajno, Marcin Fiutak, and Maciej Niedostatkiewicz

Contents

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Other Issues: Wind Engineering, Innovative Solutions, Reconstruction, and Organization of Building Site Wind-Based Form-Finding of a Tensegrity Pavilion . . . . . . . . . . . . . . . . . . . . . . . . 245 Lenka Kabošová, Tomáš Baroš, Stanislav Kmetˇ, Eva Kormaníková, and Dušan Katunský Energy-Efficient Reinforced Heating System Implemented as a Carbon Concrete Formwork . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 258 Ralf Gliniorz, Carolin Petzoldt, Kristin Mandel, and Sandra Gelbrich Originality and Safety as a Priority in the Revitalization of Immovable Monuments? . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 273 Dariusz Bajno, Agnieszka Grzybowska, and Magdalena Chylewska-Szabat Shaping the Course of Costs Curves Generated in Diversified Investment Sectors . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 283 Jarosław Konior and Mariusz Szóstak Economic Conditions of Leaving the Construction Site by the Contractor at Different Stages of its Implementation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 290 Daniel Przywara and Adam Rak

About the Editors

Assoc. Prof. Zbigniew Perkowski Ph.D., D.Sc., Faculty of Civil Engineering and Architecture, Opole University of Technology, Katowicka 48, 45-061 Opole, Poland E-mail: [email protected] He has been working since 1999 at the Faculty of Civil Engineering and Architecture of Opole University of Technology (Poland) where he is currently the head of the Department of Physics of Materials and teaches “Theory of Elasticity and Plasticity”, “Thermomechanics”, “Strength of Materials”, and “Mechanics of Concrete”. He is an author or the co-author of 3 books and about 70 publications, including proceedings of the domestic and international conferences as well as articles in the peer-reviewed scientific journals including: International Journal of Heat and Mass Transfer, Journal of Building Physics, Engineering Structures, Materials, Energies, and Applied Sciences. He was a member of scientific boards of several international and domestic conferences and belongs to the Section of Building Physics of the Committee of Civil Engineering and Hydroengineering of the Polish Academy of Sciences. He is a chairman of the Commission of Civil Engineering at Katowice Branch of the Polish Academy of Sciences in the terms of office 2015–2018 and 2019–2022. ˙ In 2013, he received the Professor Wacław Zenczykowski’s Award of Polish Association of Civil Engineers and Technicians (PACET) for the achievements in research in the field of civil engineering concerning damage mechanics. Damage mechanics, mechanics of concrete, composite structures, building physics, ultrasonic testing of concrete structures, and inverse problems are his most important research interests.

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About the Editors

Assoc. Prof. Damian B˛eben Ph.D., D.Sc., Faculty of Civil Engineering and Architecture, Opole University of Technology, Katowicka 48, 45-061 Opole, Poland E-mail: [email protected] He is the chairman of the Scientific Council of the civil engineering, geodesy and transport discipline at the Faculty of Civil Engineering and Architecture, Opole University of Technology (Poland). He is an author and co-author of 3 books and over 190 publications on the national and international conferences as well as in the peer-reviewed scientific journals indexed in the Journal Citation Reports. He has an h-index of 15 according to the Web of Science Core Collection and Scopus. He is a reviewer in many scientific international journals and the National Centre for Research and Development and a member of the International Association for Bridge Maintenance and Safety (IABMAS); International Association for Life-Cycle Civil Engineering (IALCCE); International Association of Computer Science and Information Technology (IACSIT); and Transportation Research Board (TRB) of the National Academies, Committee on Subsurface Soil-Structure Interaction (AFS40). He was the scholarship holders of the Foundation of Polish Science for the young prominent scientists, scientific scholarship for outstanding young scientists awarded by the Ministry of Science and Higher Education, 2011 Outstanding Reviewer for the Journal of Bridge Engineering (ASCE), and the European Social Found for Ph.D. Analysis of soilsteel bridge, durability of engineering structures, field load tests of structures, non-destructive evaluation of structures, and environmental protection in transportation engineering are his most important research interests. Prof. Zbigniew Zembaty Ph.D., D.Sc., Faculty of Civil Engineering and Architecture, Opole University of Technology, Katowicka 48, 45-061 Opole, Poland E-mail: [email protected] He has been working at the Faculty of Civil Engineering and Architecture of Opole University of Technology (Poland) where he is currently the dean. He is an author or the co-author of a book and about 100 other publications including about 50 covered by Science Citation Index (core collection) with about 500 citations and respective hindex 15. He has been listed among top 2% researchers of

About the Editors

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the World by Stanford index (https://elsevier.digitalcommo nsdata.com/datasets/btchxktzyw/3). The main areas of scientific research cover civil engineering structural dynamics, random vibrations, inverse problems, structural health monitoring, seismic engineering, and geophysics. He is the co-editor of three Springer books: Seismic Behaviour and Design of Irregular and Complex Civil Structures II, III & IV (2016, 2020, 2022). Cooperation with Milan Polytechnic, Trento and Pavia Universities (Italy), and Israel Institute of Technology, including frequent participation in the respective doctoral schools with lectures on random vibrations, participation in scientific committees of numerous international conferences including World Conference on Earthquake Engineering (16th WCEE in Santiago, Chile in 2017), and actual active research in investigating rotational component of seismic ground motion. He is a member of special international group IWGoRS, which deals with introducing so called 6-component seismology and seismic engineering. More information: https://z.zembaty.po.opole.pl/. Prof. Maria Rossella Rita Massimino, Ph.D., D.Sc., Department of Civil Engineering and Architecture, Geotechnical Section, University of Catania, Building 13, Via Santa Sofia 64, 95125 Catania, Italy E-mail: [email protected] She is an associate professor of Geotechnics in the Department of Civil Engineering and Architecture (DICAR) at Catania University. In April 2017, she obtained the national qualification as a full professor in Geotechnics. She is the author and the co-author of over 130 publications on national and international conferences and in the peerreviewed scientific journals indexed in the Journal Citation Reports. She has an h-index of 17 and 680 citations, according to the Scopus database. She has attended as an invited speaker and a keynote speaker several national and international conferences. She has been and is the scientific leader of several national and international research projects. She is a member of the Editorial Committee of the following international journals indexed on Scopus: Bulletin of Engineering Geology and the Environment, Geotechnical and Geological Engineering, and Advances in Civil Engineering. She is a reviewer in many scientific international journals and for the Italian Ministry of University and Research. She is a member of the Italian Geotechnical Society (AGI) and the Italian National Group of Geotechnical Engineering

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About the Editors

(GNIG). She is a member of the Research Doctorate College in Assessment and Mitigation of Urban and Territorial Risks at Catania University. She won an ECOLEADER grant, the International Geosynthetics Society “IGS Student Award”, and the Italian “Troiano Award” for the best Ph.D. thesis carried out in Italy in 1999–2001. She was a member of the European Technical Committee on Application of Eurocode 8 (ERTC-12). She teaches “Foundations”, “Soil Mechanics and Soil Dynamics”, and “Geotechnical design using FEM codes”. She is the reference teacher for three Erasmus Agreements. Foundations, soil-structure interactions, soil dynamics, local site response, and environmental protection using new mixtures including soil are her most important research interests. Assoc. Prof. Oren Lavan Ph.D., Faculty of Civil and Environmental Engineering, Technion – Israel Institute of Technology, Haifa 32000, Israel E-mail: [email protected] He is an associate professor in the Faculty of Civil and Environmental Engineering at the Technion - Israel Institute of Technology. He received his Ph.D. in Civil Engineering from the Technion in 2006 and was a visiting research scientist (2005–2007) at the Department of Civil, Structural and Environmental Engineering, University at Buffalo – The State University of New York. He spent a sabbatical as a visiting associate professor at the Disaster Prevention Research Institute at Kyoto University (2014). Prof. Lavan teaches and conducts research in structural engineering under extreme events, structural earthquake engineering, structural control, and computational mechanics. He has published extensively in the area of structural engineering and mechanics, including more than sixty peer-reviewed articles. His articles have been published in premier journals in the field. Prof. Lavan serves as an associate editor for the Journal of Structural Engineering, ASCE. He is the Israeli national delegate to the International Association for Earthquake Engineering (IAEE). He is a member of the American Society of Civil Engineers (ASCE) where he participates in works of several committees. He is also a member of the European Association for Earthquake Engineering (EAEE) where he is a member of the executive committee of TG8 (Seismic behavior of irregular and complex structures). He is also active in the Israeli technical committees of standards.

Seismic Engineering and Geotechnics

Seismic Risk and Environmentally Friendly Solutions: The Geotechnical Point of View Maria Rossella Massimino1(B) , Glenda Abate1 and Dimitris Pitilakis2

, Angela Fiamingo1

,

1 Department of Civil Engineering and Architecture, University of Catania, Via Santa Sofia 64,

95125 Catania, Italy [email protected] 2 Department of Civil Engineering, Aristotle University of Thessaloniki, P.O. BOX 424, 54124 Thessaloniki, Greece

Abstract. Over the past 50 years, there has been a tremendous growth in research in the field of seismic geotechnical engineering, producing very valuable results and highlighting the decisive role of the soil filter effect and soil-structure interaction in evaluating structural seismic risk. However, the challenge in assessing and mitigating seismic risk is not entirely over, with the continual opening up of new scenarios. A need to safeguard the ecosystem is encouraging researchers to find new solutions that combine seismic risk mitigation and ecosystem protection. Soil-rubber mixtures (SRM) have emerged as a new technique to improve the soil underneath foundations, so that seismic energy will be partially dissipated within the SRM. The rubber grains are manufactured from scrap tyres, disposal of which has become a severe environmental problem worldwide. SRMs are characterized by low specific weight, high elasticity, low shear modulus, and high damping. The present paper reports some interesting results obtained by the authors as regards seismic microzonation, local site response, and soil-structure interaction. Finally, the paper reports the main results of the first tests performed on a full-scale soil-SRM-structure system and a FEM simulation of these tests. Keywords: Local site response · Dynamic soil-structure interaction · Geotechnical seismic isolation · Scrap tires · Full-scale tests · FEM modelling

1 Introduction As demonstrated by an analysis of the damage caused by many earthquakes, the first fundamental steps to be taken towards a realistic seismic design of new structures and seismic retrofitting of existing ones are an investigation of the soil filter effect and the complex soil-structure interaction phenomena. Reducing the seismic risk of structures and infrastructures is still an environmental challenge that has not been thoroughly addressed. In all parts of the world, the soil filter effect is computed by means of free-field (FF) local site response (LSR) analyses [1–3]. This approach represents a significant step forward compared to the use of the response acceleration spectra furnished by technical codes. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 3–22, 2023. https://doi.org/10.1007/978-3-031-26879-3_1

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M. R. Massimino et al.

LSR considers the specific conditions of the subsoil (geometry, physical properties, nonlinearity) and, therefore, its fundamental filtering effect in terms of PHA, PHV, PHD, predominant frequencies, Arias intensity, and so on [4–8]. It allows us to develop seismic microzonation maps for a rational use of the territory and to appropriately estimate the seismic inputs that affect structures and infrastructures. Nevertheless, the dynamic response at the foundation level of structures can significantly deviate from the FF site response due to kinematic and inertial interaction [9–22], and in some cases, the Dynamic Soil-Structure Interaction (DSSI) could be detrimental to the structures involved [23]. In this regard, experimental tests and numerical modelling of fully-coupled soil-structure systems are the most valuable approaches, in that they produce results that most nearly reflect the actual configurations. Moreover, in recent years, a need to safeguard the ecosystem has encouraged researchers to find new solutions that combine seismic risk mitigation and ecosystem protection. Among others [24–26], soil-rubber mixtures (SRM) have emerged as a new, valuable technique for protecting structures in earthquake-prone areas. The main idea is to improve the soil immediately underneath the foundations using SRMs so that seismic energy will be partially dissipated within the SRMs before being transmitted to the structures. Rubber grains for the mixtures are manufactured from scrap tyres, the disposal of which has become a severe environmental problem worldwide. SRMs are characterized by low specific weight, high elasticity, low shear modulus, and high damping. Experimental and numerical analyses have shown good performance of SRMs in static and dynamic conditions as geotechnical seismic isolation (GSI) solutions [27–34]. The present paper reports some interesting geotechnical works developed by the authors regarding seismic risk analysis and mitigation, including new eco-sustainable solutions. In particular, the paper reports the main results of: i)

An extensive LSR analysis for the development of seismic microzonation maps in the city of Noto (south-eastern Sicily, Italy). These maps are fundamental for urban and spatial planning with the idea of sustainable development, including the reconstruction and renovation of historic buildings. Noto is famous worldwide for its unique Baroque cultural heritage, but is characterized by a very high seismic vulnerability in an area with medium-high seismicity. ii) A DSSI analysis of a fully coupled soil-structure system regarding the strategic headquarters of the National Institute of Geophysics and Volcanology (INGV), Catania Section. The building is a valuable masonry structure, built at the end of the 1800s and characterized by a critical seismic vulnerability. iii) The first experimental campaign on a full-scale prototype structure resting on gravel-rubber mixtures (GRM) and a FEM simulation of the same. Thus, the use of GRM underneath structures aims to provide a desirable new solution that combines seismic risk mitigation and ecosystem protection.

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2 An Interesting Example of Seismic Microzonation The reduction of seismic risk in urban areas is an environmental matter which should be approached starting from a careful evaluation of local seismic hazard. Several studies have been conducted in earthquake-prone areas to develop seismic microzonation maps. The authors present an interesting example regarding the city of Noto (south-eastern Sicily, Italy), famous worldwide for its Baroque cultural heritage. In ancient times south-eastern Sicily was shaken by strong earthquakes that destroyed Augusta, Siracusa, and Noto and produced serious damage in Catania, all cities with a rich, historic and artistic heritage. The seismic history of this area has been marked by a few high-energy earthquakes, with eight very strong shocks occurring over the last nine centuries, all having an epicentral intensity in the VIII–XI MCS range. The last of these (VIII MCS) dates back to January 11th, 1846. The strongest earthquake to occur after this (VII-VII MCS), took place on December 13th, 1990, with an epicentre close to Augusta [35, 36]. As regards the city of Noto, it was destroyed by the earthquake that occurred on January 11th, 1693. It also sustained heavy damage during the one on January 7th, 1917. The most recent earthquake on December 13th, 1990, damaged several important buildings and highlighted the need to safeguard the city’s valuable cultural heritage. The photographs on the left of Fig. 1 show a view of some important buildings in Noto and the cathedral, which partially collapsed as a result of the 1990 seismic event.

Catania

IbleanMaltese fault

Augusta

Ionian sea

Siracusa N

Noto

(a)

(b)

(c)

Epicentre

Fig. 1. On the left: a view of the valuable cultural heritage in Noto and the partial collapse of the cathedral due to the December 13th, 1990 earthquake; on the right: the rupture propagation models along the Iblean-Maltese fault used to compute the synthetic accelerograms for Noto. The star indicates the rupture nucleation.

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Fig. 2. Borehole location for the seismic microzonation of Noto.

Due to a lack of sufficient instrumental records, the seismic hazard of the area was established using synthetic accelerograms. Based on the geological and macroseismic observations of the 1693 earthquake, Bottari et al. [37] modelled the strong motion in Noto, considering the Iblean-Maltese fault (Fig. 1) as the seismogenic source of the target earthquake. More details can be found in [37]. Three different rupture propagating models were considered: a) a unilateral rupture from South to North; b) a unilateral rupture from North to South; c) a bilateral rupture. Then, synthetic accelerograms at the depths of 30 m and 70 m were computed for the E-W, N-S and vertical directions. The maximum PHA of the input motions, equal to 1.48 m/s2 was related to rupture model 1 – E-W component, while the minimum, equal to 0.39 m/s2 , was related to rupture model 1 – N-S component. Of all the synthetic accelerograms, those at a depth of 30 m were used here because the in-situ tests performed for the whole of Noto showed that the conventional bedrock (Vs ≥ 800 m/s) is located at variable depths of between approximately 9.00 m and 40.00 m (Figs. 2 and 3). The soil is mainly constituted by: talus material up to a depth equal to 13.00–15.00 m; sand and gravel at a depth equal to 2.00–25.00 m; clay-marly clay at a depth equal to 19.00–27.00 m and sand and silty clay up to a depth of 34 m. Laboratory tests also allowed the soil nonlinearity to be estimated [38]. The local seismic response was evaluated using the 1-D nonlinear GEODIN code [39], applying the input synthetic accelerograms at the conventional bedrock estimated for all the 16 boreholes in Fig. 2. The local seismic response was analyzed in terms of maximum horizontal acceleration, velocity and displacement at the soil surface (amax,surf , vmax,surf , d max,surf ), as well as in terms of amplification ratio (R), which is the ratio between the maximum acceleration at the soil surface and the maximum acceleration at the bedrock. Response spectra at the soil surface were also computed. Based on the values of amax,surf and R for each geological zone determined by the curves reported in Fig. 2 different microzonation maps were drawn up. Due to a lack of space, only the microzonation map in terms of R, for the N-S component of the input synthetic

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accelerograms related to source rupture model 1 is reported in this paper (Fig. 4) because it represents the most severe response in terms of R (R = 2.73–4.05).

Fig. 3. Vs vs z profiles for all the boreholes reported in Fig. 2.

Globally, the most severe zones in terms of R are: a) the small area in which the thickness of the superficial geological formation is equal to or more than 40 m, where R varies between 1.70 (source rupture model 1 – E-W component) and 4.05 (source rupture model 1 – N-S component); b) the area characterized by a thickness of the superficial geological formation in the range of 20–30 m, where R varies from 1.18 (source rupture model 3 – N-S component) to 3.38 (source rupture model 1 – N-S component). The most severe response in terms of R is reached considering model 1 and the N-S component, for which the relative input synthetic accelerogram is characterized by the lowest peak value of amplitude but by a predominant period, equal to 0.15 s, which is very close to the natural vibration period of most of the soil profiles. This confirms, once more, the critical role played not only by the peak value of amplitude but also by the frequency content of the input accelerograms. More details can be found in [40].

3 From Local Site Response to Soil-Structure Interaction In structural design/retrofitting, the design spectra and the ground motion acceleration time-histories are given by technical regulations or, more appropriately, derived by freefield (FF) local site response (LSR) analyses, such as that discussed in the previous section. The latter approach is a significant step forward, in that it considers the specific conditions of the subsoil. Nevertheless, the dynamic response at the foundation level of structures can deviate from the FF site response due to kinematic and inertial interaction and in some cases, the DSSI could be detrimental to the structures involved. In this section, the authors report a DSSI study performed in 2017 for the headquarters of the National Institute of Geophysics and Volcanology (INGV), Catania Section (Fig. 5a). The building and its subsoil were investigated during the POR-FESR Sicilia

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Fig. 4. Seismic microzonation map of Noto in terms of acceleration amplification ratio (R) using the synthetic accelerogram related to source rupture model 1 – N-S component.

2007–2013 research project, aimed at reducing seismic risk in Eastern Sicily. The seismic response of the fully-coupled system was investigated using a 2D FEM modelling (Fig. 6), considering soil-nonlinearity according to [41]. The results of the full-coupled system analyses (see the SSI alignment in Fig. 6) were compared with those related to the free-field site response (see the FFleft and FFright alignments in Fig. 6) in the time and frequency domains in terms of soil amplification ratio, Fourier and response spectra, and amplification functions. The resulting soil amplification ratio and response spectra were also compared with those given by the Italian technical regulation in force in 2017 [42]. Finally, the maximum shear forces at each level of the structure were compared with those given by the fixed-base structure configuration, traditionally used in engineering practice. In the present paper, the authors report the main results in terms of acceleration amplification ratio (Fig. 7) and acceleration response spectra (Fig. 8). The other results are reported in [44]. The INGV building is a masonry structure built at the end of the 1800s, whose bearing walls are built of lava stone; the foundations are enlargements of these walls, and they are embedded to a depth equal to 2.5 m. The floors are in brick and concrete, downloading on curbs in reinforced concrete resting on the walls. The bedrock is more than 200 m below ground level as regards the soil. However, it was fixed at 40 m (conventional bedrock) for the FEM analyses presented here, according to previous 1D analyses, which showed no significant amplification from z = 200 m to z = 40 m. The soil is of class C according to [42]. The main geotechnical properties are summarized in Table 1. The values reported in the table were modified in the FEM analyses according to [41] to take soil-nonlinearity into account. Seismic inputs were applied at the base of the FEM model. The inputs chosen were: six synthetic accelerograms evaluated assuming the source to be along the HybleanMaltese fault and generating the 1693 seismic ground motion scenario (1693(1), f = 2.06 Hz; 1693(2), f = 2.31 Hz, 1693(3), f = 2.31 Hz) and the 1818 seismic ground motion scenario (1818(1), f = 0.66 Hz; 1818(2), f = 0.52 Hz, 1818(3), f = 0.58 Hz); the

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(a)

9

(b)

Fig. 5. a) View of the INGV building; b) plan view of the INGV building (in red the boreholes performed in 2010, in blue the boreholes performed in 2014, in green the section investigated with the 2D FEM analysis).

Table 1. Main geotechnical properties – INGV building case-history. Layer

H (m)

V s (m/s)

ν

ρ (t/m3 )

G0 (kPa)

E 0 (kPa)

sandy silt (S1)

6.7

179

0.25

2.04

65363

163409

clay1 (A1)

2.8

193

0.25

2.01

74870

187176

clay2–1 (A2–1)

1.6

197

0.25

2.01

78006

195015

clay2–2 (A2–2)

6.6

222

0.25

2.01

99061

247652

clay2–3 (A2–3)

7.2

253

0.25

2.01

128658

321645

clay2–4 (A2–4)

7.2

277

0.25

2.01

154225

385563

clay2–5 (A2–5)

7.2

310

0.25

2.01

193161

482903

Fig. 6. FEM model of the 2D section of INGV building highlighted in green in Fig. 5b.

accelerogram recorded at the Sortino station during the 1990 earthquake (f = 2.38 Hz). To fit the accelerograms to the reference area, they were scaled at the same maximum expected acceleration (PHA = 0.282 g), corresponding to the SLV state (i.e. the limit state for the safety of human life) and considering the building as “strategic” type (corresponding to a return period of 975 years), in accordance with [42]. The synthetic

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seismograms were also scaled using attenuation relations, considering the epicentral distances for both the earthquake scenarios. More details can be found in [44]. Figure 7 reports the amplification ratio R along the free-field alignments on the left (FFleft ) and on the right (FFright ) and along the alignment crossing the structure (SSI), see also Fig. 6. Comparing the FFleft and FFright alignment results, it is possible to observe some differences due to the different soil profiles. Comparing the results for both the FF and the SSI alignments, it is possible to observe a good agreement for the 1693 inputs, definitively greater R values along the SSI alignment for the 1818 inputs and lower values of R along the SSI alignment for the 1990 input. The different response in terms of R along the FF and SSI alignments depends on DSSI phenomena and the relation between the frequency content of the input applied at the base of the FEM model and the natural frequency of the system, equal to 0.94 Hz considering only the soil (FF alignments) and to 0.88 Hz including the building (SSI alignment). For the 1818 inputs, their predominant frequencies are closer to the natural frequency of the entire soil-building system, so the acceleration at the base of the building is higher than those in FF conditions. In the latter case, the DSSI is detrimental. Moreover, the amplification ratios computed are generally significantly higher than that suggested by [42], equal to 1.29.

Fig. 7. Acceleration amplification ratio along the FFleft , SSI and FFright alignments shown in Fig. 6 obtained with the FEM modelling compared to the value prescribed by NTC,2008.

Figure 8 shows the average response spectra at the soil surface obtained by the FEM-2D modelling and, in accordance with [42], for a structural damping ratio equal to 8%. It can be noticed that between the average spectra obtained for the free-field conditions (FFleft and FFright ), there are no substantial differences. On the contrary, a notable difference exists between the average response spectra obtained for the free-field conditions and that concerning the SSI alignment. The FF response spectra present the highest peak for T = 0.42 s and a second, less evident peak for T = 1.27 s, while the SSI response spectrum gives the highest peak for T = 1.27 s and a second, less noticeable peak for T = 0.42 s. The natural period of the structure, including the subsoil, is equal to 1.13 s. Thus, the SSI response spectrum indeed reports the most dangerous situation. In this case, the DSSI is detrimental, and neglecting it, would produce an unsafe design. It is also important to underline that the NTC2008 response spectrum [42] does not cover the computed SSI average response spectrum for periods greater than 0.7 s. The latter is often the period of structures on the subsoil (no fixed-base), which is the realistic configuration of structures.

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Fig. 8. Comparison of the average elastic response spectra obtained by the FEM analyses and that prescribed by [42] at the soil surface.

4 A New Eco-sustainable Geotechnical Solution for Reducing the Seismic Risk of Structures To reduce the seismic energy transmitted to structures, new low-cost and eco-sustainable seismic isolation systems have been investigated, consisting of soil-rubber-mixtures (SRM) underneath the foundations. Seismic energy will be partially dissipated within this mixture before being transmitted to the structures. The first experimental campaign on a full-scale prototype structure resting on gravelrubber mixtures (GRMs) was performed in 2019 in Thessaloniki, Greece [45]. The EuroProteas prototype structure (Fig. 9a) was chosen for the experiment. This is a perfectly symmetric steel frame on a RC slab and supports two RC slabs. Steel X-braces connect the four steel columns in both directions. Over the last few years, many tests have been performed on this structure. More details can be found in [46–48]. In 2019, it was used within the framework of the European SOFIA-SERA project, after replacing the foundation soil with three different GRMs to a thickness of 0.5 m. The study investigated the rubber content effect of the GRM layer on the dynamic response of the EuroProteas structure and the overall performance of the GRMs as a geotechnical seismic isolation (GSI) system. Both free-vibration and forced-vibration tests were performed at different excitation levels, with and without rubberised soil underneath the foundation. This paper deals with the sets of forced-vibration tests. The uppermost 0.5 m of the foundation soil was replaced with three different GRM backfills (Fig. 9b). The rubber content per mixture weight (p.w.) was fixed equal to 0%, 10%, and 30% for the three foundation mixtures labelled in the following as GRM 100/0, GRM 90/10, and GRM 70/30, respectively. The gravel used was characterized by D50,G = 20.76 mm and Gs,G = 2.67. The rubber was recycled granulated rubber from car tyres characterized by D50,R = 3.27 mm, and Gs,R = 1.10. Figure 10a shows their grain size curves. GRM 100/0 was characterized by Gs,G = 2.67, Dr = 98%, γ d = 16.2 kN/m3 ; GRM 90/10 was characterized by Gs,G = 2.51, Dr = 98%, γ d = 15.2 kN/m3 ; GRM 70/30 was characterized by Gs,G = 2.19, Dr = 59–71%, γ d = 11.8 kN/m3 . The three GRMs were located in three different 3.2 x 3.2 x 0.5 m square-plan pits. A thin geotextile layer was placed to cover the pit to prevent the GSI from mixing with the underlying

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and surrounding fine-grained soil material. The GRM foundation soil was spread in two layers both 0.25 m in height. Each layer was compacted using a soil compactor.

Fig. 9. a) View of the Europroteas structure and the team involved in the European SOFIA-SERA project [https://sera-ta.eucentre.it/sera-ta-project-40/]; b) Section and plan view of the Europroteas structure including the gravel-rubber mixture.

Fig. 10. a) Grain size distribution curves of the gravel and the rubber used in the full-scale tests; b) the soil pit filled with GRM100/0, GRM90/10 and GRM70/30 and details of the mixtures in each soil pit (modified by [45]).

The soil below the structure was investigated using extensive geotechnical and geophysical surveys, including static and dynamic in-situ and laboratory tests [49]. The subsoil consisted of a 7 m thick upper layer of silty clayey sand, which overlay a layer of clayey to silty sand with gravels between 7 and 22 m and, after that, a layer of marly silt to silty sand up to a depth of 30 m. The shear wave velocity of the uppermost 5 m varied from 100 to 150 m/s and then increased to more than 250 m/s at a depth of 25 m (Fig. 11). The input force was applied to the roof by an eccentric mass shaker. The eccentricity and the operating frequency adjusted the force amplitude according to the equation F = E · (2π f )2 , where F is the shaker force (in N), E is the total eccentricity of the shaker (in kg-m), and f is the rotational velocity of the shaker (in Hz). Only the tests characterized by an eccentricity equal to 6.93 kg-m are shown here due to lack of space.

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The structure was shaken for a time window of 25 s at each excitation frequency to reach a steady state. The input frequency range was 1–10 Hz. Ambient noise measurements made it possible to estimate the natural frequencies of the fixed-base structure (f 1 = 9.13 Hz) and the flexible-base structure resting on GRM 100/0 (f 1 = 4.4 Hz; f 2 = 4.9 Hz; f 3 = 9.7 Hz), GRM 90/10 (f 1 = 4.3 Hz; f 2 = 4.6 Hz; f 3 = 9.4 Hz), and GRM 70/30 (f 1 = 2.4 Hz; f 2 = 2.7 Hz; f 3 = 7.1 Hz). GRM 100/0 and GRM 90/10 led to very close natural frequencies, thus showing that a 10% rubber content is not so influential. The natural frequencies substantially decreased when changing from GRM 90/10 to GRM 70/30. As expected, the increase in the rubber content produced a reduction in the natural frequencies.

Fig. 11. Soil stratigraphy and Vs vs z profile underneath the Europroteas structure.

Many instruments were used to record the response of the structure, the foundation, the GRM layer, and the soil in the vertical “z” direction and the two horizontal NS “N” and EW “E” directions (Fig. 12). The structure was instrumented with accelerometers installed: on the roof, both along the direction of shaking (that is the “N” direction) and at the opposite corners of the roof to capture possible out-of-plane motion; and on the foundation, forming a cross shape to capture its translational, rocking, and possible outof-plane motion. Seismometers were located on the soil surface, both along the loading axis and on the perpendicular axis. The distance between those close to the structure was defined at 0.50 m = B/6, where B is the foundation width. The most distant instrument was installed at 5 m = 5B/3. Finally, a shape-acceleration array equipped with sensors every 0.15 cm was installed immediately below the geometrical center of the foundation to capture the GRM layer response. Accelerometers were buried under the foundation to monitor the full response of the GRMs. More details about the tests can be found in [45]. The experimental test was simulated using the FEM MIDAS FEA-NX code [50]. A comparison between the experimental and the FEM results allowed the authors to validate the FEM model and to use it in future predictions of real case histories. Figure 13

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Fig. 12. Plan view of the instrumentation used for the full-scale tests on the Europroteas structure, including GRMs at the base.

shows the mesh, which was 20 m in total depth (z-direction), 30 m long (x-direction, corresponding to the input direction) and 20 m wide (y-direction). Free field edges were defined along the lateral boundaries of the 3D model. 3D solid elements were used, and the sizes of the mesh elements were fixed so as to reproduce all the waveforms of the whole frequency range under examination. To model foundation uplifting and sliding phenomena, contact surfaces were used between the foundation and the GRMs. A linear-elastic behavior was assumed for the structure, while a hardening soil model was used for the soil and the GRMs. The 3D model was fixed at the base, and a harmonic displacement input was applied to the roof in the x-direction according to the shaker force F used during the tests. Due to a lack of space, only some experimental and numerical results are presented in this paper. Figure 14 shows the experimental results regarding the motion decay along with the structure and soil in terms of maxima accelerations for the input frequencies equal to 2.5 Hz and 4 Hz, close to the resonance frequencies of the GSI-structure systems. For the input frequency f = 2.5 Hz, the highest maxima accelerations occurred in the presence of GRM 70/30 since resonance occurs for this frequency. Comparing the acceleration at the roof with that at the foundation, an almost constant decrease was found for all the GRMs. However, the accelerations differed considerably for GRM 100/10 and GRM 90/10 on the one hand and GRM 70/30 on the other. It depends on the input frequencies and the natural frequencies of the system. Similar results were obtained comparing the accelerations at the foundation and the soil surface (0.5 m from

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the foundation). Finally, moving through the soil surface (from 0.5 m up to 2.0 m from the foundation), there was a different acceleration decrease: the small value obtained for GRM 100/0 may be due to its higher stiffness and more negligible damping. Similar results were found for the input frequency f = 4 Hz. It is important to stress the critical decrease in the acceleration when going from the foundation to the soil next to the foundation, especially in the case of GRM 70/30. This confirms the supposed ability of the recycled rubber included in GRMs to trap the dynamic wave, thus avoiding its passage from the foundation to the soil. So, the GSI of the structure, represented by the GRM layer underneath the foundation, is optimized by increasing the rubber content of the soil rubber mixture up to 30% per mixture weight. For all the GRMs used and the input frequencies 2.0, 2.5, 3.0, 4.0, 5.0 and 7.0 Hz, Figs. 15 and 16 report a comparison between the experimental results and the numerical ones in terms of acceleration in the “N” direction at the roof and the foundation, respectively. The experimental results refer to accelerometer N. 102650 on the roof and to accelerometer N. 102605 on the foundation. The figures generally show an excellent comparison between the experimental results and the numerical ones. Some moderate differences exist at the foundation for GRM 70/30 and the input frequency equal to 7.0 Hz and GRM 100/0 and GRM 90/10 when the input frequency is ≥ 4.0 Hz. It is important to stress that 7.0 Hz is the estimated third natural frequency of the structure resting on GRM 70/30, and 4.0 Hz is very close to the estimated first natural frequencies of the structure resting on GRM 100/0 and GRM 90/10. When approaching the resonance frequency, some differences between experimental and numerical results are unavoidable. From Figs. 15 and 16, it is also possible to observe an increase in the signal for the input frequencies around 2.5 Hz and 7.0 Hz for GRM 70/30 and around 4.0 Hz for GRM 100/0 and GRM 90/10, according to the estimated natural frequencies of the GSI-structure systems previously discussed. Focusing on GRM 70/30 and the input frequency of 2.5 Hz, Fig. 17 reports a comparison between the experimental results and the numerical ones for the roof, the foundation and the soil. The numerical modelling allowed us to capture the results in the structure and the soil near the structure very well. Some discrepancies exist considering soilstructure distances greater than 1.00 m; see the velocity transducers N. 985F, N. 9086 and N. 908E. However, the numerical simulation completely captures the reduction of the signal from the roof (source point) to the foundation and from the foundation to the soil next to the structure. Finally, Fig. 18 reports the experimental and numerical results in terms of acceleration response factor Ra versus the input frequency. Ra was calculated as the ratio of the recorded acceleration on the roof over the shaker force divided by the superstructure mass. The experimental results refer to accelerometer N. 102650 in the “N” direction. It is possible to observe that the frequency of the system shifted to 4 Hz for both GRM 100/0 and GRM 90/10, but the values of Ra using GRM 90/10 were slightly smaller than those using GRM100/0, indicating a modest increase in the damping due to the presence of rubber in GRM 90/10. The resonant frequency of the structure resting on GRM 70/30 presented a pronounced shift to 2.5 Hz, underlining the strong effect of the rubber on the stiffness reduction of the GSI-structure system. Considering the structure resting on GRM 70/30 the Ra values were much lower than those for the structure resting on GRM

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100/0 and GRM 90/10, indicating a significant increase in the damping due to the greater rubber content. The numerical results precisely confirm the experimental ones.

Fig. 13. Mesh made with the FEM MIDAS FEA-NX code for simulating the full-scale tests on the Europroteas structure.

Fig. 14. Maxima acceleration decay along the structure and on the soil recorded during the Europroteas tests for input frequencies equal to 2.5 Hz and 4.0 Hz.

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Fig. 15. Experimental versus numerical accelerations on the roof along the “N” direction.

Fig. 16. Experimental versus numerical accelerations at the foundation along the “N” direction.

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Fig. 17. Comparison between the experimental and numerical results for the soil-structure system, including GRM 70/30 and an input frequency equal to 2.5 Hz (“N” direction).

Fig. 18. Experimental and numerical acceleration response factors Ra versus the input frequency (“N” direction).

5 Conclusions The seismic retrofitting of historic structures and the safeguarding of the ecosystem are two of the most pressing challenges at the present moment in time. Europe has a rich heritage of old structures of great architectonic and historical value, which unfortunately are characterized by a high seismic vulnerability and are located in areas with a mediumhigh seismic hazard. At the same time, waste management is pushing for waste to be transformed into valuable resources. This paper presents two levels of analysis for investigating structural seismic risk, including the fundamental presence of the soil. The first level is based on sophisticated local site response analyses. The second level is based on the investigation of fully

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coupled soil-structure systems, which most closely reflect the actual structural configurations. These levels of analysis are discussed with reference to two interesting case histories. Finally, the first experimental campaign performed on a full-scale prototype structure resting on gravel-rubber mixtures (GRM) and the related FEM simulation are presented. Rubber grains are manufactured from scrap tyres, and the aim is to test GRMs as a geotechnical seismic isolation (GSI) that combine seismic risk mitigation and ecosystem protection. The main results can be summarized as follows: – LSR analyses allow us to produce seismic microzonation maps, based on real soil profiles and including soil nonlinearity. LSRs are generally performed in free-field conditions. The critical role played by the soil must be investigated in the time and frequency domains. The case history discussed reports the most severe responses for the inputs characterized by low peak values of amplitude and predominant frequencies close to the natural frequencies of the soil. Possible resonant phenomena in the soil can be appropriately captured only through accurate LSR analysis. An approximate evaluation of the soil filtering effect can give a wrong evaluation of the real inputs affecting the structures. – DSSI analyses of fully coupled soil-structure systems often furnish values of spectral acceleration higher than those given by the technical regulations or obtained by LSR analysis in free-field conditions for periods higher than 0.7 s, which are typical of structures with an unrealistic fixed base. Nowadays, DSSI analyses should be used much more as part of routine design, recognizing the inadequacy of some of the simplistic assumptions that have been imposed in the past for reasons of calculational expediency. – The experimental tests on the full-scale prototype structure resting on GRMs, confirm the supposed ability of the recycled rubber included in the GRMs to trap the dynamic wave, thus avoiding its passage from the foundation to the soil and vice versa. The GSI of the structure, represented by the GRM layer underneath the foundation, is optimised by increasing the rubber content of the gravel-rubber mixture up to 30% per mixture weight.

Acknowledgements. The results that the authors have reported in this paper were made possible by financial support provided by the POR_FESR Sicilia 2007–2013 (Line 4.1.1.1) Project, funded by the European Community, and by the Transnational Access Project “SOFIA: Soil Frame-Interaction Analysis through large-scale tests and advanced numerical finite element modeling” funded under the European project “Seismology and Earthquake Engineering Research Infrastructure Alliance for Europe – SERA-TA – H2020 (Grant Agreement 730900)”.

References 1. Caruso, S., Ferraro, A., Grasso, S., Massimino, M.R.: Site response analysis in eastern Sicily based on direct and indirect vs measurements. In: 1st IMEKO TC4 International Workshop on Metrology for Geotechnics, MetroGeotechnics 2016 Proceedings, pp. 115–120 (2016)

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21. Amendola, C., de Silva, F., Vratsikidis, A., Pitilakis, D., Anastasiadis, A., Silvestri, F.: Foundation impedance functions from full-scale soil-structure interaction tests. Soil Dyn. Earthq. Eng. 141, 106523 (2021) 22. Brunelli, A., et al.: Numerical simulation of the seismic response and soil–structure interaction for a monitored masonry school building damaged by the 2016 Central Italy earthquake. Bull. Earthq. Eng. 19(2), 1181–1211 (2020). https://doi.org/10.1007/s10518-020-00980-3 23. Mylonakis, G., Gazetas, G.: Seismic soil-structure interaction: beneficial or detrimental? J. Earthq. Eng. 4(3), 277–301 (2000) 24. D’Amato, M., Gigliotti, R., Laguardia, R.: Seismic isolation for protecting historical buildings: a case study. Front. Built Environ. 5, 87 (2019). https://doi.org/10.3389/fbuil.2019. 00087 25. Gatto, M.P.A., Lentini, V., Castelli, F., Montrasio, L., Grassi, D.: The use of polyurethane injection as a geotechnical seismic isolation method in large-scale applications: a numerical study. Geosciences 11(5), 201 (2021) 26. Micozzi, F., Scozzese, F., Ragni, L., Dall’Asta, A.: Seismic reliability of base isolated systems: sensitivity to design choices. Eng. Struct. 256(5), 114056 (2022) 27. Anastasiadis, A., Senetakis, K., Pitilakis, K.: Small-strain shear modulus and damping ratio of sand-rubber and gravel-rubber mixtures. Geotech Geol. Eng. 30, 363–382 (2012) 28. Tsang, H.H., Lo, S.H., Xu, X., Sheikh, M.N.: Seismic isolation for low-to-medium-rise buildings using granulated rubber-soil mixtures: numerical study. Earthq. Eng. Struct. Dynam. 41, 2009–2024 (2012) 29. Tsang, H.H., Pitilakis, K.: Mechanism of geotechnical seismic isolation system: analytical modeling. Soil Dyn. Earthq. Eng. 122, 171–184 (2019) 30. Tsang, H., Tran, D., Hung, W., Pitilakis, K., Gad, E.F.: Performance of geotechnical seismic isolation system using rubber-soil mixtures in centrifuge testing. Earthq. Eng. Struct. Dyn. 50, 1271–1289 (2020). https://doi.org/10.1002/eqe.3398 31. Pistolas, G.A., Pitilakis, K., Anastasiadis, A.: A numerical investigation on the seismic isolation potential of rubber/soil mixtures. Earthq. Eng. Eng. Vib. 19(3), 683–704 (2020). https:// doi.org/10.1007/s11803-020-0589-3 32. Tasalloti, A., Chiaro, G., Murali, A., Banasiak, L., Palermo, A., Granello, G.: Recycling of end-of-life tires (ELTs) for sustainable geotechnical applications: a New Zealand perspective. Appl. Sci. 11(17), 7824 (2021) 33. Tasalloti, A., Chiaro, G., Banasiak, L., Palermo, A.: Experimental investigation of the mechanical behaviour of gravel-granulated tyre rubber mixtures. Constr. Build. Mater. 273, 121749 (2021) 34. Chiaro, G., Tasalloti, A., Chew, K., Vinod, J.S., Allulakshmi, K.: Macro and microscale engineering response of rigid-soft gravel-rubber inclusions: insights from detailed laboratory and DEM numerical investigations. In: Gupta, A.K., Shukla, S.K., Azamathulla, H. (eds.) Advances in Construction Materials and Sustainable Environment. LNCE, vol. 196, pp. 11– 27. Springer, Singapore (2022). https://doi.org/10.1007/978-981-16-6557-8_2 35. Boschi, E., Guidoboni, E, Ferrari, G., Mariotti, D., Valensise, G., Gasperini, P.: Catalogue of strong Italian earthquakes from 461 b.c. to 1997. Ann. Geophys. 3(4) (2000) 36. Camassi, R., Stucchi, M.: - NT4.1, a parametric catalogue of damaging earthquakes in the Italian area (release NT4.1.1). GNDT, Milano, Italy (1998) 37. Bottari, A., Saraò, A., Teramo, A., Termini, D., Carveni, P.: The January 11th, 1693 SouthEastern Sicily earthquake: macroseismic analysis and strong motion modelling at Noto. In: Maugeri, M., Nova, R. (eds.) Geotechnical Analysis of the Seismic Vulnerability of Historical Monuments, Pàtron, Bologna (2003) 38. Cavallaro, A., Maugeri, M.: Site characterisation by in-situ and laboratory tests for the microzonation of Noto. In: Maugeri, M., Nova, R. (eds.) Geotechnical Analysis of the Seismic Vulnerability of Historical Monuments, Pàtron, Bologna (2003)

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Seismic Design Using Advanced Technologies: An Environmental Point-of-View Oren Lavan(B)

and Ohad Idels

Technion – Israel Institute of Technology, 32000 Haifa, Israel [email protected]

Abstract. Humanity is now facing a climate emergency. If global warming continues to increase in its current rate, the related catastrophic breakdown of climate expected in a few decades may result in the entire eco-systems being destroyed. Thus, humanity needs to take strong and immediate actions. The built environment sector is currently responsible for 30% of global carbon emissions and other environmental effects and is a key player in mitigating the worst effects of climate breakdown. With the fruitful efforts put in operational energy consumption and operational carbon reduction, reduction of embodied carbon is now taking the focus. This is more pronounced in seismic regions, where seismic damage requires repairs that directly translate to environmental impacts (EIs). Thus, EIs of buildings should be added to life-cycle cost of buildings and minimized. Peak and residual inter-story drifts, and peak floor accelerations contribute to the EIs of buildings in seismic regions. Consequently, the selection and design of the seismic resisting system has a critical effect. The use of advanced technologies in the form of energy dissipation devices (EDDs, e.g., Fluid Viscous Dampers) may strongly assist in reducing the EIs of buildings. At the construction stage, the use of EDDs can reduce sizes of elements in the seismic resisting system and therefore material usage. In addition, throughout the lifetime of the building, the use of EDDs may mitigate future damage due to earthquakes, and the associated need for repairs that would have further EIs. Although these effects of incorporating EDDs on the EIs of buildings in seismic regions seem intuitive, they have never been examined. Thus, this paper examines and discusses these effects while considering both EIs and life-cycle cost aspects. This is done with the assistance of advanced optimization problem formulations and solution schemes. Keywords: Seismic design · Energy dissipation device · Fluid viscous damper · Environmental impact

1 Introduction Humanity is now facing a climate emergency. If global warming continues to increase in its current rate, warming is expected to reach 1.5 °C between 2030 and 2052 [1]. The catastrophic breakdown of climate is suspected to result in the severe damage to the entire eco-systems. To avoid such consequences, humanity needs to take strong and immediate actions. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 23–31, 2023. https://doi.org/10.1007/978-3-031-26879-3_2

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The built environment sector is currently responsible for up to 30% of annual global greenhouse gas emissions [2]. Thus, it is a key player in mitigating the worst effects of climate breakdown. In terms of carbon emission, or carbon emission equivalent, operational carbon emission has been dominant in buildings. However, fruitful efforts were put in operational energy consumption reduction and hence operational carbon reduction. Consequently, reduction of embodied carbon is now taking the focus [2]. Energy consumption and carbon footprint due to construction are usually thought as the dominant components in embodied carbon of buildings. However, in seismic regions, seismic events through the life of the building may cause various levels of damage to structural and non-structural components. This damage and the related repairs directly translate to environmental impacts [3]. Consequently, it has been lately shown that in seismic regions, the contribution of repairs due to seismic damage is an important factor as well, and may drastically affect the cradle-to-grave environmental impacts of buildings [4]. In some cases, depending on the climate and seismicity, the environmental impact associated to seismic risk, in terms of CO2 equivalent, could potentially be as high as the annual operational CO2 equivalent after thermal refurbishment [5]. As both inter-story drift ratios and peak floor accelerations were found to contribute considerably to the environmental impacts of buildings in seismic regions, the seismic resisting system is expected to have a critical effect. Indeed, this is highly reflected in the literature [6–10]. This probably motivated the works by [11–13] for minimizing the environmental effects of buildings and bridges. The high effect of the structural system, and the high contributions of both inter-story drifts and floor accelerations to the environmental impacts motivates the use of passive control devices [14–16]. This was also highlighted by [17–19] who suggested that the environmental performance of buildings in seismic regions could be improved using passive control devices (e.g. seismic isolation or viscous fluid dampers). These have been shown very efficient in reducing other life-cycle performances of buildings [20] as well as the probability of collapse [21] – hence the consequences of future events. This paper aims to explore the effect of incorporating Fluid Viscous Dampers (FVDs) in new buildings on their life-cycle environmental effects. This paper presents a modified version of the methodology proposed in [22], which was formulated to account for an economic loss, only. Incorporating FVDs in new buildings has been shown to reduce the total volume of the structural system hence reduce its total cost [20]. Thus, it is suspected to also assist in reducing the life-cycle environmental impacts. This hypothesis will be tested here by first formulating an optimization problem that minimizes a weighted sum of the life-cycle cost of the building and its life-cycle CO2 emission. The optimization will determine the cross sections of the structural elements as well as the locations and sizes of the dampers. Eventually, the optimal designs with and without dampers will be compared.

2 Methodology This section presents a framework for the optimal design of steel Moment-ResistingFrames (MRFs) equipped with Fluid-Viscous-Dampers (FVDs) in seismic regions. The methodology optimally designs both the structural elements and the viscous dampers.

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This is done while minimizing a weighted sum of the life-cycle (initial plus expected) cost and the life-cycle environmental impact in the form of CO2 equivalent (CO2 e). The expected losses (cost) due to future seismic events as well as the expected environmental damage due to future repairs are assessed using the PEER framework while relying on Nonlinear-Time-History-Analysis (NTHA). Furthermore, practical constraints to ensure an elastic behavior under gravity loads, the strong column-weak beam mechanism and minimum lateral strength of the MRF are considered. 2.1 Nonlinear Time History Analysis The dynamic response of the structure, that is required to assess the expected losses and the environmental impact due to seismic events, is evaluated using NTHA, where both the steel elements and FVDs nonlinear behaviour are considered. The equations of motion (Eq. (1)) are solved using NTHA which is a modified version of the one utilized by Pollini et al. 2018 [23]. This NTHA tool relies on the Newmark-beta integration scheme, and the Newton-Raphson algorithm is utilized to solve the nonlinear equations obtained within each time-step. Furthermore, second order effects (P − ) are considered using negative stiffness matrix as in [24]. ˙ + fs (t) + fd (t) = −Meag (t) ¨ + Cs u(t) Mu(t)

(1)

where M and Cs are the mass and inherent damping matrices, respectively; is vector of displacements of the degrees of freedom relative to the ground; is the influence vector; fs is a vector of the nonlinear resisting forces of the structural elements, which evaluate using a distributed plasticity element as in [25]; fd is the vector of the FVDs’ resisting forces, which are modelled according to the Maxwell model and evaluated over time as in [26] and ag is the ground acceleration. 2.2 Problem Formulation Initial Cost\Carbon Emission: The goal of the optimization is to minimize, among other things, the combined initial cost and initial CO2 e, taking into account the total cost/CO2 e of both the steel elements and dampers. To enable the use of gradient-based optimization approach the functions must be formulated as differentiable. The steel construction cost (Jstr ) is given in Eq. (2) and correlated to the total volume of steel. This also holds for the CO2 e of steel (with an appropriate βs value). The total cost/CO2 e of the dampers (Jdamp ), is the sums cost/CO2 e of all the dampers and assumed to be correlated to the square root of the peak force of each damper. This approach was presented and casually described in Pollini et al. 2018 [23]. As this cost function is based on the volume of materials comprising the damper, it is assumed that such a function also holds for the CO2 e (with an appropriate βs andβd values): Jstr = βs

Nele  i=1

   maxFd ,j 

Ndemp

Ai · Li ; Jdamp = βd

j=1

t

(2)

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where βs and βd are used to scale the cost and\or CO2 e of the construction steel and the FVDs, respectively; Ai and Li are the cross-section area and the length of the i-th member, respectively; maxFd ,j  is the absolute peak force at the j-th damper. t

Expected Loos\Carbon Emission: To evaluate the expected loss, the procedure presented by Aslani and Miranda [27] is adopted. Moreover, to reduce the computational cost, a regression analysis has been utilized as proposed by Aslani and Miranda [28]. Equation 3 utilize to evaluate the present expected loss of the building as follows: E[LT ] =

1 − e−λt ∞ ∫ E[LT |IM ]|d ν(IM )| λ 0

(3)

where λ is the discount rate per year; t represent the building design lifetime; E[LT |IM ] is the expected annual loss for a given intensity measure; IM , and ν(IM ), are the mean annual rate of exceedance for a given intensity measure (IM ), respectively. Equation 4 is utilized to assess the expected annual loss for a given intensity measure (E[LT |IM ]) as follows: E[L|IM ] =

N edp ds N  

E[L|DS = dsi ] · P(DS = dsi |EDP = edpk )

i=1 k=1

(4)

·|P(EDP ≥ edpk |IM )|

where E[L|DS = dsi ] is the expected annual loss for a given damage state (dsi ); P(DS = dsi |EDP = edpk ) is the probability for the i-th (dsi ) damage state to acquire for a given EDP and may be evaluated using fragility curves as presented by Aslani and Miranda [27]. In addition, a number of code requirements are considered – for example, the strong column weak beam capacity design requirement and a demand that members should remain elastic under both gravity and the equivalent lateral load (ELF) cases. These constraints are follows the ASCE 7-16 requirements. min αcost (ELoss + ICost) + βCo2 (ECO2 + ICO2 ) s.t. My ≥ MdELF (uELF , FELF )  ∀j = 1 . . . Nsec gr  ∀j = 1 . . . Nsec My ≥ Md ugravity , Fgravity   Mc ≥ γ · Mg ∀k = 1 . . . Njoints

(5)

αcost and βCo2 are two weights  or  that control if the optimization is more cost orientated environmental orientated. Mc is the sum of the columns yielding moments and Mg is the sum beams yielding moments framing the same joint, respectively, and γ is a safety gr factor due to the beam overstrength (e.g. 1.3). MdELF and Md are the design moments developed under the ELF and gravity load combinations, respectively; My is the yield moment of the relevant member. To enable the use of gradient-based optimization, the constraints are reformulated in a differentiable form. The reformulation is similar to that presented by Idels and Lavan [29] for the life-cycle cost optimization where it is fully described.

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2.3 Solution Scheme The optimization problem is solved using an efficient gradient-based optimization. In order to adopt an efficient first order optimization tool, the gradients are required. These are derived using the adjoint sensitivity analysis approach similarly to [29] to further reduce the computational efforts. The gradient-based optimization approach, that was adopted, is the sequential linear programming (SLP) approach that was found suitable for similar problems [29].

3 Example The example considers the five story 3-bay y 3-bay building presented in Fig. 1 that is adopted from [20]. As the building is symmetric, this example considers only one component of the ground motion in the direction depicted in the figure. Thus, torsional effects can be ignored and only one of the seismic MRFs is considered, along with its corresponding gravity loads and seismic mass. The design variables are the beams and columns sections as well as the damping coefficients of the dampers, while symmetry is assumed as can be seen from Fig. 1. The weight per unit area is assumed as 5.6 kN/m2 , which leads to a seismic mass of 90 tons per frame per floor. In addition, a Reyleigh damping matrix based on 5% damping in the first and the third modes is assumed. Note that this matrix depends on the design variables, as they affect the stiffness matrix. This is accounted for as in [29]. The seismic hazard is adopted from the LA region and is represented by the hazard curve given in Fig. 2. To assess the expected cost/CO2 e, the LA suit of ground motions was adopted and scaled to fit the hazard curve. To focus on the environmental impact of added FVDs the following example considers environmental effects only, thus αcost is set equal to zero. The initial CO2 e (ICO2 ) of the structure accounts for the steel elements and FVDs. The CO2 e of the steel frame elements is set to 3.78 ton of CO2 e per ton of steel leading to. βs = 29.6 ton · CO2 e/m3 . CO2 e of the FVDs is computed based on 5.4 tons of CO2 e for a damper with a capacity of 3,000kN and is set √ proportional to the square root of the force (Eq. 2). This leads to a βd = 0.1 ton · CO2 e/ kN. The expected carbon emission during the lifetime of the building (ECO2 ) due to seismic events is evaluated based on 1 ton of CO2 e per square meter. The numerical example compares the environmental performance of the same frame in the same region with and without FVDs. In both cases (with and without FVDs), the design is carried out by the same optimization algorithm, presented in this paper. Figure 3 shows the optimal design of the frame with FVDs (Fig. 3a) and without FVDs (Fig. 3b). The optimal design in the problem, that considered potential FVDs indeed, has dampers in the first four stories (Fig. 3a). The widths of the lines represent the cross-section properties (thicker line for heavier sections). It can be seen that the optimal design, that considers the possibility of utilizing FVDs, requires a significantly smaller total weight of the steel elements compared to the optimal design without FVDs (Fig. 3b). This is due to the positive effect of the FVDs in reducing damage in future seismic events. The environmental performance of both frames, considering the initial CO2 e (ICO2 ), expected CO2 e (ECO2 ), and the life cycle CO2 e, which is the sum of the two (LCCO2 ),

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Fig. 1. Plan view and elevation view of the considered frame.

Fig. 2. Hazard curve for a structure with a natural period of 1.0 s in LA.

(a) with FVDs

(b) without FVDs

Fig. 3. Optimal design of the frames. (a) steel distribution and FVDs (b) steel distribution.

are presented in Fig. 4. Also, evolution of each component (ICO2 , ECO2 and LCCO2 ) during the optimization process is shown in Fig. 5. The LCCO2 of the frame with FVDs is significantly lower, 96.0 tons compared to 146.7 tons without FVDs. While the ICO2

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of both frames are similar, the ECO2 of the frame with FVDs is lower due to the much higher seismic performance level. (a) with FVDs

(b) without FVDs

Fig. 4. Convergence of the optimization (a) with FVDs and (b) without FVDs.

As can be seen, Fig. 5a shows that although the optimal design heavily relies on FVDs, the environmental cost of the FVDs is relatively low compared to that of the steel elements. (a) with FVDs

(b) without FVDs

Fig. 5. Convergence of the ICO2 (a) with FVDs and (b) without FVDs.

4 Conclusions This paper presented a methodology for the optimal seismic design of steel MRFs with FVDs, accounts for both environmental and economic aspects. The methodology is formulated to minimize to life cycle cost of the building from an environmental and an economic point of view. The environmental impact of both the initial stage related to the construction and the expected CO2 e related to repair actions due to seismic events are considered. In addition, a number of code requirements are taking into account.

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The presented approach relies on NTHA, considering both the nonlinear behaviour of the MRF members and FVDs. To achieve efficiency, a gradient-based optimization that utilizes the adjoint method is adopted. The numerical example shows that optimal design using FVDs may be a promising strategy for structures with a high environmental performance level in seismic regions. The environmental impacts of the optimal frame with FVDs were compared to those of the optimal frame without FVDs and showed much smaller life cycle CO2 e.

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17. Sarkisian, M., Brunn, G., Nasr, M., Hachem, M., Hu, L.: Predicting the environmental impact of structures in regions of high seismic risk. In: AEI 2011: Building Integration Solutions, pp. 263–271 (2011). 18. Sarkisian, M.P.: Design of environmentally responsible structures in regions of high seismic risk. Struct. Infrastruct. Eng. 10(7), 849–864 (2014) 19. Chhabra, J.P., Hasik, V., Bilec, M.M., Warn, G.P.: Probabilistic assessment of the life-cycle environmental performance and functional life of buildings due to seismic events. J. Archit. Eng. 24(1), 04017035 (2018) 20. Idels, O., Lavan, O.: Optimization-based seismic design of steel moment-resisting frames with nonlinear viscous dampers. Struct. Control. Health Monit. 28(1), e2655 (2021) 21. Chalarca, B., Filiatrault, A., Perrone, D.: Seismic demand on acceleration-sensitive nonstructural components in viscously damped braced frames. J. Struct. Eng. 146(9), 04020190 (2020) 22. Idels, O., Lavan, O.: Seismic design of steel moment resisting frames with fluid viscous dampers using optimization. In: Mazzolani, F.M., Dubina, D., Stratan, A. (eds.) Behaviour of Steel Structures in Seismic Areas, pp. 709–716. Springer, Cham (2022). https://doi.org/10. 1007/978-3-031-03811-2_77 23. Pollini, N., Lavan, O., Amir, O.: Optimization-based minimum-cost seismic retrofitting of hysteretic frames with nonlinear fluid viscous dampers. Earthq. Eng. Struct. Dynam. 47(15), 2985–3005 (2018) 24. Rutenberg, A.: A direct P-delta analysis using standard plane frame computer programs. Comput. Struct. 14(1–2), 97–102 (1981) 25. Spacone, E., Ciampi, V., Filippou, F.C.: A Beam Element for Seismic Damage Analysis. Earthquake Engineering Research Center, Berkeley (1992) 26. Kasai, K., Oohara, K.: Algorithm and computer code to simulate response of nonlinear viscous dampers. In: Passively Controlled Structure Symposium, Yokohama, Japan (2001). (in Japanese) 27. Aslani, H., Miranda, E.: Probabilistic earthquake loss estimation and loss disaggregation in buildings, Report 157. PhD thesis, Ph. D. Dissertation, John A. Blume Earthquake Engineering Center, Stanford (2005) 28. Aslani, H., Miranda, E.: Probability-based seismic response analysis. Eng. Struct. 27(8), 1151–1163 (2005) 29. Idels, O., Lavan, O.: Performance-based seismic retrofitting of frame structures using negative stiffness devices and fluid viscous dampers via optimization. Earthq. Eng. Struct. Dynam. 50(12), 3116–3137 (2021)

Sustainable Building: The Role of the Soil Parameters on Earthquake Safety Glenda Abate , Angela Fiamingo(B)

, and Maria Rossella Massimino

Department of Civil Engineering and Architecture, University of Catania, Via Santa Sofia 64, 95125 Catania, Italy [email protected]

Abstract. Nowadays, one of the main goals is to make our buildings more ecosustainable. In earthquake-prone areas, such as Italy, eco-sustainability must be linked to earthquake safety, encouraging specific measures to mitigate the environmental impact of earthquakes. For the assessment of seismic risk, it is essential to look at the local soil condition, which can dramatically modify the transmitted motion to the structure. The real seismic amplification effects are often underestimated by the building regulations. Only FEM analyses of the fully coupled soil-structure systems make it possible to assess the filtering effects due to the soil conditions and the influence of DSSI. In this paper, the influence of the soil and DSSI were investigated on a typical residential Italian building, designed without anti-seismic criteria. The structure is placed in Fleri (Catania, Italy) and suffered severe damage during the earthquake of the 26th December 2018. Different 2D FEM analyses were performed, varying the seismic inputs. Based on the numerical results, new elastic response spectra are proposed and compared to the one of the in-force Italian Building Regulation. The results highlight how local soil conditions are an essential tool for evaluating the earthquake safety of existing and new structures. Keywords: Reinforced concrete building · 2D FEM analyses · Seismic risk

1 Introduction In the last years, eco-sustainability is a fundamental requirement in urban planning and structure retrofitting, with the goal to make our cities more liveable, comfortable, inclusive, resilient and safe [1]. Natural hazards, like earthquakes, severely test the safety of urban centers, highlighting the necessity to promote specific actions to mitigate the seismic risk for all buildings, especially for those with high vulnerability. Italy is one of the most earthquake-prone countries in Europe and with a century-old built environment, not designed to withstand significant seismic forces. For the assessment of seismic risk, it is essential to look not only at the structural performance but also at the local soil condition. The transmitted motion to the structures can be dramatically modified in terms of amplitude, duration, and frequency content by the stratigraphy and mechanical properties of the soil. The conventional building codes [2, 3] define a soil factor (S) to © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 32–47, 2023. https://doi.org/10.1007/978-3-031-26879-3_3

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evaluate the site effects, based on simplified approaches. But, many experimental and numerical analyses have recently shown that the soil factor could be not completely appropriate, suggesting a possible review [4–6]. A realistic evaluation of seismic risk requires the knowledge of the soil filtering effects, as well as of the effects of kinematic and inertial interaction between the soil and the structure (the so-called Dynamic Soil-Structure Interaction (DSSI)). The soil filtering effects can be evaluated through local seismic response (LSR) analyses; nevertheless, they are generally carried out in free-field conditions omitting the fundamental role played by the structures [7, 8]. The DSSI effects can be evaluated through in-situ and laboratory tests [9, 10] or numerical approaches [10–12]. Among these, FEM modelling allows a realistic evaluation of the DSSI effects, taking into account the soil properties, geometry, nonlinearities and boundary conditions [10]. Therefore, a numerical modelling of the soil-foundation-structure system appears a very useful tool for estimating the dynamic behaviour of the structure and the surrounding soil. This paper is aimed at investigating the influence of the soil and DSSI effects on a typical Italian reinforced concrete (RC) structure. The structure, intended to sustain vertical loads only, is situated in Fleri (Catania, Italy), and it suffered severe damage during a recent earthquake that hit the Mt. Etna’s eastern flank. A set of 2D FEM analyses on the soil-structure system were performed, varying the seismic inputs. In all the analyses, the equivalent visco-elastic behaviour and the visco-inelastic behaviour were adopted for the soil and the structure, respectively. Based on the achieved numerical results, new elastic response spectra are proposed and compared to the one of the inforce Italian Building Regulation (NTC, 2018) [3]. The results highlight that the local soil conditions play an essential role in preserving existing structures and planning new ones.

2 Earthquake Data for the FEM Analyses Along the Mt. Etna’s eastern flank, one of the most active volcanoes in the world, many tectonic processes take place due to the Timpe Fault zone, producing frequent shallow seismic events [14, 15]. Thus, being very shallow, they cause significant damage relative to the magnitude (epicentral macroseismic intensity up to IX) [15]. The Fiandaca Fault is one of the most seismically active shear zones, belonging to the Timpe Fault system. On 26 December 2018, a powerful earthquake (local magnitude, M L , of 4.8; moment magnitude, M w , of 4.9) was recorded, located along the Fiandaca Fault, at a depth of less than 1 km [16]. The earthquake struck numerous towns, among them Zafferana Etnea and its village Fleri (Catania). Many buildings, roads and other structures were heavily damaged. The seismic event was recorded in several stations of the Italian Strong Motion Network. The closest stations were respectively 4.5 km and 5.3 km away from the epicentre, called “SVN” and “EVRN”, placed on soil type A [17]. The structure analysed in this paper is situated in Fleri, 4.04 km from the SVN station and 4.90 km from the EVRN station. To evaluate the seismic response of the soil-structure system, nine accelerograms (Fig. 1) were considered: the two above-mentioned ground motions, recorded by SVN and EVRN stations (named ID SVN and ID EVRN), and seven spectrum-compatible

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accelerograms (named ID 004675, ID 000651, ID 007187, ID 000198xa, ID 007156, ID 000198ya, ID 000287). The main properties of the nine accelerograms are shown in Table 1, in terms of M w , PGA (peak ground acceleration), R (epicentral distance).

Fig. 1. Selected seismic inputs for the FEM analyses.

The selection of the seven spectrum-compatible accelerograms was made through the Rexel 3.5 code [18]. They were selected according to the average spectral compatibility requirements prescribed by NTC, 2018 [3]. More in-depth, they are compatible with the disaggregation of the seismic hazard [19] in terms of magnitude (M) and epicentral distance (R), referred to a specific interval (M = 4.0–7.0, R = 0–30 km; Fig. 2), and with the target response spectrum of NTC, 2018. The latter was developed according to

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the following parameters: Soil type: A, Topographic category: T1, Nominal Life (V N ): 50 years, Functional Type: II, and Limit State of interest: SLV. Moreover, each of the seven accelerograms was linearly scaled (by a scale factor, SF, see Table 1) to obtain an average spectrum with a 10% lower tolerance and a 30% upper tolerance, in the range of periods 0.15–2 s, compared to the reference response spectrum. Table 1. Main properties of the earthquakes data for the FEM analyses. ID Acc.

Seismic event

Date

Mw

PGA [g]

R [km]

SF

SVN

Sicilia

26/12/2018

4.9

0.559

4.5

1.0000

EVRN

Sicilia

26/12/2018

4.9

0.301

5.3

1.0000

004675

South Iceland

17/06/2000

6.5

0.225

13.0

1.7116

000651

Umbria Marche

06/10/1997

5.5

0.225

5.0

1.2030

007187

Avej

22/06/2002

6.5

0.225

28.0

0.5054

000198xa

Montenegro

15/04/1979

6.9

0.225

21.0

1.2459

007156

Firuzabad

20/06/1994

5.9

0.225

21.0

0.8804

000198ya

Montenegro

15/04/1979

6.9

0.225

21.0

1.0055

000287

Campano Lucano

23/11/1980

6.9

0.225

23.0

1.6216

Fig. 2. Disaggregation of PGA with a probability of exceedance of 10% in 50 years [19]. T R is the return period of the seismic motion.

Figure 3a shows the elastic response spectra (5% damping) of the accelerograms recorded by the SVN and EVRN stations. Figure 3b shows the elastic response spectra (5% damping) of the 7 accelerograms obtained by the Rexel 3.5 code. The average scale factor SF is 1.16. It is worth highlighting that the response spectrum of the SVN station was significantly higher than the elastic response spectrum provided by the NTC, 2018,

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for the same soil type (A) and damping ratio (5%). Moreover, the period range (T = 0.2–0.4 s) around the one registered (T = 0.26 s) is crucial for RC or masonry structures having one-to-three floors, such as that analysed below. As indicated by previous studies [20, 21] and confirmed by these seismic events, the analysed area has a high level of hazard and the local seismicity deserves to be properly investigated to identify the more exposed zones to seismic shaking.

(a)

(b)

Fig. 3. (a) Elastic response spectra of the two accelerograms recorded by the SVN ed EVRN stations; (b) Scaled elastic response spectra of the seven selected accelerograms achieved by the Rexel 3.5 code.

3 The Building and Its Subsoil The structure scrutinized is in Fleri (Catania, Italy); it has a RC frame with three floors, designed following the 1976 Italian Building Regulation [22] for gravity loads only. This structural typology is common throughout Italy. The chosen frame for the FEM analyses is highlighted in red in Fig. 4a. This latter frame was severely damaged during 26 December 2018 earthquake. The damage is located in the columns between the 2nd stair beam and the 2nd elevation (see the green circle in Fig. 4b). The beam and column cross-sections and reinforcements were evaluated through a simulated design, performed in line with the D.M. 1976 [22]. Having no information on the resistant characteristics of the concrete and of the rebars and having not carried out any tests on the materials, it was assumed that the characteristic cylinder strength of the concrete (f ck ) was equal to 20 MPa and the steel grade of the rebars was FeB38k, usual for the construction practice of the time. The following cross-sections were assumed: 30 × 60 cm for the superstructure beams, 50 × 70 cm for the foundation beam, 30 × 50 cm and 50 × 30 cm for the inside and outside columns on the first and second elevations, 30 × 40 cm and 40 × 30 cm for the inside and outside columns on the third elevation. The area of rebars was evaluated considering gravity loads only. More precisely, rebars with a diameter of 14 mm were assumed for the column sections, while rebars with

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a diameter of 14 mm and 20 mm were assumed for the beam sections. Rebars with a diameter of 8 mm were used for the stirrups. The characteristic values of the permanent and variable loads were considered equal to 6.0 kPa for the deck, 10.1 kPa for the stairs, 8.0 kPa for the balcony and 6.8 kPa for the infills. For more details see [23]. The structure scrutinized was not designed to withstand earthquake motions. So, when it is compared to structures properly designed to sustain earthquake motions, it would be expected to have poor inelastic capacity. But the sizes of the column cross-sections were greater than those strictly required to sustain vertical loads; so, the low axial load ratio improved the local ductility of the members slightly [23].

Fig. 4. (a) View of the building investigated, measured in meters; (b) Shear failure of the column, between the 2nd stair beam and the 2nd floor; (c) Shear wave velocity profile (V S vs z) from MASW surveys.

The structure rests on volcanic soil: scoriaceous lava and volcanoclastic in the first layers, compact lavas under the bedrock. MASW geophysical surveys were performed close to the investigated structure to obtain the shear wave velocity (V s ) along the soil profile. It was possible to identify three different layers of 2.4 m (V s = 209 m/s), 4.6 m (V s = 320 m/s) and 12 m (V s = 460 m/s) respectively (Fig. 4c). So, the soil is type B and the value of amplification factor (S = Ss · ST ) is 1.16, considering the stratigraphic amplification factor (S S ) equal to 1.16 and the topographic amplification factor (S T ) equal to 1 [3]. The peak horizontal ground acceleration (ag ) concerning a probability of exceeding 10% in 50 years is equal to 0.225 g; thus, the peak acceleration at the soil surface (ag · S) is equal to 0.26 g. To consider the soil non-linearity in the FEM analyses, due to the lacking dynamic laboratory tests, the values of the soil shear velocity (V s ), shear modulus at small strain (Gs0 ) and damping ratio at small strain (Ds0 ) were modified according to [24], depending on the ag ·S value, indicated with the symbols V s * , Gs * and Ds * . So, the abatement factors equal to 0.64 and 0.42 were evaluated for V s and Gs0 ,

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respectively, and Ds * = 8.5% was fixed. The soil’s mechanical properties are shown in Table 2. Table 2. Main geotechnical parameters of the foundation soil. Soil parameters

Layer 1

2

3

Bedrock

Unit weight

γ s [kN/m3 ]

18

18

18

22

Friction angle

ϕ’ [°]

33

33

33

40

Cohesion

c

0

0

0

0

Poisson ratio

νs

0.3

0.3

0.3

0.45

Layer thickness

h [m]

2.4

4.6

12.0

11

Shear wave velocity

V s [m/s]

209

320

460

953

Modified shear wave velocitya

V s * [m/s]

135

207

460

953

Shear modulus small strain

Gs0 [kPa]

78625

184320

380880

1998060

Modified shear modulusa

Gs * [kPa]

32875

77068

380880

1998060

Damping ratio small strain

Ds0

0.02

0.02

0.02

0.01

Modified damping ratioa

Ds *

0.085

0.085

0.085

0.01

a Values modified according to [24]

4 The FEM Model The filtering role of the soil and the DSSI effects were investigated through a 2D finite element model involving the soil-structure system, by means of the ADINA code [13]. A visco-elastic behaviour was assumed for the soil, adopting the “modified” parameters in Table 2 while, for the structure, a visco-inelastic behaviour was assumed. According to the soil layer subdivision presented in Sect. 3, the total depth of the model was 19 m. The model had an entire width of 100 m to minimize boundary effects as far as possible (Fig. 5). The soil was modelled by plane strain 4-node 2D-solid elements. As regards the structure, the beams and the columns were modelled by 2-node Hermitian beam elements, considering elastic members with finite-length plastic hinges at the two ends. Several moment-curvature (M-θ ) curves [23] were assigned within the plastic hinge region, equal to the height of the member cross-section, for prefixed values of the axial force. For more details see [23]. The deformability of the bedrock was properly considered using dashpots at the base of the mesh, according to Lysmer and Kuhlemeyer’s formulation [25]. As regards the boundary conditions, the nodes at the base of the dashpot were constrained only in the z-direction. Instead, the nodes of the soil vertical boundaries were linked by “constraint equations” that impose the same y- and z- translations at the same depths. The “constraint equations” were also used to link the horizontal displacements of

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the beams along the y-direction to simulate an axial rigid diaphragm. Contacts between the foundation and the soil were defined to model uplifting and/or sliding phenomena. The friction angle between the foundation and the soil (δ) was fixed equal to 2ϕ’/3. The size of the mesh element was chosen to ensure an efficient reproduction of all the waveforms of the whole frequency range under study (the element size should be 1/8 of the minimum wavelength) and to have a finer discretization near the structure. The material viscosity was modelled according to the Rayleigh damping. The Rayleigh damping coefficients (α R and β R ) were evaluated considering the following equations [26]:  (1) αR = 2 · (D · ω1 · ω2 ) (ω1 + ω2 );  βR = 2 · D (ω1 + ω2 );

(2)

in which D is the damping ratio of each material (assumed equal to Ds * for the soil elements and equal to 5% for the structure), ω1 and ω2 are the first and the second angular frequencies, computed according to the frequencies of the soil (f 1 = 3.52 Hz, f 2 = 10.55 Hz) and the structure (f 1 = 1.37 Hz, f 2 = 4.00 Hz). Distributed loads in the seismic design combination (according to those presented in Sect. 3) were applied to the beams. The mass of all the elements was also considered. The nine acceleration time histories, described in Sect. 2, were applied to the dashpots.

Soil-structure interface Layer 1 Layer 2 Layer 3

Seismic input

Dashpots

DSSI Alignments

19 m

FF Alignment

100 m

Fig. 5. FEM model of the soil-structure system.

5 Results 5.1 DSSI and Soil Filtering Effects In order to check possible DSSI effects, the modification of the acceleration amplitude at the foundation level with respect to the free-field was investigated for all the selected earthquakes. In Fig. 6, the peak acceleration at the foundation level (aEFM ) was compared with the peak ground acceleration at the free field (aFFM ). The comparison was made

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for each DSSI alignment (Fig. 5). All the points above and under the 1:1 line suggest as the dynamic response at the foundation level deviates from the free field motion due to the kinematic and inertial interaction. More precisely, for the points above the 1:1 line, aEFM > aFFM thus, the DSSI effects cannot be neglected. The omission of the foundation motion would lead to an underestimation of the structural response. For the points under the 1:1 line, aEFM < aFFM thus, the DSSI effects can be neglected. While, for the points along the 1:1 line, aEFM = aFFM thus, the DSSI effects do not occur. Moreover, it is worth noting that the peak acceleration at the soil surface defined by NTC, 2018 (ag · S value, see the red square in Fig. 6) is always less than aEFM for any analysed ground motion. The soil amplification phenomena predicted by NTC, 2018 are very different to those actually obtained through numerical analyses. The seismic motion ID SVN, recorded on 26 December 2018, deserves particular attention as the peak ground acceleration at the free field is much higher than the ag · S value. This is due to the high acceleration peak value at the bedrock level (see Sect. 2) and the local soil amplification phenomena.

Fig. 6. Comparison between the peak acceleration at the foundation level (aEFM ) and the peak ground acceleration at the free field (aFFM ), for each DSSI alignment and ground motion.

To investigate more in-depth the behaviour of the points under and above the 1:1 line, the first fundamental frequency of each earthquake (f input ) was compared both to the first fundamental frequency of the soil-structure system (f SSI , represented by diamonds

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in Fig. 7) and to the first fundamental frequency of the soil (f FF , represented by triangles in Fig. 7). When aEFM > aFFM (i.e. when the points are above the 1:1 line of Fig. 6), the f SSI aims to approach the f input (see the red arrow in Fig. 7). On the other hand, when aEFM < aFFM (i.e. when the points are under the 1:1 line of Fig. 6), the f FF aims to approach the f input (see the blue arrow in Fig. 7). So, the detrimental or beneficial effects of DSSI depend on the main properties of the input, the behaviour of the structure and its surrounding soil. These elements are critical to correctly evaluate the seismic risk, especially for all buildings with high vulnerability.

Fig. 7. Comparison between the first frequency of each earthquake (f input ), the first frequency of the soil-structure system (f SSI ) and the first frequency of the soil (f FF ).

The elastic response spectra of the nine accelerograms, evaluated at the foundation level (Fig. 8), are shown. The spectra were also compared with the elastic response spectrum furnished by NTC, 2018 for the soil type of the test site (soil type B, see Sect. 3) and for a return period of 475 years. Two average response spectra were evaluated: the first one was assessed as the average of the seven spectrum-compatible accelerograms (named “Average Spectrum (7 input)”), and the second one was assessed as the average of the totally 9 accelerograms (named “Average Spectrum (9 input)”). The average response spectra are far above the elastic response spectrum furnished by NTC, 2018 for an extensive range of periods. From the achieved results, it is interesting to determine new elastic response spectra that take into account the DSSI, soil filtering effects and soil heterogeneity. See the next section for more details. 5.2 The New Proposed Elastic Response Spectra In this section, the procedure followed to determine the new elastic response will be described. The proposed procedure aims to improve the current elastic response spectrum furnished by NTC, 2018, based on the previous average response spectra achieved at the foundation level (see Fig. 8) through the FEM modelling. Figure 9a shows a flow chart of the procedure followed to obtain the new spectra. It is worth remembering that the shape of the elastic response spectrum (see Fig. 9b) is defined by: ag · S (the value of the spectral acceleration at a period T = 0), T B and T C (the lower and the upper limit of

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the period of the constant spectral acceleration branch, respectively), and T D (the value defining the beginning of the constant displacement response range of the spectrum).

Fig. 8. Elastic response spectra evaluated at the foundation level.

The first step concerns the estimation of the corner periods T B and T C ; it is based on the current prescriptions of the NTC, 2018 and it depends on the investigated soil type. The second step involves the evaluation of the new plateau between the period T B and T C, evaluated in step 1. The third step concerns the determination of the new soil amplification coefficient (S*) and the new the spectral amplification coefficient (F * 0 ). In the final fourth step, the new spectra are implemented using modified equations based on the previous parameters. (a)

1. Estimation of the corner periods TB and TC, for the investigated soil type

(b) Plateau

2. Determination of the new plateau (between TB and TC) 3. Determination of the new amplification coefficients S* and F0* 4. Implentation of the new elastic response spectra

ag·S TB TC

TD

Fig. 9. (a) Flow chart of the procedure followed to obtain the new spectra; (b) main points of the elastic response spectrum.

Step 1. Estimation of the Corner Periods T B and T C for the Investigated Soil Type According to NTC, 2018, the corner periods depend on: the design working life of the structure (named “Nominal Life”, V N ), the importance class of the structure (named

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“Functional Type”, C u ), the reference seismic action and the soil type. In this paper, an ordinary building structure was investigated thus, its Nominal Life is equal to 50 years, the Functional Type is the “II” and the value of C u is 1. According to the no-collapse requirement, a reference seismic action with a 10% probability of exceedance in 50 years was chosen. Therefore, the Return Period (T R ) of the seismic action is 475 years. According to the T R value and the site coordinates, the seismic parameters, the peak ground acceleration (ag ), the spectral amplification coefficient (F 0 ) and the period of constant velocity of the elastic response spectrum (T * C ), can be determined. Finally, assessing the soil type (B, see Sect. 2), it is possible to determine T B and T C using Eqs. 3 and 4. The value of the period T D (see Fig. 9b) can be evaluated through Eq. 5. The aforementioned parameters are summarized in Table 3. TC = CC · TC∗

(3)

 TB = TC 3

(4)

 TD = 4.0 · ag g + 1.6

(5)

Table 3. Main parameters of the elastic response spectrum according to NTC, 2018. Parameters

Value

Parameters

Value

VN

50 yrs

T *C

0.305 s

Cu

1

TC

0.425 s

TR

475 yrs

TB

0.142 s

ag

0.225 g

TD

2.500 s

F0

2.638

Step 2. Determination of the New Plateau Once the T B and T C periods are known, the plateau of the new spectrum (S e (T)plateau ) was evaluated as the average of the spectral acceleration values, included in the range period T B ≤ T ≤ T C , see Fig. 10. The evaluation was made for the motion at the foundation level, considering the Average Spectrum (7 input) and the Average Spectrum (9 input) (Fig. 8). Then, it is necessary to determine the branch of the new spectra between the range period 0 ≤ T < T B . The spectral acceleration at the period T = 0, named as (ag · S)av , was assumed equal to the spectral acceleration at T = 0 of the Average Spectrum. Therefore, the branch between the range period 0 ≤ T < T B was evaluated considering the equation of the straight line through the points (ag · S)av and (S e (T)plateau ) at T = T B , previously determined (Fig. 10). The values of (ag · S)av and S e (T)plateau are reported in Table 4. Step 3. Determination of the New Amplification Coefficient S* and F0 * Using Eqs. 6 and 7, the values of S* and F 0 * were determined respectively (Table 4).

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The new amplification coefficients S * and F 0 * are implemented as multiplicative factors of S and F 0 , already provided by the NTC, 2018. This choice will allow us to maintain the same shape of the spectrum provided by NTC, 2018 but to vary its amplitude to take into account the DSSI and soil amplification phenomena. The value of ag · S, to use in Eqs. 6 and 7, was evaluated in Sect. 3.

Se(T)plateau

(ag·S)av TB TC

Fig. 10. Identification of the plateau of the new spectrum (S e (T)plateau ) between the range period TB and TC and of the spectral acceleration at the period T = 0, (ag · S)av .

 S ∗ = (ag · S)av ag · S F0∗ = Se (T )plateau

  ag · S · S ∗ · η · F 0

(6) (7)

Table 4. Main parameters of the new proposed elastic response spectra. Parameters

Effective foundation motion 7 input

9 input

S e (T)plateau

1.078 g

1.162 g

(ag · S)av

0.454 g

0.622 g

S*

1.735

2.378

F0*

0.919

0.726

Step 4. Implementation of the New Elastic Response Spectra Once the new amplification coefficients S* and F 0 * are known, it is possible to formalize the new Eqs. (8–11) that will define the new branches of the proposed spectra (see Fig. 11). The proposed spectra take into account the real soil amplification phenomena,

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DSSI effects and soil heterogeneity. These aspects are often neglected by the conventional building codes. It is possible to notice how the new spectral accelerations are much higher than the ones of the NTC, 2018 for the investigated area. So, they could provide more reliable data for designing and retrofitting of the buildings, such as the analysed one.         0 ≤ T < TB Se_new (T ) = ag · S · S ∗ · η · F0 · F0∗ T TB + 1 η · F0 · F0∗ 1 − T TB (8) TB ≤ T < TC Se_new (T ) = ag · S · S ∗ · η · F0 · F0∗

(9)

   TC ≤ T < TD Se_new (T ) = ag · S · S ∗ · η · F0 · F0∗ · TC T

(10)

  TD ≤ T Se_new (T ) = ag · S · S ∗ · η · F0 · F0∗ · TC · TD T 2

(11)

Fig. 11. Comparison between the new proposed spectra and the current elastic response spectrum of NTC, 2018 for the investigated area.

6 Final Remarks This paper aims to investigate the influence of the soil and DSSI effects on a nonseismically designed structure damaged by the earthquake of 26 December 2018, recorded in Italy. Moreover, new elastic response spectra are proposed and compared to the one of the NTC, 2018 [3]. The seismicity in the area investigated deserves specific attention, because very frequent, local volcano-tectonic earthquakes cause severe economic losses and the destruction of buildings. 2D FEM analyses of the soil-structure system were performed, varying the seismic inputs. The responses in the frequency domain highlight the importance of performing numerical analyses that take into account DSSI phenomena. For a set of ground motions, the peak acceleration at the foundation level is greater than the peak ground acceleration at the free-field. In this case, the DSSI effects

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are detrimental and they cannot be neglected in the evaluation of the structural response. For other ground motions, the peak acceleration at the foundation level could be lower than the peak ground acceleration at the free-field: the DSSI effects are beneficial. The effects of DSSI could be detrimental or beneficial depending on how close are the first fundamental frequency of the ground motion and the one of the soil-structure system or, the first fundamental frequency of the ground motion and the one of the soil. All the selected seismic motions underline that the soil amplification phenomena predicted by the NTC, 2018 [3] are much lower than the obtained ones through the numerical analyses. So, the knowledge of the real hazard level of the investigated area, the soil amplification phenomena and the DSSI effects are fundamental elements for the designing new buildings and the safeguarding of existing ones. Based on the average response spectra obtained at the foundation level, new elastic response spectra were proposed, introducing new amplification factors that modify the one of the NTC, 2018 [3]. The new spectra take into account the real soil amplification phenomena, DSSI effects and soil heterogeneity. The new spectral accelerations are much higher than the one furnished by NTC, 2018 [3] for the same area, and they could lead a more safety designing and retrofitting. The present study could be extended to other soil conditions and other structural typologies. Acknowledgements. Financial support provided by the DPC/ReLUIS 2019–2021 Research Project, funded by the Civil Protection Department, allowed the authors to achieve the results reported in this paper.

References 1. UN General Assembly, Transforming our world: the 2030 Agenda for Sustainable Development, 21 October 2015, A/RES/70/1. https://www.refworld.org/docid/57b6e3e44.html. Accessed 17 Apr 2022 2. Eurocode 8: Design of structures for earthquake resistance - Part 1: General rules, seismic actions and rules for buildings. European Committee for Standardization, Brussels 3. Norme Tecniche per le Costruzioni. Gazzetta Ufficiale Della Repubblica Italiana Suppl. Ord. (2018). (in Italian) 4. Tropeano, G., Soccodato, F.M., Silvestri, F.: Re-evaluation of code-specified stratigraphic amplification factors based on Italian experimental records and numerical seismic response analyses. Soil Dyn. Earthq. Eng. 110, 262–275 (2018) 5. Pitilakis, K., Riga, E., Anastasiadis, A., Fotopoulou, S., Karafagka, S.: Towards the revision of EC8: proposal for an alternative site classification scheme and associated intensity dependent spectral amplification factors. Soil Dyn. Earthq. Eng. 126, 105137 (2019) 6. Aimar, M., Ciancimino, A., Foti, S.: An assessment of the NTC18 stratigraphic seismic amplification factors. Rivista Italiana di Geotecnica 1, 5–21 (2020) 7. Cavallaro, A., Ferraro, A., Grasso, S., Maugeri, M.: Site response analysis of the Monte Po Hill in the City of Catania. In: AIP Conference, vol. 1020, pp. 240–251. American Institute of Physics (2008) 8. Ferraro, A., Grasso, S., Massimino, M.R., Maugeri, M.: Influence of geotechnical parameters and numerical modelling on local seismic response analysis. In: Winter, M.G., Smith, D.M., Eldred, P.J.L., Toll, D.G. (eds.) 16th European Conference on Soil Mechanics and Geotechnical Engineering (ECSMGE 2015), vol. 4, pp. 2183–2188 (2015)

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9. Abate, G., Massimino, M.R.: Dynamic soil-structure interaction by experimental and numerical modelling. Rivista Italiana di Geotecnica 50(2), 44–70 (2016) 10. Pitilakis, D., et al.: Large-scale field testing of geotechnical seismic isolation of structures using gravel-rubber mixtures. Earthq. Eng. Struct. Dyn. 50(10), 2712–2731 (2021) 11. Massimino, M.R., Abate, G., Corsico, S., Louarn, R.: Comparison between two approaches for nonlinear FEM modelling of the seismic behaviour of a coupled soil-structure system. Geotech. Geol. Eng. 37(3), 1957–1975 (2019) 12. Mercado, J.A., Arboleda-Monsalve, L.G., Mackie, K.R.: Nonlinear inelastic-degrading structural modeling approach to assess the seismic soil-structure interaction response of tall buildings. J. Geotech. Geoenvironmental Eng. 147(10), 04021101 (2021) 13. Bathe, K.J.: Nonlinear finite element analysis and ADINA. In: Bathe, K.J. (ed.) Proceedings of the 12th ADINA Conference on Computers and Structures. Elsevier Science, Oxford (1999) 14. Azzaro, R., D’Amico, S., Tuvè, T.: Estimating the magnitude of historical earthquakes from macroseismic intensity data: new relationships for the volcanic region of Mount Etna (Italy). Seismol. Res. Lett. 82(4), 533–544 (2011) 15. Gambino, S., et al.: Transtension at the northern termination of the alfeo-etna fault system (Western Ionian Sea, Italy): seismotectonic implications and relation with Mt. Etna Volcanism. Geosciences 12(3), 128 (2022) 16. Villani, F., Pucci, S., Azzaro, R., et al.: Surface ruptures database related to the 26 December 2018, MW 4.9 Mt. Etna earthquake, Southern Italy. Sci. Data 7, 42 (2020) 17. Luzi, L., Pacor, F., Puglia, R.: Italian Accelerometric Archive v3.0. Istituto Nazionale di Geofisica e Vulcanologia, Dipartimento della Protezione Civile Nazionale (2019). https:// doi.org/10.13127/itaca.3.0 18. Iervolino, I., Galasso, C., Cosenza, E.: REXEL: computer-aided record selection for codebased seismic structural analysis. Bull. Earthq. Eng. 8, 339–362 (2009) 19. Stucchi, M., Meletti, C., Montaldo, V., Crowley, H., Calvi, G.M., Boschi, E.: Seismic hazard assessment (2003–2009) for the Italian building code. Bull. Seismol. Soc. Am. 101, 1885– 1911 (2011) 20. Azzaro, R., Barbano, M.S., D’Amico, S., Tuvè, T., Albarello, D., D’Amico, V.: First studies of probabilistic seismic hazard assessment in the volcanic region of Mt. Etna (Southern Italy) by means of macroseismic intensities. Bollettino di Geofisica Teorica e Applicata 49(1), 77–91 (2008) 21. Azzaro, R., D’Amico, S., Peruzza, L., Tuvè, T.: Probabilistic seismic hazard at Mt. Etna (Italy): the contribution of local fault activity in mid-term assessment. J. Volcanol. Geotherm. Res. 251, 158–169 (2013) 22. Norme tecniche per la esecuzione delle opere in cemento armato normale e precompresso e per le strutture metalliche, Gazzetta Ufficiale Della Repubblica Italiana Suppl. Ord. (1976). (in Italian) 23. Fiamingo, A., Bosco, M., Massimino, M.R.: The role of soil for a building damaged by the 26 December 2018 earthquake in Italy. J. Rock Mech. Geotech. Eng. (2022). https://doi.org/ 10.1016/j.jrmge.2022.06.010 24. Eurocode 8: Design of structures for earthquake resistance - Part 5: Foundations, retaining structures and geotechnical aspects. European Committee for Standardization, Brussels 25. Lysmer, J., Kuhlemeyer, R.L.: Finite dynamic model for infinite media. J. Eng. Mech. 95, 859–877 (1969) 26. Chopra, A.K.: Dynamics of Structures: Theory and Applications to Earthquake Engineering, 5th edn. Pearson College, Englewood Cliffs (1999)

The Role of DSSI on the Seismic Risk Assessment of a Building Glenda Abate(B)

, Angela Fiamingo , and Maria Rossella Massimino

Department of Civil Engineering and Architecture, University of Catania, Via Santa Sofia 64, 95124 Catania, Italy [email protected]

Abstract. The seismic risk assessment is of fundamental importance in a city like Catania (Italy), where there is a medium-high seismic risk. Certainly, the seismic response of structures depends on the Dynamic Soil-Structure Interaction (DSSI), that can be analysed with different levels of details. By preliminary studies at local level, aimed at microzonation studies, or by detailed studies concerning single case histories. The present paper firstly deals with an easy-to-use approach to develop innovative seismic microzonation maps for an urban area of Catania, allowing the evaluation of DSSI effects on a large scale. This methodology is based on the spatial distribution of spectral accelerations, evaluated for the fixed-base structure configuration as well as for the flexible-base one. Then, the paper shows the main results of the numerical modelling of a school building located in the above-mentioned investigated area. The non-linear soil behaviour is modelled by modified shear modulus and soil damping ratio, firstly evaluated according to EC8 suggestions (2003) and secondly considering the reached effective strain level by the G(γ ) and D(γ ) curves. The response of these FEM models is analysed by investigating the soil filtering effect and estimating how the different modelling of the soil non-linearity can impact on the behaviour of the full-coupled system. Keywords: Seismic risk · Microzonation maps · FEM analyses

1 Introduction The structural dynamic behaviour is governed by a lot of aspects, including the Dynamic Soil-Structure Interaction (DSSI). Indeed, the frequency content of the soil-structure system can be modified compared to the frequency content of just the soil or just the structure; sliding or uplifting phenomena can occur at the soil-foundation interface; the structure can modify the expected accelerogram at the soil surface compared to the free-field conditions. DSSI should be analysed to develop not only the safest but also the most economical design possible. Preliminary analyses are precious to estimate amplification phenomena and, consequently, to develop microzonation maps useful for a careful planning; for these analyses, involving a great number of structures, simple approaches can be used. Otherwise, accurate analyses regarding single case-histories can be performed by experimental models, simplified approaches, and advanced numerical modelling. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 48–63, 2023. https://doi.org/10.1007/978-3-031-26879-3_4

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As for the preliminary analyses, seismic microzonation studies devoted to evaluating the accelerations expected on the ground surfaces have concerned various Mediterranean urban areas, but these studies have been generally related to free-field (FF) conditions. Actually, the dynamic response at the foundation level of structures can significantly deviate from the response in free-field (FF) conditions, because of kinematic and inertial interaction [1–10], and in different cases, DSSI could be detrimental to the structures [11]. So, the present paper shows preliminarily innovative seismic microzonation maps, allowing a large-scale evaluation of DSSI effects, developed for an urban area of the city of Catania (Italy), where there is a medium-high seismic risk [8]. The proposed approach combines: i) geotechnical characteristics; ii) building features; iii) results of 1-D local site response analyses. The fundamental structural periods and related spectral accelerations are evaluated, for both the fixed-base building configuration and the flexible-base one. Their respective ratios are plotted by pie-charts, developing innovative maps in the Google My Maps environment, that furnish a preliminary estimate of the DSSI effects based on the values of the spectral acceleration ratios. These maps eventually suggest more accurate analyses of DSSI phenomena, for the seismic retrofitting of “critical” situations. As for these more accurate studies, numerical models allowing to consider realistically initial and boundary conditions, soil profile, geometry, soil nonlinearity, soil-foundation interface. So, the present paper shows the FEM analysis of a full-coupled soil-structure system concerning a school building located in the above-mentioned investigated area. The building and its subsoil were deeply investigated in the framework of the POR-FESR Research Project Sicilia 2007–2013, aimed at reducing the seismic risk in Eastern Sicily [5, 6]. A 2D numerical modelling is performed, modelling the soil non-linear behaviour by means of modified shear modulus G and damping ratio D evaluated both according to the EC8 suggestions [1] and according to the effective strain level considering the G(γ ) and D(γ ) curves obtained by resonant column tests. The dynamic behaviour of the system is investigated in the time and frequency domains, so evaluating soil amplification ratios, amplification functions and response spectra.

2 A Preliminary Approach for the Evaluation of DSSI Effects: The Proposed Methodology for a New Seismic Zonation 2.1 Evaluation of DSSI Effects The proposed methodology for a large-scale evaluation of DSSI effects is applied for an urban area of the city of Catania. It is based on the combination of geotechnical and structural properties (essentially height, foundation geometry and material, i.e., concrete or masonry) with the results by 1D local site response studies. Considering both the fixed-base structure configuration and the flexible-base one, the corresponding fundamental periods and spectral accelerations are evaluated, their ratios T fixed /T SSI and S a (T fixed )/S a (T SSI ) are computed and proposed as indexes of DSSI effects at large-scale.

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These ratios are plotted by pie-charts, developing innovative maps in the Google My Maps environment, that furnish a preliminary estimate of the DSSI effects, suggesting more accurate studies [4, 5, 7] for ratios S a (T fixed )/S a (T SSI ) > 1.15. The fundamental period T fixed of each fixed-base building is evaluated by means of the formula provided by [12]: T fixed = C 1 · H 3/4 . The period T SSI of each flexible-base structure is estimated according to [2], i.e., modelling the soil-structure system by an equivalent oscillator with allowable translational and rocking motion of its base [13]:  kstr heff 2 kstr TSSI = Tfixed 1 + + (1) kh kr where k str is the stiffness of the fixed-base structure, heff is the effective height of the structure equal to 0.7H (unless for single-storey buildings for which h = H), k h and k r are the translational and rocking stiffness of the foundation. Then, the T fixed /T SSI ratio is evaluated for each structure and the spatial distribution is mapped. Meanwhile, soil response analyses under selected input motions are performed, achieving the response spectra for β fixed and for β SSI , being β fixed the damping ratio of the fixed-base structure, equal to 5% for concrete structures and 8% for masonry structures, and β SSI the damping ratio of the flexible-base structure, higher than β fixed . Finally, the spectral accelerations S a (T fixed ) and S a (T SSI ) are computed and the subsequent spatial distribution of DSSI effects as spectral acceleration ratios is mapped. 2.2 The Investigated Area The subsoil of Catania was subjected to many geotechnical investigations (both in situ and laboratory tests [14–16]), the most numerous of which were conducted in the framework of the two research projects named “Catania Project 1 and 2” [17, 18], allowing to get a rich database of over 1200 surveys [19–22].

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Fig. 1. The investigated area of Catania; (a) the area subdivided into three sub-areas; (b) Profiles of V s and corresponding three stratigraphies.

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The area chosen for the study of DSSI phenomena shown in this paper is the west zone of Catania. The area is divided into three subareas: T1w, T2w and T3w (Fig. 1a), based on the soil stratigraphy (Fig. 1b), known thanks to the above-mentioned database. 2.3 The Proposed Seismic Microzonation Map In accordance with the previously summarized procedure, the fundamental periods of all structures present in the investigated area, as well as the related spectral accelerations, are evaluated, considering both the fixed-base configuration and the flexible-base one. So the ratios T fixed /T SSI and S a (T fixed )/S a (T SSI ) are computed. These computations are performed for each building, but the results are summarized for a total of 5 blocks in which the area is further divided according to the urban morphology; they are plotted via pie charts, choosing three different ranges of ratios for a better and understandable representation (Fig. 2). As for the T fixed /T SSI ratios (Fig. 2a), T fixed /T SSI < 1.15 indicate probable negligible DSSI effects, 1.15 < T fixed /T SSI < 1.30 indicate moderate ones, and T fixed /T SSI ≥ 1.30 indicate high ones. It is evident that the assumption of the fixed-base structure faithfully captures reality, as T fixed /T SSI < 1.15 is achieved for almost every building. As for the S a (T fixed )/S a (T SSI ) ratios (Fig. 2b), S a (T fixed )/S a (T SSI ) ≤ 0.85 indicate beneficial effects, 0.85 < S a (T fixed )/S a (T SSI ) ≤ 1.15 indicate negligible ones and S a (T fixed )/S a (T SSI ) > 1.15 indicate detrimental ones. The developed maps shows that the DSSI effects are mainly negligible or beneficial. Just for the 2002 seismic input, one block has a small percentage of S a (T fixed )/S a (T SSI ) > 1.15, probably due to the poor properties of soil foundation. Figure 2c shows the ratios between the spectral accelerations S a (T SSI ) and those suggested by the in-force Italian Technical Code [23] S a (NTC), considering the same ranges previously adopted. The [23] is essentially conservative, providing S a (NTC) values greater than or almost equal to the S a (T SSI ) values, with the exception of two blocks which, for inputs 1818 and 2002, have ratios S a (T SSI )/S a (NTC) > 1.15. In conclusion, by the proposed methodology it is possible to state that the DSSI effects are negligible or beneficial for most structures included in the examined area. For these buildings, the DSSI phenomena could be considered to decrease the construction costs. On the contrary, for old masonry buildings located in soil with poor dynamic characteristics, the DSSI can be detrimental and so the correlated phenomena must be carefully investigated. So, this methodology is a good preliminary study to evaluate some critical situations for which more accurate studies can be performed, for example by means of full-coupled numerical analyses, as those shown in the following section.

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Fig. 2. Spatial distribution of: (a) T SSI / T fixed ; (b) S a (T SSI )/S a (T fixed ); (c) S a (T SSI )/S a (NTC).

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3 A Detailed Evaluation of DSSI Effects by Means of a Numerical Model of a Single Full-Coupled Soil-Structure System 3.1 The Structure and Its Subsoil This second part of the paper shows the numerical FEM modelling of a soil-structure system concerning a building hosting the Nazario Sauro school, located in the northern part of the investigated area. By means of the presented quick approach, the DSSI effects on the school building should be essentially negligible, because the following ratios are achieved: T fixed /T SSI = 1.08 (T fixed = 0.31 s and T SSI = 0.33 s) and S a (T fixed )/S a (T SSI ) = 0.8; 1.0; 0.74; 0.84 for the 1818, 1990, 2002, 2018 seismic inputs, respectively. Despite these comforting results, this soil-structure system is analysed by means of an accurate FEM modelling, due to its important use and its interesting foundation configuration, as described in the following. The design and construction of this building was in 1971–1975, in a period in which the anti-seismic code was not yet in force in Italy (the first technical code was published in 1982). It is a reinforced concrete structure resting on plinths having different embedments. Numerous geotechnical in-situ tests were carried out in the framework of the PORFESR Project Sicilia 2007–2013. The stratigraphy was known thanks to two boreholes,

Fig. 3. (a) Plan view with the indication of the boreholes; (b) Geotechnical model and individuation of the frame chosen for the FEM modelling (by yellow narrow); (c) V s profiles by SDMT tests and adopted for the FEM modelling; (d) G(γ ) and D(γ ) curves adopted for modelling the soil dynamic behaviour.

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named S1 and S2 (Fig. 3a–b). Different V s profiles were achieved by means of two seismic dilatometer tests (SDMT 2a and SDMT 2b), carried out inside the S1 borehole reaching two different depths, as shown in Fig. 3c, that also shows the V s profile adopted in the modelling (according to the SDMT 2b results). Three different layers were modelled, until a depth equal to 30 m, according to which the soil can be classified as type E for [23]. The natural frequency of the soil foundation (approximately equal to 4 Hz) was evaluated by means of a micro-tremor survey (HVSR test) performed in the school site. The G(γ ) and D(γ ) curves adopted for modelling the soil dynamic behaviour are shown in Fig. 3d: the curves for the layers 1 and 3 were evaluated by means of resonant column tests performed on samples of these layers; instead, literature curves were assumed for the layer 2. The frame chosen for the DSSI modelling is shown also in Fig. 3, indicated by the yellow narrow (Fig. 3b). 3.2 The Adopted Seismic Inputs Considering the investigated building subject to significant crowding, the expected acceleration at the bedrock is 0.245 g for the Limit State of Significant Damage (“SLD”), according to [23]. So, the accelerograms adopted for the FEM analyses are appropriately scaled to this amplitude. Seven accelerograms are chosen: one input recorded during the 1990 Eastern Sicily earthquake; three synthetic inputs obtained modelling the source on the Hyblean-Maltese fault and generating the 1818 scenario earthquake; three synthetic inputs achieved also modelling the source on the Hyblean-Maltese fault and generating the 1693 scenario earthquake. All the inputs differ in frequency content (Fig. 4). 3.3 The FEM Models The dynamic behaviour of the previously described system was investigated by a 2D finite element modelling by means of the ADINA code [24]. Three different FEM models are performed. The first model corresponds to the actual configuration (the structure based on isolated footings having different depths from the ground surface; Fig. 5). In the second model the isolated footings are replaced by beam foundations having the same embedments. In the third model the beam foundations rest at the ground surface (so there is not embedment). As for the geometry of the soil deposit, its length is assumed to avoid boundary effects, and its height is assumed considering stratigraphy achieved by the insitu tests. The mesh element size is chosen in order to ensure an efficient reproduction of all the waveforms of the whole frequency range under study: h ≤ V s,min /6–8 f max [25]. The frame is modelled by 2-node beam elements; the soil is modelled by 9-node 2D solid elements. As for the boundary conditions, for the nodes at the bottom only the horizontal displacements are allowed (the vertical displacements are prevented). Instead, the nodes of the soil vertical boundaries are linked by “constraint equations” that impose the same vertical and horizontal displacements and at the same depths. Uplifting and/or sliding phenomena can occur thanks to properly contact foundation-soil surfaces, for which a friction angle equal to 2/3 ϕ is assumed. As for the loading conditions, concentrated

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Fig. 4. Fourier spectra and fundamental frequencies of the adopted seismic inputs.

Fig. 5. The FEM model.

masses are applied on all the horizontal beams for simulating weights and non-structural loads. The previously described input motions are applied at the bottom of the models. The material viscosity is modelled according to the Rayleigh damping, computing the Rayleigh factors α and β as α = D · ω and β = D/ω [26], being D the damping ratio and ω the natural angular frequency of soil and structure, respectively. The natural frequency of the structure is equal to 2.69 Hz, evaluated as ωstr = 2π /T str with: T str = C 1 · H 3/4 [12]; as for the soil, it is equal to 2.95 Hz, evaluated as ωsoil = 2π /T soil with: T soil = V s /4H. As for the constitutive modelling, the frame is modelled considering a linear visco-elastic behaviour, assuming the typical values of the reinforced concrete (E = 28500 MPa, ν = 0.25, γ = 25 kN/m3 , D = 5%). For modelling the soil non-linearity [5, 27–29] two different linear equivalent viscoelastic constitutive models are used for the first FEM model. Firstly, the soil is modelled adopting modified shear modula G* and damping ratios D* based on the expected surface acceleration, as suggested by EC8 [1]. The expected surface acceleration is ag · S = 0.245g · 1.34 (being S = 1.34 the value provided by [23] for soil type E). So, the following modified values are assumed: G* s = 0.36 · Gmax for soil layers 1 and 3 and G* = 0.56 · Gmax for soil layer 2, i.e., a major value because it is a stiffer layer. The second constituitve model is characterized by G and D values chosen according to the G(γ ) and D(γ ) curves of Fig. 3 for the effective strain level obtained for each soil layer and for each different input, using an iterative sub-routine. In the other two FEM models, the soil is modelled only by means of the first approach, i.e., adopting modified G and D as suggested by [1].

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3.4 The Main Results The response of the soil-structure system is analysed both considering and neglecting the DSSI, referring to a vertical alignment along the structure (named SSI alignment) and a vertical alignment far from the structure (named FF alignment, i.e., representing the free-field condition; Fig. 5), respectively. The comparison between the second and third FEM models is here presented just with reference to the alignment below the structure, to analyse the influence of different foundation configurations on the soilstructure interaction. Figure 6a–b show the amplification ratio profiles Ra -z for the first FEM model (the original system, i.e., the structure having isolated footings), obtained along the two alignments using the first linear equivalent visco-elastic constitutive model for modelling the soil dynamic behaviour (according to the EC8 suggestions [1]). The seismic inputs suffer an amplification just inside the layer 1, having almost no modification at layers 2 and 3.

Fig. 6. Ra profiles achieved for the original system (FEM model 1), using the first linear equivalent visco-elastic constitutive model for modelling the soil non-linear behaviour, i.e., according to the EC8 suggestions: (a) considering the DSSI; (b) neglecting the DSSI, i.e., for the free-field condition.

Figure 7a shows the amplification ratios at the ground surface under the structure, with reference to first FEM model, comparing the two linear equivalent visco-elastic constitutive models adopted for modelling the soil non-linear behaviour. The Ra values achieved using the first linear equivalent visco-elastic constitutive model for the soil, i.e., according to the EC8 suggestions [1], are always minor than the values obtained using the second linear equivalent visco-elastic constitutive model for the soil, i.e., adopting the iterative procedure for evaluating G and D: this is due to the lower D values for this second linear equivalent visco-elastic model. This figure shows the amplification ratio S provided by [23], that is always minor than both the numerical Ra values. Figure 7c shows the amplification ratios at the ground surface for the first two FEM models, i.e., for the structure having isolated footings and for the structure having beam foundations. The comparison shows that for some inputs the Ra values at the foundation level are higher for one system and in other cases for the other one. The percentage difference between the two models is between –30% (for input 1818(2)) and +8% ((for input 1693(2)). Figure 7d shows the amplification ratios at the ground surface for the two

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FEM models having beam foundations with different embedment (i.e., FEM models 2 and 3; shown in Fig. 7b). The Ra at the foundation level is always higher for the building without embedment; for some inputs, it is more remarkable in comparison with other inputs, for which the difference is instead minimal. However, the embedment guarantees a confinement effect that dampens the acceleration that impacts the foundations.

Fig. 7. (a) Ra values at the ground surface, for the SSI alignment, for the FEM model 1 (i.e., with original foundations), achieved by the two soil constitutive modelling approaches; (b) zoomed view of the two FEM models 2 and 3, having modified foundations, with and without embedment, respectively; (c) Ra values at the ground surface, for the SSI alignment, for the two FEM models 1 and 2, i.e., with original and modified foundations, respectively; (d) Ra values at the ground surface, for the SSI alignment, for the two FEM models 2 and 3, i.e., having modified foundations, with and without embedment, respectively. All the figures also show the amplification ratio S by [23].

Figures 8a–d show the amplification functions A(f ) for the two investigated alignments for the original system (FEM model 1), using both the two linear equivalent visco-elastic constitutive models for modelling the soil non-linear behaviour. A(f ) is evaluated as ratio between the Fourier spectrum computed at the foundation level and the Fourier spectrum computed at the bedrock level, so considering just the soil. Using the first linear equivalent visco-elastic constitutive model for the soil, i.e., according to the EC8 suggestions [1], the following soil natural frequencies are achieved: 2 Hz, 4 Hz and 5.4 Hz along the SSI alignment, and an average value equal to 3.2 Hz along

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the FF alignment. Using the second linear equivalent visco-elastic constitutive model for modelling the soil non-linear behaviour, i.e., according to the iterative procedure for evaluating the modified G and D values, very similar three soil natural frequencies along the SSI alignment are achieved (2.2 Hz, 4 Hz and 5.4 Hz) and the average value along the FF alignment is equal to 4 Hz (coincident with that calculated by the HVSR test). The frequencies estimated by the iterative approach are slightly higher than those obtained by the first approach, due to the greater G decay. Similarly, greater amplitudes are obtained by the second approach, due to the adopted D values lower than 10%. For the structure having beam foundations, substantially the same results are achieved. Figure 9 shows the soil amplification functions A(f ) achieved for the adopted input motions, for the two modified soil-structure systems (FEM models 2 and 3). There are essentially two natural frequencies of the soil, equal to about 2 Hz and 4 Hz; the first is practically coincident with the frequency of the structure. This is visible in Fig. 10, which shows, on the left, the amplification functions evaluated as ratio between the Fourier spectra computed at the roof of the structure (“top”) and the Fourier spectra computed at the foundation (“bottom”). Figure 10 also shows, on the right, the amplification functions evaluated as ratio between the Fourier spectrum calculated at the roof of the structure and at the bedrock, thus considering the entire soil-structure system. Two frequencies are evident: 2 Hz and 3.6–4 Hz; it is therefore possible to observe both the contribution of the structure (f = 2 Hz) and the contribution of the soil (f = 4 Hz).

Fig. 8. Amplification functions, for the original system (FEM model 1), achieved using the first linear equivalent visco-elastic constitutive model for modelling the soil non-linear behaviour, i.e., according to the EC8 suggestions (first row): considering (a) and neglecting (b) the DSSI; and using the second linear equivalent visco-elastic constitutive model for modelling the soil nonlinear behaviour, i.e., according to the iterative procedure for evaluating G and D (second row): considering (c) and neglecting (d) the DSSI.

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The presence of the embedment causes a shift to a higher frequency of the structure. This occurs because the embedment causes a stiffening of the entire structure, consequently causing a decrease in the period and therefore an increase in frequency. In general, the soil frequencies, considering the structure resting on the ground surface, are slightly shifted to the right. So, it is possible to see how the presence of the structure and the embedment of the foundation can affect the response in the frequency domain.

Fig. 9. Amplification functions for the two modified soil-structure systems (FEM models 2 and 3), just relating to the soil deposit interacting with the structure.

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Fig. 10. Amplification functions for the two modified soil-structure systems (FEM models 2 and 3), relating to just the structure (left column) and to the entire soil-structure system (right column).

4 Conclusions The present paper deals with two different approaches for studying some aspects of the DSSI: preliminary studies at urban level, aimed at microzonation studies useful for the rational use of the territory; detailed studies concerning single case histories performed by numerical modelling. As for the preliminary studies, the paper presents a quick approach for the evaluation of DSSI effects for an area of the city of Catania (Italy), characterized by medium-high seismic risk. As results, negligible or beneficial DSSI effects are achieved for most structures of the examined area, for which so DSSI could be taken into account to minimize the construction costs. As for the detailed studies concerning single soil-structure system, the paper presents the results of a full-coupled numerical modelling regarding a school building located in the above-mentioned investigated area. According to the first quick approach for the evaluation of the DSSI effects, essentially negligible DSSI effects are achieved for this building.

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The FEM modelling underlines the importance of considering the DSSI for the seismic safety of buildings, because very different results are achieved if the DSSI is considered or not. In particular, the response of the FEM models is analysed comparing: the two different approaches adopted for modelling the soil-nonlinearity (modified G and D according to the expected surface acceleration and modified G and D according to the effective strain level reached in the soil); the different foundation configurations modelled (the actual plinths having different embedment and the hypothesized beam foundations, with and without embedment). As for the response of the original system, i.e., the structure founded on isolated footings, investigated in terms of amplification ratio, it is possible to see that the structure causes a severe amplification of some seismic inputs, that instead suffer a minor amplification in free-field conditions (so detrimental effects of the DSSI are achieved). Furthermore, the amplification ratio provided by [23] is always minor than the numerical amplification ratios at the ground surface for both the investigated alignments. As for the response in terms of amplification functions, it is evident how the structure modifies the frequency content of the soil: three soil natural frequencies are achieved along the SSI alignment. Finally, as for the different modelling of the soil non-linearity, the values achieved using the first linear equivalent visco-elastic constitutive model for modelling the soil non-linear behaviour, i.e., according to the EC8 suggestions [1], are always minor than the values obtained using the second linear equivalent visco-elastic constitutive model for modelling the soil non-linear behaviour, i.e., using an iterative procedure for estimating the updated G and D values for the effective strain level. This is due to the lower D values evaluated for the second linear equivalent visco-elastic constitutive model adopted for the soil. As for the last investigated aspect, that is the different foundations hypothesized, the modified foundations do not lead to a structural different response; instead, it is evident the effect of the embedment of the foundations, that causes a shift to a higher frequency of the structure. So, it is possible to see how the presence of the structure and the embedment of the foundation can affect the response in the frequency domain. Acknowledgement. The research reported in this paper was performed by the financial supports provided by the European Research Project eWAS “An Early Warning System for Cultural Heritage”, ID: ARS01_00926, PNR 2015–2020 and by the DPC/ReLUIS 2019–21 Research Project, funded by the Civil Protection Department.

References 1. EN 1998-5: Eurocode 8. Design of structures for earthquake resistance—part 5: foundations, retaining structures and geotechnical aspects. ENV 1998, European Committee for Standardization, Brussels (2003) 2. Building Seismic Safety Council (BSSC). National Earthquake Hazard. Reduction Program. (NEHRP): Recommended Provisions for Seismic Regulations for Buildings and Other Structures; FEMA 750; BSSC: Washington, DC, USA (2010)

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3. Massimino, M.R., Biondi, G.: Some experimental evidence on dynamic soil–structure interaction. In: Papadrakakis, M., Papadopoulos, V., Plevris, V. (eds.) Proceedings of 5th ECCOMAS Thematic Conference on Computational Methods in Structural Dynamics and Earthquake Engineering, Crete Island, Greece, 25–27 May 2015, COMPDYN 2015, pp. 2761–2774 (2015) 4. Massimino, M.R., Abate, G., Grasso, S., Pitlakis, D.: Some aspects of DSSI in the dynamic response of fully coupled soil-structure systems. Rivista Italiana di Geotecnica 1, 44–70 (2019) 5. Massimino, M.R., Abate, G., Corsico, S., Louarn, R.: Comparison between two approaches for non-linear FEM modelling of the seismic behaviour of a coupled soil-structure system. Geotech. Geol. Eng. 37(3), 1957–1975 (2019) 6. Abate, G., Corsico, S., Grasso, S.: Dynamic behaviour of a coupled soil-structure system by means of FEM analyses finalized to the seismic risk mitigation of a school in Catania (Italy). In: Earthquake Geotechnical Engineering for Protection and Development of Environment and Constructions – Proceedings of the 7th International Conference on Earthquake Geotechnical Engineering, 17–20 June 2019, Roma, Italy, pp. 977–984 (2019) 7. Abate, G., Massimino, M.R.: Dynamic soil–structure interaction analysis by experimental and numerical modelling. Rivista Italiana di Geotecnica 50(2), 44–70 (2016) 8. Abate, G., Bramante, S., Massimino, M.R.: Innovative seismic microzonation maps of urban areas for the management of building heritage: a Catania case study. Geosciences 10(12), 1–22, 480 (2020) 9. Karatzetzou, A., Pitilakis, D.: Modification of dynamic foundation response due to soilstructure interaction. J. Earth. Eng. 22(5), 861–880 (2017) 10. Brunelli, A., et al.: Numerical simulation of the seismic response and soil–structure interaction for a monitored masonry school building damaged by the 2016 Central Italy earthquake. Bull. Earthq. Eng. 19(2), 1181–1211 (2020). https://doi.org/10.1007/s10518-020-00980-3 11. Mylonakis, G., Gazetas, G.: Seismic soil-structure interaction: beneficial or detrimental? J. Earthq. Eng. 4(3), 277–301 (2000) 12. NTC 2008. D.M. 14/01/08 – New technical standards for buildings, Official Journal of the Italian Republic, 14th January 2008 (2008). (in Italian) 13. Veletsos, A.S., Meek, J.: Dynamic behavior of building-foundation systems. Earthq. Eng. Struct. Dyn. 3, 121–138 (1974) 14. Cavallaro, A., Grasso, S., Maugeri, M.: Volcanic soil characterization and site response analysis in the city of Catania. In: Proceedings of the 8th US National Conference on Earthquake Engineering 2006, San Francisco, CA, USA, 18–22 April 2006, pp. 835–844 (2006) 15. Cavallaro, A., Grasso, S., Ferraro, A.: A geotechnical engineering study for the safeguard, restoration and strengthening of historical heritage. Procedia Eng. 158, 134–139 (2016) 16. Grasso, S., Massimino, M.R.: A GIS for data mining in seismic microzonation studies. Smart Innov. Syst. Technol. 142, 191–201 (2019) 17. Faccioli, E., Pessina, V.: The Catania project: earthquake damage scenarios for high-risk area in the Mediterranean. In: Faccioli, E., Pessina, V. (eds.) CNR-Gruppo Nazionale per la Difesa dai Terremoti, Roma, Italy, 225 p. (2000) 18. Maugeri, M.: Advances in earthquake engineering. In: Seismic Prevention of Damage: A Case Study in a Mediterranean City; WIT Press, Southampton (2005) 19. Grasso, S., Maugeri, M.: Vulnerability of physical environment of the city of Catania using GIS technique. In: Maugeri, M. (ed.) Seismic Prevention of Damage: A Case Study in a Mediterranean City, pp. 155–175. WIT Press, Southampton 2005) 20. Grasso, S., Laurenzano, G., Maugeri, M., Priolo, E.: Seismic response in Catania by different methodologies. In: Maugeri, M. (ed.) Seismic Prevention of Damage: A Case Study in a Mediterranean City, pp. 63–79. WIT Press, Southampton (2005)

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21. Caruso, S., Ferraro, A., Grasso, S., Massimino, M.R.: Site response analysis in eastern sicily based on direct and indirect vs measurements. In: Proceedings of the 1st IMEKO TC4 International Workshop on Metrology for Geotechnics, MetroGeotechnics, Benevento, Italy, 17–18 March 2016, pp. 115–120 (2016) 22. Castelli, F., Cavallaro, A., Ferraro, A., Grasso, S., Lentini, V., Massimino, M.R.: Static and dynamic properties of soils in Catania (Italy). Ann. Geophys. 61, 221 (2018) 23. NTC 2018. D.M. 17/01/18. Updating of technical standards for buildings. Off. J. Ital. Repub. (2018) 24. ADINA: Automatic dynamic incremental nonlinear analysis. Theory and modelling guide. ADINA R&D, Inc., Watertown (2008) 25. Lanzo, G., Silvestri, F.: Risposta sismica locale: teorie ed esperienze. Helvius Edizioni, Napoli (1999) 26. Lanzo, G., Pagliaroli, A., D’Elia, B.: Influenza della modellazione di Rayleigh dello smorzamento viscoso nelle analisi di risposta sismica locale. In: Proceedings of XI National Conference “Seismic Engineering in Italy” Genova, 25–29 January 2004 (2004) 27. Abate, G., Caruso, C., Massimino, M.R., Maugeri, M.: Validation of a new soil constitutive model for cyclic loading by fem analysis. Solid Mech. Appl. 146, 759–768 (2007) 28. Pecker, A., Chatzigogos, C.T.: Nonlinear soil structure interaction: impact on the seismic response of structures. In: Proceedings of XIV European Conference on Earthquake Engineering. August 2010, Ohrid, FYROM, Keynote Lecture (2010) 29. Pecker, A., Paolucci, R., Chatzigogos, C., Correia, A.A., Figini, R.: The role of non-linear dynamic soil-foundation interaction on the seismic response of structures. Bull. Earthq. Eng. 12(3), 1157–1176 (2013). https://doi.org/10.1007/s10518-013-9457-0

Response of Geodesic Domes on the Seismic Excitation with Time History Analysis Dominika Bysiec(B)

, Adriana Janda, and Tomasz Maleska

Faculty of Civil Engineering and Architecture, Opole University of Technology, Katowicka 48, 45-061 Opole, Poland {d.bysiec,t.maleska}@po.edu.pl

Abstract. The paper presents the response of six geodesic domes under seismic excitation. The structures subjected to seismic analysis were created based on two different methods of subdividing spherical triangles (the original octahedron face). These structures are often used for the strategic reference point in the city e.g. in Montreal of Canada (Montreal Biosphère in Canada), in Missouri of USA (Climatron at the Missouri Botanical Garden) and in Figueres of Spain (The Glass Dome in Dali Museum). The analyzed structures were made of steel, which is recommended for the lightweight structures with large spans. It should be emphasized that the designing steel domes are currently a challenge for structural engineering, as well as architects, who take into account their aesthetic considerations. In the current research, the finite element method with numerical program was used. Using numerical analysis, six different geodesic domes shaped according to two different methods were analysed under seismic excitation (El Centro). In addition, the Time History method was used. This paper presents the results of natural frequencies, displacements and accelerations in time domain. The results from this analysis show the impact of type of used topology under seismic excitation on the response of geodesic domes. The occurred results will be helpful for these structures located on the seismic areas. Keywords: Geodesic dome · Seismic response · Natural frequency · Dynamic analysis · Seismic analysis · Steel structures

1 Introduction The domes are one of the oldest architectural forms in construction and are the leading achievements of human civilization. For centuries, the domes have exceeded the limits relative to the span of elementary parts of the structure. These structures have many advantages. The most important ones include the possibility of covering large surfaces without the need to use internal supports, as well as their durability, impeccable acoustics and excellent ventilation. It is also necessary to emphasize the aesthetic considerations of domed roofs, resulting from various methods of creating topology of strut elements. It is possible to obtain a lighter structures using optimization methods, despite the fact that domed structures are economical in terms of material consumption already in © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 64–73, 2023. https://doi.org/10.1007/978-3-031-26879-3_5

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their design and construction compared to traditional forms of structures [1]. Model solutions of the geometry scheme obtained Saka [2], Kaveh and Talatahari [3], Carvalho et al.[4], Carbas and Saka [5], Gholizadeh and Barati [6], Kaveh and Rezaei [7, 8], Kaveh et al. [9], Ye and Lu [10]. So far, there are several research papers on geodesic domes subjected to seismic loads. In the paper of Cai et al. [11], where the Time History methods and the modal method were used, the main task was to determine the response to the action of seismic loads. Takeuchi et al. [12] proposed a method based on the amplification factor. The paper of Nakazawa et al. [13] focuses on methods for assessing the reaction of metal roof structures to earthquakes. Also the papers of Kato and Nakazawa [14], Li et al. [15] and Qin et al. [16] refer to the analysis of the evaluation of light structure reactions to seismic interactions. Moreover, Li and Xu [17] presented an analysis of the dynamic stability and the probability of failure of dome structures under the conditions of random seismic excitation. It should be added that most of the scientific studies on geodesic domes relate to space frames, the basis of which is the icosahedron, which is the development of Fuller’s patent [18]. Thus, in this paper, the response of six geodesic domes generated from other polyhedra, that is regular octahedron, under seismic excitation was analysed. The structures subjected to seismic analysis were created by Pilarska [19–22] on the basis of two different methods of subdividing spherical triangles proposed by Fuli´nski [23]. As a seismic excitation, a forcing characterized by high intensity and long duration of the intensive area was used. The applied El Centro excitation is recognized as highly destructive, allowing to determine the seismic response to building structures. In this paper, it was decided to carry out a numerical experiment due to the size of the domes tested (diameter 50 m and height 25 m) and the wide application of the numerical program, which accurately, using the finite element method, allows to determine the seismic response of the tested engineering structures.

2 Description of the Performed Research 2.1 Numerical Models The analyzed geodesic domes were made of steel struts (round pipes) of S235 standard steel according to Eurocode 3 [24]. The other material properties of numerical models subjected to the analysis were recognized in paper Pilarska and Maleska [25]. Taking into account two different methods of creating topology of structures, this paper focuses on three models designed according to the first method: (i) model I1 (2888-hedron), (ii) model II1 (3872-hedron), (iii) model III1 (5408-hedron) and three models created according to the second method: (iv) model I2 (2904-hedron), (v) model II2 (4704hedron), (vi) model III2 (7776-hedron). In addition, the struts of all geodesic domes were grouped into four groups and assigned different cross-sections [21]. 2.2 Research Methods The paper presents the developed structures created on the basis of the regular octahedron. Two methods proposed by Fuli´nski [23] (Fig. 1) of subdividing the initial triangle

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of the regular octahedron were used to design six geodesic domes subjected to seismic analysis. The initial triangle of the regular octahedron was subdivided in both methods used with the appropriate frequencies to finally obtain domes with a similar number of struts. All designed domes were 50 m wide and 25 m high. The detailed description of the subdivision methods used was present in Pilarska [19–22].

Fig. 1. Methods of subdividing the initial triangle edge, according to Fuli´nski.

The Time History method was used in the research in numerical analysis. The method bases on the steps by steps analysis in time domain. This method is common used in the case of seismic response of civil engineering structures, as presented in Maleska and Beben [26], Maleska et al. [27, 28]. In the case of seismic excitation, the El Centro record from 19 May 1941 was used [28]. This record is very destructive. Furthermore, its scale was VII of Richter scale and acceleration equal of 3.43 m/s2 . It should be added that the load was applied simultaneously to all nodes of fixed supports (Fig. 2), in three directions in the same time (horizontal X and Y as well as vertical Z) [29].

Fig. 2. View of numerical models: a) model I1 and b) model I2 [25].

2.3 Numerical Results Based on the research carried out on the six analyzed strut domes, the values of the natural frequencies were obtained (Table 1), as well as the maximum displacements and accelerations (Table 2).

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Table 1. Natural frequencies of geodesic domes from numerical analysis. Mode of shape

Model I1 (2888)

I2 (2904)

II1 (3872)

II2 (4704)

III1 (5408)

III2 (7776)

1

5.40

5.72

5.54

6.02

6.21

6.29

2

5.41

5.73

5.55

6.03

6.22

6.30

3

6.10

6.81

6.33

7.03

8.22

7.09

4

6.83

7.46

7.13

7.49

8.79

7.44

5

6.94

7.84

7.22

7.91

8.85

7.85

6

6.95

7.85

7.23

7.92

8.86

7.86

7

7.14

7.87

7.38

7.95

8.87

7.91

8

7.26

7.90

7.47

7.96

8.94

7.92

9

7.27

7.92

7.48

8.01

8.95

7.98

10

7.32

8.02

7.59

8.02

9.00

8.05

1 Method 1 2 Method 2

Table 2. Maximum values of geodesic domes from numerical analysis. Model

Acceleration [m/s2 ]

Displacement [cm] Direction X

Y

Z

X

Y

Z

I1 (2888)

87.59

80.26

42.71

18.26

16.45

15.22

I2 (2904) II1 (3872)

41.76

34.14

21.42

15.72

12.94

10.52

94.51

56.23

43.85

20.12

18.78

17.16

II2 (4704) III1 (5408)

37.78

27.17

21.82

18.88

18.77

11.41

35.40

42.10

16.32

20.20

10.08

15.83

III2 (7776)

36.78

47.60

34.16

23.93

21.09

19.03

1 Method 1 2 Method 2

3 Discussion of the Obtained Results 3.1 General Remarks In this paper the natural frequencies were analyzed (Table 1). The values from the Time History analysis for displacements and accelerations were also taken into account. The obtained maximum results for displacements and accelerations are presented in Table 2. These results were presented for three directions, i.e. two horizontal (X and Y) and one vertical (Z). In turn, the natural frequencies are presented for the first 10 modes of shape. It should also be emphasized that all analyzed domes were designed in such a way that

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the use of steel cross-sections was as large as possible. For the analyzed steel struts, the use of the cross-section was close to 100% and amounted to about 95% of the load capacity due to dimensioning according to Eurocode 3. 3.2 Natural Frequencies Analyzing the natural frequencies (Fig. 3), it is not possible to mark an unequivocal tendency regarding the influence of the applied methods of shaping geodesic domes as well as the frequencies of subdividing on the obtained results. It can be seen, however, that in the case of domes with the smallest subdivision frequencies (model I1 and model I2 ) the lowest natural frequencies were obtained, 5.40 Hz (model I1 ) and 5.72 Hz (model I2 ), respectively. However, higher natural frequencies were obtained for the remaining researched domes (models: II1 , II2 , III1 and III2 ). Undoubtedly, the dome of the III1 model deserves attention (9.0 Hz – the 10th mode of shape). In this model, the values of the natural frequencies in the first two modes of shape were comparable to the values of the III2 model. In the successive modes of shape of the III1 model, a significant increase in relation to the other models can be observed. Moreover, it can be noticed that the values of the natural frequencies increased linearly between the 2nd and the 5th mode of shape for all the analyzed geodesic domes. In the remaining ranges (1–2 and 5–10), the values of natural frequencies remained rather at a level comparable to the previous eigenvalue.

Fig. 3. Natural frequencies of geodesic domes from numerical analysis.

3.3 Displacements from Numerical Analysis When analyzing the displacements results, it was noticed that the first method is more sensitive to the effect of seismic excitations, and this effect is especially noticeable in

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the models I1 and II1 , where the results obtained are clearly higher (by 60% higher) than the results for the second method for models I2 and II2 . The highest displacement result for the researched numerical models was recorded in model II1 , where its value was 94.51 cm. Compared to model II2 from the second method, the value was lower by 60% and it was equal to 37.78 cm (model II2 ). It can also be seen that the results (Fig. 4) obtained in the horizontal directions (X and Y) are much higher than in the vertical one (Z) directions. In this case, the tendency was maintained for all tested models (model I1 , I2 , II1 , II2 , III1 , III2 ), i.e. the lowest values of displacements were obtained in the vertical direction Z. Moreover, a greater dispersion of the obtained results between particular directions can be noticed, depending on the method of creating a geodesic dome (the first or the second method). In the first method, the differences were much greater on the horizontal direction and amounted to as much as 63% (model II1 – 94.51 cm and model III1 – 35.40 cm). On the other hand, in the second method, the differences reached a maximum of 37% on the vertical direction (model I2 – 21.42 cm and model III2 – 34.16 cm).

Fig. 4. Displacements for all models.

It can be concluded that the obtained displacement results are higher than for the same domes of the I1 and I2 models [25] tested under seismic loads with a lower intensity of seismic excitations. In the paper Pilarska and Maleska [25], the maximum value of the displacement was 23.4 cm (the first method), therefore the obtained value was 75% lower than in the research of this paper.

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3.4 Accelerations from Numerical Analysis In the case of accelerations (Fig. 5), slightly different trends were noted as for the displacements. In this case, the II1 model (20.12 m/s2 ) created in accordance with the first method obtained higher acceleration values than the analogous II2 model generated according to the second method (18.88 m/s2 ). The difference in this case was 14%. In the other domes, however, we can see the greater importance of creating geodesic domes of the second method. This tendency increased with the increase of subdivision frequencies, i.e. the greatest disproportion was obtained between the III1 (20.20 m/s2 ) and III2 (23.93 m/s2 ) models and it reached 16%. When analyzing the obtained results, it can also be noticed that in most of the analyzed models (except for the III1 model) the lowest values for the tested models were obtained in the vertical direction Z. In this case, as in the case of displacements, the disproportions increased with the increase of subdivision frequencies.

Fig. 5. Accelerations for all models.

4 Conclusions After the numerical analysis, it can be concluded that the methods of creating geodesic domes and the subdivision frequencies of the initial face of the regular octahedron being the basis for generating the developed structures have a significant impact on the response of the domes under a given strong seismic load. Moreover, it can be stated that:

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• the first applied method of the subdivision is more sensitive to the effects of seismic excitations based on displacements, • an increased subdivision frequency of the analyzed geodesic domes causes greater displacements and accelerations, • the lowest values of displacements and accelerations (for almost all designed domes analyzed) were obtained for the vertical direction Z, Therefore, the analysis of the effect of seismic excitations on the geodesic domes generated on the basis of the regular octahedron is correct and shows a significant impact on their seismic response. In the future, it is planned to carry out numerical studies for a larger number of octahedron – based geodesic domes, i.e. for two different methods of creating a structure and for a different subdivision frequencies of the initial face of the regular octahedron, and therefore to investigate the impact of the mass of geodesic domes on the seismic response. The conducted research will be helpful in designing and assessing the seismic vulnerability of geodesic domes (methods of their creation) in seismic areas. Moreover, the obtained results show that there is no unequivocal trend determining the effect of seismic excitations on geodesic domes, so further numerical and experimental studies in this area are necessary.

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28. Maleska, T., Beben, D., Nowacka, J.: Seismic vulnerability of a soil-steel composite tunnel – Norway Tolpinrud railway tunnel case study. Tunn. Undergr. Space Technol. 110, 103808 (2021). https://doi.org/10.1016/j.tust.2020.103808 29. Maleska, T., Bonkowski, P., Beben, D., Zembaty, Z.: Transverse and longitudinal seismic effects on soil-steel bridges. In: Köber, D., De Stefano, M., Zembaty, Z. (eds.) Seismic Behaviour and Design of Irregular and Complex Civil Structures III. GGEE, vol. 48, pp. 23–36. Springer, Cham (2020). https://doi.org/10.1007/978-3-030-33532-8_3

Algorithms for the Near-Real Time Identification and Classification of Landslide Events Detected by Automatic Monitoring Tools Alessandro Valletta1(B)

, Andrea Carri2

, Roberto Savi2 , and Andrea Segalini1

1 Department of Engineering and Architecture, University of Parma, Parco area delle Scienze

181/a, 43124 Parma, Italy [email protected] 2 ASE – Advanced Slope Engineering S.r.l., Via Robert Koch 53/a, Fraz. Pilastrello, 43123 Parma, Italy

Abstract. Early Warning Systems (EWS) represent one of the most effective activities for risk management and mitigation associated to landslides occurrence, allowing to reduce the possibility of human losses and damages to involved structures. One of the most challenging aspects of early warning activities is the reliable detection of potentially critical events starting from monitoring outcomes. In fact, a timely assessment of the landslide evolution would allow to undertake the most appropriate mitigation measures for a correct management of the ongoing event. In this context, it is especially important to avoid the occurrence of false alarms, which could induce significant issues from a social and economic point of view. The algorithm described in this study was conceived with the intent to identify an increasing pattern in landslide displacement rates and provide a classification of the detected event. The analysis is fully automated and relies on the elaboration of monitoring datasets sampled by automatic instrumentation to define the onset of acceleration of the event and assess its alert level. The proposed procedure was integrated in an automatic software developed for the elaboration of monitoring data sampled by automatic inclinometers, analyzing several datasets coming from different sites of interest with the objective of detecting the occurrence of unusual behaviors in the monitored landslides evolution. Keywords: Landslides · Monitoring · Early Warning Systems

1 Introduction Among the different measures and interventions aimed at the reduction of risks related to landslide events, Early Warning Systems (EWS) represent a relevant option especially in those cases where structural measures are not viable for economical or practical reasons, or if they are unable to guarantee the safety of areas interested by instability phenomena. Taking as reference the definition provided by the United Nations Office for Disaster Risk Reduction (UNDRR), an EWS can be described as “an integrated system of hazard monitoring, forecasting and prediction, disaster risk assessment, communication and © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 74–84, 2023. https://doi.org/10.1007/978-3-031-26879-3_6

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preparedness activities systems and processes that enables individuals, communities, governments, businesses and others to take timely action to reduce disaster risks in advance of hazardous events” [1]. Several authors over the years have proposed different EWS general schemes, trying to highlight the key activities that should always be included in the design process of a landslide-oriented EWS [2–6]. On this basis, it is possible to state that a complete and effective EWS should include four main components: risk knowledge, monitoring and warning service, dissemination and communication, and response capability. The correct functioning of the entire system depends on these elements, and a weakness in a single one of these could result in the failure of the entire system [7]. It is also possible to define a series of features and criteria that any EWS must fulfil, including ease of implementation and comprehension, redundancy, precision, autonomy, robustness, and affordability [8]. One of the most crucial and complex aspects in early warning activities is the identification and evaluation of a potentially critical event, represented by a significant acceleration of slope displacements. Traditionally, the identification of the onset-of-acceleration (OOA) and the following classification of the detected event have been performed manually on the available monitoring dataset [9–11]. This approach, involving the first-hand evaluation of an expert, while potentially leading to accurate results, is definitely a suboptimal solution when a timely warning is needed. This is mainly due to the lack of automation, which is a key component for a near-real time approach to perform early warning activities [12, 13].

2 Materials and Methods The methodology described in this study was designed to provide a procedure to identify the presence of an increasing trend in landslide displacements and determine an alert level connected to the event. The elaboration process relies on two separate algorithms, which are carried out in series to generate the outcome of the analysis. 2.1 OOA Identification Algorithm The first stage of the elaboration process aims to determine the presence of an accelerating pattern to assess the onset-of-acceleration, representing the beginning of the potential slope instability. The OOA identification algorithm here discussed relies on four different criteria, which the authors firstly introduced and discussed in detail in a previously published paper [14]. This analysis is performed through a multi-level algorithm designed with a “dropdown” structure, which allows to define a number of sub-conditions to verify specific features of the dataset. The fundamental hypothesis of the method is based on the presence of a transition from constant to increasing displacement rates, corresponding respectively to a linear and a non-linear behavior according to the creep theory [15]. Taking as a reference the reading j and the corresponding displacement velocity vj , the following conditions are imposed:

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• Criterion 0: The dataset must include four consecutive positive velocities. This is intended as a preliminary and mandatory step, needed to obtain meaningful results from the application of forecasting models, such as the Inverse Velocity Method [16], to the selected dataset. • Criterion 1: Displacement rate values should increase as the monitoring activity progresses over time, i.e., available data are displaying a positive acceleration over time. • Criterion 2: The displacement rate trend should fit a parabolic trend with an upward concavity (this operation is performed on a dataset composed of 10 values). The identification of this type of pattern refers to one of the features of the tertiary creep phase, represented by a non-linear evolution. • Criterion 3: The curve interpolating the monitoring data should present a more pronounced concavity as the monitoring activity progresses over time. This is intended to express another characteristic of a potentially critical acceleration phase, i.e., a curve which increases its slope as it gets close to the theoretical failure.

Fig. 1. Schematic representation of the structure of the OOA identification algorithm, evidencing each sub-condition and the corresponding success rates.

The procedure is summarized in Fig. 1 where it is also reported the success rate of each sub-criterion, i.e., the number of elements that must fulfil the corresponding condition in order to continue the analysis. The introduction of a tolerance margin is intended to prevent errors deriving from anomalies in monitoring data. For example, a too strict rate could lead to the interruption of the analysis due a single negative result in one of the conditions, which would be interpreted as an actual deceleration in the available dataset. If the last step returns a positive outcome, it could be assumed that an accelerating phase is taking place, and the j − 3 reading could be taken as the OOA of the monitored event. 2.2 Event Classification Algorithm The second part of the algorithm was developed to classify the hazard level posed by an event previously identified from monitoring data. The methodology takes advantage of

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a multi-level system that allows to take into account both trends presenting a potentially critical behavior and minor events that could be still taken into account and analyzed even without leading to the slope collapse. The main advantage coming from this approach lies on the possibility to provide an adequate dissemination of information related on the ongoing phenomenon, depending on the level reached by the observed event. This system was designed to work in synergy with the OOA identification algorithm, which has the task to define the dataset that should be processed through the multi-level procedure. The basis of the classification algorithm relies on a set of three parameters, which were selected to be taken as a reference for the definition of the alert level reached by the analysis: • The determination coefficient R2 obtained from the application of the Inverse Velocity Method (IVM) under the hypothesis of linear behavior, which has been considered a good approximation to describe the landslide behavior as the slope gets closer to the final collapse [17, 18]. Therefore, it could be assumed that the achievement of a good results in the linear interpolation of the dataset (i.e., a high determination coefficient value) could be indicative of a potentially hazardous event. • The number of monitoring data included in the IVM analysis. The choice of this parameter come from the assumption that the landslide behavior approximates a tertiary creep curve when approaching its failure. Therefore, a dataset containing a high number of monitoring points could represents a trend consistently accelerating over time, which could be a sign of a critical developing event. • The number of sensors included in the IVM analysis after fulfilling all conditions imposed by the OOA identification algorithm. This parameter comes into plays since the methodology was originally designed for a monitoring system composed of automatic inclinometers integrating several sensors in the same array. Because of this, a higher number of sensors displaying an accelerating trend could be an indication of the relevant magnitude of the detected event. Each time a new displacement measurement reaches the elaboration center, the automatic software activates the OOA identification routine in order to verify if the displacement rate and acceleration data display a potentially critical trend. A positive outcome should be obtained from the analysis to continue the elaboration process and execute the classification algorithm, which depends on the three parameters previously described. At the end of this analysis, the model is able to assign to the detected event a level ranging from 1 to 5 depending on the parameters value. In this context, levels from 1 to 3 identify different ranges of activity on the monitored slope, even if they do not represent an immediately critical situation. On the other hand, the achievement of higher levels is usually related to a situation where the monitored phenomenon is showing several different signs of unusual activity at the same time. Therefore, Level 4 is generally observed in correspondence of substantial movements recorded by monitoring tools installed on site and could be considered as a sort of pre-alert condition. Level 5, the highest level achievable by the classification algorithm, could be related to a situation where the monitoring instrumentation highlight an unstable condition of the monitored slope, characterized by an extremely active phenomenon (Fig. 2).

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Fig. 2. Block diagram summarizing the procedure to assess the alert level connected to an observed event according to the parameters featured by the collected dataset.

The present version of the model was included in the software designed for the analysis of displacement data recorded by automatic inclinometers based on MUMS (Modular Underground Monitoring System) technology. The instrumentation is composed of a series of epoxy resin nodes, called Links, connected by a quadrupole electrical battle and an aramid fiber cable. The result is an arbitrary long array of sensors, which can be located at specific distances according to specific requirements and can measure different parameters depending on the monitoring activity [19, 20]. Starting from October 2019, the algorithm has been continuously applied to several monitored sites, elaborating up to 23,000 new datasets every single day. The outcomes were used to calibrate the models and provided several case studies for the validation process.

3 Results and Discussion This chapter includes two examples of the application and outcomes deriving from the application of the OOA identification and event classification algorithms. The case study here considered regards a monitoring site located in a Northern Italy region, characterized by the presence of a construction site of a new transport infrastructure crossing a mountainous and hilly area. In order to control the interaction between the structure and the surrounding environment, the site was equipped with a MUMS-based monitoring system. In particular, a total of 4 Vertical Arrays (i.e., automatic inclinometers, Fig. 3) was installed in different parts of the area of interest to control both displacements and water level variations. Each single Array was produced with specific features, integrating a different number and typology of sensors according to the monitoring plan (Table 1). After their installation and the subsequent definition of the reference date, all Vertical Arrays have continued the monitoring activity to the present day, with a sampling frequency of 6 readings every day.

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Fig. 3. a) Vertical Array elements and working principle; b) Tilt Link HR 3D V components.

Table 1. ID and features of the monitored instrumentation installed on site. Array ID [-]

Installation date [dd/mm/yyyy]

Array typology

Sensors number and typology

Array length [m]

DT0099

06/03/2019

Vertical Array

20 × Tilt Link HR 3D V 2 × Piezo Link

20.00

DT0100

05/12/2018

Vertical Array

20 × Tilt Link HR 3D V 2 × Piezo Link 1 × Baro Link

20.00

DT0101

06/03/2019

Vertical Array

20 × Tilt Link HR 3D V 1 × Piezo Link 1 × Baro Link

20.00

DT0102

06/03/2019

Vertical Array

20 × Tilt Link HR 3D V 1 × Piezo Link

20.00

Despite never reaching the highest level of the classification algorithm, the site generated some activations of the automatic procedure achieving lower alert levels. The following section presents two examples of these occurrences, both measured by Vertical

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Array DT0101, describing the displacement trend and the corresponding level achieved by the analysis. 3.1 Event #1: March 2020 The first example refers to a movement observed by the monitoring device in March 2020 when the instrumentation reported an increase of displacement measured by Tilt Link 31, located at 4 m of depth. After the activation of the OOA identification algorithm, which identified the beginning of the accelerating trend on 5 March 2020 11:27 AM, the automatic software processed the selected dataset in order to perform the failure forecasting analysis and evaluate the parameters needed to classify the event. The outcome of this analysis showed that only a single node of the Array (i.e., Tilt Link 31) detected a trend compatible with the requirements imposed to perform the classification analysis. Moreover, the corresponding displacement dataset consisted of 5 monitoring points. The failure forecasting analysis returned a determination coefficient equal to 0.9419, indicating a very good correspondence between inverse of velocity values and the line interpolating the monitoring data. Consequently, according to the value of the three parameters, the automatic software classified this event as a Level 3 (configuration C3) (Fig. 4).

Fig. 4. Displacement measured over time by Tilt Link 31, with a detail regarding the event of March 2020 detected by the automatic software.

Following the results provided by the analysis, the monitored site was kept under closer observation in order to study the monitoring data evolution over time. However, both displacement and velocity trends did not show a particularly alarming pattern, and following measurements evidenced a stable configuration. 3.2 Event #2: December 2019 The second example regarding this specific site involves another event, detected once again by Vertical Array DT0101 in December 2019. The instrumentation measured a steep increase of slope displacements on Tilt Links 31 and 35, located respectively 4 and 2

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m below the ground level. In both cases, the automatic software processed the datasets to identify the beginning of the acceleration phase, locating the OOA date on 21 December 2019 10:05 AM, and proceeded to the level evaluation. In this case, since two datasets displayed an accelerating trend starting from the same date, the software performed two separate analyses to compute a single classification level for the whole event. Monitoring data provided by Tilt Link 31 provided a fairly non-linear trend in the inverse-velocity vs time plot, with the failure forecasting process returning a determination coefficient R2 = 0.8777 evaluated on a 5-point dataset. On the other hand, the application of the IVM algorithm on values recorded by Tilt Link 35 showed a result more consistent with the linear interpolation, generating a determination coefficient of 0.9375. As well as for the other Link, this analysis was also performed on a dataset composed of 5 monitoring points. At the end of the elaboration process, based on the parameters during the analysis, the automatic software classified the event as a Level 4 (specifically a C1 +C3 configuration). Compared to the example previously reported, the higher level achieved by this event reflects a more serious situation where the movement identified by the monitoring devices involves different parts of the landslide. Given the fact, the two nodes interested by this phenomenon are only 2 m apart, and considering the similarities between the monitoring data (Figs. 5 and 6), it could be assumed that the displacements measured by the two sensors are strictly related.

Fig. 5. Displacement measured over time by Tilt Link 31, with a detail regarding the event of December 2019 detected by the automatic software.

Additionally, by observing the slope displacement vs time plots, it is noticeable how the magnitude of this event is higher compared to the previous example. In fact, the total displacement generated by the event resulted to be 23.4 mm and 25.6 mm for Tilt Link 31 and Tilt Link 35, respectively. Even if this specific parameter is not involved in the level assessment procedure, it is safe to assume that it played an indirect role in the achievement of this alert level.

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Fig. 6. Displacement measured over time by Tilt Link 35, with a detail regarding the event of December 2019 detected by the automatic software.

Upon receiving the notification of the achievement of this level, surveillance of monitoring data was increased to identify further signs of potential instabilities. Moreover, data collected by the piezometer showed a slight and sudden variation in the water table level, evidencing a possible correlation with the movement detected by the Array. Following recording evidenced once again the reaching of a stable conditions, with slope displacements following a horizontal trend. The integration of a redundant sensor allowed also to perform a check on the measured movements validity, which were confirmed by the second sensor placed in each Tilt Link presenting the same trend both during and after the event. Considering all these information, no further actions were undertaken, and subsequent on-site inspections performed for maintenance reasons did not display any relevant damages on the slope surface.

4 Conclusions The monitoring activity of slope instabilities and landslides has gained increasing importance in recent years, becoming a fundamental component in any landslide Early Warning System. Notably, recent studies have focused their attention on improving automation and data elaboration procedures thanks to the introduction and widespread of latestgeneration technologies. This paper presents a methodology designed to tackle one of the most critical aspects of the data analysis process, aimed to the detection and classification of potentially critical events. The first step relies on a multi-criteria algorithm designed to identify automatically an accelerating trend starting from displacement monitoring data. Specifically, the proposed approach involves a drop-down sequence composed of four steps applied to each single data sample, allowing to identify the OOA (onset-of-acceleration) point referred to the accelerating trend. Following the critical trend identification, a classification procedure is performed to define the alert level for the event of interest. The evaluation depends on three different parameters representing specific aspects of the landslide behavior. All these features are arranged in a 5-level structure to automatically provide a classification of the potentially critical events, according to the analysis outcomes.

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The present work includes two events displaying a practical application of the procedure previously described. Both examples refer to a case study where the site activity led to the observation of several displacements during the monitoring activity. For each one of these datasets, the onset of acceleration of the event was correctly identified, and the corresponding alert level was evaluated. In the first example, the software detected an event that resulted in a minor movement at a specific depth, achieving a Level 3 in the classification analysis. The second event involved two sensors placed at different depth and generated a notable increase in slope displacements, resulting in a Level 4 according to the elaboration software. In both cases, the event classification allowed to disseminate appropriate messages to inform the authorities responsible for the monitoring activity regarding the behavior of the studied area.

References 1. UNDRR: Report of the open-ended intergovernmental expert working group on indicators and terminology relating to disaster risk reduction. United Nations General Assembly (2016) 2. UNISDR: Developing early warning systems, a checklist. In: Third International Conference on Early Warning (EWC III), United Nations, Bonn, Germany (2006) 3. Di Biagio, E., Kjekstad, O.: Early warning, instrumentation and monitoring landslides. In: Proceedings of the 2nd Regional Training Course, RECLAIM II, Phuket, Thailand, pp. 1–24. Asian Disaster Preparedness Centre (2007) 4. Intrieri, E., Gigli, G., Casagli, N., Nadim, F.: Brief communication “Landslide Early Warning System: toolbox and general concepts.” Nat. Hazards Earth Syst. Sci. 13, 85–90 (2013). https://doi.org/10.5194/nhess-13-85-2013 5. Calvello, M., d’Orsi, R.N., Piciullo, L., Paes, N., Magalhaes, M., Lacerda, W.A.: The Rio de Janeiro early warning system for rainfall-induced landslides: analysis of performance for the years 2010–2013. Int. J. Disaster Risk Reduction 12, 3–15 (2015). https://doi.org/10.1016/j. ijdrr.2014.10.005 6. Fathani, T.F., Karnawati, D., Wilopo, W.: An integrated methodology to develop a standard for landslide early warning systems. Nat. Hazards Earth Syst. Sci. 16, 2123–2135 (2016). https://doi.org/10.5194/nhess-16-2123-2016 7. de León, J.V.D., Bogardi, J., Dannenmann, S., Basher, R.: Early warning systems in the context of disaster risk management. Entwicklung and Ländlicher Raum 2, 23–25 (2006) 8. Stähli, M., et al.: Monitoring and prediction in early warning systems for rapid mass movements. Nat. Hazards Earth Syst. Sci. 15, 905–917 (2015). https://doi.org/10.5194/nhess-15905-2015 9. Voight, B., Kennedy, B.A.: Slope failure of 1967–1969, Chuquicamata mine, Chile. In: Voight, B. (ed.) Developments in Geotechnical Engineering, pp. 595–632. Elsevier (1979) 10. Osasan, K.S., Stacey, T.R.: Automatic prediction of time to failure of open pit mine slopes based on radar monitoring and inverse velocity method. Int. J. Min. Sci. Technol. 24, 275–280 (2014). https://doi.org/10.1016/j.ijmst.2014.01.021 11. Mazzanti, P., Bozzano, F., Cipriani, I., Prestininzi, A.: New insights into the temporal prediction of landslides by a terrestrial SAR interferometry monitoring case study. Landslides 12, 55–68 (2015). https://doi.org/10.1007/s10346-014-0469-x 12. Allasia, P., Manconi, A., Giordan, D., Baldo, M., Lollino, G.: ADVICE: a new approach for near-real-time monitoring of surface displacements in landslide hazard scenarios. Sensors 13, 8285–8302 (2013). https://doi.org/10.3390/s130708285

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13. Valletta, A., Carri, A., Segalini, A.: Innovative monitoring instruments as support tools for natural risks management. Rendiconti Online Della Soc. Geol. Ital. 48/2019 (2019). https:// doi.org/10.3301/ROL.2019.44 14. Valletta, A., Carri, A., Segalini, A.: Definition and application of a multi-criteria algorithm to identify landslide acceleration phases. Georisk: Assess. Manag. Risk Eng. Syst. Geohazards 16, 555–569 (2021). https://doi.org/10.1080/17499518.2021.1952610 15. Tavenas, F., Leroueil, S.: Creep and failure of slopes in clays. Can. Geotech. J. 18(1), 106–120 (2011). https://doi.org/10.1139/t81-010 16. Fukuzono, T.: A new method for predicting the failure time of a slope. In: Proceedings of the Fourth International Conference and Field Workshop on Landslides, pp. 145–150. Tokyo University Press, Tokio (1985) 17. Rose, N.D., Hungr, O.: Forecasting potential rock slope failure in open pit mines using the inverse-velocity method. Int. J. Rock Mech. Min. Sci. 44, 308–320 (2007). https://doi.org/10. 1016/j.ijrmms.2006.07.014 18. Dick, G.J., Eberhardt, E., Cabrejo-Liévano, A.G., Stead, D., Rose, N.D.: Development of an early-warning time-of-failure analysis methodology for open-pit mine slopes utilizing ground-based slope stability radar monitoring data. Can. Geotech. J. 52, 515–529 (2015). https://doi.org/10.1139/cgj-2014-0028 19. Segalini, A., Chiapponi, L., Pastarini, B., Carini, C.: Automated inclinometer monitoring based on micro electro-mechanical system technology: applications and verification. In: Sassa, K., Canuti, P., Yin, Y. (eds.) Landslide Science for a Safer Geoenvironment, pp. 595–600. Springer, Cham (2014). https://doi.org/10.1007/978-3-319-05050-8_92 20. Carri, A., Chiapponi, L., Giovanelli, R., Spaggiari, L., Segalini, A.: Improving landslide displacement measurement through automatic recording and statistical analysis. Procedia Earth Planet. Sci. 15, 536–541 (2015). https://doi.org/10.1016/j.proeps.2015.08.091

Analysis of Dynamic Compaction Effects – Tavira Case Study Elisa M. J. da Silva(B)

and Miguel Oliveira

Institute of Engineering, University of Algarve, Faro, Portugal {esilva,mjolivei}@ualg.pt

Abstract. Ground vibrations are associated with different types of elastic waves propagating on the ground, and wave propagation depends on the soil stiffness, dumping and distance from the source. Urban activities also generate vibration, which are mainly associated to traffic of heavy vehicles, railway compositions, infrastructures construction, and maintenance. On site construction, ground wave propagation occurs when pile driving, dynamic compaction, blasting, and operation of heavy construction equipment is required. Regarding vibration energy analysis in structures, the Peak Particle Velocity (PPV) is the reference parameter and according to the Portuguese standard NP 2074 (2004), “Evaluation of the vibration impact on buildings, caused by explosions or similar construction procedures”, the PPV can range from 5 mm/s up to 20 mm/s, for rigid structures founded, respectively, on sandy soils and soft clays, or rocks and hard clays. The study hereby presented is a real case study where the effect of heavy vibratory tandem compaction, associated to the construction of an industrial parking in the vicinity, caused extensive wall cracks on a reinforced concrete structure, located 60 m from the site construction. The structure is an unifamilial building, located on the northern urban area limit of Tavira city, in the Algarve, Portugal. Keywords: Ground vibrations · Vibratory compaction · Structure’s damage

1 Introduction Ground vibrations are associated with different types of elastic waves (i.e., bulk, shear, and superficial waves) propagating on the ground. Wave propagation depends on the soil stiffness, dumping, and distance from the source [1]. Artificial vibratory energy from urban activities generates waves in the frequency range of 1 Hz to 150 Hz, whilst for natural vibration sources, such as earthquakes, the predominant energy transport is associated with waves of the frequency between 0.1 Hz and 30 Hz, and for the wind load effect in tall buildings, it is from 0.1 Hz to 2 Hz [2]. The common sources of vibration in urban areas are mainly the traffic of heavy vehicles, railway compositions (trains and undergrounds), infrastructures construction, and maintenance. The main sources of ground vibrations associated to site construction are pile driving, dynamic compaction, blasting, and operation of heavy construction equipment. These vibrations may harmfully affect surrounding buildings, and their effect © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 85–95, 2023. https://doi.org/10.1007/978-3-031-26879-3_7

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ranges from disturbance on residents, aesthetical cracks on walls to visible structural damage. They are usually felt by the habitants, leading to discomfort on users, due to slab, windows or doors movements, lamps displacements, as well as vibration of object inside cabinets. The waves are propagated by the superficial soil/rock layer and primarily received by the building foundation, inducing a displacement on it, which, in turn, will be transmitted to the structure. The response of the building to this excitation depends on the natural frequencies of the structure, its modes of vibration, and damping [2]. If the resonance frequency of the building structure is similar to the frequency of propagation of the vibratory wave, there will be an amplification of soil/foundation displacement, causing an unwanted impact. For rigid structures founded on medium to soft soils, the walls displacements are neglectable due to the stiff structure response at lower frequencies. For flexible structures founded on stiff soils and rocks, the soil-structure displacements at the interface are low, however the structure will transmit the vibration, and consequently cracks may occur on masonry walls [3]. In this last case, the wave frequency has high impact on the lower dumping building components (walls, ceilings, modular lining, divisions), since they have higher natural frequencies and therefore are more influenced by the continuous vibrations (fatigue) [4]. Regarding vibration energy analysis in structures, the Peak Particle Velocity (PPV) is the parameter used to measure it. According to the Dutch standard DIN 4150-3, there are three building categories, and they can support different limit levels of PPV [5]. For steel or reinforced concrete structures, they tend to have better responses to vibration effects. Nevertheless, even with small PPV, these structures can suffer damage when founded on stiff soils, as mentioned earlier. The Portuguese standard NP 2074, “Evaluation of the vibration impact on buildings, caused by explosions or similar construction procedures”, define PPV limits according to the soil foundation, structure type, and constructive typology [6]. For conventional structure, the PPV can range from 5 mm/s for sandy soils and soft clays up to 20 mm/s for rocks and hard clays. The difference vibratory wave velocity in different types of soils is associated to the elastic impedance of each one.

2 Case Study Description The study hereby presented is a real case study where the effect of heavy vibratory tandem compaction was felt on an unifamilial building, causing vibration on the reinforced concrete structure (RCS) and extensive wall cracks. During the compaction works the residents captured, on video, lamps bouncing on ceilings, dishes vibrating inside cupboards, and doors moving. The compaction works were associated to the construction of an industrial fruit packaging and distribution park, located 60 m from the unifamilial building. The site construction parking place, for the compaction equipment’s, was only 41 m away (Fig. 1).

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Fig. 1. Distances from the damaged unifamilial building to the industrial park under construction and to the compaction equipment’s parking place (Goggle Earth aerial image, taken on February 2020).

The compactions works were interrupted immediately after the alarm, being followed by an assessment of the structure pathologies and by the project revision. The characteristics of the compaction equipment’s, as well as the structure, local geology, lithology, and geotechnical characterization were fundamental for the comprehension of the phenomena. 2.1 Unifamilial Building Characterization The building is a three-floor house, composed of a basement, ground floor, and 1st floor, built in 2007, and with 304.90 m2 of constructed area. The building is a reinforced concrete structure, founded on shallow foundations and with stiff concrete walls on the basement contour. The superstructure consists of columns, beams, and massive slabs. The concrete and steel characteristics are: C16/20 for foundations; C20/25 for all the others concrete elements (slabs, beams, pillars, and walls); S400 in all structural elements. Considering the information gathered on the Housing data sheet, the surrounding exterior walls have a thickness of 33 cm and are composed of two cloths of 11 cm bored brick masonry. In between, there is an air box, partially filled with 30 mm thick thermal insulation wall mat, fixed on the interior cloth. All façade walls are lined outwardly by plastering and inwardly by designed stucco. The hillside walls are simple of perforated brick with a total thickness of 25 cm, coated inside with projected stucco. All interior walls consist of simple cloths of 11 cm bored brick masonry, coated with designed stucco, except for the kitchen and sanitary facilities where ceramic tile was applied.

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According to the proprietary, the exterior walls were repainted on 2015/2016 and the interior between May to July of 2018. 2.2 Geographical Location and Geomorphological Landscape The structure is located on the northern limit of Tavira city urban area, in the Algarve, Portugal (Fig. 2).

Fig. 2. Geographical placement of the unifamilial building in the Algarve region (source: Google Earth, 2022).

The city extends from the south coastline, baith by the Atlantic Ocean, known as “Litoral”, to the northern smooth hills, designated as “Barrocal”. Algarve’s morphology is a mixture of different landscape. The “Serra” region is characterized by steep slopes with rounded hills and dispersed rocky outcrops. “Barrocal” as a slightly rugged landscape, similar to “Serra”, and the hills are covered by Mediterranean vegetation, forming a unique landscape. Regarding the “Litoral” coastline, the Western part is characterized by rocky cliffs and the Eastern part by sandy dunes, islands, and long beaches (Fig. 3).

Fig. 3. Algarve region geographical division: “Litoral” = coastline, “Barrocal” = in between zone, “Serra” = mountains (source: [7]).

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2.3 Geological and Lithological Site Characterization Algarve is a complex Mesozoic sedimentary basin. The most recent formations (Jurassic, Cretaceous, Miocene, Pleistocene, Pliocene and Holocene) are located in the South and Southeast, whilst the oldest formation (Paleozoic) is on the West and North of the Algarve (Fig. 4). Algarve’s lithology is also very rich. The sandstones, conglomerates, and sands, associated to sedimentary formations, are the dominant type of rock massifs along the “Litoral” coastline. Clay shales and graywacke prevail on the “Serra”, however, an igneous formation intrusion, constituted by nepheline syenite rocks, is the “Serra de Monchique” birthplace. “Barrocal”, the in-between zone, is mainly constituted by limestones, sand stones, and dolomites (Fig. 5).

Western Algarve Western Sub-basin

High-BoƩom of Budens-Lagoa

Eastern Sub-basin

Eastern Algarve

Middle-Lower Jurassic

Igneous Syenite Intrusion

Holocene

Upper Triassic and Lower Jurassic

LocaliƟes

Lower Cretaceous

Soco Paleozoic

Cenozoic:

Miocene-

Faults

Fig. 4. Simplified geological map of Algarve (source: [8], based upon [9]).

The unifamilial structure is on the fringe of “Litoral” and “Barrocal”, where conglomerates, sandstones, clayey sandstones, limestones usually dolomitic, marls and marly limestones from the Upper Triassic-Lower Jurassic, and dolerites from the Middle-Lower Jurassic are the predominant formations. In this specific zone, the massif is relatively close to the soil surface, and rocky outcrops are common.

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Alluvions Sands, rolled pebbles, poorly consolidated sandstones, clays

Dolerites

Sands and gravels

Dunes and eolic sands

Sandstones, limestones somewhat marly, sands, gravels, clays

Gabbros

Sandstones, conglomerates, limestones, dolomiƟc, marls,…

Basalts

Clayey sandstone, conglomerate, marls, limestones usually dolomiƟc

Nepheline syenite

Limestones, dolomiƟc and marly limestones, marls Conglomerates, sandstones, limestones, dolomiƟc, marls,…

Clay

sales,

graywackes,

sandstones

Fig. 5. Algarve lithological map (source: [8], adapted from [10]).

2.4 Pavement Characterization and Compaction The pavement of the loading/unloading parking zone was not considered on the initial design project, and therefore, this was an additional work. The pavement characteristics adopted were proposed by the contractors and defined after visual inspection of the soil foundation, after scarification. Fragments of rocks were detected on the soil surface and no “in-situ” tests were performed. The pavement characteristics prescribe were: 1. Granular sub-base with variable thickness, aiming to level the parking pavement. 2. Well grade base, preferably tout-venant, with 30 cm (15 cm + 15 cm). 3. Asphalt layer, in bituminous material, with 10 cm. After compaction the sub-base and base course materials must met the following criteria: 1. At least 95% of the Modified Proctor test. 2. The layers must be perfectly stable. Additional criteria:

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1. Moisture: A uniform distribution of water must be carried out, using pressure whose jet tanks, if possible, in order to cover the total width of the compacted area. The water spreading should be continuous and maintaining the same speed, however soil saturation cannot occur. If case of excessive moisture, compaction works must be delay, giving time for the material to dry. 2. The upper surface layer should present a uniform texture, free of cracks, undulations or loose material and may not, at any point, present differences greater than 2.5 cm in relation to the longitudinal and transverse profiles established. 3. The total thickness of the sub-base, after compaction, must respect the valor prescribe. If the projected thickness is not attended, then scarification and correction will be executed. However, if supervisors agree, the compensation thickness of sub-base layer can be added to the base layer and increase its thickness. A motor leveling machine or other similar equipment was used for the spreading of the material. Spreading will be done regularly to warranting a perfectly homogeneous layer. If sulcus, wheel ridges or other inconvenient marks were perceptible on the layer surface, and correction of the deformations are imposed. 2.5 Compaction Equipment and Procedure Due to heavy raining on the beginning of the compaction works, the tout-venant already applied had to be completely removed, since it showed high water content. It was also remarked excessive percentage of fine particles on the aggregates – therefore, it was rejected. After this incident only “first category” tout-venant was used on compaction. The first compaction equipment choice falls within a heavy vibratory single drum smooth roller – ABG Alpha 190V cylinder with an operating weight of 11 ton and frequencies ranging from 30 Hz to 35 Hz [11] (Fig. 6a). However, after the heavy vibration effects, felt on the surroundings, the contractor opted for a light tandem smooth roller – Ammann ARX36 with an operating weight of 4ton and frequencies ranging from 41 Hz to 55 Hz [12] (Fig. 6b). The vibrations reduced significantly, but the number of cylinder passages necessary to ensure the compaction criteria increased, taking more time.

a)

b)

Fig. 6. Images of the compaction equipment: a) vibratory single drum smooth roller, b) vibratory tandem smooth roller.

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2.6 Non-structural Pathologies A technical assessment was carried out to determine the extension of the damages on the building, affected by the vibrations, and no structural anomalies were founded. However, extensive non-structural pathologies were easily perceptible and visible, especially on the wall coverings and ceilings, both exterior (Fig. 7a, b) and interior (Fig. 7c, d).

a) Joint displacement with implication to other elements of the contiguous dwelling.

b) Cracking on the balcony wall.

c) Cracks on the living room near the celling.

d) Tiles from the bathroom fractured.

Fig. 7. Images reporting visible cracking of the exterior and interior covering walls.

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3 Proposed Solution Since the causes of the pathologies on the unifamilial building were caused by the compactions works, they were immediately interrupted after the alarm, followed by the decision of contacting an expertise team, to advised them on the type of compaction machinery to use and for the building repairing. 3.1 Compaction Equipment Proposed The major implications of using a vibrator compaction equipment on the vicinity of this urban area, where the upper soil is rather stiff, is the vibration propagation. In the case study, the effect on the structure building was amplified due to its stiffness. A rigid reinforced concrete structure will vibrate as a part of the ground itself and the basement moves along with the ground, passing the waves to the above structure [13]. The solution to minimize the vibrations passed by choosing a static tandem roller or an oscillation compaction equipment. This last one was considered the most adequate. Oscillation works by using exciters to move the drum back and forward, rather than directing the forces downwards as a conventional drum. This effect is achieved by having twin, coupled out of balance exciters synchronized, that rotate at the same speed but in opposite directions [14] (Fig. 8). The direction of the forces is a key issue for oscillation since they act horizontally and generate shear forces. These causes the soil to move both downwards due to the weight (load) applied, but also laterally. So, the drum remains in contact with the layer surface, delivering both dynamic as well as static loads to squeeze out any voids. The aggregates are redistributed within the mat, instead of being damaged and this avoids the risk of over-compaction.

Fig. 8. Forces and moments of dynamic drums: vibrating drum (left), oscillating drum (right). (source: [14]).

A crucial advantage associated to this type of compaction is the 85% reduction of vibration, to obtain the same compaction effect. By comparison, when using conventional vibration techniques, the out of balance (up and down) weights act only vertically, and these forces can be reflected off underlying layers, travelling some distance through the ground. In urban areas, this can be of concern as it may affect surface structures of all types in the vicinity.

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Both soil and asphalt compactors are available with oscillating type drums in the market, but the technology is of particular benefit for asphalt applications, and therefore its common to only find them being used by pavements contractors. The oscillating compaction method can be used effectively for base, binder and wearing courses. 3.2 Walls Repairing Solution and Procedure Regarding the interior wall coverings repair, it was proposed the replacement of the cracked ceramic elements, as well as the ceiling/crown molding, followed by a complete painting. For the external elements, a slimy armed mortar was prescribed. However, the criteria requests for waterproofing, permeability to water vapor, support compatibility, and durability had to be taken into consideration when choosing the coating. The repairing procedure for the affected regions, was: 1. Pressure water washing to remove non-adherent paint, dirt, and other contaminants. Surfaces should be dry and without residues of aged paints afterwards. 2. Treatment of fungi with Antibiosis Decontaminant, N/Ref. 18-220, and wait 24 h before painting. 3. Plastering, with proper mortar, the bigger cracks, and for the smaller ones application of “Tapa cracks, N/Ref. 18-110”. 4. Application of Primary Cynolite HP, N/Ref. 10-850 on the repaired areas. 5. Application of a general fiber mortar Princol, N/Ref. 29-573. 6. Application of a sandy mortar, N/Ref. 29-575, for plaster regularization. The painting should only be done after 10 to 12 days after the mortar. 7. Application of a primary layer of painting Cinolite HP, N/Ref. 10-850. 8. Application of two to three coats of Novatex HD, N/Ref. 10-175.

4 Final Considerations Intense vibrations were video recorded once the compaction works of an industrial fruit packaging and distribution park started. Windows vibrating, doors and lamps movements, as well as dishes vibration inside cabinets (equivalent to a seismic phenomenon) were registered and observed for several days by the owner of the unifamilial affected building. The equipment, initially used for the parking sub-base and base aggregates compaction, could be considered adequate and efficient for the purpose if the rocky substrate was not so close to the surface. Also, the compaction roller drums frequencies were within the range of values usually used in this type of soil. This parameter is not the most determinant in the compaction effect, but it is for vibrations propagation. Thus, given the proximity to other structures built in the area, it can be stated that the initial compaction equipment was not the most appropriate, having transmitted the vibration into the rocky substrate, which were subsequently transmitted to the building foundations and superstructure. The major influence on compaction is the value of the wave amplitude (Peak Particle Velocity) transmitted by the equipment. Unfortunately, this parameter was not measured

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during the compaction works and consequently no comparison to the maximum PPV standards could be made. Despite the significant pathologies observed on the wall’s coatings, ceiling mouldings and tiles, especially on the ground floor divisions, no structural pathologies were found after the expertise assessment. The cracks on the exterior and interior walls are extensive and some present a considerable dimension. The reparation works had to be carried out immediately and a slimy armed coated was prescribed.

References 1. Assis Chaves, G.V., Pimentel, R.L., de Melo, R.A., de Farias, J.P.: Faixa de domínio e sua relação com a redução de vibrações produzidas por trens de superfície em áreas urbanas. Transportes 17 (2009) 2. ISO 4866:1990: Mechanical vibration and shock – Vibration of buildings – Guidelines for the measurement of vibrations and evaluation of their effects on buildings. https://www.iso. org/standard/10845.html. Accessed 15 May 2022 3. François, S., Pyl, L., Masoumi, H.R., Degrande, G.: The influence of dynamic soil-structure interaction on traffic induced vibrations in buildings. Soil Dyn. Earthq. Eng. 27, 655–674 (2007). https://doi.org/10.1016/J.SOILDYN.2006.11.008 4. BS 7385-2: Evaluation and measurement for vibration in buildings – Part 2: Guide to damage levels from groundborne vibration, Engineering 360. https://standards.globalspec.com/std/ 596323/BS7385-2. Accessed 15 May 2022 5. DIN 4150-3:1999: Structural vibration – Part 3: Effects of vibration on structures, Building CodeHub. https://codehub.building.govt.nz/resources/din-4150-31999/. Accessed 15 May 2022 6. Instituto Português da Qualidade (IPQ): Norma Portuguesa NP 2074: avaliação da influência em construções de vibrações provocadas por explosões ou solicitações similares, Lisboa (2004) 7. Duna, Sapal, Barrocal, Serra – Vida de Planta. https://vidadeplanta.wordpress.com/dunasapal-barrocal-serra/. Accessed 26 Apr 2022 8. Feroldi, A., da Silva, E., Gonçalves, M.M.: Drystone walls in the Algarve, Portugal. Characterization and interconnection with the geology and lithology. In: IOP Conference Series: Materials Science and Engineering, vol. 1203 (2021). https://doi.org/10.1088/1757-899x/ 1203/2/022129 9. Programa Reordenamento e Gestão da Paisagem das Serras de Monchique e Silves (PRGPSMS), DGT. https://www.dgterritorio.gov.pt/Programa-Reordenamento-e-Gestao-daPaisagem-das-Serras-de-Monchique-e-Silves-PRGPSMS. Accessed 15 May 2022 10. Sapientia: Geologia e génese do relevo da Rocha da Pena (Algarve, Portugal) e o seu enquadramento educativo. https://sapientia.ualg.pt/handle/10400.1/657. Accessed 15 May 2022 11. IR-ABG Alpha 190 V Fiches techniques & données techniques (1987–1995), LECTURA Specs. https://www.lectura-specs.fr/fr/modele/machines-de-chantier/rouleaux-rouleaux-mon ocylindres-ir-abg/alpha-190-v-21774#techSpecs. Accessed 15 May 2022 12. TANDEM ROLLERS, ARX 36-2, Ammann. https://www.ammann.com/en/machines/soiland-asphalt-compactors/tandem-rollers/arx-36-2. Accessed 15 May 2022 13. Taranath, B.S.: Reinforce Concrete Design of Tall Buildings. CRC Press, Boca Raton (2010) 14. Pistrol, J., Kopf, F., Adam, D., Villwock, S., Völkel, W.: Ambient vibration of oscillating and vibrating rollers. In: Adam, C., Heuer, R., Lenhardt, W., Schranz, C. (eds.) Vienna Congress on Recent Advances in Earthquake Engineering and Structural Dynamics 2013 (VEESD 2013), Vienna, Austria, 28–30 August 2013 (2013). Paper no. 167

Bridge Engineering

Stress State and Several Problems of Estimating the Actual Operation of Effective Bridge Superstructures Oleksandr Shymanovskyi1(B)

, Valerij Shalinskyi1 and Wiesław Baran2

, Maryna Shymanovska1

,

1 V. Shymanovskyi Ukrainian Institute of Steel Construction, Kyiv, Ukraine

[email protected]

2 Faculty of Civil Engineering and Architecture, Opole University of Technology,

Katowicka 48, 45-061 Opole, Poland [email protected]

Abstract. The paper presents the results of investigating the under-load behaviour and some estimates of the mutual influence of the stress state of structural elements during the actual operation of effective bridge superstructures with orthotropic deck plates. A brief description of the design solution of orthotropic deck plates is presented. It is stated that, from a structural point of view, significant saving of steel in the case of orthotropic deck plates is due to the fact that the longitudinal stiffeners and decking of such plate are included in joint work with the main girders and, thus, are included in the upper flange cross section. Therefore, the normal stresses in the plate near the webs are always greater than at any distance from them. Estimating this irregularity is very important for determining the actual operating conditions of the orthotropic deck as a whole and is carried out in each case by performing calculations of the superstructure. The results of calculating the real bridge deck of the carriageway at the Dnieper Hydroelectric Power Station dam in Zaporozhye are described. Relevant recommendations are provided. Keywords: Bridge structure · Superstructure · Orthotropic deck plate · Computational model · Reduction factor · Stress

1 Introduction Recently, efficient orthotropic steel plate systems (by “orthotropic” we mean deck plates which consist of longitudinal stiffeners and crossbeams welded to the decking plate, intersecting with each other; since the rigidity of such a plate is different in perpendicular directions, it is called orthotropic, or orthogonally anisotropic one) have become widespread in many bridge structures as one of the main load-bearing structural elements of the carriageway (Fig. 1). The almost widespread use of this design solution is due not only to its exceptional technical characteristics [24−26, 31, 33], but also to many other important advantages over commonly used solutions. First, orthotropic deck plates can increase the bearing capacity and reliability of the carriageway when resisting transport © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 99–117, 2023. https://doi.org/10.1007/978-3-031-26879-3_8

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(static and dynamic) loads. Secondly, it can increase rigidity of thin plated structural elements under compression, and finally thirdly, probably the most important thing, orthotropic deck plates can achieve significant saving of steel and reduce construction time, what, of course, leads to lower costs of bridge structures.

Fig. 1. Standard structural element of the orthotropic steel plate system. General view: 1 − main girder; 2 − crossbeam; 3 − longitudinal stiffener; 4 − steel plate.

Let’s pay attention to the fact that initially the cases of using orthotropic steel deck plates in bridge construction were sporadic, but further they are becoming more common. In confirmation of the above, we note that the design of the first bridge using orthotropic deck plates was developed by German engineers in the 1930s, and the first such bridge was built by them in 1936 [26]. German engineers are also ascribed creating the very term “orthotropic” and registering the corresponding patent in 1948 [30]. It is also obvious that similar work was simultaneously carried out in other countries. For example, in the United States in the State of California, across the San Francisco Bay between the cities of San Francisco and Oakland, the Bay Bridge was built in 1936, which is an example of the first suspension bridges using orthotropic steel deck plates. By the way, American engineers introduced the term “battle deck”, which differs from the European one, to name orthotropic deck plates, apparently by analogy with the name of an armed capital ship (in a more common abbreviated version – a battleship) that has long been used in shipbuilding – “battleship” [26]. Orthotropic deck plates were especially widely used in bridge construction in Germany after World War II due to the fact that almost all bridges in the country at that time were destroyed and the problems of saving steel became very urgent. In addition, the expanded introduction of orthotropic deck plates was additionally facilitated by two significant circumstances: the improvement of welding technology and the development of the calculation theory [4, 26, 28, 30].

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Subsequently, in many countries around the world, a large number (hundreds and thousands) of various bridge structures were built using orthotropic steel deck plates. Therefore, it makes no sense to dwell on the description of all these bridges in this paper due to their large number. However, we would like to mention only one of a lot of these wonderful structures of the world’s transport infrastructure using orthotropic deck plates, which is rightfully among the most famous bridge structures. It is about the cable-stayed bridge (more precisely, the viaduct) of Millau, passing through the valley of the Tarn River near the city of Millau, located in the south of France in the Aveyron department (Fig. 2).

a

b

c

d

Fig. 2. Panoramic views of the Millau viaduct from the eastern (a), north-eastern (b), southwestern (c) and south-eastern (d) sides.

The mentioned bridge attracts attention by the fact that at commissioning in 2004 it held three world records in bridge construction (for the height of reinforced concrete supports, for the height of reinforced concrete supports with a crowning metal pylon and for the carriageway height). However, in order to save space, we will not give more detailed information about this remarkable engineering structure, since it can be easily obtained using its official website capabilities Le Viaduc de Millau [27]. As follows from the above, the history of using orthotropic steel deck plates in bridge construction goes back almost a century. During this time, these structural elements of the bridge carriageway have found the widest application. Moreover, the latter was facilitated not only by their naturally inherent numerous positive properties (efficiency, reliability, increased bearing capacity, steel economy and much more), but also

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by constant improvement of both the theory and research methods, as well as engineering solutions. If we talk about improving the theory and methods of studying orthotropic deck plates, it should be noted that at first, analytical methods for their calculation were actively developed. In 1947, German engineer W. Cornelius [14] proposed to apply the wellknown theory of M.T. Huber for calculation of plates [20−22] with some clarification. Moreover, the refinement was based on the fact that the thickness of the plate was determined as a result of taking into account the metal plate and additional “smearing” on it of the longitudinal stiffeners and crossbeams, proceeding from equivalent cylindrical rigidity of the orthotropic deck plate and a smooth thin plate. The theory of calculating orthotropic deck plates was further developed in 1957, when engineers V. Pelikan and M. Esslinger [29] proposed to consider longitudinal stiffeners as continuous beams on elastically subsiding supports, which made it possible to somewhat simplify the calculated ratios due to neglected tensional rigidity of open stiffeners. In the early 1960s, G. Gomberg and K. Trenks [19] recommended using the girder grillage method for calculating orthotropic deck plates, based on the orthogonalization of unknown quantities with presentation of external load and internal forces in the form of group factors, the change of which is described by trigonometric functions. As for the girder grillage design model, it was adopted in the form of a freely supported contour grillage made as a system of crossbeams resting on infinitely large number of elastic yielding rotating supports, or, in other words, a decking plate with longitudinal stiffeners. In the 1990s, the era for applying computer technology and numerical calculation methods began, which were implemented in high-performance software and computing systems, as a rule, on the basis of the finite element method. The use of these complexes removed almost all the difficulties inherent for analytical methods in terms of setting, detailing and dimensioning the research problem, the introduction of assumptions and simplifications in creating design models of structures, adequate presentation of existing static and dynamic loads, and much more. Moreover, it became possible to obtain reliable results not only by creating design models that fully reflect the design features of real structures, but also by performing calculations (geometrically and physically) both in a linear and nonlinear setting. The latter made it possible to assess the stress–strain state (SSS) of orthotropic steel deck plates of complex configuration with almost maximum detailing, taking into account the features of the design solution, various formulations of the integral mode for material state and random external loads and influences. And as a confirmation of what has been said, let us mention only a few of the most indicative publications from a very large number of available examples. In particular, the papers [15, 16] describe the design model of an orthotropic deck with detailed modeling of cuts in the webs of crossbeams for passing longitudinal stiffeners, the results of determining the stresses in the beam webs at the cuts and changes in their values depending on the fatigue strength are given, which allowed to reveal the reasons for growth of fatigue cracks. Moreover, a visual method for calculating stresses in the webs at the cuts is proposed, which not only allows us to understand the origin of all stress components, but also simplifies the comparison of results for cases of cuts of various geometry. By the way, the authors of the paper [32] draw attention to the rather thorough development of

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structural models of all elements of orthotropic deck plates in the analysis of secondary bending stresses in webs of enclosed trapezoidal stiffeners. The principles and features of the design model are described, and the calculation results are illustrated by a numerical example. It is noted that secondary stresses are caused by the transverse expansion of the lower parts of the stiffeners, and in combination with primary bending stresses, they can contribute to the growth of fatigue cracks. Recommendations are given for decreasing the values of these stresses. The authors of a very interesting paper [12] proposed an idealized design model of an orthotropic deck plate, the equivalent physical and mechanical properties of which in perpendicular directions are established by taking into account the stiffeners by analogy with reinforcing fiber beams in composite materials. The study results are presented concerning the joint behavior of stiffeners, beams and the decking plate under transverse load, based on which the authors formulated a fairly simple analytical and engineering approach based on Ritz method for determining the deformations of an orthotropic deck plate. In the article of the journal [13], in order to improve the design of bridge superstructures, several options for calculating the effective width of beam flanges be f f in reinforced concrete single-span and multi-span bridges were considered, depending on three main parameters: the distance between the beams, the span length and the inclination angle, as well as more universal criteria for determining the effective width of the flanges in comparison with those used in engineering practice. The studies were carried out using the finite element method, the results of which were not only compared, but also confirmed by experimental data on test samples in scale 1/4 and 1/2, and the criteria were extended to the areas of positive and negative bending moments. In terms of improving the engineering and technical solutions for orthotropic steel deck plates, we only point out that this improvement, in fact, has always been the dominant component of their use. This is evidenced by the fact that even the first experience of using orthotropic deck plates revealed two disadvantages in them: low fatigue resistance (cracking at the intersection of longitudinal stiffeners and crossbeams) and destruction of the asphalt concrete pavement on the deck plate. Further, a lot of studies were devoted to this issue, and at present a number of recommendations have been developed that can significantly improve the situation. In a similar way, the parameters of longitudinal stiffeners were improved, such as: construction (open, closed), geometry (flat, corner, tee, strip etc.), configuration (trapezoidal, V-shaped, circular), with or without diaphragms, and much more. The results of all the above long-term and numerous studies have been implemented in the norms and guidelines in many countries specifying the design of orthotropic steel deck plates for bridge superstructures. Among them are the most wellknown norms and manuals of USA [8–10, 18], Europe [17], United Kingdom [11], Japan [23], and Ukraine [1–3]. It should also be emphasized that these regulatory documents, in addition to the norms themselves, guidelines and manuals for design of orthotropic deck plates, contain extensive bibliographies and large-scale analyzes of the problem state, what, in the opinion of the authors of this paper, is very useful for acquainting the reader with the topics under consideration herein.

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Thus, summarizing the above analysis of literature sources, we can conclude that over a relatively short historical period of using orthotropic steel deck plates in bridge construction, the theory and methods of their research, as well as improvement of structural and technical solutions, have got really impressive development. However, both in the works discussed above and in other more numerous articles and monographs, due attention was not paid to the study of the operation of orthotropic steel deck plates under conditions of long-term (decades of years, and especially many decades) use under conditions of development and accumulation of various kinds of operational damage. Let’s note that with regard to bridge structures, the nature and consequences of these damages are as follows. First, most often they are caused by temperature actions, seismic and aggressive influences, collisions of vehicles and accidental impacts of extraneous objects, and secondly, this leads to distortion of structural elements, occurrence of fatigue cracks, destruction of protective paint coatings and subsequent corrosion of metal and decreasing the design thickness. Meanwhile, it seems quite obvious and absolutely indisputable that only a correct understanding of the actual under-load operation of orthotropic steel deck plates of bridge superstructures, when there are operational damages, provides the ability to maintain the required level of their strength and reliability as a result of timely implementation of the necessary repair measures. If, for one reason or another, the repair work cannot be completed on time, then only knowledge about the actual technical condition of orthotropic deck plates can help the matter, on the basis of which schemes for organizing temporary traffic on the bridge superstructure are to be developed in order to ensure its limited operation (as a result reducing the load on the bridge superstructure) until the completion of the repair work. However, it should be noted once again that this traffic schema is temporary one and is used exclusively to facilitate the operation of the bridge superstructure. And it is for this reason that it contains a set of interrelated proposals aimed both at direct reduction of the load on the bridge structure (especially from oversized, freight and public transport), and at redistribution of this load over the carriageway area (taking into account the places of damage) due to changes of the number and width of traffic lanes, as well as the traffic intensity and speed of vehicles. Finally, we note that the study of the actual technical condition of bridge superstructures with the roadway orthotropic steel plates, the operation of which lasts a long time, is an important and at the same time complex engineering problem. The solution to this problem is of great importance and contributes to timely extension of operation of bridge superstructures and ensuring the traffic on them without restrictions. As an example of solving the indicated problem, let us consider the existing driveway at the Dnieper Hydroelectric Power Station dam, which was built in 1977 with use of the driveway orthotropic steel deck plates and which is one component of the complex of bridge crossing structures over the hydraulic structures of the said HPS.

2 General Characteristics of the Bridge Superstructure of the Driveway at the Dnieper Hydroelectric Power Station Dam It is known that the Dnieper Hydroelectric Power Station dam in the city of Zaporozhye was commissioned in 1932. In addition to actual generation of electricity, its distinctive

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feature is that it simultaneously served and now serves as an important traffic link between the right bank and left bank of the Dnieper river due to the arrangement of a two-lane highway crossing on its entire length (Fig. 3). The specified highway crossing contains several structures (from left to right): the bridge across the fore chamber of 319.76 m length (section I–II); the driveway on the dam structures of 666.00 m length, which includes superstructures executed in 1932 and 1977 (section II–III, Fig. 4); 111.50 m long overpass connecting the dam with the left bank (section III–IV), 136.57 m long soil insert-section (section IV–V) and 352.00 m long overpass over gateways (section V–VI). Thus, the total length of the highway crossing is 1,585.82 m. The driveway located on piers of the dam with a curvature radius of about 600 m before the reconstruction of 1977 had a width of 7.5 m. Its load-bearing structures were split spans that overlapped the span girders between the dam piers and consisted of four steel riveted solid beams and reinforced concrete carriageway slabs placed on them. And now, running a little ahead, we will point out that these spans were used during reconstruction of the driveway, as a result of which the width of the latter increased to 15.9 m (Fig. 5). At the same time, a 2.25 m wide sidewalk was installed on the downstream side of the dam and a 0.75 m wide inspection gangway on the upstream side. The reason for the reconstruction was quite simple and obvious: it was caused by a significant increase in traffic intensity in 1977. That is why the reconstruction of the driveway was carried out, one of the main objectives of which was the expansion of the carriageway and the arrangement of four traffic lanes in both directions instead of the existed two lanes. At the same time, two traffic lanes remained along the spans of 1932, and the other two lanes were organized along the spans of 1977 (Fig. 6).

Fig. 3. Highway crossing on the superstructures at the Dnieper HPS dam. General view.

During the reconstruction, steel cantilever girders were brought under the existing spans (as shown in Fig. 6). Then they were installed on the dam piers and attached to the dam by means of anchoring tie rods made from wire ropes Ø 68 mm passed in the openings drilled in advance, which were injected with cement mortar after pretensioning the wire ropes. The box-section steel girders are taken without field joints.

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Fig. 4. Highway crossing on the superstructures at the Dnieper HPS dam. Layout: blue color – spans executed in 1932; yellow color – spans executed in 1977; gray color – Hydroelectric Power Station building-2; II – driveway beginning; III – driveway end.

Fig. 5. Cross section of the dam after reconstruction of the carriageway in 1977: 1 − frame girder; 2 – side brace.

And the steel spans of 1977 are executed as split ones consisting of two main girders, each of them 1,000 mm high, crossbeams – with a pitch of 2,000 mm, longitudinal stiffeners – with a pitch of 400–575 mm with a steel plate located on top of them. Let’s note that the crossbeams and longitudinal stiffeners, commonly with the decking, form a steel orthotropic deck plate of the carriageway, and the main girders have transverse stiffeners in the places where the orthotropic deck plate crossbeams rest on them. To reduce the number of transverse expansion joints, the top plates of the deck were welded during assembly with a pitch of about 100 m, thus forming a number of separate enlarged blocks of superstructures with a continuous plane of the carriageway. Taking into account the above, and also the fact that the length of the driveway on the dam superstructures makes 666.00 m, it becomes clear that in total, during the construction, six such span blocks were created. It should also be noted another significant design property of the driveway caused by the fact that the span of 1932 and the span of 1977 have a significant

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difference in rigidity. Since this difference could contribute to the negative action from one superstructure to another, in order to avoid such an effect, a continuous longitudinal expansion joint was arranged between these two constructs (Fig. 6), which additionally serves as a central reservation area for two traffic directions. As for the structural solution of the joints, both the longitudinal expansion joint and the transverse joints between the enlarged blocks are of the spring type.

Fig. 6. Cross section of frame girder structures after reconstruction of the driveway in 1977: 1 – frame girder; 2 – side brace; 3 – lifting bar; 4 – tie rod; 5 – tie bar; 6 – movable inspection platform; 7 – inspection gangway; 8 – monorail; ➀, ➁, ➂ – frame girder bearing axis.

In the 90s of the last century, it became necessary to repair steel frame girders, which was caused by three important factors. The first of them was a significant (several times) increase of vertical mobile loads from vehicles (especially large-capacity vehicles). The second one was associated with corrosion damage of the frame girders themselves, which led to partial loss of their cross-section. And the third, probably the most important one, consisted in progressive dynamics of grows of through cracks in concrete piers of the dam. Therefore, after development of several variants of repair work, it was decided to reinforce the frame girders by placing rigid side braces under their consoles (Fig. 6).

3 Design Model of Superstructure Taking into account the configuration, structure and topology of the superstructure, a finite element model was taken as a design model, built by modeling six enlarged blocks of the driveway superstructures at the Dnieper HPS dam. In the model, all the constituent parts of the orthotropic deck plate and main girders are approximated by a set of thin shell triangular and quadrangular finite elements. In places where the main girders are supported on steel cantilever beams (Fig. 6), bracing elements are applied

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on corresponding displacements of the joints. The material of the structural components of the orthotropic deck plate and the main girders is steel grade 09G2S-12 with elastic modulus E = 2.1·105 MPa, Poisson’s ratio ν = 0.3 and design resistance R = 340– 400 MPa, depending on structural element thickness. In terms of specifying and further taking into account in the finite element model the actual values of thicknesses of the finite elements of the steel plate, crossbeams and longitudinal stiffeners of the orthotropic deck plate, as well as the lower flange, web and transverse stiffeners of the main girders, we note the following. During the last special surveys of the driveway at the Dnieper HPS dam, a number of defects in bridge structures were identified (their exhaustive list and description is given in article [7], which not only negatively affect the durability of the latter, and in some cases even lead to a decrease in the total bearing capacity of the driveway. Therefore, to assess the influence of these factors on the stress–strain state of bridge structures, the existing defects were represented in the finite element model of the driveway superstructures, which was created by selecting all the initial data from the detailed drawings when carrying out priority strength and dynamic calculations of the driveway at the Dnieper HPS dam with design values geometric parameters. However, due to the significant number of identified defects, before their introduction into the finite element model of the driveway superstructures, their systematization and a certain typical averaging were performed. As a result, it was assumed that the defects can be reproduced in the model by adjusting the thickness of the elements of driveway spans, depending on the depth of their corrosion damage while maintaining other geometric and rigidity parameters. In Fig. 7, the constructed finite element model of the enlarged superstructure block of the driveway at the Dnieper HPS dam is shown, which has 9,938 joints and 9,920 finite elements. And in Fig. 8 the general finite element model of the driveway superstructure at the Dnieper HPS dam is shown, which has 59,047 joints and 59,588 finite elements.

4 Load on the Superstructure It should be noted that the permanent loads and imposed loads on the superstructure are accepted in full compliance with the requirements of the norms [3, 6]. The permanent load on the driveway superstructure is the dead weight of its structural elements and it was taken into account by two types of uniformly distributed design load – along the line (designated by symbol P with corresponding subscript) and over the area (designated by symbol q with corresponding subscript) – depending on the type of structural element. Taking into account the above, nine components of the permanent load were taken as a whole, namely: carriageway plates (q2 = 2.4 kPa), main girders (P1 = 2.6 kN/m), barrier railings (P2 = 0.5 kN/m) and pedestrian railings (P3 = 0.9 kN/m), monorails with a movable inspection platform (P4 = 1.0 kN/m), lighting poles (P5 = 0.9 kN/m), inspection gangway (P6 = 1.0 kN/m), carriageway deck surfacing (q2 = 2.4 kPa) and sidewalk pavement (q3 = 1.0 kPa). The load case for application of permanent load components to the cross-section of the superstructure is shown in Fig. 9.

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Fig. 7. Finite-element model of enlarged block of the driveway superstructure at the Dnieper HPS dam: a – model of plate decking; b – model of longitudinal stiffeners; c – model of crossbeams; d – model of main girders; e – general model of enlarged block.

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Fig. 8. General finite-element model of the driveway superstructure at the Dnieper HPS dam.

As for the imposed load on the driveway superstructure, it consists of three components, namely: wheel four-axle load from one motor vehicle weighing 0.8 MN (HK-80), wheeled vehicle load of a convoy of cars weighing 0.3 MN each (H-30) and uniformly distributed design load from human crowd on the sidewalk (q4 = 11.7 kN/m). The load drawing concerning application of imposed load components to the cross-section of the superstructure is given in Fig. 10, which also shows the position of the resultant (indicated by R letter with corresponding subscript) from motor vehicles. From the above information, it follows that the calculation took into account two groups of loads: permanent and imposed. Consequently, the total number of design loads was twelve, and combinations of the loads taken according to requirements of regulatory documents [3, 5, 6] was fourteen, depending on the application place of imposed wheeled vehicle loads on the span length in accordance with actual operating conditions of the driveway, and permanent loads were included in all combinations. Comparison of the calculation results obtained for all fourteen combinations of loads made it possible to determine the most unfavorable combinations for each characteristic design point of the driveway superstructure – on the supports and in span quarters. At the same time, the most interesting thing concerning these results was that all considered fourteen load combinations (once again we emphasize – all of them!) turned out to be the most unfavorable. However, thirteen of them are of a local nature caused by acting permanent and temporary loads (HK-80 wheeled vehicle load and human crowd load on the sidewalk), since they were formed by sequentially rearranging the HK-80 wheeled vehicle load along the driveway, which led to occurrence of vertical displacements and normal stresses exclusively within limited areas around this wheeled vehicle load – from one four-axle wheeled vehicle weighing 0.8 MN. And only one combination, containing both permanent load and imposed load (from H-30 wheeled vehicle and human crowd on the sidewalk), turned out to be the most unfavorable combination as a whole for the entire driveway, an attribute of which was occurrence of maximum values for vertical displacements and normal stresses along the entire plane of its superstructure.

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Fig. 9. Load case for application of permanent load components to the cross-section of the driveway superstructure: A, B – main girders.

Fig. 10. Load drawing for application of imposed load components to the cross-section of the driveway superstructure: A, B – main girder; I – wheel four-axle load from motor vehicle HK-80; II – wheeled vehicle load H-30; III – load from human crowd on sidewalk.

5 Results of Numerical Calculations Made for Superstructure Bearing Strength Taking into account the above, we’ll consider behavior of the driveway superstructure at the Dnieper HPS dam under the most unfavorable load combination, namely: permanent load and imposed load as a part of the H-30 wheeled vehicle load from the convoy of cars weighing 0.3 MN each and the uniformly distributed design load from human crowd on the sidewalk. The results of numerical calculations in form of isofields for vertical displacements and normal stresses in the decking of an orthotropic plate and tangent stresses in the webs of main girders are shown in Figs. 11, 12 and 13.

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As follows from the analysis of the results obtained, the decking of the orthotropic plate on the entire length of the driveway superstructure has two types of fundamentally different zones of vertical displacement (Fig. 11). The first zone is located above the places where the main girders are supported on the cantilever girders, it covers almost the entire width of the carriageway of the orthotropic plate with maximum positive displacement values of 3.51 mm. The second zone is located in the spans between the cantilever girders, where the decking deflects downward with a maximum value of 36.22 mm. As for distribution of normal stresses in the decking of the orthotropic plate (Fig. 12), we note that above the places where the main girders are supported on the cantilever girders, the decking resists tensile stresses in the range of σx = 43–48 MPa, and in span centers between the cantilever girders, the decking as a whole is more or less uniformly compressed within the range of σx = 45–80 MPa. As for behavior of the webs of main girders, a multidirectional stress state can be traced in them (Fig. 13): there are compression stresses that are insignificant in terms of values (up to τxy = 82.1 MPa) in some fragments of these webs, in others of them there are tensile stresses (up to τxy = 82.1 MPa), which is caused by peculiarities inherent for the support of main girders on cantilever girders.

Fig. 11. Isofields of vertical displacements w in decking of orthotropic plate for the most unfavorable combination of loads, mm.

Fig. 12. Isofields of normal stresses σx in decking of orthotropic plate for the most unfavorable combination of loads, MPa.

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Fig. 13. Isofields of tangent stresses τxy in webs of main girders for the most unfavorable combination of loads, MPa.

Comprehending now the presented results, it is possible to understand that the orthotropic plate under the most unfavorable combination of loads is in a somewhat understressed state. The above can be explained by the fact that the design and construction of the driveway superstructure was primarily aimed at accelerating the solution of urgent city transport problems in Zaporozhye, which was then carried out in a very short time, and the main attention was paid exclusively to executing several accompanying organizational and technological measures, and economic indicators were practically not taken into account.

Fig. 14. Graphs of changes in stress–strain state parameters along local longitudinal axis of the second enlarged block of driveway superstructure: a − vertical displacements; b − normal stresses.

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In addition to the above information, for one of typical enlarged blocks of the driveway superstructure at the Dnieper HPS dam (in this case, the second one), Figs. 14 and 15 show graphs of changes in stress–strain state parameters along the local longitudinal axis connecting the joints of the orthotropic plate decking with the largest positive displacement values and the local transverse axis in the middle of this block. Let us draw attention to the fact that the solid black curve on these graphs corresponds to the actual operation of the superstructure, taking into account defects and damage (including corrosion) of bridge structures revealed during the last special surveys of the driveway at the Dnieper HPS dam [7], and the solid blue curve presents its behavior in design position. If we carefully analyze the graphs presented in Figs. 14 and 15, we can reach the expected qualitative conclusion that the defects and damage of the driveway superstructure arisen during the operation period affect the parameters of the stress–strain state of bridge structures. As for the quantitative assessment of this effect, we point out that discrepancy between displacements and stresses of design state and actual state is in the range from 8 to 11% for almost all structural elements due to actually uniform longterm corrosion of metal with slight weakening of cross-sections (simultaneously from Fig. 15b, it also implies that in the decking above the main girders of the orthotropic plate, there are insignificant undulating fluctuations in the values of normal stresses caused by conditions of interaction between the decking and the girder).

Fig. 15. Graphs of changes in stress–strain state parameters along local longitudinal axis of the second enlarged block of driveway superstructure: a − vertical displacements; b − normal stresses; O − zone at longitudinal expansion joint.

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The only exception to this is a zone of about 0.6–0.7 m width, adjacent to the longitudinal expansion joint between the superstructure of 1932 and superstructure of 1977, which extends along the driveway entire length (Fig. 3). And the peculiarity of this zone is that, in addition to corrosion damage, there are also many through-burns in metal of the orthotropic deck plate elements, which happened in 2010 during disassembling and installation of longitudinal expansion joint, which is expounded in detail in the previously mentioned paper [7]. As a result in this zone, the difference between displacements and stresses of design state and actual state is increased by about 2–2.5 times and reaches 22–25%, which is clearly shown in Fig. 15.

6 Conclusions The paper discusses the methodological aspects of an integrated approach to the calculation and analysis of operation of bridge superstructures with orthotropic deck plates of the carriageway. The paper presents a technique for studying the uneven distribution of normal stresses in the orthotropic plate decking. As a matter for research, the superstructure of existing driveway at the Dnieper HPS dam was considered, the design model for which was created taking into account the actual technical state of structural elements based on the results of the performed engineering surveys. The results obtained using the numerical calculation method are presented and analyzed in form of vertical displacements and normal stresses in the orthotropic plate decking, as well as tangent stresses in webs of the main girders. The reduction factors have been determined by applying engineering-and-analytical (normative) method and numerical calculation method. Comparison of the obtained results is carried out and recommendations are given regarding application of calculation methods. The results and, as a consequence, recommendations formulated on their basis are as follows. 1. Calculations of bridge structures must be carried out taking into account their actual technical condition determined on the basis of the results of engineering surveys. The analysis of the data obtained shows that defects, damage and destruction in structural elements that have been in operation for many decades significantly affect under-load operation of these elements. 2. In order to obtain the numerous results most adequate to the actual technical condition of investigated bridge superstructure, one should proceed as follows. Namely: accounting for defects, damage and destruction in the finite element model of bridge structures due to their significant number (as a rule!) should be carried out not by presenting these “imperfections” directly, but indirectly – by adjusting thicknesses of finite elements depending on presence of corresponding “imperfection”, while maintaining other geometrical and rigidity parameters. Moreover, the magnitude of this correction is to be determined by averaging the existing “imperfections” for the whole surface of the structural element. 3. When conducting research of bridge structures that have defects, damage and destruction acquired during operation, and in particular those that, in addition to this, are characterized by non-standard design solutions, it is necessary to use numerical

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calculation methods for two reasons. The first one is that only numerical calculation methods make it possible to achieve the maximum possible structure detailing when creating a numerical model and, as a consequence, to obtain better, more reliable and accurate calculation results. The second reason is associated with the current regulatory documentation, where methods are absent that allow accounting in the standard engineering-and-analytical calculation method neither for engineering survey results for structural elements of bridge structures, nor for the applied non-standard structural solutions, and therefore the results determined on its basis contain a significant error. 4. Taking into account that the current regulatory documents do not contain provisions regarding the calculation and design of bridge structures accounting for data of engineering surveys and non-standard design solutions, it is recommended to supplement these regulatory documents with appropriate content sections.

References 1. DBH B.2.3-26:2010 Cpopydi tpancpopty. Mocti ta tpybi. Ctalevi konctpykci|. Ppavila ppoektyvann, 195 p. Minpegion Ukpa|ni, Ki|v (Transport construction. Bridges and culverts. Steel construction design code. Design rules) (2011) 2. DBH B.2.3-22:2009 Cpopydi tpancpopty. Mocti ta tpybi. Ocnovni vimogi ppoektyvann, 73 p. Minpegion Ukpa|ni, Ki|v (Transport construction. Bridges and culverts. Basic design requirements) (2009) 3. DBH B.1.2-15:2009 Cpopydi tpancpopty. Mocti ta tpybi. Havantaenn i vplivi, 66 p. Minpegion Ukpa|ni, Ki|v (Transport construction. Bridges and culverts. Loads and actions) (2009) 4. Kopneev, M.M.: Ctalnye mocty: Teopetiqeckoe i ppaktiqeckoe pocobie po ppoektipovani moctov v dvyx tomax, vol. 1, 532 p. Akademppec, Kiev (Steel Bridges: Theoretical and Practical Guide to Bridge Design in Two Volumes) (2010) 5. CHiP 2.05.03-84*. Mocty i tpyby, 239 p. Gocctpo Poccii, Mockva (Building regulations. Bridges and culverts) (2005) 6. CH 200-62. Texniqeckie yclovi ppoektipovani eleznodoponyx, avtodoponyx i gopodckix moctov i tpyb, 328 p. Gocctpo CCCP, Mockva (Technical Specifications for Design of Railway, Road and City Bridges and Culverts) (1962) 7. Ximanovcki, O.B., Xalincki, B.B.: Obcteenn konctpykci avtoppo|zdy po ppogonovix bydovax gpebli DnippoGEC, zvedenix y 1977 p. In: Ppomiclove bydivnictvo ta inenepni cpopydi, vol. 2, pp. 2–7 (Inspection of Driveway Structures at Dnieper HPS Dam Superstructures Installed in 1977) (2020) 8. AASHTO LRFD: Bridge Construction Specifications, 3rd edn. American Association of State Highway and Transportation Officials, Washington (2010) 9. AASHTO LRFD: Bridge Design Specifications. American Association of State Highway and Transportation Officials, Washington (2012) 10. AISC: Design Manual for Orthotropic Steel Plate Deck Bridges. American Institute of Steel Construction, New York (1963) 11. BS 5400-3: Steel, concrete and composite bridges. Part 3: code of practice for design of steel bridges. British Standards Institution, London (2006) 12. Chen, M., Xue, J., Li, P., Jin, F.: Orthotropic analysis of steel Deck-Girder-Rib systems subjected to transverse load. Int. J. Steel Struct. 19, 1010–1022 (2019). https://doi.org/10. 1007/s13296-018-0180-1

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13. Chen, S.S., Aref, A.J., Chiewanichakorn, M., Ahn, I.-S.: Proposed effective width criteria for composite bridge girders. J. Bridge Eng. 12(3), 325–338 (2007) 14. Cornelius, W.: Die Berechnung der ebenen Flaechentragwerke mit Hilfe der Theorie der orthogonal-anisotropen Platte. Der Stahlbau 21(2), 21–24; (3), 43–48; (4), 60–64 (1952) 15. De Corte, W., Delesie, C., Van Bogaert, P.: Examination of local stresses in relation to fatigue failure at the rib to floorbeam joint of orthotropic plated bridge decks. Bridge Struct. Assess. Des. Constr. 3(3), 183–191 (2007) 16. De Corte, W., Van Bogaert, P.: Improvements for the analysis of floorbeams with additional webs cutouts for orthotropic plated decks with closed continuous ribs. Steel Compos. Struct. 7(1), 1–18 (2007) 17. Eurocode 1993-2: Design of steel structures. Part 2: steel bridges. European Committee for Standardization, Brussels (2009) 18. FHWA: Manual for Design, Construction, and Maintenance of Orthotropic Steel Deck Bridges. US Department of Transportation Federal Highway Administration, Washington (2012) 19. Homberg, H., Trenks, K.: Drehsteife Kreuzwerke. Springer, Berlin (1962). https://doi.org/10. 1007/978-3-642-50271-2 20. Huber, M.T.: Zur Theorie der Berührung fester elastischer Körper. Ann. Phys. 319(6), 153–163 (1904) 21. Huber, M.T.: Die Theorie der kreuzweise bewehrten Eisenbetonplatten nebst Anwendungen auf mehrere bautechnisch wichtige Aufgaben über rechteckige Platten. Der Bauingenier 4(12), 354–360; 13, 392–395 (1923) 22. Huber, M.T.: Probleme der Statik technisch wichtiger orthotroper Platten. In: Timoshenko Anniversary Volume. Macmillan Company, New York (1938) 23. JRA Specifications for Highwaybridges. Part 2: Steel Bridges. Japan Road Association, Tokyo (2017) 24. Korniev, M.: Orthotropic deck bridges in Ukraine. In: Orthotropic Bridge Conference, Sacramento, pp. 113–131 (2004) 25. Korniev, M.: Bridge engineering in Ukraine. In: Handbook of Bridge Engineering, pp. 865– 905. CRC Press, London (2014) 26. Kurrer, K.E.: The History of the Theory of Structures: Searching for Equilibrium, 2nd edn. Wiley, Berlin (2018) 27. Le Viaduc de Millau: https://www.leviaducdemillau.com/fr. Assessed 10 Nov 2022 28. Mangus, A.R., Sun, S.: Orthotropic deck bridges. In: Bridge Engineering Handbook. CRC Press, Boca Raton (2000) 29. Pelikan, W., Esslinger, M.: Die Stahlfahrbahn-Berechnung und Konstruktion. In: MANForschungsheft, vol. 7, pp. 23–46 (1957) 30. Sadlacek, G.: Orthotropic plate deck bridges. In: Constructional Steel Design: An International Guide, pp. 227–247. Elsevier Applied Science Publishers, London and New York (1992) 31. Troitsky, M.S.: Orthotropic Bridges: Theory and Design, 2nd edn. Published by The James F. Lincoln Arc Welding Foundation, Cleveland (1987) 32. Wolchuk, R., Ostapenko, A.: Secondary stresses in closed orthotropic deck ribs at floor beams. J. Struct. Eng. 118(2), 582–595 (1992) 33. Wolchuk, R.: Steel orthotropic decks: development in 1990s. Transp. Res. Rec. 1688(1), 30–37 (1999)

Durability and Assessment of Early Post-tensioned Bridges Petra Bujˇnáková1(B)

, Jakub Kraˇlovanec1 , Martin Moravˇcík1 and František Bahleda2

,

1 Faculty of Civil Engineering, Department of Structures and Bridges, University of Žilina,

Univerzitna 8215/1, 010 26 Žilina, Slovakia [email protected] 2 Faculty of Civil Engineering, Laboratory of Civil Engineering, University of Žilina, Univerzitna 8215/1, 010 26 Žilina, Slovakia

Abstract. Post-tensioned technology has been widely applied to bridges in the early 1960s in former Czechoslovakia. Nowadays, the considerable effect of aggressive environment, environmental conditions, and inadequate maintenance lead to the poor condition of early post-tensioned bridges. Tendons and anchorages are failing by corrosion and with excessive traffic load, some bridges are close to partial or total collapse. Inspection of several early post-tensioned concrete bridges indicates a similar problem. Understanding the processes leading to the deterioration of materials and bridge components, information about the residual prestressing force is the key how to predict a potential risk and remaining service life of post-tensioned concrete bridges. The various methods for bridge assessment have been practiced worldwide, providing valuable information in this field. Generally, visual inspection, destructive or non-destructive testing, probing, and structural health monitoring are preferable. The paper summarizes an overview of selected assessment methods applied to some early-built post-tensioned concrete bridges reaching their service life limit. Data should help the administrators to decide about the future of problematic structures in terms of strengthening or inevitable replacement by the new one. Keywords: Assessment · Prestressed concrete · Testing · Residual prestressing · Diagnostic survey

1 Introduction Over the past decades, attention to durability has been more concerned than ever before. The bridge structures are placed in a wide range of aggressive environments (chloride attack, freezing action, humidity) for a long service period and are overloaded by traffic. In recent years, the number of collapses or partial collapses of post-tensioned bridges has rapidly increased, either in Slovakia or worldwide. In 2020, sudden unexpected collapses of following Slovak bridges happened, e.g. the bridge in Trstená, the bridge in Spišská Nová Ves, and the bridge in Kysak. The demand to inspect and assess aging bridges has arisen [1, 2]. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 118–126, 2023. https://doi.org/10.1007/978-3-031-26879-3_9

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In the case of the post-tensioning system, prestressing force is transferred into the structure through built-in anchorages. Its durability is significantly affected by the condition of the prestressing steel and anchorages. Various anchorage devices and methods patented by manufacturers and prestressing companies have been adapted for posttensioning around the world, especially in the first half of the 20th century. In former Czechoslovakia, the Horel wedge anchorage system based on the Freyssinet principle was the most popular system in the early post-tensioned bridges. The first post-tensioning system consists of metal corrugated ducts, prestressing wires and steel wedge anchorages. Predominantly, each tendon consists of smooth parallel patented wires P (120/165) with a diameter of 4.5 mm or 7 mm. A varying number of wires were used per one anchorage, usually from 9 to 24. Each anchorage consists of a steel bearing plate with a conical hole and a steel wedge cone. The prestressing force was applied with the help of a small “A12” (later A24) jack reacting against an already-cast member at one end of the concrete structure. All wires are stretched at the same time. After post-tensioning action, ducts were filled with grouting. The bond was provided using cement mortar grout. Grout was pumped into the duct at one end of a structure under pressure via small gaps between wires and cone (later through an injection hole in a steel wedge). The system was very laborious. Different practice design and stressing technology have also been adopted for post-tensioning systems in Europe. Later, the wires were replaced by the multi-strand system. Generally, the post-tensioning system facilitates the design of long-span structures. Moreover, the application of curved tendons can deliver suitable redistribution of internal forces as the eccentricity of the prestressing force varies [3, 4].

2 Post-tensioned Bridges in Slovakia Currently, the number of Slovak bridges on the road network is over 8 200. Reinforced concrete and prestressed concrete dominate as a construction material in road bridges (about 90%). A significant group is made of ordinary and reinforced concrete approximately 65% and 25% of bridges are constructed of prestressed concrete. Early posttensioned bridges are approaching about 60 to 70 years of service life. Predominantly, the fully precast technology for the bearing structure was used. The superstructure consists of a series of precast girders placed side by side. After erection on the place, girders were transformed into a deck by transversal post-tensioning [5]. The post-tensioning system was made of bonded tendons. Each tendon was composed of several smooth patented wires and anchored using the original steel wedge system. The problem of early post-tensioned bridges is often linked to design philosophy and inadequate consideration for long-term maintenance. Generally, the early design specifications overlooked design details. A simply supported deck was mainly designed instead of a more durable continuous structure. The deck expansion joints have allowed water leakage through the structure to piers and abutments, Fig. 1a. Visions for future strengthening were also lacking. Some damaged bridge structures have to be demolished and replaced by new ones [6].

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The inspection of selected bridge structures built in the 1960s indicates similar problems. It has been noticed that the corrosion of the anchorage zone contributed to the prestress loss. Anchorages have been exposed to aggressive agents from years of leaking expansion joints, Fig. 1b. Approximately 50% of anchorages were not protected against corrosion in some cases [7]. Part of the prestressing wires was broken due to the weakened wire profile at the intersection of a conical hole in the bearing plate and the duct. Ducts were made from metal strips. Some ducts were totally dysfunctional with no grouts. The most damaged tendons were anchored in the top flange and were in poor condition, essentially due to the absence of a waterproofing layer and the position of transverse joints. Concrete spalling and other deficiencies were observed by visual inspection, Fig. 2. Low thickness concrete cover allows more rapid ingress of moisture and carbon dioxide. The widespread corrosion of reinforcement and tendons was observed. On the contrary, concrete was in a relatively good condition. Concrete hardness testing, ultrasonic velocity measurements, and concrete strength testing were used. The concrete strength follows the concrete class from C30/37 to C40/50. Patented wires with a diameter of 4.5 mm were used as a basic material for post-tensioning. Based on the tensile test, the mean measured tensile strength of the prestressing wires was 1647 MPa. Tensile test of corroded wires showed that the strain decreases with corrosion. Wires are more brittle and are sensitive to mechanical surface damage.

Fig. 1. Views on: a) the end region and dysfunctional expansion joint, b) severe corrosion of anchorage with prestressing wires.

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Fig. 2. Concrete spalling on the bottom part of the bridge structure. Corroded wires.

The consequences of a poor condition of prestressing, inadequate conceptual details and excessive traffic were the major reason for the brittle failure of the precast girder bridge in the Orava region. Critical vertical crack and broken girders have proofed this statement, Fig. 3a. The absence of conventional reinforcement also negatively influenced the ductility of post-tensioned girders.

Fig. 3. Major failures on investigated bridges: a) wide vertical crack on the bridge structure, b) joint opening of precast segments.

Some early post-tensioned bridges were designed from several precast segments joined together, especially for a long span. For example, the bridge in Trstená was built in 1957. The five precast segments were joined together, creating the single-span bridge. During a visual inspection in 2019, a joint opening between precast components was observed [8]. The critical joint was opened approximately 50 mm, Fig. 3b. The bridge was immediately closed. The low level of prestressing with the abovementioned problem caused the bridge spontaneously collapse in 2020, Fig. 4.

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Fig. 4. Collapse of the road bridge in Trstená (2020).

3 Monitoring of Prestressing Force Civil engineers have developed and tested various methods for determining the actual state of prestressing level, which is essential for the reliable evaluation of existing prestressed concrete structures. In the case of the diagnostic survey of the bridge in practice, the choice of evaluation method should consider the type of prestressed bridge structure (a girder bridge, a box girder bridge, slab structure, etc.), access to structure and traffic on the bridge. The essence of the diagnostic survey is to ensure applicability, availability, and affordability of accurate determination of the load-carrying capacity of investigated structure. Moreover, the diagnostic team should also consider the influence of the technique on the structure and the difficulty of repair resulting from its application [9, 10]. The most common classification method for prestressing force monitoring is according to their influence on the structure. This classification distinguishes the techniques as non-destructive, destructive, and sometimes semi-destructive. Nondestructive methods have only a negligible effect on the investigated structure. Usually, the intervention is only local in the area adjacent to testing and can be properly and easily repaired. Such methods do not have an impact on the integrity of the investigated structure. On the other hand, destructive methods cause significant damage that can be difficult to repair. However, destructive testing typically provides more reliable results. An overview of some residual prestressing force monitoring methods is listed in Table 1 [11]. Some of the above-mentioned methods were successfully applied on existing bridges in practice [6, 12]. Such applications have provided crucial information for determining the load-carrying capacity of investigated structures and enabled the evaluation of the length of the remaining service life. Application of the Structural Response Method and Saw-cut Method has become a standard element of the diagnostic survey used Department of Structures at Bridges. The Structural Response Method is based on the observation of a structure’s behavior as a result of a known external load applied to the structure. The prestress force is derived

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Table 1. Several methods for residual prestressing force monitoring [11]. Method

Influence on the structure

Saw-cut Method

The local intervention – possible proper repair

Structural Response Method

The initiation or reopening of crack which closes after the test

Barkhausen Noise Technique

Local intervention – the need for strand (wire) exposure

Wire (Strand) – Cutting Method

Local intervention – the need for bonded strand (wire) exposure

Drilling Method (Stress-Relief Coring Technique)

Local intervention – possible proper repair

Exposed Strand Method

Local intervention – the need for strand (wire) exposure

Magnetoelastic Method

Local intervention – the need for strand (wire) exposure

from the change of the structural response (strain, deflection, crack width, etc.) to the loading effect [6]. Saw-cut Method is usually considered a non-destructive indirect method based on the so-called normal stress release in the area adjacent to two or more saw-cuts. These saw-cuts isolate “concrete block” from the acting forces and thus, the recorded normal stress change represents the effect of all acting loads at the time of testing. Therefore, the absence of traffic is suitable because it simplifies the calculation of residual prestressing force. The support of reliable numerical analysis is significant in the case of all listed techniques [13]. 3.1 Examples of Bridge Assessment The Saw-cut method (Normal stress release method) was applied as part of diagnostic surveys of bridges in Banska Bystrica (Central Slovakia) in 2022 and Nizna (Northern Slovakia) in 2016. This method was used for the indirect determination of residual prestressing force value. The bridge in Nizna (Fig. 5) was built in 1956, so it was sixty years old at the time of testing. The structure consisted of four spans simply supported beams with an effective length of 17.18 m. The total length of ten T-shape post-tensioned beams was 17.88 m and they were 1.05 m deep. The spacing of beams was 0.97 m, so the bridge width was 10.05 m. Transversal stiffness was provided by five diaphragms – two end diaphragms and three intermediate diaphragms. Prestressing consisted of longitudinal and transversal tendons. The transversal tendons were located in the top flange with a spacing of 0.70 m and the diaphragms (transversal beams) with overall elven tendons. Longitudinal prestressing was represented by thirteen tendons consisting of nine patented wires with a diameter of 4.5 mm [12].

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Fig. 5. View on the bridge in Nizna while it was still in service.

In the case of the bridge in Nizna, two pairs of saw-cuts with an axial distance of 0.120 m were applied on the bottom edge of the edge post-tensioned beam. The pair of saw-cuts SC1 was 23 mm deep, while the depth of saw-cuts SC2 was 31 mm, Fig. 6. All these parameters were determined by the in-situ measurement of the saw-cuts. For the recording of normal stress change, the linear foil strain gauges were installed at the halfway point of the axial distance between the saw-cuts, whereas transversely, they were in the middle of the beam’s flange. Recorded normal stress relief showed that the maximum value of the normal stress change measured between saw-cuts SC1 was approximately 3.1 MPa and between saw-cuts SC2 was 4.2 MPa. Such analysis suggested that the residual prestressing force after the assumed length of service and neglected maintenance has caused a decrease of nearly 14% compared to the expected value after sixty years [12].

Fig. 6. In-situ measurement – view on saw-cuts.

The Saw-cut method was also used for the monolithic post-tensioned bridge in Banska Bystrica, which was analyzed in 2021 after exposure to a significant corroded anchorage area, Fig. 7. The bridge in question was built in 1962. The structure consisted of three post-tensioned beams with a depth of 0.750 to 1.800 m and a width of 0.400 m. The

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beams were connected by a reinforced concrete slab with a height of 0.160 to 0.260 m and stiffened by two diaphragms. The bridge was supported by struts and post-tensioned ties at the end of the beams so the structure can be considered as a frame. The effective length of the bridge was approximately 31.0 m.

Fig. 7. Longitudinal section of the bridge in Banska Bystrica.

The in-situ analysis of the bridge in Banska Bystrica consisted of the application of one pair of saw-cuts with an axial distance of 0.120 m and a depth of 0.030 m. Sawing was provided at a distance of 2.50 m from the strut of the bridge. For the measurement, two linear foil strain gauges were used, Fig. 8. Similarly, as in the previous bridge, they were installed in the middle of the axial distance between the saw-cuts. However, the reading of one of the strain gauges could not be used because it was damaged. Finally, the value of the normal stress relief was 4.1 MPa. Consequently, the diagnostic survey and analysis proved a good state of prestressing tendons and good agreement with the expected level of residual prestressing.

Fig. 8. Saw-cut testing in Banska Bystrica.

4 Conclusions Durability is influenced by design deficiencies, type of materials, quality of construction, and an aggressive environment. Corrosion of patented wires, anchorages, ducts and

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the presence of broken wires cause a decreasing number of active tendons in control structures. In the case of fully prestressed structures built and designed in the last century, the failure of prestressing wires can be hazardous and trigger bridge collapse. That’s why the actual level of the prestressing force is a key element for bridge safety. The Saw-cut method was successfully applied in situ to bridges in practice to assess the residual value of prestressing force. The effort of applying indirect methods depends on detailed visual inspection, NDT methods, consideration of relevant mechanical properties, FE models and a high level of personal skills. Acknowledgements. This work was supported by the Slovak Grant Agency under contracts No. 1/0048/22 and No. 1/0306/21, Cultural and Educational Grant Agency (KEGA) under contract No. 020ŽU-4/2021 and by the project of Operational Program Interreg V-A Slovak Republic – Czech Republic No. 304011Y277.

References 1. Koteš, P., Brodˇnan, M., Bahleda, F.: Diagnostics of corrosion on a real bridge structure. Adv. Mater. Sci. Eng. 2016, 10 (2016) 2. Odrobiˇnák, J., Gocál, J.: Experimental measurement of structural steel corrosion. In: 22nd European Conference on Fracture, ECF22-Loading and Environmental Effects on Structural Integrity, Proceedings, Belgrade, Serbia, 26–31 August 2018, vol. 13, pp. 1947–1954 (2018) 3. Hurst, M.K.: Prestressed Concrete Design, 2nd edn. E & FN SPON, An Imprint of Routledge, London (1998) 4. Abdel-Jaber, H., Glisic, B.: Monitoring of prestressing forces in prestressed concrete structures – an overview. Struct. Control Health Monit. 26, e2374 (2019). https://doi.org/10.1002/ stc.2374 5. Moravˇcík, M., Bujˇnáková, P., Bahleda, F.: Conceptual problems of first generation of precast bridges. In: Proceedings of the International fib Symposium on Conceptual Design of Structures, Madrid, Spain, 26–28 September 2019, pp. 301–308 (2019). ISBN: 978-2-940643-02-8 6. Moravˇcík, M., Bujˇnáková, P., Bahleda, F.: Failure and damage of a first-generation precast prestressed bridge in Slovakia. Struct. Concr. 21(6), 2353–2362 (2020) 7. Bujˇnáková, P.: Anchorage system in old post-tensioned precast bridges. Civ. Environ. Eng. 16(2), 379–387 (2020) 8. Bujˇnáková, P., Moravˇcík, M., Kraˇlovanec, J.: Condition assessment of a first generation precast prestressed bridges in Slovakia. In: Pellegrino, C., Faleschini, F., Zanini, M.A., Matos, J.C., Casas, J.R., Strauss, A. (eds.) EUROSTRUCT 2021. LNCE, vol. 200, pp. 108–115. Springer, Cham (2022). https://doi.org/10.1007/978-3-030-91877-4_13 9. Farbák, M., Jošt, J., Hlinka, R., Rosmanit, M.: Numerical analysis of the load-displacement behaviour of cast-in-place progressive anchorage in reinforced concrete members. Appl. Sci. 11, 2343 (2021) 10. Bagge, N., Nilimaa, J., Elfgren, L.: In-situ methods to determine residual prestress forces in concrete bridges. Eng. Struct. 135, 41–52 (2017) 11. Kraˇlovanec, J., Prokop, J.: Indirect methods for determining the state of prestressing. Transp. Res. Procedia 55, 1236–1243 (2021) 12. Kraˇlovanec, J., Moravˇcík, M., Bujˇnáková, P., Jošt, J.: Indirect determination of residual prestressing force in post-tensioned concrete beam. Materials 14, 1338 (2021) 13. Kralovanec, J., Bahleda, F., Prokop, J., Moravcik, M., Neslusan, M.: Verification of actual prestressing in existing pre-tensioned members. Appl. Sci. 11(13), 5971 (2021)

Impact of Duration and Intensity of Seismic Records on the Soil-Steel Bridge Behavior Tomasz Maleska1(B)

, Nashwan Al Zubairi2 , Adriana Janda1 and Kareem Embaby3

, Damian B˛eben1

,

1 Faculty of Civil Engineering and Architecture, Opole University of Technology, Katowicka

48, 45-061 Opole, Poland {t.maleska,d.beben}@po.edu.pl 2 Faculty of Engineering and Natural Science, Konya Technical University, Ardıçlı, Mahallesi, 42250 Selçuklu, Turkey 3 AECOM, New York City, NY 10004, USA [email protected]

Abstract. The design bridge standards only determine the maximum acceleration of the seismic excitation during the analysis of their influence on the soil-steel structure. It is well known that in seismic analysis the duration and intensity are very important, not just the ground acceleration caused by the seismic excitation. In addition, these aspects have a significant impact on the load capacity of the structure, so their response under seismic excitations. Therefore, this problem was analyzed using a numerical program based on the finite element method. The real soil-steel bridge located in Trzebaw (Poland) was analyzed. The bridge span is over 17 m. In this study, three seismic records (Taft, El Centro and San Fernando) were investigated. Moreover, during the numerical analysis, the aspect of intensity and duration of seismic records was taken into account. Displacements, velocities, and accelerations in the soil-steel bridge are analyzed in detail in the non-linear response range. Research with the use of the computational numerical model can find application in engineering practice, especially in the design of this type of bridges located in seismic areas. Keywords: Soil-steel bridge · Culverts · FEM · Corrugated steel plate · Seismic excitation · Intensity of seismic record · Duration of seismic record

1 Introduction Due to their simple structure and low construction costs, soil-steel bridges have gained considerable popularity in various parts of the world, especially in Western and Central Europe, Scandinavian countries, and North America [1–3]. Bridges made of corrugated steel plates (CSP) most often have a span of several meters. The largest soil-steel bridge was built in the United Arab Emirates with a span of 32.67 m and was entered in the Guinness Book of Records [4, 5].

© The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 127–138, 2023. https://doi.org/10.1007/978-3-031-26879-3_10

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Despite the increasing popularity of these structures, the number of standards and guidelines is relatively modest compared to traditional bridge structures. In many countries, where such bridges are used, there are no standards for these structures. However, you can find guidelines for the design of these objects [6–8]. So far, these objects have been analyzed mainly in the area of static loads where the answer to the question was sought: what will be the deformation of the steel shell caused by the arrangement and compaction of the backfill (construction stage [9–11]), and what will be the impact of loads from vehicles [12, 13]. In the case of dynamic loads, several studies focused on determining the dynamic identification of these structures and their response to moving loads (vehicle journeys [14–18]). Particularly, the modest knowledge is in the field of seismic loads where only the CHBDC standard [19] allows determining the response of a soil-steel object to seismic forcing. So far, work has been carried out only in the area of numerical analyzes or at the post-earthquake impact assessment stage [20, 21]. The numerical analysis focused on the assessment of the impact of the height of the soil over the soil-steel bridges [22] and the assessment of the impact of longitudinal ribs stiffening the steel shell of the tunnel [23]. It was found that the resulting discrepancy between the CHBDC standard and the results of numerical calculations results from a simple standardized approach to the assessment of the impact of seismic tremors on soil-steel bridges. This standard does not take into account the duration of the recording and the intensity of the shock, which, in the opinion of the authors of the cited works, is the cause of discrepancies between the standard and the numerical results. In addition, the effect of the EPS layer on the damping of seismic quakes was also analyzed [24]. Therefore, the aim of this study is to analyze the response of a soil-steel bridge subjected to seismic tremors of varying strength and recording duration. The MIDAS GTS NX numerical program was used to assess this impact using three seismic records (Taft, El Centro and San Fernando). The Time History method was used in the numerical analysis (analysis of the structure behavior in successive time steps). The obtained results will allow evaluating the behavior of the tested soil and steel objects in seismic areas, and therefore these tests can be used to better dimension the steel coating subjected to strong seismic excitations. As a result, these tests have a real impact on the environment by optimizing the steel shell of the soil-steel bridge and thus translating into lower consumption of structural steel.

2 Applied Method of Analysis 2.1 Description of Real Bridge The soil-steel bridge analyzed was located in Poland of Trzebaw over National Road no. 5 in Wielkopolski National Park (Fig. 1). After the construction, this bridge was one of the largest objects of this type in Europe. The load-carrying structure was made of CSPs shell with the maximal span of 17.67 m and a vertical height of 6.05 m. In addition, the total length of the top of shell structure was 40.39 m, while in the lower part of the structure was of 53.83 m. The soil cover depth above the shell crown was of 1.80 m. Furthermore, the shell structure was made from CSP with depth of 0.14 m, a pitch of

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0.38 m and a plate thickness of 0.007 m. Detailed description of the analyzed bridge can be found in [9]. In this research, the numerical models were created without additional steel ribs.

Fig. 1. View of the real bridge from Trzebaw (Poland).

2.2 Description of Numerical Model In the numerical model (Fig. 2) the shell structure was modelled according simplifications [1, 25] with steel of S235. The backfill was modelled using the Duncan-Chang nonlinear elastic hyperbolic model with following properties: (i) elastic modulus of 100000 kN/m2 , (ii) Poisson’s ratio of 0.2, (iii) unit of weight of 20.5 kN/m2 , (iv) angle of internal friction 39°, (v) cohesion 3 kPa, (vi) failure ratio 0.8, (vii) initial loading modulus of 47, (viii) minimum tangential modulus 100 kN/m2 , (ix) minimum confining stress of 10 kN/m2 , (x) atmosphere pressure 101313 N/m2 . The boundary conditions were modelled as pinned support on the base and side wall of the numerical model. These type of boundary conditions allow to correct seismic wave propagation in numerical model.

Fig. 2. View of numerical model with mesh.

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Fig. 3. Maximal soil acceleration of seismic records: a) Taft, b) El Centro, and c) San Fernando of x-direction (perpendicular to cross-section of bridge).

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2.3 Description of Interface The connection between the backfill and shell steel was modelled as an automatic interface using a “Coulomb friction” function with: (i) angle of internal friction 25.3°, (ii) normal stiffness modulus of 2979166.67 kN/m3 , (iii) cohesion of 19.5 kN/m2 , (iv) shear stiffness modulus of 270833.33 kN/m3 and (v) dilatancy angle 0°. 2.4 Seismic Excitations Three seismic records were used in the study, i.e.: Taft, El Centro, and San Fernando (Fig. 3). These records differ in the length of the entire record and the duration of the intensive zone. The El Centro earthquake record is of particular importance as it is a reference record that is used to evaluate the seismic response of various types of engineering structures [23, 24]. This recording lasts 53.72 s and is characterized by a maximum acceleration of 3.49 m/s2 (this record comes from a different database than in the work [23]). The Taft record is characterized by ground acceleration equal to 1.38 m/s2 and a total duration of 54.38 s. On the other hand, the San Fernando record is the longest of the tested records, lasting 61.84 s, and its maximum acceleration is 3.09 m/s2 . Table 1. Maximum values from numerical analysis. Direction

Seismic records Taft

El Centro

San Fernando

Displacements [×10−3 m] Dz

7.35

13.02

4.83

Dy

0.24

0.34

0.32

Dz

5.17

8.76

4.38

Vz

0.58

0.99

0.47

Vy

0.03

0.04

0.04

Vz

0.48

0.84

0.50

Az

5.31

9.95

5.16

Ay

0.37

0.84

0.49

Az

5.23

10.76

6.59

Velocities [m]

Accelerations [m/s2 ]

3 Results 3.1 General Remarks The study analyzed the values from the Time History analysis for displacements, velocities, and accelerations (Table 1). The mentioned results were presented for three directions, i.e. two horizontal ones (X and Y) and for the vertical direction (Z) [26]. In the

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Time History analysis, 12 modes of natural vibrations were used, which is in line with Eurocode 8 [27]. 3.2 Displacements The maximum displacement values were observed in the model where the El Centro excitation was used (Fig. 4). In this case, the maximum displacement was 13.02 × 10–3 m in the horizontal direction (Dx). For the other two excitations, lower values were obtained and amounted to 7.35 × 10−3 m (Taft) and 4.82 × 10−3 m (San Fernando). As for El Centro and for the Taft and San Fernando excitations, the maximum values were obtained in the horizontal direction (perpendicular to the bridge structure (Dx)). Thus, in the vertical direction (Dz), the obtained values of vertical displacements were lower in the range of 10–49% than in the horizontal direction (Dx) for the three analyzed records. In turn, the lowest values were obtained for the horizontal direction (Dy), i.e. longitudinal to the tested bridge. The greatest horizontal displacements recorded for the El Centro record were located near the inlet and outlet of the bridge, at the height of the quarter points (Fig. 5b). In the models where the Taft (Fig. 5a) and San Fernando (Fig. 5c) excitations were used, the maximum values of displacements were recorded in the same places.

Fig. 4. Maximum displacements from seismic analysis.

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Fig. 5. Place of maximum horizontal displacements in models with seismic records: a) Taft, b) El Centro, and c) San Fernando.

Fig. 6. Maximum velocities from seismic analysis.

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3.3 Velocities After the numerical analysis of Time History, it can be observed that the highest velocities in the steel shell were recorded for the model where the El Centro excitation was applied (Fig. 6). In this case, the maximum value was 0.99 m/s (Fig. 7b) and was recorded in the horizontal direction (Dx). In the remaining analyzed cases, lower values were obtained. For the Taft recording, the maximum velocity value was 0.58 m/s (Fig. 7a), and it was also recorded in the horizontal direction Dx. In turn, for the San Fernando record, the maximum value was equal to 0.50 m/s (Fig. 7c), and it was obtained in the vertical direction (Dz).

Fig. 7. Maximum velocities from seismic analysis for excitation: a) Taft, b) El Centro, and c) San Fernando.

The place of occurrence of the maximum velocities was varied, which resulted from different directions of occurrence of the maximum velocities. In the San Fernando record, maximum velocities are recorded in the vertical direction (Dz), and the highest velocities are recorded around the shell crown at the inlet and outlet of the bridge. In the case of other records, where the maximum velocities were recorded in the horizontal direction, these values can be observed near the quarter points.

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3.4 Accelerations The highest acceleration values in the steel shell were recorded for the El Centro record, and it was 10.76 m/s2 (Fig. 8). In the remaining analyzed seismic events, the maximum values were respectively 5.31 m/s2 (Taft) and 6.59 m/s2 (San Fernando). In this case, the same trend was noted as for velocity, i.e. the maximum values were recorded in the same directions. For the Taft recording, the maximum acceleration was noted in the horizontal direction (Ax). In the other records (El Centro and San Fernando), the maximum values were recorded in the vertical direction (Az). The location of the acceleration peaks varied as they depended on the direction in which the peaks appeared. A similar trend was also noted in the case of velocity. For the Taft recording (Fig. 9a), the maximum values were recorded at the height of the quarter points (shell inlet and outlet). In the case of El Centro (Fig. 9b) and San Fernando (Fig. 9c) records, the maximum values were recorded in the vicinity of the shell crown (bridge inlet and outlet).

Fig. 8. Maximum accelerations from seismic analysis.

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Fig. 9. Maximum accelerations from seismic analysis for excitation: a) Taft, b) El Centro, and c) San Fernando.

4 Conclusions After the numerical analysis, it can be concluded that the impact of the intensity and length of the recording is visible, which is shown in the values of displacements, velocities and accelerations. It was noticed that the decisive factor for the maximum values is the length of the intense seismic zone, and not its maximum acceleration or the length of the excitation. This theory is confirmed by the highest values recorded for the El Centro record. In addition, it was noted that: • in the analyzed cases, the maximum values appeared near the inlet and outlet (displacements, velocities, accelerations), • the greatest displacement and velocities were recorded in the horizontal direction Dx, • the greatest accelerations were recorded in the vertical direction of Dz. Summing up, it can be concluded that further research in this area is necessary, as the conducted research did not show a clear tendency to determine the impact of the duration of the shock or its intensity. The conducted research will be helpful in designing and assessing the seismic sensitivity of soil-steel bridges in seismic areas. In addition, the obtained results show that the

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duration of the recording and the intense zone are important in determining the seismic response, so further numerical and experimental studies in this area are necessary.

References 1. Beben, D.: Soil-Steel Bridges. Design, Maintenance and Durability. Springer, Cham (2020). https://doi.org/10.1007/978-3-030-34788-8 2. Janusz, L., Madaj, A.: Soil-Shell Structures with Corrugated Plates. Design, Construction and Maintenance. Transport and Communication Publishers, Warsaw (2019) 3. Machelski, C.: Construction of Soil-Shell Structures. The Lower Silesian Educational Publishers, Wroclaw (2013) 4. Embaby, K., El Naggar, K.H., El Sharnouby, M.: Investigation of bevel-ended large-span soilsteel structures. Eng. Struct. 267, 114658 (2022). https://doi.org/10.1016/j.engstruct.2022. 114658 5. Embaby, K., El Naggar, K.H., El Sharnouby, M.: Performance of large-span arched soil– steel structures under soil loading. Thin-Walled Struct. 172, 108884 (2022). https://doi.org/ 10.1016/j.engstruct.2022.114658 6. Rowinska, W., Wysokowski, A., Pryga, A.: Design and technology recommendations for flexible structures with corrugated steel plates. Road Bridge Research Institute, Warsaw (2004) 7. Handbook of Steel Drainage and Highway Construction Products. Corrugation Steel Plate Institute, Cambridge, Ontario (2007) 8. Guidelines for the Design of Buried Steel Pipes. American Society of Civil Engineers, American Lifelines Alliance (2001) 9. Maleska, T., Beben, D.: Numerical analysis of a soil-steel bridge during backfilling using various shell models. Eng. Struct. 196, 109358 (2019). https://doi.org/10.1016/j.engstruct. 2019.109358 10. Flener, E.B.: Soil-steel interaction of long-span box culverts—performance during backfilling. J. Geotech. Geoenviron. Eng. 136(6), 823–832 (2010) 11. Koruszewicz, L., Kunecki, B.: Behaviour of the steel box-type culvert during backfilling. Arch. Civ. Mech. Eng. 11(3), 637–650 (2011) 12. Manko, Z., Beben, D.: Tests during three stages of construction of a road bridge with a flexible load-carrying structure made of Super Cor type steel corrugated plates interacting with soil. J. Bridge Eng. 10(5), 570–591 (2005). https://doi.org/10.1061/(ASCE)1084-0702(2005)10: 5(570) 13. Beben, D., Stryczek, A.: Numerical analysis of corrugated steel plate bridge with reinforced concrete relieving slab. J. Civ. Eng. Manag. 22(5), 585–596 (2015). https://doi.org/10.3846/ 13923730.2014.914092 14. Beben, D.: Dynamic amplification factors of corrugated steel plate culverts. Eng. Struct. 46, 193–2014 (2013). https://doi.org/10.1016/j.engstruct.2012.07.034 15. Beben, D.: Field performance of corrugated steel plate road culvert under normal live load conditions. J. Perform. Constr. Facil. 27(6), 807–817 (2013). https://doi.org/10.1061/(ASC E)CF.1943-5509.0000389 16. Beben, D.: Experimental study on dynamic impacts of service train loads on a corrugated steel plate culvert. J. Bridge Eng. 18(4), 339–346 (2013). https://doi.org/10.1061/(ASCE)BE. 1943-5592.0000395 17. Beben, D.: Corrugated steel plate (CSP) culvert response to service train loads. J. Perform. Constr. Facil. 28(2), 376–390 (2014). https://doi.org/10.1061/(ASCE)CF.1943-5509.0000422 18. Flener, E.B., Karoumi, R.: Dynamic testing of a soil-steel composite railway bridge. Eng. Struct. 31, 2803–2811 (2009)

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19. Canadian Highway Bridge Design Code CAN/CSA-S6-17. Canadian Standards Association International, Mississauga (2017). 20. Katona, M.G.: Seismic design and analysis of buried culverts and structures. J. Pipeline Syst. Eng. Pract. 1(3), 111–119 (2010) 21. Fairless, G.J., Kirkaldie, D.: Earthquake performance of long-span arch culverts. Report 366. NZ Transport Agency Research, Wellington (2008) 22. Maleska, T., Beben, D.: Behaviour of soil-steel composite bridge with various cover depths under seismic excitation. Steel Compos. Struct. 42(6), 747–764 (2022). https://doi.org/10. 12989/scs.2022.42.6.747 23. Maleska, T., Beben, D., Nowacka, J.: Seismic vulnerability of a soil-steel composite tunnel – Norway Tolpinrud Railway Tunnel Case Study. Tunn. Undergr. Space Technol. 110, 103808 (2021). https://doi.org/10.1016/j.tust.2020.103808 24. Maleska, T., Nowacka, J., Beben, D.: Application of EPS geofoam to a soil-steel bridge to reduce seismic excitations. Geosciences 9(10), 448 (2019). https://doi.org/10.3390/geoscienc es9100448 25. Maleska, T., Bonkowski, P., Beben, D., Zembaty, Z.: Transverse and longitudinal seismic effects on soil-steel bridges. In: Köber, D., De Stefano, M., Zembaty, Z. (eds.) Seismic Behaviour and Design of Irregular and Complex Civil Structures III. GGEE, vol. 48, pp. 23–36. Springer, Cham (2020). https://doi.org/10.1007/978-3-030-33532-8_3 26. Pilarska, D., Maleska, T.: Numerical analysis of steel geodesic dome under seismic excitations. Materials 14(16), 4493 (2021). https://doi.org/10.3390/ma14164493 27. CEN: Actions design of structures for earthquake resistance. Eurocode 8, European Committee for Standardization, Brussels (2005)

Impact of Reinforcement Layer in Soil-Steel Culvert on Laboratory and Numerical Tests Tomasz Maleska1

, Adam Wysokowski2

, and Damian B˛eben1(B)

1 Faculty of Civil Engineering and Architecture, Opole University of Technology, Katowicka

48, 45-061 Opole, Poland {t.maleska,d.beben}@po.edu.pl 2 Faculty of Civil Engineering, Architecture and Environmental Engineering, University of Zielona Góra, Prof. Z. Szafran St. 1, 65-417 Zielona Góra, Poland [email protected]

Abstract. Soil-steel structures, such as culverts, bridges, or tunnels have recently been gaining popularity. This state of affairs is undoubtedly influenced by their simplicity of construction and low construction costs in relation to traditional bridge structures made of concrete, steel, or wood. Despite the high popularity of these objects, knowledge about their behavior under various load is relatively limited. An even bigger problem can be noticed when there is a need to strengthen an already existing structure due to insufficient soil cover depth or increased load on a given structure. Only a few scientific works can be found in this area. Therefore, the main aim of this study is to evaluate the use of geotextile for the reinforcement of soil cover over the culvert shell. The analyzed object has a span of 3.55 m and a clear height of 1.42 m. The culvert shell is made of corrugated steel plate with dimensions of 150 × 50 × 5 mm. The paper proposes a numerical model of the soil-steel culvert made in the DIANA FEA program based on the finite element method (FEM). A static load of 800 kN was applied in accordance with the program of previous laboratory tests. The results of the FEM calculations were verified on the basis of experimental tests. The study analyzed the values of displacements, axial forces, and bending moments. The obtained results will be helpful for designers who will have to face the problem of insufficient soil cover or increased load on the structure. Keywords: Soil-steel culvert · Geotextile reinforcement · FEM · Laboratory tests · Static loads

1 Introduction Soil-steel structures, such as bridges, culverts, tunnels, thanks to their numerous advantages, are becoming more and more popular in various parts of the world (Fig. 1). The largest structure of this type was built in the United Arab Emirates, with a span exceeding 32 m. However, it should be noted that the most common are objects whose span does not exceed a few meters [1]. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 139–148, 2023. https://doi.org/10.1007/978-3-031-26879-3_11

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Fig. 1. Example of a box-shaped soil-steel road bridge.

Despite the considerable popularity of these structures in bridge engineering, there is little literature [2–4], standards [5, 6], and recommendation [7] related to this subject. In the literature, the dominant issue is the analysis of soil-steel structures under static loads. In particular, research works were carried out during the construction stage of these structures (the stage of backfilling the steel shell [8–10]) and at the stage of real loads using trucks [11, 12]. The aspect of the influence of dynamic loads on soil-steel bridges and culverts can be found less frequently. In this area, the works were carried out under the influence of passing trucks and trains over the bridge or culvert [13–15]. It should also be added that the tests were carried out in the field of strong dynamic loads (seismic excitations) [16–19]. In many cases, soil-steel structures require the reinforcement of the backfill above the shell as a result of low soil cover or dictated by changes in the design load (change of the road class). At present, there are no clear guidelines for this type of reinforcement. In the literature, you can find works where. For concrete structures laid in the ground, for example, EPS layers over the load-bearing structure were used [20, 21]. Taking into account the interaction of soil-steel structures, the relief of the steel shell can be obtained as a result of the redistribution of stresses arising from live loads over a larger area of soil. The work [22] shows that such a solution makes sense and works in practice. On the other hand, in the work [23], the experimental studies with the use of geotextile were carried out. It was placed over the steel shell of the soil-steel culvert. The result of these studies was the reduction of maximum displacements. It should be emphasized that there is no method of modeling the relief material in the form of geotextile in the ground in the literature. Therefore, the aim of this study is to perform a numerical analysis of the soil-steel culvert with the use of a geotextile relief layer under a static load of 800 kN. The paper presents two numerical models: (i)

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I (without geotextile) and (ii) II (with geotextile) based on experimental studies from Wysokowski [23]. It was found that the influence of geotextile on the obtained results is significant and of practical importance. The numerical program DIANA FEA based on the finite element method was used for the numerical analysis.

Fig. 2. Model of soil-steel culvert in laboratory test: a) general view of test site, b) cross-section with measurement points 1–3 (dimensions are given in mm) [23].

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2 Description of Experimental and Numerical Analysis 2.1 Properties of Laboratory and Numerical Models The numerical models were based on the laboratory model of the culvert with a height of 1.42 m, a span of 3.55 m and the length of 13.70 m (Fig. 2). The numerical models (Fig. 3) had the same dimension as the laboratory model. Other parameters of the tested structure were as follows: (i) corrugation steel plate: 150 × 50 mm, (ii) thickness of steel plate: 5.00 mm, (iii) type of profile – box (open cross-section), (iv) type of foundations – concrete footing with dimension of 0.50 × 0.50 m, (v) the steel structure was additionally reinforced (by special ribs made of steel plates located on the top section of perimeter – in the crown).

Fig. 3. Numerical models: a) without geotextile (model I), b) with geotextile (model II).

2.2 Properties of Steel Shell The steel shell structure was characterized by parameters which were in accordance with European Standard EN 10025 [24] with minimum yield stress of 235 MPa. In numerical analysis, the following properties were used: (i) Young modulus of 2.1 × 108 kN/m2 , (ii) Poisson’s ratio of 0.3, (iii) density of 7.85 T/m3 , (iv) area cross-section of 6.27 mm2 /mm, and (v) moment of inertia of 1914.5 mm6 /mm. In numerical analysis, the corrugated steel plate was modelled as flat plate with orthotropic characteristics. An orthotropic shell (Fig. 3) was used to calculate the equivalent parameters [2, 17, 19, 25]. 2.3 Properties of Backfill The backfill material was well-graded soil with a maximum grain size of 32 mm. The backfill was placed in layers with a maximum thickness of 0.30 m prior to compaction. The required degree of compaction was 97% Standard Proctor density. The overall depth of the soil cover over the structure was 0.60 m. In numerical model, the soil was modelled as: (i) elastic modulus 100000 kN/m2 , (ii) Poisson’s ratio 0.2, (iii) unit of weight 20.5 kN/m2 , (iv) angle of internal friction 39°, (v) cohesion 3 kPa, (vi) failure ratio 0.8, (vii) minimum tangential modulus 100 kN/m2 , (viii) minimum confining stress of 10 kN/m2 , (ix) atmosphere pressure 101.31 kN/m2 .

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2.4 Properties of Geosynthetic Reinforcing In both the numerical and laboratory testing, geosynthetic reinforcing (geotextile – Figs. 2 and 3) was used with a grammage 500 g/m2 to strength the soil cover over the shell. In addition, the Poisson’s ratio was equal 0.1 and density 2.49 × 10–8 T/m3 . 2.5 Properties of Interface In the numerical analysis of soil-steel bridges, the aspect of taking into account the SoilStructure Interaction (SSI) phenomenon is very important. SSI was used between steel shell and soil as well as backfill and geotextile. Thus, for this numerical research the connection between various materials was modelled as an automatic interface using a “nonlinear elastic friction” function with: (i) normal stiffness and shear stiffness modulus 2.8 × 106 kN/m3 (between backfill and steel shell) and 250 kN/m3 (between backfill and geotextile), and (ii) R-factor for nonlinear elastic friction of 1.0 (for both cases).

3 Results Analysis 3.1 General Remarks The paper presents a numerical analysis for the soil-steel culvert, which determines the impact of the application of the bridge structure reinforcement through the use of geotextile. The numerical program DIANA FEA based on the finite element method was used for the numerical analysis. For the numerical model, the mesh density was not greater than 0.2 m. The numerical studies were based on the experimental studies presented in [23]. The results of the tests carried out for the displacements of the culvert are presented in Table 1. In addition, the work uses a static load of 800 kN, which corresponds to the standard load for class A according to Polish Standards of bridge load [26]. Table 1. Maximal displacements from numerical analysis and experimental testing. Measurement points

Displacements [mm] Without geotextile

With geotextile

Model I

Experiment

Model II

Experiment

1

– 0.52

0.93

– 0.41

0.74

2

– 4.56

– 4.46

– 3.29

– 3.31

3

– 0.52

0.84

– 0.41

0.63

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3.2 Displacements Analyzing the obtained results of displacements for the numerical models and the data experiment, it was noticed that the use of reinforcement of the geotextile structure makes sense and reduces displacements. The maximum values obtained in model I (without geotextile) are similar to the values obtained during the experiment. In the numerical model (model I), the result was –4.56 mm (Fig. 4a), while the experiment value was –4.46 mm, so the difference was about 2%.

Fig. 4. Displacements from numerical analysis for: a) model I (without geotextile layer), b) model II (with geotextile layer).

In the case of model II (with geotextile), the displacements in relation to model I (without geotexile) were smaller by 28%, which amounted to –3.29 mm (Fig. 4b). However, during the experiment (with geotextile), displacements equal to –3.31 mm were obtained. This value was lower than the laboratory model without geotextile by 26%. In turn, the difference in relation to the numerical model II was only 1%, i.e. at the level of the measurement error. The location of the maximum displacements was the same for both the experimental and numerical tests, i.e. they were obtained in the steel shell crown under an applied load. In other places, lower values were obtained as shown in Fig. 4. It should also be added that in the remaining measurement points (# 1 and 3), which are located above the spread footings, greater differences were noted between the numerical model and the experiment. In the case of model I (without geotextile), the maximum values for points 1 and 3 were 0.52 mm, while in the experiment the values of 0.93 mm and 0.84 mm were obtained, respectively. In the case of model II (with geotextile), a displacement of 0.41 mm was obtained over the footing, so this value was lower by 21% compared to model I (without geotextile). However, for the values from the experiment, they were 0.74 mm and 0.63 mm for points 1 and 3, respectively, so these values were greater than the model II (with geotextile) by 54–80%. The conducted analyzes also show that the use of geotextile in the structure of the soil-steel culvert influences its structure by the visible redistribution of the load over the steel shell. A greater distribution of the applied load on the steel shell structure is visible (Fig. 4).

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3.3 Axial Forces In the case of axial forces, the tendency to reduce the maximum values is more pronounced than in the case of displacements. The highest values were recorded in model I (without geotextile), i.e. –281.33 kN/m. In the case of model II (with geotextile), the maximum axial force was lower by 62% compared to model I and was equal to –107.98 kN/m. It should be added that in both cases, the obtained maximum values were of a compressive nature. In addition, it is worth noting that the application of the relief layer (model II with geotextile) resulted in the maximum values being obtained in a different direction than in the case of model I (without geotextile), the maximum values were obtained in the Nxx direction, and in the case of model II in the direction of Nyy. The aftermath of this phenomenon is the fact that the location of the maximum axial forces has changed. In the case of model I (without geotextile), the maximum axial forces were recorded under the given loads, while in model II (with geotextile), the maximum forces were obtained near the quarter points. On the other hand, in the places where the load was applied over the steel shell (model II), the axial forces were lower by about 20–30% compared to the maximum value (Fig. 5).

Fig. 5. Axial forces from numerical analysis for: a) model I (without geotextile), b) model II (with geotextile).

3.4 Bending Moment In the bending moments, similar values were recorded in both numerical models with a slight increase in the value in model II. The obtained difference was small in relation to model I (without geotextile) and amounted to 7%. In these models, the values were 18.13 kN (model I) and 19.40 kN (model II). In the case of bending moments, a redistribution of forces is visible as a result of a given load. In model II (with geotextile), the maximum values of bending moments were obtained over a larger area of the corrugated steel shell than in the case of model I (without geotextile). It can also be seen that the maximum values appeared only in the vicinity of the crown of the steel shell, in its central part as shown in Fig. 6. In the remaining parts of the steel shell, the bending moments were negligible and did not exceed the value of 2.5 kN for model I (without geotextile) and model II (with geotextile).

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Fig. 6. Bending moments from numerical analysis for: a) model I (without geotextile), b) model II (with geotextile).

4 Conclusions After the numerical analysis, it can be concluded that the use of geotextile to strengthen the structure of the soil-steel culvert has a positive effect on the behavior of the steel shell under static load. Moreover, based on the conducted experimental studies and numerical analyzes, it can be found that: • significant reductions in the maximum values of displacements and axial forces have been noticed, • the application of the geotextile layer changes the location of the maximum values of the axial forces, • the use of a relieving layer (geotextile) changes the occurrence of the directions of maximum values (axial forces and bending moments), • the application of the reinforcement leads to a redistribution of the applied load over a larger area of the steel shell (displacements and bending moments); model II shows a larger area in the steel shell where the maximum values appear, • the results from the numerical model were consistent with the actual behavior of the tested bridge, which was confirmed by the displacements in the steel shell. The conducted tests will be helpful in designing and determining the influence of external loads on soil-steel structures. In addition, the research shows a solution to easily increase the load-bearing capacity of a soil-steel culvert or bridge used. In addition, these studies show that the use of relieving materials for soil-steel structures makes sense and constitutes a new research area for further numerical and laboratory tests. The aspect of the influence of dynamic loads on reinforced soil-steel structures seems to be of particular interest.

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References 1. Embaby, K., El Naggar, K.H., El Sharnouby, M.: Investigation of bevel-ended large-span soilsteel structures. Eng. Struct. 267, 114658 (2022). https://doi.org/10.1016/j.engstruct.2022. 114658 2. Beben, D.: Soil-Steel Bridges. Design, Maintenance and Durability. Springer, Cham (2020). https://doi.org/10.1007/978-3-030-34788-8 3. Machelski, C.: Construction of Soil-Shell Structures. The Lower Silesian Educational Publishers, Wroclaw (2013) 4. Janusz, L., Madaj, A.: Soil-Shell Structures with Corrugated Plates. Design, Construction and Maintenance. Transport and Communication Publishers, Warsaw (2019) 5. Canadian Highway Bridge Design Code CAN/CSA-S6-17. Canadian Standards Association International, Mississauga (2017) 6. AASHTO. Bridge LRFD Design Specifications. American Association of State Highway and Transportation Officials, Washington (2017) 7. Rowi´nska, W., Wysokowski, A., Pryga, A.: Design and technological recommendations for ˙ flexible engineering structures made of corrugated sheets. IBDiM, Zmigród, Poland (2004) 8. Maleska, T., Beben, D.: Numerical analysis of a soil-steel bridge during backfilling using various shell models. Eng. Struct. 196, 109358 (2019). https://doi.org/10.1016/j.engstruct. 2019.109358 9. Koruszewicz, L., Kunecki, B.: Behaviour of the steel box-type culvert during backfilling. Arch. Civil Mech. Eng. 1(3), 637–650 (2011) 10. Machelski, C., Koruszewicz, L.: Contact interaction between corrugated steel shell and the soil backfill determined based on the measurements of shell deformations. Arch. Civ. Eng. 1, 57–79 (2021) 11. El-Sawy, K.M.: Three-dimensional modeling of soil-steel culverts under the effect of truckloads. Thin-Walled Struct. 41(8), 747–768 (2003). https://doi.org/10.1016/S0263-823 1(03)00022-3 12. Manko, Z., Beben, D.: Tests during three stages of construction of a road bridge with a flexible load-carrying structure made of Super Cor type steel corrugated plates interacting with soil. J. Bridge Eng. 10(5), 570–591 (2005). https://doi.org/10.1061/(ASCE)1084-0702(2005)10: 5(570) 13. Beben, D.: Experimental study on dynamic impacts of service train loads on a corrugated steel plate culvert. J. Bridge Eng. 18(4), 339–346 (2013). https://doi.org/10.1061/(ASCE)BE. 1943-5592.0000395 14. Beben, D.: Corrugated steel plate (CSP) culvert response to service train loads. J. Perform. Constr. Facil. 28(2), 376–390 (2014). https://doi.org/10.1061/(ASCE)CF.1943-5509.0000422 15. Flener, E.B., Karoumi, R.: Dynamic testing of a soil-steel composite railway bridge. Eng. Struct. 31, 2803–2811 (2009) 16. Maleska, T., Beben, D.: Behaviour of soil-steel composite bridge with various cover depths under seismic excitation. Steel Compos. Struct. 42(6), 747–764 (2022). https://doi.org/10. 12989/scs.2022.42.6.747 17. Maleska, T., Beben, D., Nowacka, J.: Seismic vulnerability of a soil-steel composite tunnel– Norway Tolpinrud Railway tunnel case study. Tunn. Undergr. Space Technol. 110, 103808 (2021). https://doi.org/10.1016/j.tust.2020.103808 18. Maleska, T., Nowacka, J., Beben, D.: Application of EPS geofoam to a soil–steel bridge to reduce seismic excitations. Geosciences 9(10), 448 (2019). https://doi.org/10.3390/geoscienc es9100448 19. Maleska, T., Bonkowski, P., Beben, D., Zembaty, Z.: Transverse and longitudinal seismic effects on soil-steel bridges. In: Köber, D., De Stefano, M., Zembaty, Z. (eds.) Seismic

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Concrete Engineering

Comparative Analysis of the Use of Different Concrete Models in the Evaluation of the Propagation of Damage in the Reinforced Concrete Sample Kseniya Yurkova

and Tomasz Krykowski(B)

Department of Mechanics and Bridges, Faculty of Civil Engineering, Silesian University of Technology, Akademicka 5, 44-100 Gliwice, Poland {kseniya.yurkova,tomasz.krykowski}@polsl.pl

Abstract. The paper compares elastic-plastic material models with hardening/softening used to estimate the evolution of degradation of concrete cover in reinforced concrete structural elements, available among others in the ANSYS system. To avoid the solution’s dependence on the FEM mesh, regularization using fracture energy was used. In addition, the effect of bond slip in the steel-concrete contact area on the evolution of degradation of concrete cover was analyzed. The paper point out the large discrepancy of the results depending on the concrete model and relatively low impact of steel-concrete contact region. Keywords: Corrosion · Cover damage · FEM · Computational plasticity

1 Introduction Modeling of concrete cover cracking caused by reinforcement corrosion is a complex issue and has been the subject of several scientific studies. The description of the evolution of cracking depends on several factors, such as the choice of material model, the method of describing the contact interactions steel-concrete-corrosion products, the physicochemical transformations that contribute to corrosion process or the way the corrosion products impact on the structure of the cover. In the literature, one can find several more and less advanced models describing the cracking of the cover of reinforced concrete elements. It is necessary to mention here both simple analytical models [1, 2] as well as more advanced complex continuum models of concrete taking into account the nonlinear behavior of concrete [3–6] or the degradation of the contact region of steel and concrete [3, 7]. Using different approaches to the description of the degradation of concrete cover is burdened with a certain error, which is overlooked in relation to the issues of reinforcement corrosion. There are few works in which you can find a comparison of the effectiveness of using various continual models describing the propagation of damage in the cover, as well as contact interaction models in relation to reinforcement corrosion tests. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 151–164, 2023. https://doi.org/10.1007/978-3-031-26879-3_12

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This work compares the effectiveness of using two models of concrete to assess the propagation of damage in the elements of reinforced concrete structures in which reinforcement corrosion occurs. For the purposes of calculations, it was assumed that the models used in computational analysis will be the elastic-plastic models based on the yielding surface of Menetrey-Willam and Drucker-Prager, available in the ANSYS computational program. In addition, the influence of the adopted model of contact interactions of reinforcing steel-concrete on the degradation of concrete cover was analyzed. Moreover, two models of contact that consider both slip and ideal bounding of steel and concrete were examined. In both cases, the parameters of contact interactions were estimated in a theoretical way. The obtained results were compared with the results taken from literature, as well as the ones obtained for a different model of concrete.

2 Models and Methods 2.1 Analyzed Problem and the Target of Analysis The subject of computer analysis of the paper was the estimation of changes in the geometry of reinforced concrete sample (100 × 100 × 80 mm) in the case of accelerated corrosion test of reinforcement (φ = 20 mm using various computational models). The results of the calculations were related to the results of experimental research and computer calculations obtained for the elastic-plastic-fracturing model available in the ATENA program [8–10]. The paper presents the data of measurements of the displacements of the corners of the sample, as well as the validation of the obtained tests using a computer model. The calculations were made considering the contact at the steel-concrete interface, although the adopted material parameters of the contact layer excluded the possibility of the bound slip. The research presented in this paper compares the results obtained on the basis of the computational approach similar in concept to the material model (modified elasticplastic models of Menetrey-Willam [11] and Drucker-Prager [11]). The impact of the contact slip between a corroded rebar and concrete was also analyzed. The model was made in the ANSYS system using eight-node CPT215 elements (concrete), Solid185 elements (steel), while contact interactions were modeled using TARGE170 and CONTA174 elements. The calculation model of the analyzed sample together with the generated finite element mesh was presented in Fig. 1. Figure 1a shows the analyzed reinforced concrete sample subjected to the electrolysis process using an externally connected power source along with the measured edge of the sample (distance between the corner points A and B). Figures 1b and 1d show the FEM model of test specimen, Fig. 1b shows the discretized concrete part of the tested element, while Fig. 1d discretized steel rebar element. It was assumed that the impact of corrosion products on the concrete part of the sample would take place through contact elements (marked schematically in the drawing), as the result of the substitute increase in the volume of the rebar. The schematic diagram of the interaction between corrosion products and the cover of concrete was shown in Fig. 1c.

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Fig. 1. Computational model of the analyzed reinforced concrete sample (dimensions in mm, description in the text).

2.2 Constitutive Relationships The elastic-plastic hardening/softening material model taking into account material volumetric strains caused by corrosion products was adopted, i.e.: σ˙ = Ce : ˙ e , ˙ = ˙ e + ˙ p + ˙ v , ˙ p = γ˙

∂Q ∂σ

(1)

where σ is stress tensor, Ce is elastic stiffness tensor, ε is total strain tensor, εκ is strain tensor: elastic κ = e, plastic κ = p, volumetric κ = v, γ is plastic strain rate multiplier, Q is plastic potential. 2.3 Contact The impact of reinforcing steel and corrosion products on the concrete cover was described through the introduction of the contact elements [12]. The Coulomb friction model and a flexible surface-to-surface contact model were used, including among others a normal interaction to the contact surface, a tangential interaction associated with friction, and the possibility of contact surface separation. The problem of contact was analyzed using an augmented Lagrange algorithm [12]:  (2) τlim = μP + c, τ = τ21 + τ22 ≤ τlim ;  P=

if |un | > 0 0 Knn un + λk+1 if |un | ≤ 0

(3)

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where P is normal contact pressure, un is contact gap size, λk+1 is Lagrange multiplier component at iteration k + 1, Knn is normal contact stiffness. Sliding of the contact elements will take place when τ exceeds limit value τlim according to Eq. (2). When the equivalent friction stress of the maximum equivalent friction τ is reached, regardless of the amount of contact pressure τmax , slippage will also occur. Lagrange multiplier component can be calculated using the following equation:  λk + Knn un if |un | > δ (4) λk+1 = if |un | ≤ δ λk where δ is compatibility tolerance. Tangential stresses in the contact plane (isotropic friction) can be determined according to the relationship:  n−1 + Ks ui , for τ − μiso P < 0 (no sliding) τ τi = i (5) i for τ − μiso P = 0 μiso P u (sliding) u , where Ks is tangential contact stiffness, ui is slip increment in direction i over the current substep (in the case of sliding, ui is the elastic slip increment, which represents the reversible tangential motion within the current substep in the direction i), u is a equivalent slip increment over the current substep, μiso is coefficient of friction, τn−1 i frictional stress in direction i at the end of previous time step. 2.4 Corrosion Interaction Model It was assumed that the interactions in the transition layer of steel concrete interface caused by corrosion products can be defined by the velocity tensor of volumetric strains [8]. It will be assumed that non-zero tensor coordinates occur only in a plane perpendicular to the rebar axis:   (1 − β) α−1 ϑ − 1 kcomp I V V δij ; i, j = 1, 2; kcomp = χkeff ; (6) ε˙ ij = ε˙ δij = ηρ Fe2+ V0 ⎧ ⎪  dla t ≤ t0 ⎨ 1 3β + 2(1 − β) dla t ≤ tcr −t dla t0 ≤ t ≤ tcr η= ; β = ttcrcr−t (7) ⎪ 2 dla t > tcr ⎩ 0 0 dla t > t cr where δij is Kronecker delta; β is a parameter that describes the changes in the intensity of the exposure of corrosion products to the structure of the cover; t0 is corrosion activation time (time between initiation of corrosion products formation and initiation of mechanical interaction); α, ϑ are parameters depending on the composition of corrosion products; kcomp and keff are respectively the calculated and effective electrochemical equivalents of iron; χ is the coefficient of reduction of interaction associated with the transfer of corrosion products outside the cover; I is the intensity of electric current; tcr is critical time; V0 is the initial volume.

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2.5 Menetrey-Willam Model Flow surface in the Menetrey-Willam model (M-W model) [11, 12] is a function of the invariants and deviators of the stress tensor in the form: 

√ 2 c2 1 (8) fMW = √ I1 + rρ + ρ2 − , ρ = 3J2 c3 c3 3 where I1 is the first invariant of the stress tensor; J2 is the second invariant of the deviator of the stress tensor; r is a function of invariants of the stress deviator and material parameters; c2 and c3 are quantities depending on the compression hardening/softening function c , tension softening function t , and material parameters as follows:

 1 1 1 1 c fb − tc ft 3 ; c3 = c2 = √ − + ; (9) 2 f f 2 ( c fc ) ( c fc )2 6 c c c b   4 1 − e2 cos2 θ + (2e − 1)2 (10) r(θ, e) =      0,5 ; 2 1 − e2 cos θ + (2e − 1) 4 1 − e2 cos2 θ + 5e2 − 4e  tc ft ( c fb )2 − ( c fc )2 1+ε t for κc ≤ κcm ;ε = e= ; = . (11) tc t c for κc > κcm 2−ε c fb ( c fc )2 − ( tc ft )2 In the Eqs. (9–11) κc is compression internal parameter; θ is Lode angle (deviatoric angle); ft is uniaxial tensile strength; fc is compressive strength; fb is bidimensional compressive strength. It was assumed that the flow rule in the M-W model is non associated and defined according to the relationship (1) by the plastic potential Q = QMW [12]: I1 QMW = ρ2 + Bg ρ + Cg √ ; 3   √ 2 c fc tan ψ − 2( tc ft ) Bg 2( tc ft ) Bg = . ; Cg = √ + √    √ 2 3 3 1 − 2 tan ψ

(12)

(13)

2.6 Drucker-Prager Composite Model with Rankine Tension Failure Surface In the composite Drucker-Prager model with Rankine tension failure surface (D-P-R) material damage for loading in tension is described by associative flow, where Rankine tension failure fR [13] is given by:   2 I1 2 + (14) fR = 3J2 sin θ + π − ft t . 3 3 3 Materials behavior in compression and tension-compression was described by DruckerPrager failure surface fDPc : √ 3(fb − fc ) βc I1 fb fc fDPc = J2 + − σγ c c ; σγ c = √ ; βc = (15) 3 (2fb − fc ) 3(2fb − fc )

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where βc and σγ c are functions of material parameters. Plastic potential under compression can be accepted in the form of the relationship QDPc =



J2 + δc βc

I1 3

(16)

where δc is compression dilatancy parameter. 2.7 Hardening and Softening Hardening and softening in the M-W and D-P-R model were defined by using functions c and t , which describe material behavior under tension and compression (the functions depend on internal parameters κ analyzed in the model in the compression κ = κc and tension regime κ = κt ) [12]. As the internal parameter κ the unidimensional plastic strains under tension and compression were used. • Hardening/softening function in compression c , κ = κc :  ⎧ 2 κ ⎪ + − 2 κcm − κκ2 ; κ < κcm (1 ) ⎪ ci ci ⎪ ⎨ 2 cm  cm c = 1 − (1 − cu ) κκ−κ ; κcm ≤ κ ≤ κcu cu −κcm ⎪   ⎪ ⎪ −1 κ−κcu ⎩ + ( − )exp 2 cu cr cu cr κcu −κcm · cu − cr ; κcu < κ • Exponential softening function in tension t , κ = κt :     κ g Gft f2t ; at = ft ; gft = max t = exp − , at ft Li E

(17)

(18)

where Li is effective element length, E is Young’s moduli, Gft is fracture energy. The meaning of parameters κcm , κcu , cr , and tr , characterizing the evolution of hardening/softening function under tension and compression c and t , are presented in Fig. 2

Fig. 2. Hardening/softening functions under compression c and tension t .

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3 Computational Example 3.1 Determination of the Effective Values of the Components of Volumetric Strain Tensor Computer simulation was carried out using the measurement data presented in the paper [8]. The analyzed reinforced concrete element was subject to the influence of electric current causing dissolution of reinforcing steel and the deposition of corrosion products on the bar circuit. The electrical parameters of the analyzed system together with theoretically determined values of interactions are presented in Fig. 3. It was assumed [8] that the effective value of the electrochemical iron equivalent is keff = 0.006271(g/μA · year) while the corrosion products are made entirely of iron hydroxides: Fe(OH)2 (parameters α = 0.523, ϑ = 2.09) and Fe(OH)3 (parameters α = 0.622, ϑ = 2.24). In addition, the values of the following parameters, that are necessary to determine the increments of volumetric strain tensor, were assumed: porosity of transition layer ε = 0.55, transient layer thickness wws = 7.5 · 10−3 cm, critical time tcr = 53.83 h. 120

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Fig. 3. The changes of intensity of corrosion current and substitute increments of corrosion volumetric strains εV ii , i = 1, 2, in the analyzed sample.

3.2 Parameters Describing the Behavior of Steel and Concrete in Elastic and Inelastic Regions The material constants describing the behavior of concrete in the elastic region along with the strength parameters are summarized in Table 1, while the material parameters used to describe the concrete behavior in the inelastic region for both the M-W and D-P-R model with HSD2 are summarized in Table 2. The parameters of contact of steel and concrete are summarized in Tables 3 and 4. Based on the Model Code 90, the value of the limit contact actions at the joint of the reinforcing bar and the concrete cover was √ assumed τmax = 2.5 fc , and the value √ of cohesion at the interface between reinforcing bar and concrete cover was c = 0.05 fc . It was assumed that reinforcing steel is described by the elastic-plastic Huber-MissesHencky (H-M-H) material without hardening. The material constants are Young’s modulus E = 200 GPa, Poisson coefficient v = 0.3, and steel yield strength fy = 235 N/mm2 .

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Table 1. The elastic and strength material properties of concrete accepted in the analysis. Material parameters

Value

Initial Young’s moduli, E0 (MPa)

38,280

Initial Poisson coefficient, ν0 (−)

0.2

Tensile strength, ft (MPa)

3.99

Compressive strength, fc (MPa)

56.4

Biaxial compressive strength, fb = 1.15fc (MPa)

64.86

Fracture energy, Gft (N/m)

151

Table 2. List of material constants used to determine the parameters of the M-W model (M-W only *) and D-P with HSD2 (**D-P model with HSD2 only). Material parameters

Value

Plastic strain at uniaxial compressive strength, κcm (−)

0.00151

Plastic strain at transition from power law to exponential softening, κcu (−)

0.00175

Relative stress at start of nonlinear hardening, ci (−)

0.33

Residual relative stress at κcu , cu (−)

0.85

Residual compressive relative stress, cr (−)

0.2

Residual tensile relative stress, tr (−)

0.1

*) Dilatancy angle, ψ(Deg) **) Compression dilatancy parameter, δc (−)

20 1

Table 3. Steel – concrete contact material parameters (flexible contact). Description

Value

Friction coefficient, μ(−) Cohesion coefficient, c(MPa)

0.2 



Stiffness in normal direction knn · 10−8 MN/m3   Stiffness in tangential direction ktt · 10−8 MN/m3 Maximum equivalent frictional stress, τmax (MPa)

0.375 1.795 0.375 18.77

3.3 The Results of Computer Simulation The calculations were made for two values of the reduction coefficient of interaction χ, which is related to the transport of corrosion products outside the concrete element through the crack [8, 14]. The case under consideration was one where accordingly 40%

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Table 4. Steel-concrete contact material parameters (*fixed contact layer). Description

Value

Friction coefficient, μ(−)



Stiffness in normal direction knn · 10−8 MN/m3

1.0





Stiffness in tangential direction ktt · 10−8 MN/m3

2.2  2.2

*Contact interaction model

Bonded

(χ = 0.4) and 50% (χ = 0.5) of corrosion products affect the concrete of the cover. The results of computer analysis, the increase of test element edge length LAB , according to Fig. 1, for the fully bounded contact of steel and concrete and respectively 40% (50%) of corrosion products impacting on the concrete cover were shown in the Fig. 4 (Fig. 5). The comparison of the results for the fully bounded (stiff) and the flexible contact of steel and concrete for the 40% (50%) of corrosion products impacting on the concrete cover was shown in the Fig. 6 (Fig. 7), respectively. The curves show two extreme situation were the corrosion products are composed purely from Fe(OH)2 or Fe(OH)3 . 120

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Fig. 4. The results of calculations of changes of geometry of the test specimen LAB for the assumed fixed model of steel-concrete contact, impact parameter of corrosion products χ = 0.4(40%) and different material models (description in text).

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Fig. 5. The results of calculations of changes of geometry of the test specimen LAB for the assumed fixed model of steel-concrete contact, impact parameter of corrosion products χ = 0.5(50%) and different material models (description in text).

120 9

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Fig. 6. The results of calculations of changes of geometry of the test specimen LAB for the assumed fixed and flexible model of steel-concrete contact, impact parameter of corrosion products χ = 0.4(40%) and different material models (description in text).

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Fig. 7. The results of calculations of changes of geometry of the test specimen LAB for the assumed fixed and flexible model of steel-concrete contact, impact parameter of corrosion products χ = 0.5(50%) and different material models (description in text).

The following denotations (Figs. 4, 5, 6 and 7) were adopted for the test results obtained (computational model, software and chemical composition): 1) elastic-plasticfracturing model, ATENA, Fe(OH)2 ; 2) elastic-plastic-fracturing model, ATENA, Fe(OH)3 ; 3) elastic-plastic hardening/softening M-W model, Fe(OH)2 , flexible contact; 4) elastic-plastic hardening/softening M-W model, Fe(OH)2 , fixed contact; 5) elastic-plastic hardening/softening D-P with HSD2 model, Fe(OH)2 , flexible contact; 6) elastic-plastic hardening/softening D-P with HSD2 model, Fe(OH)2 , fixed contact; 7) elastic-plastic hardening/softening M-W model, Fe(OH)3 , flexible contact; 8) elasticplastic hardening/softening M-W model, Fe(OH)3 , fixed contact; 9) elastic-plastic hardening/softening D-P with HSD2 model, Fe(OH)3 , flexible contact; 10) elastic-plastic hardening/softening D-P with HSD2 model, Fe(OH)3 , fixed contact; 11) experimental results [8]. The results of calculations obtained in time t = 384 h for the models analyzed in the paper and the result of the experimental study obtained on the basis of literature were compared, depending on the coefficient of impact of corrosion products χ (40% and 50% respectively), and summarized in Tables 5, 6, 7 and 8. In addition, the deviations of the calculation results from the results obtained by the ATENA program model (percentage deviation) were determined. While analyzing the obtained results, significant discrepancies can be noticed between individual models that allow forecasting the degradation of concrete. The most conservative results of the calculations were obtained using the elastic-plastic-fracturing model implemented in the ATENA program.

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Table 5. Comparison of the displacement of test specimen (element edge elongation) LAB by using different material and contact models (interaction coefficient χ = 0.4, chemical composition Fe(OH)2 time 384 h). Model

Contact

LAB (mm)

(%)

ATENA, elastic-plastic-fracturing, Fe(OH)2

Fixed

0.95

0

M-W, elastic-plastic, hardening/softening, Fe(OH2 )

Fixed

1.05

11

M-W, elastic-plastic, hardening/softening, Fe(OH2 )

Flexible

1.03

8

D-P, elastic-plastic hardening/softening, Fe(OH)2

Fixed

1.15

22

D-P, elastic-plastic hardening/softening, Fe(OH)2

Flexible

1.14

20

Table 6. Comparison of the displacement of test specimen (element edge elongation) LAB by using different material and contact models (interaction coefficient χ = 0.4, chemical composition Fe(OH)3 , time 384 h) Model

Contact

LAB (mm)

(%)

ATENA, elastic-plastic-fracturing, Fe(OH)3

Fixed

1.12

0

M-W, elastic-plastic, hardening/softening, Fe(OH3 )

Fixed

1.25

12

M-W, elastic-plastic, hardening/softening, Fe(OH3 )

Flexible

1.22

9

D-P, elastic-plastic hardening/softening, Fe(OH)3

Fixed

1.34

20

D-P, elastic-plastic hardening/softening, Fe(OH)3

Flexible

1.36

21

Table 7. Comparison of the displacement of test specimen (element edge elongation) LAB by using different material and contact models (interaction coefficient χ = 0.5, chemical composition Fe(OH)2 time 384 h). Model

Contact

LAB (mm)

(%)

ATENA, elastic-plastic-fracturing, Fe(OH)2

Fixed

1.21

0

M-W, elastic-plastic, hardening/softening, Fe(OH2 )

Fixed

1.36

12

M-W, elastic-plastic, hardening/softening, Fe(OH2 )

Flexible

1.35

12

D-P, elastic-plastic hardening/softening, Fe(OH)2

Fixed

1.56

29

D-P, elastic-plastic hardening/softening, Fe(OH)2

Flexible

1.46

21

The results of calculations obtained using models available in the ANSYS program, based on the M-W surface with fixed contact and different chemical composition of corrosion products Fe(OH)2 and Fe(OH)3 (results in brackets), differ from those obtained by the ATENA program by about 8% (12%), assuming 40% impact of corrosion products and 11% (12%) for 50% impact. The results obtained using the D-P-R model with

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163

Table 8. Comparison of the displacement of test specimen (element edge elongation) LAB by using different material and contact models (interaction coefficient χ = 0.5, chemical composition Fe(OH)3 , time 384 h). Model

Contact

LAB (mm)

(%)

ATENA, elastic-plastic-fracturing, Fe(OH)3

Fixed

1.41

M-W, elastic-plastic, hardening/softening, Fe(OH3 )

Fixed

1.57

11

M-W, elastic-plastic, hardening/softening, Fe(OH3 )

Flexible

1.54

10

D-P, elastic-plastic hardening/softening, Fe(OH)3

Fixed

1.74

23

D-P, elastic-plastic hardening/softening, Fe(OH)3

Flexible

1.62

15

0

HSD2 were even more deviated from the solution obtained with the ATENA program. The results obtained here, in the case of rigid contact, differed from the solution obtained using the ATENA program by about 20 (22%) with 40% of corrosion products contribution in the mechanical impact on the cover and 29% (23%) in the case of 50% of impact. As can be seen by analyzing Tables 5, 6, 7 and 8, the discrepancies between the results obtained on the basis of a fixed and a flexible contact model in the case of 40% impact are small and do not exceed 3%. Larger discrepancies appeared for the increased contribution of corrosion products (50%) in the case of the D-P-R model (discrepancies in the M-W model are negligible and do not exceed 1%). The differences for this model, depending on the composition of the corrosion products, were 8% (9%). However, it should be emphasized that during the calculations in the case of the analyzed model problems with convergence appeared.

4 Conclusions 4.1 Conclusion No. 1 The analysis of the results shows the deviation of the results obtained with the use of different, elastic-plastic material models. These differences, in the case of the models analyzed in the work, can vary about 29% compared to the results obtained in the work [8] for the elastic-plastic-brittle model implemented in the ATENA program. These differences are so important that they can lead to significant errors in the assessment of the fracture time of the concrete cover. The obtained research results also point out the need of critical evaluation of the results based on different theoretical models which can be found in published research papers related to the field of modeling of cover cracking. 4.2 Conclusion No. 2 The obtained results of comparison of various analyzed contact models, taking into account the slippage of the contact layer (flexible) and the fixed, ideally rigid, contact of steel and concrete, allow to obtain results differing by a value not exceeding 3% in

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the case of M-W model and 9% in the case of D-P-R. This is not much and initially the conclusion can be formulated that the impact of slippage of the contact layer for degradation of the concrete cover and the evolution of the crack width can be ignored. This remark will not probably be correct in the case of a very advanced corrosion process of reinforcement. Acknowledgements. This research project was financially supported by the National Centre for Research and Development, NCBR, Poland, the competition for research projects with the participation of scientists from Belarus entitled “Solidarity with scientists”, “Modeling the durability and degradation of reinforced concrete elements in conditions of reinforcement corrosion”.

References 1. Balafas, I., Burgoyne, C.J.: Environmental effects on cover cracking due to corrosion. Cem. Concr. Res. 40, 1429–1440 (2010). https://doi.org/10.1016/j.cemconres.2010.05.003 2. Bhargava, K., Ghosh, A.K., Mori, Y., Ramanujam, S.: Model for cover cracking due to rebar corrosion in RC structures. Eng. Struct. (2006). https://doi.org/10.1016/j.engstruct.2005. 11.014 3. German, M., Pamin, J.: FEM simulations of cracking in RC beams due to corrosion progress. Arch. Civil Mech. Eng. 15(4), 1160–1172 (2015). https://doi.org/10.1016/j.acme.2014.12.010 4. Cao, C., Cheung, M.M.S., Chan, B.Y.B.: Modelling of interaction between corrosion-induced concrete cover crack and steel corrosion rate. Corros. Sci. 69, 97–109 (2013). https://doi.org/ 10.1016/j.corsci.2012.11.028 5. Chen, D., Mahadevan, S.: Chloride-induced reinforcement corrosion and concrete cracking simulation. Cement Concr. Compos. 30, 227–238 (2008). https://doi.org/10.1016/j.cemcon comp.2006.10.007 6. Du, Y.G., Chan, A.H.C., Clark, L.A.: Finite element analysis of the effects of radial expansion of corroded reinforcement. Comput. Struct. (2006). https://doi.org/10.1016/j.compstruc.2006. 02.012 7. Berto, L., Simioni, P., Saetta, A.: Numerical modelling of bond behaviour in RC structures affected by reinforcement corrosion. Eng. Struct. (2008). https://doi.org/10.1016/j.engstruct. 2007.08.003 8. Krykowski, T., Ja´sniok, T., Recha, F., Karolak, M.: A cracking model for reinforced concrete cover, taking account of the accumulation of corrosion products in the ITZ Layer, and including computational and experimental verification. Materials 13, 5375 (2020). https://doi.org/ 10.3390/ma13235375 ˇ 9. Cervenka, J., Papanikolaou, V.K.: Three dimensional combined fracture-plastic material model for concrete. Int. J. Plast (2008). https://doi.org/10.1016/j.ijplas.2008.01.004 ˇ ˇ 10. Cervenka, V., Jendele, L., Cervenka, J.: Atena Program Documentation, Part 1, Theory. ˇ CERVENKA CONSULTING, Prague (2016) 11. Menetrey, P., Willam, K.J.: Triaxial failure criterion for concrete and its generalization. ACI Struct. J. (1995). https://doi.org/10.14359/1132 12. ANSYS Inc.: Material Reference, Canonsburg, PA 15317 (2021) 13. Chen, W.F.: Constitutive Equations for Engineering Materials. Plasticity and Modeling, vol. 2. Elsevier Publishing, Amsterdam (1994) 14. Krykowski, T., Jasniok, T., Recha, F.: Modeling the time of concrete cover damage taking into account the uncertainty of model parameters. Ochrona Przed Korozj˛a, vol. 62 (2019). https://doi.org/10.15199/40.2019.3.1

Diagnostic and Design of Reconstruction of Building Váhostav Peter Koteš(B)

, Michal Zahuranec , and Martin Vavruš

University of Žilina, Univerzitna 8215/1, 010 26 Žilina, Slovakia [email protected]

Abstract. The paper deals with diagnostics and design of repairs and reconstruction of the multi-story administrative building Váhostav in Žilina, Slovakia. This is the first high-rise office building in Žilina, which was built in 1965. The height of the building from the ground is 48 m. The modifications should lead to a predominant change in the use of the building – it is planned to convert it into a residential complex. First, a visual and tactile diagnostics of the building was performed in order to verify the dimensions of the building and its individual elements and to determine their material characteristics with a focus on determining the quality of concrete and the quality and type of reinforcement. Subsequently, the building was recalculated to new loads according to Eurocodes. The recalculation showed that some elements were not suitable for the new loads and therefore needed to be strengthened – these were mainly vertical columns and some horizontal girders. The latest research activities of the Department Structures and Bridges, University of Žilina, focused on the use of fiber concrete for the reinforcement of vertical elements (columns), were also used in the strengthening. Keywords: Multi-story building · Reconstruction · Strengthening

1 Introduction Žilina is a town in the north-west part in Slovakia. One of the first high-rise buildings in the city was the Váhostav building (named after the Váhostav company), which was built in 1965. The building is used as office space. It is currently the third tallest building in Žilina. The building has fourteen floors and one underground floor and is rectangular in shape (ground plan) with dimensions of 25.74 m × 20.94 m. The height of building is about 48 m (Fig. 1). The building is located south of the center of Žilina by the main road Rajecká. The selected place is characterized by its position. From the north side it connects to solitary apartment buildings, from the west it connects to the Rajecká road, the southern part connects through the park part of the land to the communication junction – roundabout and to the east there is currently undeveloped land and grammar school building. Since the building no longer belongs to the Váhostav Company, the requirement is to redesign the building into a residential complex, increase the energy efficiency of the building, ensure a suitable technical solution for installations, and increase the comfort of using the building. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 165–174, 2023. https://doi.org/10.1007/978-3-031-26879-3_13

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Fig. 1. A view on the building Vahostav in Žilina.

2 Building Description – Existing State The load-bearing system (superstructure) is a reinforced concrete skeleton made of monolithic as well as prefabricated elements. On the first underground and first aboveground floor, there are monolithic elements as columns, girders and a floor slab, and their connection is considered rigid. On the remaining floors, the columns and girders are prefabricated, where the girders are continuous and the columns are connected using hinges (see Figs. 2 and 3). Slab floor panels and stiffeners are connected to the girders’ flanges also using hinges. The bracing of the building is provided by a circumferential stiffener which is prefabricated in one direction and connected to the girders and monolithic in the other direction and rigidly connected to the perpendicular girders. The overall bracing of the building is ensured by a monolithic core, which is continuous across all floors along the entire height of the building.

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Rez B-B'

B12

B12

B12

B01

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B01

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ROZMERY A POLOHA Z¡ KLADOV S⁄ PREDPOKLADAN…!!!

Fig. 3. A typical vertical section of the building with basic dimension.

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3 Building Diagnostics 3.1 General Description The diagnostic works were focused on a detailed inspection of the building and its basic dimensions, as well as on determining the quality of concrete and the reinforcement of Reinforced Concrete (RC) elements [1–3]. The main purpose of diagnostics is to determine the actual condition of the object and identify its possible failures [4–6]. 3.2 Failure Detection and Analysis A detailed inspection was carried out on the solved building as part of a diagnostic survey, where damaged “PZD” roof panels were discovered on the thirteenth floor. As a result of the damaged roof cladding occurs the leakage, thus the formation of mold, corrosion of the reinforcement and, in some places, the concrete cover layer spalling (see Fig. 4).

Fig. 4. The views of a roof panel failure due to leaks.

3.3 Evaluation of the Detailed Inspection During a detailed inspection of the building, no other defects of the main load-bearing members of the structure, such as columns, floor slabs, girders and walls, were visible. No moisture, corrosion or cracks were visible on these elements. The diagnostic survey and a detailed inspection of the building were associated with the measurement of the basic dimensions of the load-bearing elements and the position of the load-bearing walls. The diagnostics was performed during the month of March in 2021. All these obtained data serves as a basis for the development of project documentation of the current state of the building and subsequently for the development of project documentation of the reconstruction. The photo documentation of the construction was realized using the Xiaomi Redmi 4 device.

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3.4 Results of Diagnostics The concrete class C25/30 of the columns was determined by using non-destructive testing (NDT) – Schmidt hammer (see Fig. 5a). It is recommended to verify the results obtained by the Schmidt hammer with core drills, but in this case the building manager (administrator) was not allowed to drill into the elements due to their dimensions, therefore only the results from the Schmidt hammer were used. The dimensions of the precast columns are 500 × 500 mm and they are reinforced by reinforcement quality B 420B (old denotation 10425-V) of diameter Ø18 as mail longitudinal reinforcement and by reinforcement quality B 210B (old denotation 10216-E) of diameter Ø6 as stirrups. The position of the reinforcements was also verified using a Hilty scanner (Figs. 5b and 6).

Fig. 5. The diagnostics a) Schimdt hammer, b) position of reinforcement by Hilty Scanner.

Fig. 6. The position of the reinforcement in columns – vertical scan (stirrups).

The precast frame girders are inverted “T” shape with double-sided cantilevers (see Fig. 7). The concrete of frame girders C25/30 complies with the regulations of the Type document [7]. The main longitudinal reinforcement consists of the reinforcements Ø18 and Ø14 (B 490B) and the stirrups are Ø8 (10429-T).

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Fig. 7. The detail of placing the floor panels on the prefabricated girders of inverted “T” shape.

The dimensions of the precast stiffeners around building are 250 × 500 mm of rectangular cross-section. The concrete is class C25/30 and also complies with the regulations of the Type document [7]. The main longitudinal reinforcement consists of reinforcements of diameter Ø18 (B 420B) and the stirrups are Ø8 (10216-E). The slabs were made of prefabricated floor panels of type PZD 120/570 with dimensions 1190 × 240 × 5670 mm. These are standardized reinforced concrete lightweight floor slabs.

4 Load-Carrying Capacity Calculation For the global analysis of the structure, a numerical model was developed using members, slab and wall elements in the program Scia Engineer 19.1, which works on the principles of Finite Element Method theory (FEM) (Fig. 8). The supports of the reinforced concrete columns in the model were considered to be fixed to the foundation structure, as well as the stiffener walls of the building core and the underground wall around the building (line support, fixed). The core walls are not reinforced, they are only from unreinforced concrete. The inner columns and the wall of the core were interconnected, along the entire height of the building. The first underground floor and the first above-ground floor are monolithic, and therefore the girders were modeled as rib of slab, aligned to the lower surface. The remaining floors are prefabricated, where the transverse frame of the building is formed by beams with cantilevers and in the middle span there is intermediate girders connected with hinges. The joint between the girder and the column was considered as rigid. The prefabricated floor panel was modeled as a slab. They were connected to the girders by the hinges. A joint was created at the contact of the longitudinal stiffener of the building and girders. The longitudinal stiffener was modeled as a member and was connected to the cantilevers of the girders with loose rotation in the direction of the y-axis and z-axis.

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monolithic core of building (monolithic walls)

floor slabs from precast floor panels

prefabricated inverted Tshape girders

prefabricated part of building prefabricated columns

monolithic part of building

monolithic floor slab monolithic columns

Fig. 8. The numerical model of the building – view.

Fig. 9. The influence of seismicity in the direction of the Y axis, view of the whole building.

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The permanent loads and combinations of variable loads (imposed, wind and snow) were modeled as loads on structure. The structure was also verified for seismic loading. The effects of seismicity in the Y-axis direction achieved that 89.35% of the structure oscillated and the most unfavorable frequency for the structure was achieved in the first step. In the Fig. 9, it can be seen the behavior of the building on seismicity in the Y direction. From the results follows that the core walls do not satisfied for the seismic loads and have to be strengthened in the underground floor and first above-ground floor. The core walls can be strengthened by new reinforced concrete walls connected to the existing unreinforced concrete walls. Other possibility to strengthen them is also using CFRP lamellas [8–10]. The reinforced concrete columns in 13th floor also do not satisfy to the load – not due to seismicity, but due to Ultimate Limit States (ULS). In this case, two types of strengthening were advised – first type using the typical strengthening with new reinforced concrete layer of concrete class C40/50 and reinforcement of type B 500B (Fig. 10a). The second type of strengthening was using new layer of fiber concrete of concrete class C40/50 with steel fibers Dramix 3D (Fig. 10b). The possibility of using that type of strengthening was investigated in research works [11–13]. The advantage of this method of strengthening is that reinforcements do not have to be used (less labored on the construction site), it is possible to use thin layers of fiber concrete, and it is fast to apply. Another members that did not fulfil the new loads were the PZD floor panels on the top floor under the roof, due to the fact that the new roof is considered a “green roof”, so with a significantly higher load. In this case, reinforcement with CFRP lamellas CarboDur S (50 × 1.2) in the number of 2 per plate were proposed. (Fig. 11). By recalculation, it was proved that the prefabricated beams (inverted T-section) also suited the new load, and it was not necessary to reinforce them.

Fig. 10. The strengthening of columns a) with now reinforced concrete layer, b) with new layer of fiber concrete.

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Fig. 11. The strengthening of the floor panels PZD.

The project of reconstruction is still in the preparation period, so it is not clear whether the modifications and methods of reinforcement proposed by us will be used by the owner of the building, or whether there will be any changes to the project.

5 Conclusions The aim of the diagnostics and calculation was to verify the load-carrying capacity of the main load-bearing elements of the Váhostav high-rise building. The building will be used for new multifunctional use for offices and apartments in the city of Žilina. Due to the nature of the building under consideration, a global model was created in Scia Engineer 19.1, thanks to which the internal forces acting in the structural elements were obtained. All structural elements in the building were assessed by analytical calculation, or software Scia Engineer 19.1 or IdeaStaciCa 20.1. Unsatisfactory structural members in the building were strengthened with reinforced concrete, CFRP lamellas and an alternative experimental method using fiber concrete. After strengthening the members, it can be stated that all load-bearing elements meet the conditions of safety and reliability. Acknowledgements. This research was supported by Research Projects No. 1/0623/21 of the Slovak Grant Agency. This work was supported under the project of Operational Programme Interreg V-A Slovak Republic – Czech Republic: Assessment of the impact of the environmental load on the bridge structures condition of the cross-border transportation network, No. 304011Y277. The project is co-funding by European Regional Development Fund .

References 1. Zybura, A., Ja´sniok, M., Ja´sniok, T.: Diagnostyka konstrukcji z˙ elbetowych tom 2. Badania korozji zbrojenia i wła´sciwo´sci ochronnych betonu. Wydawnictwo Naukowe PWN, Warszawa (2011) 2. Rehacek, S., Citek, D., Citek, A., Krystov, M.: The quality control of the concrete of the supporting structure of bridge Reg. No. V-32 in Vrchlabi. In: 17th International Conference on Special Concrete and Composites, vol. 2322, p. 020022 (2021)

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3. Moravˇcík, M., Bujˇnáková, P. New precast bridge girder with combined prestressing. Commun. Sci. Lett. Univ. Zilina 13(3), 19–23 (2011) 4. Bacharz, K., Raczkiewicz, W., Bacharz, M., Grzmil, W.: Manufacturing errors of concrete cover as a reason of reinforcement corrosion in a precast element-case study. Coatings 9(11), 702 (2019) 5. Blikharskyy, Y., Selejdak, J., Kopiika, N. Specifics of corrosion processes in thermally strengthened rebar. Case Stud. Constr. Mater. 15, e00646 (2021) 6. Blikharskyy, Y., Selejdak, J., Kopiika, N., Vashkevych, R.: Study of concrete under combined action of aggressive environment and long-term loading. Materials 14(21), 6612 (2021) 7. PRIEMSTAV N.P. Bratislava - Project Department Enterprise, Type document (1963) 8. Brozda, K., Selejdak, J., Koteš, P.: The analysis of beam reinforced with FRP bars in bending. In: 12th International Scientific Conference of Young Scientists on Sustainable, Modern and Safe Transport. Procedia Engineering, vol. 192, pp. 64–68 (2017) ˇ 9. Koteš, P., Farbak, M., Kotula, P., Brodˇnan, M., Cavojcova, A.: Using CFRP lamellas for strengthening of dynamically loaded beams. In: Concrete and Concrete Structures 2013 – 6th International Conference. Procedia Engineering, vol. 65, pp. 302–310 (2013) 10. Kotula, P., Koteš, P., Brodnan, M.: Experimental and numerical analysis of anchorage zone of CFRP sheet. In: Concrete and Concrete Structures 2013 – 6th International Conference. Procedia Engineering, vol. 65, pp. 176–185 (2013) 11. Vavruš, M.: Use of high performance concrete to strengthening existing load-bearing elements. Ph.D. thesis, University of Žilina, p. 152, EDIS Žilina (2020) 12. Koteš, P., Vavruš, M., Jošt, J., Prokop, J.: Strengthening of concrete column by using the wrapper layer of fibre reinforced concrete. Materials 13(23), 5432 (2020) 13. Vavruš, M., Koteš, P.: Numerical comparison of concrete columns strengthened with layer of fiber concrete and reinforced concrete. In: 13th International Scientific Conference on Sustainable, Modern and Safe Transport (Transcom 2019), vol. 40, pp. 920–926 (2019)

Rheological Properties of Concrete Based on Waste Materials Mateusz Zakrzewski(B) , Artur Sanok, and Jacek Domski Faculty of Civil Engineering, Environmental and Geodetic Sciences, Koszalin University ´ of Technology, Sniadeckich 2, 75-453 Koszalin, Poland {mateusz.zakrzewski,artur.sanok,jacek.domski}@tu.koszalin.pl

Abstract. This report presents the results of laboratory research on the shrinkage and creep of concrete based on waste materials. Used waste aggregate was from the factory producing products from red ceramics. In total, five mixtures were tested. One without and four with the addition of fibers. Steel cord from recycled car tires and steel fibers in the amount of 0.5% and 1.0% by volume of fiberconcrete was used. The research was conducted for a period of one year. As a result of the measurements, concrete deformation values comparable to ordinary concrete with similar strength characteristics were obtained. The obtained results were compared with the results obtained by other scientists and with the values determined using selected calculation methods. For example, the addition of steel fibers increases the value of deformation due to concrete shrinkage by about 31% for both mixtures. Concrete creep values are lower by 12% also for both mixtures. The addition of steel cord increases the value of deformation due to concrete shrinkage by about 55% and 64% for the steel cord content of 0.5% and 1.0%. Concrete creep values are lower by about 1% and 4% respectively for the fiber content of 0.5% and 1.0% Individual results show differences of selected fibers on concrete parameters. In the conclusions, the authors of the report state that the red ceramic aggregate and the steel cord can be substitute of traditional materials to produce of steel fiber- reinforced concrete. Keywords: Creep · Shrinkage · Waste aggregate · Fibers · Steel cord · Fiber-reinforced concrete · Rheological properties

1 Introduction The study of rheological features such as shrinkage and creep plays an important role in the design of structures made of concrete based on waste materials. These features are not yet sufficiently tested to be able to design optimized structures. Knowledge of these features will allow for more accurate predictions of the behavior of the structure over time. The share of building materials using recycled waste will increase taking into account current trends. The use of waste as a substitute for aggregates of natural origin is currently a popular topic taken up by researchers from around the world. Currently, there is information that sand, like other natural resources, is too consumed [12]. Hence, it is reasonable to look for replacement materials. A logical and good solution for the © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 175–184, 2023. https://doi.org/10.1007/978-3-031-26879-3_14

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environment is to use crushed red ceramics as a replacement. There are many publications on the replacement of natural aggregate with crushed bricks derived from the demolition of buildings [2, 16, 17]. In this paper, another type of ceramic waste is used as a substitute, namely waste from the production of ceramic hollow blocks. The influence of various types of fibers on the properties of concretes is also an ongoing topic [4, 9, 13, 14]. In this paper, the fibers from the recycling of used car tires – steel cord – were used as dispersed reinforcement. As already proven, these fibers improve some properties of concrete [13]. Nevertheless, their influence on the rheological properties is not still sufficiently determined.

2 Materials and Methods As a replacement for natural aggregate, waste from a factory producing red ceramic blocks was used (Fig. 1). The waste was crushed in a jaw crusher and then sifted on sieves in accordance with standard [7]. Samples were made using the recommendations of standard [7] regarding the sieve curve “A” for standard concrete.

Fig. 1. a) Waste from a factory producing red ceramic blocks. b) Waste crushed in a jaw crusher prior to sieving.

Cement CEM I 42.5 R in the amount of 400 kg/m3 was used for the research. The plasticizer Sika S-600 was used, and w/c ratio was 0.45. As an addition, steel fibers with a slenderness of 50/0.8 (mixtures E0.5 and E1.0) and steel cord from recycled end-of-life car tires (mixtures K0.5 and K1.0) in the amount of 0.5% and 1.0% of the volume of fibre-reinforcement concrete were used (Fig. 2). Steel cord is obtained during recycling process of end-of-life car tires. These are steel fibers of irregular shape, length, and diameter (Table 1).The properties of the steel cord were determined by the authors during previous tests [10, 13]. The tensile strength was determined on the fibers with a diameter of 0.20–0.35 The value of tensile strength was finally estimated at 1788.53 MPa. The results obtained in the tests were compared with the results for concretes in which traditional aggregate was used and flat or similar steel fibers [3, 11]. Recipes of mixtures of these concretes are presented in Table 2. The research program was as follows. 12 cylindrical samples with a diameter of 150 mm and a height of 300 mm were made for each mixture. Three samples were

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Fig. 2. Fibers used in mixtures: a) hooked-end steel fibres 50/0.8, b) hooked-end steel fibers 30/0.55, c) steel cord. Table 1. Steel cord parameters. Parameter Diameter Length

Average (mm)

Standard deviation (mm)

Median (mm)

Minimum value (mm)

Maximum value(mm)

0.31

0.18

0.25

0.05

1.35

13.29

5.80

12.57

2.90

27.90

used to determine each of listed parameters: the compressive strength, the modulus of elasticity, shrinkage, and creep. The specimens were unmoulded at the age of 24 h and given to laboratory cellar with constant temperature of 20 ± 2 °C. The samples were tested at age of 28 days, except shrinkage samples. Fiber-reinforced concrete shrinkage measurements were carried out from the first day after the unmoulding of the samples (Fig. 3). All samples were incised to level the surface. The measurement base was 250 mm and 100 mm for the measurements of deformation caused by shrinkage and creep of concrete, respectively. The measurement was made with a extensometer with range 5 mm and resolution 0.001 mm. For the first two months, the deformation was measured once a week. In the following months, the interval between measurements was about 30–50 days (Fig. 4). The creep test specimens were placed in a loading frame. Three samples were placed in one frame which is in accordance with standards [1] and [6]. The loading frame is capable of applying and maintaining the required load on the group of specimens for duration of the test, irrespective of any change in the dimension of the specimens. On the basis of the average compressive strength, the level of long-term load was assumed. The standard [1] allows the load the specimens at an intensity of not more than 40% of the compressive strength at the age of loading. Due to the slight differences in load, a constant load value of 17 MPa was assumed for all mixtures.

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Mixtures Cement type

0 E0.5

CEM I 42,5 R

Cement Water Aggregate Fibers type (kg)

(dm3 ) (kg)

-

(%)

-

400

180

-

-

4.00 dm3

50/0.8

0.5

1,100

E1.0

1.0

K0.5 K1.0 B9 [3]

Fibers Plasticizer reinforcement and others ratio

B10 [3]

CEM-II/B-V 374 32,5R 378

JL-B [11]

CEM-II/AV 42,5R

420

Steel cord

1.0

0.5

150

1,835

50/0.8

0.42

3.47 dm3

140

1,835

30/0.55 0.43

3.51 dm3

160

1,570

50/0.8

Silica dust 21 kg

1.2

Fig. 3. Shrinkage samples with 250 mm measurement base (bright color).

Fig. 4. Concrete creep test. Specimens in loading frames.

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Shrinkage and creep tests were carried out in an air-conditioning chamber under constant thermal and humidity conditions. According to standards [5, 6] the air humidity should be 50% ± 4 and the air temperature 20 ± 2 °C [6]. Air conditions were measured every hour with a sensor. The measured values are shown in Fig. 5.

3 Results and Comparison with the Results of Other Scientists The following are the results of the compression strength test and the secant elastic modulus tests. The results of the compressive strength and elastic modulus test are shown in Table 3. Table 3 also includes the parameters of mixtures made by Domski [3] and Laskowska-Bury [11]. Table 3. Compressive strength and modulus of elasticitytest results. Mixture

Average compressive strength fcm

Standard deviation

Average modulus of elasticity E

Standard deviation

(MPa)

(MPa)

(GPa)

(GPa)

0

47.55

1.55

20.05

0.32

E0,5

48.91

0.69

20.55

0.11

E1,0

50.70

0.92

18.21

0.31

K0,5

49.40

1.91

18.74

1.83

K1,0

50.52

2.12

18.65

0.12

B9 [3]

31.03

2.55

29.31

0.94

B10 [3]

45.80

5.09

29.44

0.82

JL-B [11]

64.40

4.10

36.70

2.47

Figures 5, 6, 7, 8, 9, and 10 show the results of the research on rheological features. The shrinkage results were compared to the calculations made in accordance with the standards [8] and [15] (Figs. 5 and 6). The addition of steel fibers to the mixture with analyzed aggregate increases the deformation caused by shrinkage. For mixtures E0.5 and E1.0, shrinkage increased about 30%. The difference between samples with fiber reinforcement ratio 0.5 and 1.0 is very small. For K0.5 and K1.0 mixes with steel cord, there was an increase by 55.6% and 64.4%, respectively. The addition of fibers slightly reduces the creep of analyzed concrete. The creep value for all samples is close to the values calculated according to the standards [8, 15]. After 365 days, the greatest deviation from the standard curve (0.2%) was recorded for the sample without the addition of fibers. For the analyzed samples, the creep curves calculated on the basis of both standards are almost identical. Mixtures E0.5 and E0.1 were compared with the results obtained by Domski [3] for series B9 and with the results obtained by Laskowska-Bury [11] (Figs. 7 and 8). Identical steel fibers were used in all of these mixtures.

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0 E0,5 acc. to PN

1.60

K0,5 E1,0

K1,0 acc. to EC2

1.40 Shrinkage [‰]

1.20 1.00 0.80 0.60 0.40 0.20 0.00

0

50

100

150

200

250

300

350

400

Time [days] Fig. 5. Shrinkage deformations of tested mixtures compared with standards [8, 15].

0

K0,5

K1,0

E0,5

E1,0

acc. to EC2/PN

2.00 1.80 1.60 Creep [‰]

1.40 1.20 1.00 0.80 0.60 0.40 0.20 0.00

0

50

100

150

200

250

300

350

Time [days] Fig. 6. Creep deformations of tested mixtures compared with mixture B10 [3].

400

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1.60 1.40 1.20 Shinkage[‰]

1.00 0.80 0.60 0.40 E0,5 B9

0.20 0.00

0

50

100

150

200

250

E1,0 JL-B 300

350

400

Time [days] Fig. 7. Shrinkage deformations of mixtures E0.5 and E1.0 compared with mixture B9 [3] and JL-B [11].

1.80 1.60 1.40 Creep[‰]

1.20 1.00 0.80 0.60 0.40

0.00

JL-B E1,0

B9 E0,5

0.20 0

50

100

150

200

250

300

350

400

Time [days] Fig. 8. Creep deformations of mixtures E0.5 and E1.0 compared with mixture B9 [3] and JL-B [11].

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Samples made with waste red ceramic aggregate are characterized by much greater shrinkage and creep than concrete samples based on natural aggregate. At the end of the test, the analyzed samples achieved over twice bigger creep value then samples of B9 mixture and 58.8% bigger value the samples of JL-B mixture. Mixtures K0.5 and K1.0 with steel cord were compared to the results obtained by Domski [3] for series B10 with steel fibres with slenderness of 30/0.55 (Figs. 9 and 10).

1.60 1.40 Shrinkage[‰]

1.20 1.00 0.80 0.60 0.40 0.20 0.00

K0,5 0

50

100

150

200 250 Time [days]

K1,0 300

B10 350

400

Fig. 9. Shrinkage deformations of mixtures K0.5 and K1.0 compared with mixture B10 [3].

K0.5 and K1.0 mixtures showed significantly higher shrinkage and creep values than B10 mixture. After 365 days, shrinkage is 145% and 159% higher for the fiber reinforcement ratio 1.0% and 0.5% than for mixture B10. The creep coefficient was calculated in accordance with the procedure contained in EC2 [8] and on the basis of the obtained results using the following equation: ϕ=

εcc εce

(1)

where: ϕ – creep coefficient, εcc – strain of the specimen under the load at the time 365 days, εce – initial strain, measured immediately after the load was applied. The values of the creep coefficients are presented in Table 4. The influence of the fibers has a negative impact on the value of the coefficient calculated on the basis of the performed tests. The less fiber, the lower the creep coefficient. The values obtained depending on the fiber-reinforcement ratio are very similar for both tested types of fibers. Creep coefficient calculated in according to EC is very similar for mixtures 0; E0.5; E1.0; K0.5; K1.0; B10. The tested mixtures are characterized by a lower coefficient than a mixture B10. Much better values creep of the coefficient were obtained in the research by Laskowska-Bury [11] for the mixture JL-B.

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2.00 1.80 1.60 Creep [‰]

1.40 1.20 1.00 0.80 0.60 0.40 0.20 0.00

K0,5 0

50

100

150

200 250 Time [days]

K1,0 300

B10 350

400

Fig. 10. Shrinkage deformations of mixtures E0.5 and K1.0 compared with mixture B10 [11]. Table 4. Creep coefficient. Mixture

εcc [‰]

εce [‰]

Tests

Acc. to EC

0

1.84

1.37

1.34

1.98

E0,5

1.67

1.07

1.57

1.92

E1,0

1.65

0.93

1.78

1.85

K0,5

1.74

1.06

1.64

1.89

K1,0

1.82

1.00

1.82

1.89

B9 [3]

0.43

0.36

1.19

2.63

B10 [3]

0.63

0.48

1.31

1.89

JL-B [11]

0.17

0.25

0.68

1.32

4 Conclusions The results obtained during the research and the analyzes carried out led to the following conclusions: 1) By using aggregate obtained from waste from the production of red ceramic blocks, you can obtain concrete with strength parameters similar to traditional concrete. 2) Creep deformations of tested mixtures are close to the values obtained using the calculation procedure included in the EC2 standard. 3) The disadvantage of aggregate-based concrete is the increased deformation caused by shrinkage. Guidelines for increased concrete care should be developed and additives reducing rheological deformations should be applied.

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4) The rheological properties of the tested mixtures depend mainly on the used aggregate. The influence of the fiber type is much less important. 5) Shrinkage and creep deformations have not stabilized yet. Continuation of the research is necessary. The tested mixtures require further research and analysis in order to adapt the calculation procedures to their unusual properties.

References 1. ASTM C512-02 Standard Test Method for Creep of Concrete in Compression 2. Awoyera, P.O., Ndambuki, J.M., Akinmusuru, J.O., Omole, D.O.: Characterization of ceramic waste aggregate concrete. HBRC J. 14(3), 282–287 (2018). https://doi.org/10.1016/j.hbrcj. 2016.11.003 3. Domski, J.: No´sno´sc´ , ugi˛ecie i zarysowanie belek piaskobetonowych z włóknami stalowymi pod obci˛az˙ eniem dora´znym. Dissertation (2005) 4. Domski, J., LaskowskaBury, J., Zakrzewski, M.: Bending moment of the waste fine aggregate concrete beams. ROS 21(2), 1505–1514 (2020) 5. EN 12390-16: Testing hardened concrete. Determination of the shrinkage of concrete (2020) 6. EN 12390-17: Testing hardened concrete. Determination of creep of concrete in compression (2020) 7. EN 1766: Products and systems for the protection and repair of concrete structures. Test methods. Reference concretes for testing (2017) 8. EN 1992-1-1 Eurocode 2: Design of concrete structures - Part 1–1: General rules and rules for buildings 9. Głodkowska, W., Ziarkiewicz, M.: Estimation of load bearing capacity of bending fibrocom´ posite elements. Rocznik Ochrona Srodowiska. 21(1), 294–315 (2019) 10. Katzer, J., Domski, J., Zakrzewski, M., Ponikniewski, T.: Comparison of the mechanical characteristics of engineered and waste steel fiber used as reinforcement for concrete. J. Clean. Prod. 158, 18–281 (2017). https://doi.org/10.1016/j.jclepro.2017.04.165 11. Laskowska-Bury, J.: Wybrane cechy fizyko-mechaniczne fibrokompozytu wytworzonego na bazie kruszywa odpadowego. Dissertation (2017) 12. Marschke, M., Rousseau J.-F.: Sand ecologies, livelihoods and governance in Asia: a systematic scoping review. Resour. Policy. 77, 102671 (2022). https://doi.org/10.1016/j.resourpol. 2022.102671 13. Paj˛ak, M., Krystek, M., Zakrzewski, M., Domski, J.: Laboratory investigation and numerical modelling of concrete reinforced with recycled fibers. Materials 14(8), 1828 (2021). https:// doi.org/10.3390/ma14081828 14. Paj˛ak, M., Wandzik, G.: Laboratory tests of concrete beams reinforced with recycled steel fibres and steel bars. Materials 14(22), 6752 (2022). https://doi.org/10.3390/ma14226752 15. PN-B-03264 Konstrukcje betonowe, z˙ elbetowe i spr˛ez˙ one. Obliczenia statyczne i projektowanie. PKN (2002) 16. Vieira, T., Alves, A., de Brito, J., Correia, J.R., Silva, R.V.: Durability-related performance of concrete containing fine recycled aggregates from crushed bricks and sanitary ware. Mater. Des. 90, 767–776 (2016). https://doi.org/10.1016/j.matdes.2015.11.023 17. Zhao, Y., Gao, J., Chen, F., Liu, C., Chen, X.: Utilization of waste clay bricks as coarse and fine aggregates for the preparation of lightweight aggregate concrete. J. Clean. Prod. 201, 706–715 (2018). https://doi.org/10.1016/j.jclepro.2018.08.103

Effect of the Addition of Waste Fibers on Some Properties of Concrete Artur Sanok(B) , Mateusz Zakrzewski, Marek Lehmann, and Jacek Domski Faculty of Civil Engineering, Environmental and Geodetic Sciences, Koszalin University ´ of Technology, Sniadeckich 2, 75-453 Koszalin, Poland {artur.sanok,mateusz.zakrzewski,marek.lehmann, jacek.domski}@tu.koszalin.pl

Abstract. This paper presents the results of laboratory research on effect of the addition different amount of waste fibers on some properties of concrete. Used waste fibers are produced during recycling process of the end-of-life car tires. Those fibers are called steel cord. Standard concrete with different amounts of waste fibers was tested. The applied amounts of fiber reinforcement were accordingly 0.5%, 0.75%, 1.0%, 1.25%, and 1.5%. In the tests, compressive strength, limit of proportionality, and residual flexural strength were determined. The compressive strength test proved that the addition of fibers increases the strength from 13% (for 0.5% fiber value) to 21% (for 1.5% fiber value). Residual research has shown that even the smallest value of fibers significantly changes the strength properties of concrete. The results showed slight differences between the fiber content from 0.5% to 1.0%. The samples with 1.25% of fibers showed higher rest strength, however, there was no increase in forces after the appearance of the crack. The samples with the highest amount of fibers showed only a temporary strengthening after the appearance of cracks. Obtained results shows that steel cord can be used as fiber-reinforcement in concrete. Keywords: Waste fibers · Steel cord · Fiber-reinforced concrete · Residual flexural strength · CMOD

1 Introduction Recycling is an important aspect nowadays in every area of life and ecology – also in construction [2, 14]. Economic development and demographic growth in the world contributed to the increased demand for building structures, also for concrete structures. Concrete is a widely used material that has no substitute [6]. Its production consumes significant amounts of natural resources [18] and emits carbon dioxide [3]. In order to prevent the consumption of natural materials, researchers are looking for various materials that can replace them, including construction waste [16, 27, 29] and ceramic waste [15].

© The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 185–196, 2023. https://doi.org/10.1007/978-3-031-26879-3_15

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An unfavorable feature of ordinary concrete is a relatively low tensile strength value, therefore many scientists use various admixtures and additives to improve its properties. The basic solution is the use of steel fibers as dispersed reinforcement of concrete [4, 7, 12, 19, 24]. Steel fibers come in various lengths and shapes. These fibers can be replaced with their waste counterparts, one of them is the so-called steel cord. The cord is one of the materials obtained during the disposal of end-of-life car tires. This material is heterogeneous, and individual fibers have different lengths, diameters, and shapes [8]. This randomness makes the fibers better adapt to the aggregate in the concrete mix and may show better mechanical parameters in concrete than many different commercial fibers [28]. Research conducted by many scientists [1, 5, 8, 13, 20–23, 25, 26] prove the possibility of using steel cord in concrete mix. The research on this material is not finished, however, it is necessary to continue with the more effective use of this material in construction. When talking about structural elements made of fiber-reinforced concrete, one should bear in mind the way they are designed. For civil engineers, the basic information needed to design a reinforced concrete construction is the compressive and tensile stresses of the material in post-cracking phase. Compressive strength testing is not very complicated and is described in international standards but in the last 20 years, several proposals have been made to describe the behavior of post-cracked fiber-reinforced concrete in tension. The most popular way to define this property is included in Model Code 2010 [11]. This method consists in measuring the ratio between the crack width (CMOD) and the force applied to the tested element. The result of the test carried out with this method is the load-CMOD relationship, from which the so-called residual flexural tensile strengths are determined. This paper investigates the effect of waste steel fibers on the tensile and compressive strength of concrete. These studies are part of a broader study by the authors.

2 Materials and Methods To make samples of the reference concrete, the guidelines of the [9] standard were used. Aggregate was sifted on sieves to obtain the “A” screening curve. Cement CEM I 42.5 R in the amount of 350 kg/m3 was used for the research, the plasticizer Sika S-600 was used in the amount of 1% of the cement mass, and the assumed c/w ratio was 0.55 (Table 1). Table 1. The amount of individual aggregate fractions in the tested mixtures. Amount of aggregate [%]

Sieve maximum size [mm] 0.25

0.50

1.00

2.00

4.00

8.00

Amount of fraction

10

15

12

15

20

28

Amount that passed through the sieve

10

25

37

52

72

100

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Fig. 1. Waste fibers – steel cord.

Steel cord, was used as fiber reinforcement (Fig. 1). The steel cord is obtained during the end-of-life car tire recycling process. This process involves mechanical separation of the rubber from the steel wires in the tyres. The product of this treatment is steel cord, rubber granulate and other materials. The amount of steel obtained is from 15% to 25%. This amount depends on the type of tires subjected to the recycling process [17]. The obtained steel cord varies in terms of the shape, length, and diameter of the fibers. The properties of the steel cord were determined by the authors during previous tests [8]. The value of tensile strength was finally estimated at 1,788.53 MPa (Table 2). Table 2. Steel cord parameters. Parameter Diameter Length

Average [mm]

Standard deviation [mm]

Median [mm]

Minimum value [mm]

Maximum value [mm]

0.31

0.18

0.25

0.05

1.35

13.29

5.80

12.57

2.90

27.90

Figure 2 shows the number of results with in the slenderness ranges for the test sample. An important parameter of fibers is their slenderness. Due to the different dimensions of the recycled fibers, a random sample of fibers was tested.

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35 Number of results

30 25 20

25 19

17

15 10

8

5

1

1

2

0 20-100 101-180 181-260 261-340 341-420 421-500 501-580 581-660 Slenderness of steel cord

Fig. 2. Number of fibers in selected slenderness ranges.

The research program was as follows. 6 cubic samples (150 mm × 150 mm × 150 mm) and 3 beams (150 mm × 150 mm × 600 mm) were made for each mixture. Compressive strength tests were performed on cubic samples, and the bending strength tests were performed on the prepared beams. These tests were also used to determine the residual bending tensile strength and the limit of proportionality (LOP). 24 h after preparing, the samples were demolded and stored in an air-conditioning chamber at a constant temperature of 20 ± 2 °C and humidity of 50 ± 5% until the day of testing. The tests were carried out on samples aged 28 days.

Fig. 3. Static scheme of the residual strength tests.

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The samples subjected to the residual bending strength test were prepared in accordance with the guidelines of [10]. The geometries of the beam specimen are shown in Fig. 3. The residual bending tensile strength test was performed as a three-point bending test. A variable speed load was applied in the middle of the span of the beam. Initially, the speed of applying the load was 0.5 mm/min, and after obtaining the value of CMOD = 0.1 mm, the value of the speed of applying the force decreased to 0.2 mm/min. The study was completed at CMOD = 4.0 mm. Crack width, deflection and force values were measured using a sensor system (Fig. 3). According to [10], the residual flexural tensile strengths are determined from the load-CMOD relationship (Fig. 4). 4 residual strengths (fR.1 –fR.4 ) are define for 4 CMOD values as shown in Fig. 4 (CMOD = 0.5 mm; 1.5 mm; 2.5 mm; 3.5 mm). The second parameter that has been determined is the proportionality limit (fL ), which depends on the load FL . The value of the residual bending tensile strength and the proportionality limit according to [10] are determined on the basis of the following formulas: fR.i =

3 · FR.i · l 3 · FL · l , fL = 2 · b · h2sp 2 · b · h2sp

(1)

where: hsp – the distance between the notch tip and the top of the specimen, FR.i – load corresponding to CMODi , FL – load corresponding to the limit of proportionality, l – span of the beam, b – width of the specimen. The FL value is determined on the basis of the load-CMOD graph, it is the highest load value in the range from 0 to 0.05 of the CMOD value.

20

Load [kN]

F fL FR,1L

fR,1 fR,2

FR,2

fR,3

FR,3

fR,4

FR,4

0

0,05

0.0

1.0

2.0 CMOD [mm]

3.0

4.0

Fig. 4. Method for determining the bending residual strength based on the force-CMOD relationship.

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3 Tests Results This chapter presents the results of laboratory tests. Compressive strength test results are presented in Table 3. The samples show a relatively small scatter of results. Already a small amount of fibers 0.5%, improves the compressive strength of concrete. The increase in strength was noted for each subsequent mixture, except for the mixture with fibers in the amount of 1.25%. Nevertheless, the increase for this mixture over the nofiber concrete was 14%. The increase in compressive strength may result from the fact that steel fibers may, to some extent, obscure coarse aggregate in the tested material, which was also confirmed by other authors [7, 12, 19]. Table 3. Compressive strength of tested mixtures. Mixture

Average compressive strength fcm [MPa]

Standard deviation [MPa]

Coefficient of variation [%]

0.0%

35.08

1.80

5.13

0.5%

39.88

0.62

1.56

0.75%

40.29

0.60

1.49

1.0%

41.48

1.08

2.60

1.25%

40.08

1.16

2.90

1.5%

42.79

1.00

2.34

Table 4 shows the results of residual flexural tensile strength. Figure 5 shows a set of curves for all mixes. One curve was selected for each of them. Table 4. Residual flexural tensile strength of tested mixtures. Mixture

Residual flexural tensile strength [MPa] fR.1

fR.2

fR.3

fR.4

0.5%

2.97

1.79

1.52

1.36

0.75%

2.94

2.17

1.93

1.65

1.0%

3.55

2.37

2.09

1.94

1.25%

3.97

3.43

3.03

2.72

1.5%

5.13

4.60

4.13

3.67

Load [kN]

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20 18 16 14 12 10 8 6 4 2 0

191

1,5% 1,25% 1,0% 0,75% 0,5%

0.0

1.0

2.0 CMOD [mm]

3.0

4.0

Fig. 5. Selected load – CMOD diagrams of tested mixtures.

The lowest values of flexural residual strengths were determined for the 0.5% mixture. As the amount of fibers in the mixture increases, the strength values increase but the increase is uneven. The course of the curves for the mixtures 0.5%, 0.75%, and 1.00% is very similar. Also the average values of residual strengths of the mixtures 0.75% and 1.0% are similar. A clear increase in value of flexural residual strengths and a higher course of the curve were recorded for the mixtures 1.25% and 1.50%. The highest values of flexural residual strengths were determined for the 1.5% mixture. Figure 6 shows that all the mixtures were characterized by a slight discrepancy in the diagrams of curves. The fib Model Code 2010 [11] recommendation allows to classify fiber- reinforced concrete based on the knowledge of residual strengths. The grade of fiber-reinforced concrete is a numeral-letter symbol. The number corresponds to the characteristic residual strength fR1 and is taken from the range 1 to 8. The letter is determined on the basis of the ratio of the two residual strengths fR3 /fR1 . The letters ‘a’ and ‘b’ for which fR3 /fR1 < 0.9 characterize material with post crack softening, the letters ‘d’ and ‘e’ where fR3 /fR1 > 1.1 classify the material with post crack hardening. The letter ‘c’ characterizes a composite with an average constant residual strength after cracking. It should be noted that the fib Model Code [11] imposes minimum requirements for residual strength in the event that a given fiber reinforced concrete would partially or completely replace traditional reinforcement in the ultimate limit state, these are: fR1 k/fLOP > 0.4 and fR3 /fR1 > 0, 5. By analyzing the results from Table 5, it can be concluded that even a small content of fibers (0.5%) allows to meet the requirements of replacing traditional reinforcement with fibers. In addition the fiber content of 0.5% to 1.0% allows the material to be classified as ‘a’ and with a content of 1.25 to 1.5% as ‘b’.

Load [kN]

A. Sanok et al.

20 18 16 14 12 10 8 6 4 2 0

A0% B0% C0%

Load [kN]

0.0

0.2

20 18 16 14 12 10 8 6 4 2 0

0.4 0.6 CMOD [mm]

0.8

A0.5% B0.5% C0.5%

0.0

Load [kN]

192

1.0

20 18 16 14 12 10 8 6 4 2 0

2.0 3.0 CMOD [mm]

4.0

A0.75% B0.75% C0.75%

0.0

1.0

2.0 3.0 CMOD [mm]

Fig. 6. Load-CMOD curves of tested mixtures.

4.0

Effect of the Addition of Waste Fibers on Some Properties of Concrete

Load [kN]

20 18 16 14 12 10 8 6 4 2 0

A1.0% B1.0% C1.0%

Load [kN]

0.0

1.0

2.0 3.0 CMOD [mm]

20 18 16 14 12 10 8 6 4 2 0

A1.25% B1.25% C1.25%

0.0

Load [kN]

4.0

1.0

2.0 3.0 CMOD [mm]

20 18 16 14 12 10 8 6 4 2 0

4.0

A1.5% B1.5% C1.5%

0.0

1.0

2.0 3.0 CMOD [mm]

Fig. 6. (continued)

4.0

193

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Mixture

Limit of proportionality [MPa] fLOP

fR.3 /fR.1

fR.1 /fLOP

0.0%

4.03

–

–

0.5%

4.31

0.51

0.69

0.75%

4.33

0.66

0.68

1.0%

5.03

0.59

0.71

1.25%

4.47

0.76

0.89

1.5%

4.85

0.81

1.06

4 Conclusions The paper presents the results of the basic mechanical properties of concrete reinforced with steel cord in various amounts. The compressive strength test proved that the addition of fibers increases the strength from 13% (for 0.5% fiber value) to 21% (for 1.5% fiber value). The residual strength test was also performed. Low fiber content significantly changes the tensile strength properties of the material under test. Despite the fact that post crack softening was found for all variants of fiber content, it should be emphasized that the addition of fibers in the amount of 0.5% allows the material to be classified as substitutes for conventional reinforced concrete. Finally, class 2a, 3a, 3b and 5b fiber reinforced concrete were obtained, according to the fib Model Code 2010. It should be noted that the use of fiber reinforcement in the form of steel cord is an important ecological aspect. The possibility of using steel cord for the production of construction material would largely solve the problem of managing post-recycling products.

References 1. Aiello, M.A., Leuzzi, F., Centonze, G., Maffezzoli, A.: Use of steel fibres recovered from waste tyres as reinforcement in concrete: pull-out behaviour, compressive and flexural strength. Waste Manage. 29(6), 1960–1970 (2009). https://doi.org/10.1016/j.wasman.2008.12.002 2. Babiak, M., Błaszczy´nski, T., Ratajczak, A., W˛egli´nski, S.: Przydatno´sc´ kruszyw z recyclingu do produkcji betonu. Przegl˛ad Budowlany 10, 60–63 (2017) 3. Błaszczy´nski, T., Król, M.: Produkcja betonu, a problem redukcji emisji dwutlenku w˛egla. Izolacje, 3 (2014) 4. Brandt, A.M.: Fibre reinforced cement-based (FRC) composites after over 40 years of development in building and civil engineering. Comp. Struct. 86(1–3), 3–9 (2008). https://doi.org/ 10.1016/j.compstruct.2008.03.006 5. Caggiano, A., Folino, P., Lima, C., Martinell, E., Pepe, M.: On the mechanical response of hybrid fiber reinforced concrete with recycled and industrial steel fibers. Constr. Build. Mater. 147, 286–295 (2017). https://doi.org/10.1016/j.conbuildmat.2017.04.160 6. Czarnecki, L., Kurdowski, W.: Tendencje kształtuj˛ace przyszło´sc´ betonu. Budownictwo, Technol. Archit. 1, 50–55 (2007)

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Architecture and Urban Planning

Ecological Aspect of Water Retention Through Algae Based Hydraulic Systems in Interactive Architecture Anna Grajper-Dobiesz1(B) and Sebastian Dobiesz2 1 Faculty of Civil Engineering and Architecture, Department of Architecture and Urban

Planning, Opole University of Technology, Katowicka 48, 45-061 Opole, Poland [email protected] 2 LAX Laboratory for Architectural Experiment, Ul. Pełczy´nska 4/324, 50-950 Wrocław, Poland

Abstract. This paper discusses the potential usage of algae in designing interactive architecture devoted to dealing with climate change in the cities. It highlights its diverse functionality through a set of industrial applications in architecture and design experiments made by various artists and researchers across the globe. It specifically focuses on the design proposal by Laboratory for Architectural Experiments investigating the ways of merging algae into a water management system dedicated for green areas in the cities. This paper demonstrates the potential of architectural form that can both influence the shape of the green areas of future cities as well as it could improve their resilience to climate fluctuations in an environmentally friendly manner. Ongoing research projects will give an indication of further research questions. Keywords: Architecture · Algae · Rainwater

1 Introduction The present contribution reviews a set of design studies based on examples which use algae as the main actor of its operation. The scale of the designed elements vary, from facades or rooftops, to small elements of architecture entwined across civic space. It shows growing interest in harnessing algae power and characteristics to shape our urban environment. Moreover, this trend is particularly visible in interactive architectural (iA) and has strongly been evaluated in recent years by moving projects towards solutions, dependent less on human oriented variables, but more on environmental ones [5]. Thus, design studies presented in this article are based on examples which use algae – and its natural responses to physical and chemical traits of the surroundings –as the main actor of the interactive system. Interactive architecture in its basic assumptions is “characterized by constant reconfiguration as part of the interaction between physical elements and a human being in real time, through an applied interactive system, actuators, and designed scenarios of interaction” [3]. Currently the projects of iA are still based primarily on the cybernetic © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 199–210, 2023. https://doi.org/10.1007/978-3-031-26879-3_16

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assumptions from 1960s and are using the technical innovations [2] to build the human oriented environment. Yet there is a branch to those speculative designs that are going to the direction of a new form of system which derive its interactiveness mainly from natural processes [14]. One of the fundamental path of naturalistic connections between iA and processes observed in environment was developed by architect and researcher Philip Beesley with his manifesto of the livingness of architecture, as the circumstances which create an environment with dynamic, open, and evolving system of language. His series of “The Hylozoic” projects that started in 2010 at Venice Architecture Biennale, tried to find the solution to the new kind of architecture based on evolving systems like in synthetic organic environments [2]. As the author emphasise, “the new paradigms are implied where humanity is seen as a participant in the complex negotiations between nature, culture and technology” [2]. This direction is also visible in the book of Rachel Amstrong where the author states: “calls for an unmediated connection between the two fields, one in which architecture literally behaves like nature; that is, in which buildings will be able to grow, adapt and mutate just like plants do” [12]. These studies of implementation the systems based on nature processes, have been widely developed towards the use of specific elements of nature such as algae as the most oldest organisms on Earth which can be bred in all aquatic conditions [6]. Thus, the algae based interactivity can fulfil the attitudes toward designing future forms of photosynthetic interactive architecture turning the whole buildings and architecture elements into more like “the living organism with its intelligent ability to sense the surrounding environment and adapt to it” [6].

2 Research Aim and Objectives Purpose of the study is to develop solutions and possibilities from algae as a main material of iA which create the interactive system based on nature process. For this purpose, the study focuses on analysis of realized projects and designed conceptual case study based on the collected guidelines from analyzed examples. The following research questions were raised in the study: a. Can algae be harnessed in the efforts to increase the resilience of cities to climate change by improving the water retention systems? b. Does the system of algae implicate into the shape of designed elements?

3 Methodology The established methodology is divided into two main phases: Phase 1: The theoretical study searching for main characteristics and design guidelines of iA using algae. The phase consists of following steps: • Exploring the identification of algae as the main material of interactive system. • Exploring two layers of functionality: • Wetware – selection and management of microalgae cultures.

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• Hardware – Materials and systems for algae growth environment. • Exploring the possibilities of form and interactions based on algae. Phase 2: The case study of speculative design “Aureola” realised by LAX Laboratory for Architectural Experiments as the solutions for cities in water retention and an increase in biodiversity of existing green squares. Aureola is the method of experimenting with a directional setup yet without strict assumptions, what can be achieved. The design was constructed without a predefined knowledge of where the process can lead the project. The next stage would be a prototype of the structure that could test proposed solutions. Such methodology is well established in the field of iA, specifically dealing with algae hydrosystems such as EcoLogisStudio [6].

4 Micro-algae Purifying Water Treatment 4.1 Biological and Technical Aspects The application of microalgae as a form of treatment for water purification was investigated not only by architecture research groups designing conceptual projects, but it mainly gained attention from the wide scientific body [17]. As the sources say, the systems of water purification based on microalgae have been showed since the 1960’s, but till now it is still under investigating, struggling with many aspects including its effectiveness and pace that require up to 11 days for algae to clean a desired amount of water [1]. The research topics discussed, concern a wide spectrum of algae properties focusing on their resilience to grow depending of types of water, processing the biomass and economic profitability of the whole system. The algae system of water treatment need the following aspect to be taken under consideration during design process to make it feasible [1]: 1. The contamination of water which allow to estimate the demand of organic matter (COD), nitrogen, phosphorus, and CO2 or pollutants. 2. Setting up the proper pH of the hydrogel because the high pH has a negative effect onto growth of the bacteria and microalgae. According to researchers from the University of Almería: “The composition of wastewater varies mainly as a function of location and the predominant activities in the surrounding area (agriculture, industry, farms, etc.). Moreover, inside the wastewater treatment plant, three different types of wastewater are also identified: (i) after primary treatment when the solids and fats are removed, (ii) after secondary treatment once most of the organic matter is removed, and (iii) the centrate from anaerobic digestion, which contains a high contaminant concentration.” In the case of microalgae water treatment, all those processes are performed at once [1]. There is however a range of other applications of algae in built urban environment. The very recent project in Hamburg investigated the potential of using algae in enclosed façade system as a shading that can also produce electricity for the inhabitants of the

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building. It was proven by implementation of such system, that algae can efficiently grow in an enclosed environment [7]. All the achievements made on the academic and industrial ground referring to algae’s possible application has started also a range of speculative projects in the field of design and architecture (especially Interactive Architecture). Designers such as EcoLogisStudio are testing the potential of algae in air purification in the range of their projects such as H.O.R.T.U.S. XL Astaxanthin.g[20], or AirBubble [21]. In those projects, the authors were trying to explore the right formal structure for efficient algae growth. Despite those attributes, like water treatment, energy production, and air purification, that are widely exploited in the projects worldwide, there is also a crucial characteristics of algae as an efficient fertilizer in gardening industry. Algae biomass can increase the biodiversity of the treated green areas by producing plant growth hormones and biocide, in an environmentally friendly processes [13]. Yet such characteristic have not been implemented in architectural designs to date. In this research paper, there are two project being investigated, that are referring to those less widely exploited features of algae. There are projects investigating the use of algae in water treatment to be used by humans (project 01) or to be used for green areas, and by that, increasing small retention in the cities (project 02). 4.2 Speculative Design Project 01: The Indus Authors: Bio I-D LAB, the research let by Dr Brenda Parker, the professor Marcos Cruz, and Shneel Malik, from UCL. Location: Rural areas with limited resources. Type of pollution: Heavy metals, e.g. cadmium. Form of the algae system: A seaweed-based hydrogel. Main role: Cleaning water. The designed and constructed wall is a response to the local problem of rural areas in India which struggling with the problem of polluted water caused by manufacturing of textiles or jewellery – as the statistic shows almost up to 80% of India’s surface and ground waters are polluted. Construction of the wall is based on ceramic tiles design in parametric process which was used to achieve the appearance of a leaf. The micro structure of tiles consists of vein like channels. These micro-gaps are filled with a seaweed-based hydrogel. The hydrogel is the main material in “Indus”, which mimics nature’s ability to distribute water, is composed of modular tile units with vein like channels and inspired by the structure of leaves. The tiles contain a preparation of micro-algae suspended with a biological scaff old of a seaweed-derived hydrogel. This enables to keep the algae alive and also makes it recyclable and biodegradable. The steps of operation of the water purification system: Step 1: “Users are required to simply pour the heavy metal polluted water into the inlets present at the top of the wall”. Step 2: “The water then enters the system through the branched channels running onto the algae-laden hydrogel allowing for heavy metal uptake”.

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Step 3: “Water can be recirculated through the wall for further treatment, depending upon the level of contamination” [11] (Fig. 1).

Fig. 1. The Indus project is a modular bioreactor system composed of parametric wall tiles, cleaning water through the process of bioremediation [18].

Analysis layers of functionality: Wetware: A seaweed-based hydrogel spread in channels. Hardware: The system of tiles with channels. Positive effect: The modularity of the wall gives the ability to be efficient and cost reduced system which could be built on-site areas with contaminated water sources. Moreover, the system is expected to be adaptable to various countries with similar problems at rural communities [4]. Speculative observations: The project does not consider the durability of the whole system, especially the hydrogel. Moreover the concentrations of heavy metals such as cooper can impede photosynthesis and modify the microalgae cell walls [1]. 4.3 The Case Study Project 02: Aureola Authors: LAX Laboratory of Architectural Experiments, Anna Grajper, Sebastian Dobiesz. Location: Green squares in cities (on the example of Wrocław). Type of pollution: Phosphorous, heavy metals, e.g. cadmium. Form of the algae system: A seaweed-based hydrogel (a nutrient liquid for algae growth). Main role: Water retention, water cleaning, plants fertilizing.

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This project is a conceptual case study devoted to the problem of small retention in green urban areas. The perceived problem is the situation in which there is an unfavourable difference in terrain between the existing paved areas and green, biologically absorbable, and permeable areas, that does not allow for surface runoff from paved areas to green areas. It is a common situation in cities that this situation is conducive to accelerated surface runoff contributing to the drainage of cities and the reduction of their resistance to climatic changes, illustrated on the example of selected green squares in the city of Wrocław. It is nowadays a common trend to design new green areas in the cities as a prospective water storage in a form of rain-gardens that can accumulate stormwater from the streets. To achieve such functionality they need to be designed below the street level so the water can easily flow into its basins [15]. They are therefore fitted with special plants that can bear the water level fluctuations and neutralize the water pollution [16]. However, cities have plenty of green areas that are existing and are designed above the level of neighbouring streets and consists of vulnerable old trees, thus cannot be redesigned to meet the requirements of rain-gardens. Thus, if such areas are about to retain any stormwater from the streets, they require an additional system that can leverage the waterflow and increase the water quality before it can be used by plants. The experimental installation “Aureola” by LAX, is the example of such system. The setup of “Aureola” presented in this paper is a hydraulic system in the form of a freeform bench surrounding trees in green areas. The project involves the use of algae in a passive hydraulic conditions where algae convert pollutants contained in rainwater to a clean water, which can then be reused for irrigation and fertilization of the soil near plants. The project exploits a feature of algae requiring phosphorus to be grown and the fact that stormwater from the city streets has been proven to contain the large amount of it [8]. The principle of operation of the system is based on the pressure and the capillary rise commonly found in nature [9]. It has a form of anamorphic translucent acrylic containers with the porous structure running on the inside, filled with algae hydrogel. The whole vessel rests on pairs of antagonistic supports, inside of which there is an exchange of water between the inside of the vessel and the environment. The antagonism of the system relies on the fact that one of the supports in the pair is shaped in a way that increases the capillary pressure allowing for the uptake of excess water from the streets. The operation of the system is also based on the algal growth through photosynthesis when the system is exposed to solar radiation with the help of carbon dioxide from the atmosphere. The growth of algae increases their volume in the tank, allowing the removal of their excess through the second pairs of antagonistic supports. The properties of algae matter in fertilizing the soil are used both as an element providing plants with the necessary micronutrients, but also as a source of water accumulated in the structures of their cells. Thus, this system allows for the retention of water and the simultaneous improvement of its soil fertility properties for use in dry landscapes correlated with periods of greater exposure to solar radiation (Fig. 2). “Aureola” can work in a synergy with a macro-pervious (MP) pavement systems, that are one of the most effective water collectors, which can be installed under the pavements as a bearing structure. Such systems can be used as a source of still water needed for “Aureola” to operate coherently in the urban settings. While those PM systems

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Fig. 2. “Aureola” is a hydraulic system composed of the water container with microalgae cultures and the inlet/outlet fixtures allowing the transportation of the stormwater into a high-leveled urban greenery [22].

require a mechanical removal of all the pollutants accumulated during exploitation [10], “Aureola” can therefore operate as an automatic catalyst to transmit that water stored in MP-systems towards green area of city squares before impurities sediment in the settling tanks; then automatically clean it through biological processes, reducing the need of manual interventions. By those actions “Aureola” improves the maintenance of these under pavement water collection systems, making a composition of those two systems a prospective urban utility able to retain more water and thus, contributing to increase the city’s resilience to climate change. The essence of the designed system is an appropriate balance that allows the formation of a capillary thrust causing constant water uptake and organic matter ejection. The capillary pressure is therefore used to release the algae matter into the soil. Such connection allows water storing during the dry periods and successive water release thanks to algae growth that increases the pressure in the “Aureola” hydro system. The steps of operation of the water purification system are as follows [11]: Step 1: The “Aureola” system is mounted on site, with connection to the macro-pervious pavement system. Step 2: The polluted water collected in the system is transported by capillary rise to the “Aureola” tanks. Step 3: While water enters the system, it is being pushed by capillary pressure with the algae hydrogel towards the other end of the “Aureola” porous channels. Step 4: After up to 11 days [1] of algae growth and treating the polluted water, the clean water with algae biomas is pushed out of the “Aureola” system into the soil around the greenery.

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Analysis layers of functionality: Wetware: A seaweed-based hydrogel condensed in. Software: Optional – dedicated to measure water intake and the quality of water coming in and out of the system. Hardware: A tank with a porous structure that increase the length of the water channels. The inlet tube with capillary nanotubes and the outlet tube with valves and seals regulating the water flow. In the Project there were used a 0.2 to 0.3 mm radius fibreglass tubes stacked in a metal body of the legs/supports. Based on the Jurin’s law (Eq. 1 in which: h is the liquid height, γ is the surface tension, θ is the contact angle of the liquid on the capillary wall, ρ is the liquid mass density, r is the capillary radius, and g is the gravitational acceleration), this size allows the water capillary rise of 40 to 70 cm up to the acrylic container where the algae are stored. The “Aureola” can be mounted as a part of the road water harvesting system, and use the polluted water to clean it and be ready to be returned to the planting areas sensitive to water quality. h=

2γ cosθ ρgr

(1)

The form of the “Aureola” refers to the morphic shapes of nature, freely entwining the green urban elements. Inner structure is porous to extend the path of the water between the inlet and the outlet to extend the water purification process. The shape waves at different heights, allowing passers-by to sit, lean back or touch “Aureola”. Built in this way, the luminous rim creates a diverse spatial composition in various places, which on the one hand contrasts with the surrounding architecture and on the other hand is a harmonized background for fauna and flora. Interactivity: The idea of “Aureola” as the light ring is to create an interactive element that will not only protect green areas from devastation and increase water retention, but will also strengthen the connection between city users and nature, showing that greenery feels and reacts as they do. It can be achieved due to the algae hydrogel used in the project that is sensitive to light condition. During the night it can glow with ambient light, with a constant movement of the liquid, symbolizing the breathing of the trees giving the passersby the magical feeling of the true nature liveliness in the city. “Aureola” therefore creates a fairytale world of flora and fauna, responding to the environmental conditions in real time, engaging in dialogue with people and encouraging them to interaction and wistfulness. Positive effect: The modules of “Aureola” can be implemented in the cityscape independently, or can form wider systems. Thus are economically feasible for testing, and improving urban conditions in a sense of “acupuncture” interventions. The system does not require extensive groundwork preparation and can be shaped into a desired urban furniture functionality. The system can be easily shipped and adjusted to various urban climatic conditions.

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5 Observations In the examples of projects using algae as the main catalyst for the variability of the processes taking place in architecture, the authors’ emphasis on the self-sufficiency of the designed installations and their beneficial impact on the environment in which their applications can be observed. Examples of implemented industrial technologies in housing constructions indicate that it is a technology that can be widely used and proven in terms of utility. They also indicate that their potential use has a strong economic support in terms of costs of implementing technologies based on algae. Experimental installations implemented by such studios as “ecoLogisStudio” show the directions of weaving tissues and architectural forms into the structure of modern city buildings, which can strengthen the self-sufficiency of urban space and reduce its environmental impact on the climate in a wider context. This is evidenced by, for example, experiments using algae to reduce air pollution or purification of highly polluted water in India. The conceptual projects of the LAX studio show that algae can also be allies in efforts to increase urban water retention in symbiosis with increasing biodiversity and the resilience of green urban areas to changes in the hydrological structure of cities and regions. Thus, LAX’s design stays in line with the priorities of the EU climate policy concerning increasing the climate resistance of cities to droughts. The concept called “Aureola” is a kind of network of self-sustaining installations filled with algae that both help to keep more water in the city, but as a result also promote biodiversity by increasing the fertility of the soil in which they are installed. The Laboratory for Architectural Experiments project shows that similar systems have the important feature that they can be shaped in a way that meets the basic utilitarian needs of participants in life in public space thanks to adaptation to various objects such as benches or stands in public spaces. They can be implemented as isolated objects or create wider systems supporting each other depending on the specificity of climatic conditions. Thus, their implementation does not require significant financial outlays in order to gradually implement the proposed solution on a mass scale. The freedom and plasticity of forms allows for freedom in shaping a product tailored to the requirements of specific urban spaces in the context of aesthetic requirements.

6 Conclusions The use of algae has the potential to shape interactive architecture due to the variety of their applications and proven resistance to conditions in an artificially created urban environment. Their natural properties, channelled in hydraulic structures with appropriate parameters, allow to obtain real environmental benefits on a local scale (electricity supply to architectural objects in which they are used, or purification of drinking water) as well as on a global scale (air purification or retention of the water in cities). They are selfsufficient and mostly maintenance-free, and the systems themselves are, in most cases, completely passive. The factor of natural biological processes occurring in algae aggregates is the main driver of the variability of the discussed systems and their interaction

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with humans and the environment, which is usually based on the reactions of algae bioculture to physical or chemical stimuli supplied from the environment and generating in natural processes the result desired in a given interactive system (e.g. the reaction of algae to pollutants is their neutralization). Thus, algae have properties desirable in the process of shaping interactive architecture, because their properties can influence the variability of the architecture in which they are implemented. Proving that algae can have an impact on how the architecture is shaped. The strength of algae confirmed by the implementation of industrial and systemic projects, using algae to produce electricity for the needs of the community in the abovementioned housing project, also allows to claim that their durability and maintenancefree operation allows for a wide range of applications also in the installations and experimental projects cited in this article (case of “Aureola”), with a similar effectiveness in selecting the appropriate structure of the hydrosystems in which the algae will function. Thus such projects as “Aureola” proves that algae can be harnessed in the efforts to increase the resilience of cities to climate change by improving the water retention systems.

7 Guidelines for Further Research The possibilities of using algae in architecture that reacts directly with humans or with the environment, indicate numerous environmental benefits resulting from their use. However, it should be noted that the cited examples correspond to the diagnosed and confirmed problems of modern cities, such as limited access to electricity, air pollution or disturbances in water management causing droughts (the “Aureola” design discussed in the article). It would be important in future design initiatives and research to take into account factors that may emerge in the discourse on city resilience to climate change. Future models should also take into account the problems of reduced mobility of inhabitants and goods resulting from energy crises, problems of hunger, and mass migrations, or even increasing cases of cataclysms that leave a mark on agronomy, strongly associated with cities and the city structure itself. Algae appear to be an important factor offering a variety of uses as active organic matter. As such, it requires the creation of appropriate conditions for the functioning of algae in the form of tanks and various types of supporting structures. Therefore, it requires the use of not always fully ecological materials, such as plastics (even if their production fully uses secondary raw materials). In shaping architecture that actively adapts to the needs of man and the environment, it is also worth exploring the possibilities of using the matter obtained in the production of algae after its physicochemical treatment for the purpose of shaping this inorganic part serving as the environment for the functioning of algae. In the cited analysis of the literature, the authors pay attention to factors that may negatively affect the hermetic environment formed in the cited industrial and experimental projects. Researchers dealing with the use of algae in architecture should carefully follow the achievements of biotechnologists in increasing the resistance of algae to adverse environmental conditions.

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In the context of the use of algae as a catalyst increasing retention in green urban areas, at the same time favoring the increase of their biodiversity and resilience, which was presented by the “Aureola” project of the LAX studio, this direction should be continued in cooperation with biotechnologists, which would enable empirical measurement of the effectiveness of fertilizing the areas in which it is implemented. Moreover the deeply study of data of the exact composition of rainwater is the clue to determine the whole system considering the type of algae, hydrogel, and any extra element which allow to control the quality of algae. In research on increasing the productiveness of the proposed system, the feature of algae for adaptation to unfavorable living conditions should be taken into account, which may, in laboratory conditions, allow for the creation of a variety that can be used particularly in the proposed system.

References 1. Acién, F.G., GómezSerrano, C., MoralesAmaral, M.M., FernándezSevilla, J.M., MolinaGrima, E.: Wastewater treatment using microalgae: how realistic a contribution might it be to significant urban wastewater treatment? Appl. Microbiol. Biotechnol. 100(21), 9013–9022 (2016). https://doi.org/10.1007/s00253-016-7835-7 2. Beesly, P.: Near-living architecture. Work in Progress from the Hylozoic Ground Collaboration 2011–2013. Riverside Architectural Press (2014) 3. Grajper, A.: Ludic Context of Interactive Architecture in Public Space. Politechnika Wrocławska, Wrocław (2019) 4. Lane, K.: Algae Tiles Offer Affordable Purification of Polluted Water Architecture & Design. https://www.springwise.com/sustainability-innovation/architecture-design/ indus-algae-wall-wastewater. Accessed 12 May 2022 5. Nora, S.: Interactive architecture extending the Kansei Engineering approach to real-time interactive spatial systems, In: KEER 2010, Paris (2010) 6. Pasquero, C., Poletto, M., Greskova, T.: Photosynthetic architecture in times of climate change and other global disruptions. In: Proceedings of 38th conference on D1.T6.S1. BIO DATA / BIO Tectonics For Architectural Design - Volume 1, by eCAADe, pp. 583–592 (2020) 7. Hanafi, W.H.H.: Bio-algae: a study of an interactive facade for commercial buildings in populated cities. J. Eng. Appl. Sci. 68(1), 1–16 (2021). https://doi.org/10.1186/s44147-02100037-5 8. Using leaf collection and street cleaning to reduce nutrients in urban stormwater, Upper Midwest Water Science Center (2019). https://www.usgs.gov/centers/upper-midwest-water-sci ence-center/science/using-leaf-collection-and-street-cleaning-reduce. Accessed 29 Apr 2022 9. de Gennes, P.-G., Brochard-Wyart, F., Quéré, D.: Capillarity and Wetting Phenomena. Springer, New York (2004). https://doi.org/10.1007/978-0-387-21656-0 10. Newman, A.L., Aitken, D., Antizar-Ladislao, B.: Stormwater quality performance of a macropervious pavement car park installation equipped with channel drain based oil and silt retention devices. Water Res. 47, 7327–7336 (2013) 11. Dorda, K.: Woda, Design w przestrzeni publicznej, Zamek Cieszyn, Cieszyn (2019) 12. Bottazzi R.: Living Architecture. https://www.architectural-review.com/essays/books/livingarchitecture. Accessed 14 May 2022 13. Das, P., Khan, S., Chaudhary, A.K., AbdulQuadir, M., Thaher, M.I., Al-Jabri, H.: Potential applications of algae-based bio-fertilizer. In: Giri, B., Prasad, R., Wu, Q.-S., Varma, A. (eds.) Biofertilizers for Sustainable Agriculture and Environment. SB, vol. 55, pp. 41–65. Springer, Cham (2019). https://doi.org/10.1007/978-3-030-18933-4_3

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14. Peters, T.: Architectural Design, Text. John Wiley & Sons Ltd., vol. 89(5), pp. 110–115 15. Yuan, J., Dunnett, D.: Plant selection for rain gardens: response to simulated cyclical flooding of 15 perennial species. Urban For. Urban Greening. 35, 57–65 (2018). https://doi.org/10. 1016/j.ufug.2018.08.005. (ISSN 1618–8667) 16. Sharma, R., Malaviya, P.: Management of stormwater pollution using green infrastructure: the role of rain gardens. WIREs Water 8(2), e1507 (2021) 17. Pacheco, D., Rocha, A.C., Pereira, L., Verdelhos, T.: Microalgae water bioremediation: trends and hot topics. Appl. Sci. 10(5), 1886 (2020) 18. Drawing: Connelly, L.: https://www.materialsource.co.uk/the-indus-project-tile-based-bio remediation. Accessed 05 May 2022 19. http://lax.com.pl/portfolio_page/aureola/. Accessed 05 May 2022 20. https://www.ecologicstudio.com/projects/h-o-r-t-u-s-xl-astaxanthin-g. Accessed 05 May 2022 21. https://www.ecologicstudio.com/projects/air-bubble-air-purifying-eco-machine. Accessed 05 May 2022 22. Drawing: Grajper-Dobiesz, A., Dobiesz, S.: Aureola http://lax.com.pl/portfolio_page/aur eola/. Accessed 05 May 2022

Summer City: Campsites as a New Ecological Approach to Sustainable Living Luca Trabattoni1(B) , Carlo Berizzi2 , Margherita Capotorto2 Gaia Nerea Terlicher2 , and Marta Mazurkiewicz2

,

1 Faculty of Civil Engineering and Architecture, Opole University of Technology,

Katowicka 48, 45-061 Opole, Poland [email protected] 2 Pavia University, 27100 Pavia, PA, Italy [email protected], {margherita.capotorto01, gaianerea.terlicher01,marta.mazurkiewicz01}@universitadipavia.it

Abstract. Today the dream of sustainable development applied to the urban context needs a redefinition. In the search for new answers, we intend to shift our gaze from the traditional forms of the urban fabric to that of the accommodation facilities of outdoor tourism, which for their characteristics combine the issues of living in contexts characterized by the prevalence of natural elements. The sustainability of these structures seems an intrinsic value, but the growing awareness of environmental issues related to climate change and the protection of ecosystems leads to design experiences based on new models of land use and landscape design. Through three case studies defined through research by design method, we intend to promote new forms of settlements that, starting from the experience of outdoor tourism, can represent new urban structures capable of protecting, enhancing and caring for the environment and the landscape in which they are located. The projects are based on the use of maxi-caravans as housing units, the most used solution to date to promote green issues that differ in terms of themes addressed and the contexts in which they are inserted. Keywords: Sustainable tourism · Campsite · Architectural landscape

1 The City of Desires Between Nature and Artifice. The Hypothesis of Intervention on Open-Air Accommodation Facilities Taking on the inadequacy of the Smart City model to find solutions for the sustainability of urban space, we choose to start from the study of the settlements of outdoor tourism, in which we see the possibility of enhancing the natural element characterizing these places. The study of the settlement systems of open-air accommodation facilities highlights a parallel between these and the urban space. These “Summer Cities” are built on an idea of anthropized landscape declined according to receptivity and tourism. However, these spaces must respond to clear functional needs, considering environmental impact mitigation. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 211–225, 2023. https://doi.org/10.1007/978-3-031-26879-3_17

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In this framework, it is possible to define a design solution that synthesizes the technical and natural aspects to ensure comfort and quality in the accommodation facility, keeping the protection and promotion of the places a priority. As Joseph Rykwert writes “If the city itself must be related to physiology, more than anything else, it resembles a dream” [6]. The shape of the contemporary city has always clashed and confronted with ambitions of citizens and their ways of living and expresses an ideal model that can be traced back to the dream. However, the contemporary city is the first city in history that is centred on the demand of the citizen [1] but that perhaps, also for this reason, eschews a dreamlike and perhaps utopian vision. Among the prevailing ideologies of contemporary culture, the myth of sustainability clashes with the idea of living in urban contexts, which are increasingly perceived as agglomerations that cause harm to the environment and man. So, the “dream” of sustainable development applied to the urban context today needs a redefinition not only limited to the responses of a technical, energetic and productive nature but in a more complex sense, which concerns the community, thus including social, cultural, psycho-physical health aspects linked to the most profound reasons for the well-being of those who live in the city. If the theme of the Smart City has proved to be limited and inadequate1 to respond to the demands of contemporaneity, especially concerning the environment [3], it is essential to look for alternative proposals. Searching for the definition of the image of the sustainable city, the technical solutions must thus be mediated with the quality of the urban environment. It can be important to start these reflections by considering some famous examples of urban development related to sustainability approach to neighbourhood and town planning. For example, the Bed Zed in London, respond coherently to the question, using the architectural element as a compensatory system for urban impact thanks to technology development. Others, such as the Solar City of Linz, decline the sustainability of public space with widespread plant systems or with landscape elements with a technical background, as in the case of the Bo01 district of Malmö. In both cases, the technique imposes itself from the formal architecture language and public space point of view. From these reflections, the hypothesis that is being considered within the AUDe laboratory of the University of Pavia with Camping Design and Architecture2 is to promote new forms of settlement development based more on the emphasis on the natural element of the places than on the technical factor starting from the study of the settlements of outdoor tourism. These “seasonal” citadels originate in the natural territory seeking integration with the landscape. Among their settlement characteristics, 1 See the interesting speech by Rem Koolhaas to the European Commission on 24 September

2014: “The rhetoric of smart cities would be more persuasive if the environment that the technology companies create was actually a compelling one that offered models for what the city can be… Smart cities and politics have been diverging, growing in separate worlds. It is absolutely critical that the two converge again”. (from Rem Koolhaas, “My thoughts on the smart city” at the High Level Group meeting on Smart Cities, Brussels, 24 September 2014.). 2 Design studio born from the synergy between the architect Luca Trabattoni, the AUDe laboratory of the University of Pavia and the CrippaConcept company, which experiments in the field of outdoor tourism and, in particular, offers solutions related to the insertion of maxi caravans in receptive contexts. https://www.campdesign.it/

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determined by functional and market needs, there is the mediation between density and the sustainable development of open space. In these places, where the landscape image is linked to the holiday value, the coding of the sustainable landscape is linked to its transmissibility. Therefore, a careful look at these places can represent an opportunity to try to understand which architectural and settlement solutions can represent virtuous and exportable models, also capable of interpreting the ecological “dream” of contemporary living. Through the design experimentation on three different case studies, we intend to identify strategies and techniques for the enhancement and construction of the anthropized landscape that include different aspects of sustainability, from the promotion of the natural element and biodiversity to the recovery of resources, to energy saving, and the use of renewable resources. Applying the technique leads to a new coding of the language that puts on the same level elements of architecture and landscape decoding the overall image; so, while plant systems are declined as a formal part of architecture and landscape, natural elements3 are interpreted and iconized as technical elements. The proposed design approach thus seeks a synthesis of the two aspects, technical and natural, according to the comfort and quality of the living space concerning the site.

2 Organization and Characteristics of Open-Air Accommodation Facilities In order to define the scope of study and design experimentation, it is essential to identify the elements that characterize the open-air accommodation facilities. Even if these places live a temporality of seasonal use, their spatiality must respond to the expectations of “users”, just like urban spaces. The open-air accommodation facilities are characterized by recognizable foundational elements (fence, vegetation, paths and services) and welcome a finite number of accommodation types (tents, caravans, maxi-caravans, bungalows). The definition of the relationship with the external context and the organization of the internal landscape passes through the management of these two macro-categories. Therefore, the integration of the building and the removable receptive with the natural element becomes the central focus of the design experimentation. The urban sociologist Giandomenico Amendola in his studies on the organization of contemporary cities identifies ten “urban aspirations” in the current collective imagination: the sustainable city, the business city, the show city, the cosmopolitan city, the à la carte city, the ubiquitous city, the beautiful city, the safe city, the friendly city, the city of citizens [1]. Many of these aspirations are found in the city of outdoor tourism (campsite and tourist villages). Some projections of positivity accompany the aspiration to a place other than everyday life: comfort first of all, which is codified in a balance between 3 As George Monbiot and Greta Thunberg tell in the famous video "A tree saves the planet", the

tree is intended as a sustainability machine, which intrinsically manages to generate quality and compensation and, in this case, becomes a manifesto of an architecture that is overwhelmed by it. Following this principle, some famous reforestation campaigns such as “One billion trees” have been launched over the years, spread worldwide.

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security, privacy and conviviality; the provision of physical and entertainment services; the beauty declined not only on the built, but on the primary landscape (sea, mountain and lake) and interior one; and in recent years also sustainability, brought as an ethical attitude towards the world and, often, brought back to the natural presence. The landscape of these places is an intrinsic condition, which is almost always related only indirectly to the prevailing surrounding landscape, introjecting some peculiar elements. Instead, the direct relationship is defined by a fence, which defines a specific safe, protected, exclusive area. The codification of what happens inside the open-air accommodation facilities passes through the definition of hierarchies and some macro-elements: – the set of anthropized natural elements (internal vegetation, paths and soil); – the system of permanent artefacts (services); – the system of accommodation units (tents, caravans, maxi-caravans, bungalows). All this defines an autonomous citadel that allows you to live during your holiday, usually 7–14 days, without having to leave the fence. The tourist model most similar to these accommodation facilities is cruise ships, both in organization and services the construction system, however, has numerous differences (the cabins of the boats are characterized by fixed prefabricated furniture consisting of a single block). However, over the years, the shape of the villages has led to the definition of a unique and repeated model, indifferent to the nature of the places. Like shopping centres and airports, the village has defined a reassuring and standard model paradoxically closer to the non-place [2] than to the place, understood as a geographical context of belonging. Everything is reassuring and easily understood, and it is, in this dimension, that a part of the dream is reached. In the short time of stay, this city of the holiday [7] must be easily decoded. I must know where I can move, where I can find the essential services, the centres for fun, and where to do the shopping. The supermarket must also be able to provide all the products that can be served on a holiday. Even nature undergoes this alteration that makes it understandable and noticeable; the intensive use of pine forests, suitable for creating shade without preventing the visual relationship with the surrounding environment, is an emblematic example. However, the immersion in nature is evident, in the presence of vegetation and the soil almost always unchanged, the heritage of the culture of the first informal campsites but also the need to respect areas almost always protected.

3 Environmental Friendliness of Open-Air Accommodation Facilities Given the intrinsic connection with the natural landscape, the actual environmental impact of open-air accommodation facilities is now defined, considering the risks and potential of ecotourism as defined by the OMT. Although outdoor tourism has a strong presence of natural elements, the substantial influx of people impacts the environment. The theme of “overtourism”, which remained frozen during the health emergency, returns

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to be contingent. Alongside the management responses of open-air accommodation facilities design strategies can be identified to mitigate the environmental impact, integrate virtuous plant systems, and work with tree essences and architectural language. Within the tourist context, the transmissibility of the value of the landscape is fundamental. The theme of sustainability, such as the local landscape or vernacular architecture, must be integrated into the overall design vision of the open space. The proliferation of open-air accommodation facilities, especially in the Italian coastal territory, poses important questions regarding the environmental compatibility of these structures; in 2018, ISTAT recorded 2,612 tourist villages and campsites for 1,346,536 beds, capable of absorbing about 8% of Italian tourist presence. Although the settlement system of open-air accommodation facilities is characterized by removable elements and low soil consumption [5], the prevalence of the natural element, precisely because of the anthropic mass, is not a sufficient condition to compensate for its environmental impact4 . The OMT (World Tourism Organization) has defined three main categories of the environmental impact of the tourism sector: the economic, the socio-cultural, and the ecological. The ecological impact is mainly linked to the phenomenon of “Overtourism”, or “the impact of tourism on a destination, or parts of it, which excessively and negatively affects citizens’ perceived quality of life and/or the quality of visitors’ experiences.” [8]. Four parameters define the impact of “Overtourism”: – pollution: of air (linked to car traffic), water (discharge of sewage, discharge of solid waste, discharge of hydrocarbons from boats), waste, acoustic; – the loss of natural landscapes: the construction of buildings leads to invading open spaces and leads to the disappearance of entire wooded areas; – the destruction of flora and fauna: the large influx of tourists can lead to the disappearance of some species; – congestion: the concentration of tourists on holiday in a specific place damages the landscape (traffic congestion on the roads in periods coinciding with mass holidays increases pollution, loss of time and massive fuel consumption and increased C02 emissions)5 .

4 First national report on land consumption, DiAP Politecnico of Milano - Legambiente - I.N.U.,

National Observatory on Land Consumption, 2009, Maggioli Editore. “In general, it can be defined as an anthropogenic process that involves the progressive transformation of natural or agricultural areas through the construction of buildings and infrastructures, and where it is assumed that the restoration of the pre-existing environmental state is challenging, if not impossible, due to the nature of the upheaval of the earth matrix. This definition is characterized negatively, since the problem of the subtraction of natural or agricultural surfaces is negatively perceived considering the finiteness of the earth’s surface; and it would therefore be more correct to speak of soil transformations.”. 5 The World Code of Ethics for Tourism, adopted by the resolution of the General Assembly of the World Tourism Organization in Santiago de Chile (27 September - 1 October 1999), has its fundamental objective to promote responsible, sustainable, and accessible tourism for all.

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Tourist overcrowding is, in fact, “that situation in which the impact of tourism, at a certain time and in a certain locality, exceeds the threshold of physical, ecological, social, economic, psychological and/or political capacity” [4]. Therefore, it is clear that the relationship with the natural or landscape vocation of the accommodation facilities is not enough to declare their environmental compatibility. From the point of view of space design, it is possible to codify some generalizable approaches through direct experimentation and the design of some receptive areas or parts of them. For example, the themes of density, the relationship with the plant and architectural context, and the use of natural environmental mitigation strategies are, among others, intervention systems that lend themselves to direct interaction with users or to their explication. First of all, the open-air accommodation facility is to be recognized as a settlement system in all respects, which suffers from the absence of specific and overall planning. Often these places are made for parts and annexations. Therefore, beyond some indications for controlling the flows and security of the sector, there are no development plans that are neither potential nor implemented. All the case studies that will be analyzed refer to interventions on existing structures, regeneration or expansion and are based on a design approach for clusters, which, like urban acupuncturists, punctually modify the context, ensuring an evolutionary possibility consistent with the needs of the structure. Working on a small scale with repeatable elements, it is possible to build sustainable, comfortable and “didactic” landscapes, that is, repeatable. The goal should be to construct spaces of the city of tourism so that they can dialogue with tourists, refining the link between signifier and meaning in the definition of these “urban spaces” sui generis. The use of artificial and plant elements, the integration of plant technologies, and the search for an architectural language are all necessary topics for constructing these places according to the specificity of the context to which they belong. Through the design of some areas, it has been possible to apply the reflections made so far by defining three macro-strategies of intervention: – environmental mitigation, combining light infrastructures and vegetation (order and homogeneity of the landscape); – The Externalization of Virtuous Plant Systems for the Optimization and Use of Natural Resources (Technical Landscape); – Coding of the Elements of Vernacular Architecture for the Definition and Organization of the Comfortable and Habitable Landscape. Concerning this last point, it should be noted that the difficult question of the language of the anthropized natural landscape, as a function of tourism promotion, has already been addressed in the past: this is the case, for example, of the Costa Smeralda (in Italy), which in 1969 adopted an architectural plan to define “the Sardinian landscape” imposing it as a constraint on the construction of accommodation facilities; it is also the case, although less incisive, of the Club Med, conceived in 1950, but spread enormously only since 1963, which built holiday spaces codified on the architecture and landscape of the place, creating a false vernacular.

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Today, the world of outdoor tourism in Italy works mainly through temporary structures (maxi-caravans), and, for this reason, the question of language is entrusted to the organization and management of natural elements and removable structures. It is a matter of embarking on a new path that interprets the architecture of the existing landscape in the same way as that of the local vernacular construction tradition.

4 Case Studies Three case studies are introduced relating to open-air accommodation facilities: Camping Concadoro, Lake Maggiore; The International Village in Bibione, Adriatic Sea; Camping 4Mori in Muravera, Sardinia. The three projects are an opportunity to define specific intervention strategies, highlighting some contingent themes. The case studies aim to define the latter’s thinking, synthesis, and concretization, providing in-depth information and outlining ways to deal with problems. The AUDE laboratory and the Camping Design and Architecture studio has developed an experimental design reflection that has led to the definition of some paradigmatic case studies. Environmental Mitigation is done by working in concert with plant essences and architecture, seeking compatibility with the existing context and identifying a recognizable identity for the intervention in Camping Concadoro. The externalization of the plant systems serves to build a portion of the International Village of Bibione. The construction of a sustainable landscape takes place by placing water at the centre of the design, as a technical element and as an urban furniture system. The codification of the vernacular elements of architecture for the definition and organization of the landscape serves to build a new sector of the 4Mori Camping, finding a recognizable identity linked to the local context. 4.1 Environmental Mitigation The first case study concerns the reorganization of the Camping Village Conca d’Oro on Lake Maggiore. The campsite is located within the Toce River Delta Park, known as the Fondo Toce Nature Reserve and characterized by a large poplar area, constrained and protected. The accommodation facility is affected by a precarious situation: the acquisition in successive temporary phases of adjacent areas without unitary planning has generated an inconsistent whole, defined by three different areas developed along the coast of the lake. Furthermore, the coexistence of anthropic elements and protected natural elements contributes to the need for a reorganization intervention, that does not distort the existing in compliance with the constraints of the park. The campsite can be read as a succession of three macro areas partially sectored and detached from each other. The entrance area is a typical “campsite”, with development mainly with free pitches and usable by tents and caravans. On edge away from the lake are installed maxi-caravans in line; the next sector is organized as a village, occupied only by maxi-caravans, surrounded by a high hedge and accessible only through a road “at the back” or from the beach; the last area sees a mixture of maxi-caravans and free pitches but intended only for campers.

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This organizational rigidity precludes relational and landscape opportunities between the disconnected and not integrated areas. As a result, the relational possibilities are lost, and above all, the continuity of the camping space is lost. Increasing this fragmentation and this sense of disorientation contributes to a productive agricultural area (poplar grove) that interrupts the development of the campsite, generating a void with excellent landscape value, but unusable by the accommodation facility. The design objective is represented by the search for a formal order and spatial homogeneity using strategies common to urban design, which facilitate the use of spaces, making them more understandable, accessible and easy. An ordering system is designed that links the different situations, using light infrastructures and vegetation. The pedestrian paths and tree essences give the rhythm to the public space and mark the areas, distributing them harmoniously. The intervention focuses on arranging landscape architecture designed ad hoc, mainly pergolas, seats and covered paths. An orthogonal rib is designed that combines a system of punctual “events”, enhances the existing potential and implements the transversal connections. The infrastructure will consist of equipped paths and rest areas flanked by plant elements. The area occupied by the campsite is about 50,000 square meters with a rectangular extension in favour of the lakeside. The project enhances the optical cones towards the lake, identifying two main distribution equipped paths that give the measure of space. These are put into a system by creating a long transversal path that develops parallel to the coast of the lake and becomes a backbone, that unites the three areas of intervention. Strategic points of the existing receptive context are activated using particular furnishings. As in a regenerative acupuncture action, by intervening with lines and points, it is possible to change the perception of the entire context and give it the missing unity. The preparation of these two large arteries also guarantees the maintenance of escape routes, which represent one of the few planning constraints to which campsites are subjected and which, very often, determine their distribution, remaining indifferent to the existing topography. In Camping Village Conca d’Oro infrastructure becomes landscape, designed with simple and easy-to-maintain elements, favouring integration with vegetation. In general, the issue of environmental compatibility is central precisely because of the constraints imposed by the Toce River Delta Park. The choice of furnishing materials aims to bring a solid approach to the natural environment, so iron (for maintainability) and metal gabions filled with local stones are mainly used. There are, therefore, five hot spots of intervention and three avenues. The paths are characterized by defining the beginning and end of the infrastructure with recognizable rest areas, designed for shadow situations with metal elements, prepared for climbing plants, and gabions of stones that can be used as seats. A focal point of the intervention concerns the passage through the village area. First, a crossing point is defined to avoid excessive isolation of this space. Next, a cut is made in the existing hedge, taking advantage of the passage between two maxi-caravans. An entrance square is then built, a small staircase covering the difference in level, and an internal path. Similarly, work is done at the exit of the village area. In this way, the connecting ridge continues without interruption, ensuring a long promenade through all campsite areas.

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Alongside the architectural and infrastructural intervention, the use of vegetation and the material connection to the context is fundamental. The constraint of the park requires the choice of native plant essences or compatibility with the native ones so as not to cause damage to the existing ecosystem. If the architecture is integrated with seats made of wood resting on stone gabions, proposing the use of local stone, granite of Baveno, the study of vegetation is relevant for the homogeneity of the intervention. Three large families of plants have therefore been identified to be inserted: – Plants of reduced height to create the low flower beds and generate a visual order that guarantees the reading of the various pitches. The low plants and shrubs are used both in isolation, as a partition between the pitches, and side by side with the seats, to enhance the paths. The list of these essences includes: Stipa Tenuissima, Stuca Glauca, Elijah Blue, Verbena Bonariensis, Artemisia Powis Castle, Pervoskia Atriplicifolia, Calamagrostis Acutiflora, Lavandula Angustifollia, Rosmarinus Officinalis. – Larger vegetation to create the high hedge and identify the points of intervention within the existing. High essences are used as visual dividers in situations where it is necessary to guarantee privacy. The list of these essences includes: Carpinus Betulus, Acanthus Mollis, Fargesia Robusta Campbell. – Climbing plants to be integrated with the light architectures of the infrastructure. The list of these essences includes: Hedera Helix, Parthenocissus Quinquefolia, Wisteria Sp., Buxus Sempervirens. The tree species were selected for climatic adequacy, aesthetic pleasantness, and environmental benefits. The selection of the essences was accompanied by a study of the colour so that the plants’ chromatic effects also had aesthetic feedback in the sensory system of insertion (Fig. 1). 4.2 Externalization of Virtuous Plant Systems for the Optimization and Use of Natural Resources The second case study is the expansion of the International Family Camping campsite in Bibione. This accommodation area is located on the Venetian coast, in one of the Italian areas with the greatest tourist attraction6 . The campsite covers an area of about 12 hectares, divided into two large sectors: one directly connected to the landscape system of the equipped beach and the sea, which continues throughout the city of Bibione, and a second strip located away from the sea, behind the coastal state road that develops for kilometres along the coast. The two areas have two different natures: the one towards the sea is mainly intended for temporary accommodation systems (tents, caravans, and a few maxi caravans); the one far from the sea houses instead of a large area of bungalows and an empty area of expansion, the subject of this study. 6 4.4 million arrivals, 24.4 million visitors and 230,901 beds in 2019 in Veneto, XXIV Italian

Tourism Report.

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Fig. 1. Infrastructure over the actual situation within evidence the hot points (Credit: Camping Design and Architecture).

The redevelopment of the accommodation area away from the primary landscape system passes through a “resemantization” of the soil and the natural environment and the integration of mobile units with the landscape. The project aims to build a sustainable landscape, that declares the environmental compensation strategies adopted by operating according to the principles of ecology and the protection of biodiversity. Living in nature means creating a direct relationship between the experience of living, in this case on holiday, and the surrounding natural environment. Water, soil, and trees become the protagonists of the fun landscape, creating remarkable and ventilated places and mitigating the environmental impact. The project, therefore, involves the sector’s construction of about 40 maxi-caravans, organized in an area of about 5,000 square meters (about 125 sqm. Per pitch). Given the climatic characteristics of the Veneto area of Bibione, the search for the environmental comfort of the open space passes through a double attitude: the first concerns the search for plant solutions that compensate for the impact of consumption of mobile units; the second concerns the definition of passive control strategies of air temperature through the arrangement of maxi-caravans and vegetation. Since the area is mainly warm (average summer temperatures 26–29 degrees) and humid (significant rainfall throughout the year), the objective of the interventions in the open space is to promote cooling and natural ventilation, using shaded areas, plant elements, and water. Water, therefore, becomes the theme of the project, declined functionally and aesthetically, linked to sustainability and fun. As in the Bo01 district of Malmö, the technical approach is made explicit and declined to characterize the public space. In Malmö, among

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the numerous and virtuous environmental mitigation strategies7 , the one related to water occupies a prominent place. Here the rainwater is diverted through gutters above ground and collected in exposed tanks, which characterize the urban dimension. Drainage and collection systems become aesthetically designed elements. As in the famous Glenn Marcutt’s boreal projects, the technological solution becomes manifest in the ethical approach. Similarly, the project of the new area of the International Family Village in Bibione defines a landscape integrated with plant solutions. The maxi-caravans are organized on a double system of pedestrian paths while the cars are left on the side lines. The main paths are ordered and parallel to each other, guarantee orientation and functionality, and define the housing units’ accesses. On the other hand, the secondary paths are sinuous and allow a dynamic perception of the natural environment, articulating the landscape and connecting the water elements. Water is a symbol of sustainability in the context of insertion, sustainability that can be declined through numerous actions: collection, accumulation, purification, and reuse. The use of active and passive sustainable strategies can give the correct value to the natural element and educate the user to recognize its preciousness and possible uses. The principle of recovery of natural resources and optimization of consumption is the basis of the project. The exploitation of water, as an element of compensation for the ecological impact of the settlement, passes through two distinct strategies: reuse and environmental mitigation. The reuse of water implies its collection and cleaning to be usable again. Excluding potabilization (as it is too expensive and cumbersome), strategies of natural purification (phytodepuration) and reuse in the domestic and recreational environment have been developed. The widespread “phytodepuration” system, organized in wet tanks and dry tanks, potentially allows covering the water needs of 200 equivalent inhabitants. The system selected for wet tanks is the compact Fitofluver model8 . The water treatment takes place inside the tanks using a pump that continuously recirculates sewage from top to bottom through a complex layer of septa and various draining materials. The oxygenation coming from the plants and the combined purification action of the volcanic lapilli and the zeolite filter complete the purification up to the parameters required for compliance. The system guarantees that, with an area of over 5 m2 , it is possible to treat wastewater up to 20 inhabitants equivalent, with the cultivation of herbaceous varieties not necessarily specific for this kind of water treatment. 7 Bo01 (also known as the "City of Tomorrow") is a neighbourhood in the southern city of Malmö,

Sweden known for its sustainable development and design. Bo01 began as part of the European Housing Exposition in 2001 and served as a prototype to help later design Västra Hamnen. Today, Bo01 is known for its holistic approach to incorporate sustainable design into highquality living and serves as one of the first Swedish models for sustainable urban planning. Givan, Kate. What does good leadership look like? Lessons from Bo01, Sweden. Architecture + Design Scotland. 8 The FITOFLUVER Natural Compact Phytodepuration Plant of the Company Edil Impianti2 consists of prefabricated tanks and represents a good solution for wastewater treatment up to 20 A.E. Its efficiency, higher than the activated sludge plants usually proposed, guarantees continuity of yield over time, with almost no maintenance.

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Phytodepuration is a black and grey water cleaning system that includes: rainwater, wastewater from showers and sinks, and toilet wastewater. These solutions are applied to public space generating two types of elements: the dry tanks appear as vegetable masses, encumbrances that condition the passages, defining the sinuous paths that cross the clusters; the exposed collection tanks are instead used as an aesthetic element that defines the entrance to the compartments of houses that make up the entire agglomeration. Finally, the purified water is also conveyed inside the bio-pools positioned in the cluster. The visible water element also contributes to improving the quality and comfort of the outdoor space. For example, if associated with the creation of convective motions, water can regulate summer temperatures, helping to limit heat islands, giving moisture to the air, cooling it, and thus, creating a natural air conditioning system. The trees are positioned to optimize their passive effects: the natural shade, which allows first protection from the sun’s rays; air cooling; visual mitigation; CO2 compensation. Furthermore, the project proposes to exploit the natural characteristics of the maritime pine (local essence already present in the area), which is very high with a free trunk and wide and flat fronds. It is possible to exploit natural convective motions that ventilate the external environment, using its shadow effects, making the campsite more comfortable in the summer seasons. The study of the shading and irradiation of the areas under the trees was carried out employing parameterization software that evaluated the effects of shade and the amount of solar radiation to which the project area is subjected. A helpful tool for architecturaldesign purposes was the analysis of solar radiation, particularly the consequent shading effect, generated by the surfaces of the volumes and vegetation included within the project and adjacent. The study made it possible to evaluate the incidence of the shadows projected on the area in question, estimate what kind of spaces they give life to, and, consequently, evaluate where best to place the bodies of water, both playful and technical. During the project’s development phase, the irradiation analysis was also carried out, which made it possible to evaluate how much solar energy affects the surfaces of the volumes and vegetation included in the project. Furthermore, this analysis made it possible to control tree species’ positioning and consequently optimize their passive effects integrated into the water system (Fig. 2).

Fig. 2. 1) Maxi-caravan system. 2) Accessibility. 3) Water system (Credit: Camping Design and Architecture).

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4.3 Coding of the Elements of Vernacular Architecture for the Definition and Organization of the Landscape The latest case study deals with the 4Mori Camping, Muravera, south of Sardinia (Italy), along the coast and close to some of the most important settlements of the region. It has a considerable extension and a distribution on the territory organized by areas with different uses. A central strip houses the services for catering, sports and fun/entertainment, where the housing units are distributed, or the pitches that host campers and caravans are arranged. The slice closest to the sea is home to numerous maxi caravans and a large buffer area for future recreational sports destinations. The expansion area affected by the project is far from the sea and separated from the rest of the campsite by a public road, generating the need to create quality spaces despite the non-contiguity with the main natural element. The need to build a sector inside the campsite with its own recognizable identity starts from reading the landscape. The Sardinian landscape of Muravera is characterized by large agricultural land, separated by low dry stone walls, which, especially in mountainous areas, are integrated with sheepfolds and other service elements for pastoral activities. The theme of the wall, in general, is also a theme dear to the historic Nuragic villages. The Sardinian Nuragic villages, in fact, have the housing units leaning against a wall of the enclosure that groups them into courtyards. The reasons for this design choice are many: the house and the fence constitute a continuous line, and the living space is incorporated into the line that separates the public from the private sphere. Moreover, this solution saves the construction of a wall since the fence and perimeter of the dwelling merge. In defining the founding elements of the settlement system of the new area, we try, as far as possible, to use and interpret vernacular systems. However, given that maxicaravans are strongly constrained in customisation as an industrial product, we choose to work both on the aggregation system and the infrastructural elements of aggregation. The accommodation units are collected in clusters of a few units within a large enclosure that manages the internal flows. The wall remains, and, according to its variations, it becomes an element of dialogue with the project area and the tourists themselves. It is chosen to maintain the materiality of the wall, using local stone as a coating or as a filling. The wall defines the fence and the internal infrastructure and generates aggregative contexts of the maxi-caravans. 64 maxi-caravans have been arranged on an area of about 31,000 m2 , and the wall constitutes the settlement project’s principle defining the identity of the designed sector. As in the previous case, also in Muravera, the parking lots are set up near the entrance but outside the settled area to amplify the holiday experience in nature and encourage travel through the spaces designed sustainably. The wall dialogues with the space changing direction, size, and nature, each change of direction coincides with an entrance or a systemic design development at the macroscale; the wall bends, for example, to generate the primary and secondary entrance to the area and then develops within the area without changing size or height. The distributive macrosystem linked to the fence and the main path is defined by the road/wall relationship, along which it is possible to sit and discover the internal landscape through openings arranged in the wall itself.

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In this way, a distributive infrastructural system is built that will condense both the visual and technical elements of the use of the area (lights, signals, sounds). The heart of this macro system is a water tank, developed with the technology of bio pools, for both recreational and thermoregulatory use. When it detaches from the primary distribution system, the wall changes nature, and it becomes a system of metal gabions filled with local stones. This short element accompanies the avenues that lead to the clusters of mobile homes, tangling the houses themselves and their appurtenances. The short wall flanked by local essences, such as myrtle and eucalyptus, becomes the visual system that encloses the domestic sector, as in the Nuragic villages, while the high wall is the main road that distributes the main street and services and organizes the area (Fig. 3).

Fig. 3. 1) Actual situation. 2) The high wall folding. 3) Short wall clustering. (Credit: Camping Design and Architecture).

5 Conclusions The design experimentation has demonstrated the potential for research on the open space of open-air accommodation facilities. The language of the landscape, full of the expectations of tourists, looking for a synthesis between naturalness and comfort, finds different potential declinations. The sustainability of open space is declined simultaneously in the two directions of efficiency and living comfort. The habitat of open-air accommodation facilities thus approaches a high-performance urban space. Also, because of its particular temporality, the design choices propose virtuous strategies, which cone complementary and can also be application models for “ordinary” situations. Environmental mitigation, the coding of a technical landscape and the search for a language similar to the vernacular are design approaches aimed at preserving the identity of the place without sacrificing the environmental compatibility of the intervention. The summer city is the place of potential mediation between the local landscape and the atopic one of tourism (between the specificities of the territory in which it is inserted and the landscape to which it refers) and the needs for comfort and luxury of outdoor tourism.

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The image of the landscape of these summer cities, which wants to be the promoter of a socio/economic phenomenon, can become a manifesto of an ethical value of living without pedagogical ambitions, creating efficient collective spatiality.

References 1. Amendola, G.: Tra dedalo e Icaro. La nuova domanda di città. Laterza (2010) 2. Augè, M.: Nonluoghi. Introduzione a una antropologia della surmodernità. Elèuthera (2009) 3. Koolhaas, R.: My thoughts on the smart city. In: High Level Group meeting on Smart Cities, 24 September 2014, Brussels (2014) 4. Peeters, P., et al.: Research for TRAN Committee – Overtourism: Impact and possible policy responses, p.19 (2018) 5. Pileri, P., La questione «consumo di suolo». In: AA VV, Osservatorio Nazionale sui Consumi di Suolo. Primo rapporto 2009, pp. 10–15. Maggioli Editore, Rimini (2009) 6. Rykwet, J.: L’idea di città. Antropologia della forma urbana nel mondo antico, p. 7. Adelphi (1963) 7. Trillo, C.: Territori del turismo tra utopia e atapia. Alinea editrice, Firenze (2003) 8. World Tourism Organisation (UNWTO), Executive Summary. In: Overtourism? Understanding and Managing Urban Tourism Growth beyond Perceptions, p. 4 (2018)

The Use of a Prefabricated Mass Timber Structure in the Design of Single-Family Houses in Terms of Sustainable Development Kamila Wilk(B) Faculty of Civil Engineering and Architecture, Department of Architecture and Urban Planning, Opole University of Technology, Katowicka 48, 45-061 Opole, Poland [email protected]

Abstract. Prefabricated wood components is not a new idea in architecture, but still not very common in the construction of single-family houses. Both of the analyzed construction methods, wooden frames and mass timber constructions, helped solve many design and engineering solutions. Prefabricated solid wood structures can help to reduce emissions of the embodied carbon during the construction of a building on a site and while the items are produced. Prefabricated wood elements offer the advantages of high performing and energy efficient passive designs. An additional advantage of the use of wooden technology are biophilic attributes which helps improve human well-being, as well as promotes environmental consciousness. The research methodology involves a comparative case study the analysis and synthesis of the similarities, differences and patterns across those two approaches. The case study was based on a comparison of two uses of wood structures. Single-family house with usable area of a 100 m2 , prefabricated with a mass timber structure was compared to a single-family house with the same amount of square meters in a frame structure. Research shows a strong advantage of mass timber structure. Tested approach turns out to be the future of building in terms of sustainable development. Keywords: Wood structure · Mass timber structure · Sustainability · Architecture · Wooden houses · Wooden single-family houses

1 Introduction Buildings that are created on the basis of architectural projects, live longer than their authors, which is why we should strongly consider the materials we use to erect them. These buildings will not only affect the sense of aesthetics of future generations, but will directly affect the pollution of the environment around them. A common aspect in the implementation of single-family and individual house projects is the need to create something unique. This need is realized both on the part of the architect and the investor, who, realizing his dream of a perfect house, wants it to be exceptional. Unfortunately, this desire is often realized with the involvement of © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 226–233, 2023. https://doi.org/10.1007/978-3-031-26879-3_18

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excessive means, which have a huge impact on the degradation of the natural environment. Leveling the influence on the natural environment, is still not at the top of the pyramid of needs realized in the construction of a dream house. Reducing the carbon footprint should be a topic already considered at the level of conceptual decisions when designing a single-family house [1]. The aim of the publication is to try to compare, two wooden structures in order to determine which one better corresponds to the principle of sustainable development. Carbon footprint does not currently exist in building regulations in Poland. Before we get into how to determine emissions, it is important to provide a definition of carbon footprint (CF). CF is defined as the total emission of Greenhouse Gas, GHGs during the full life cycle of a production of a construction material. It is expressed as the equivalent of carbon dioxide per functional unit of the product (CO2 e/functional unit). The size of CF, e.g. is calculated for a fixed functional unit throughout the supply chain (cradle to grave). A limited scope is also acceptable, e.g. only the production process (cradle to gate In many European countries, tools are being prepared to check the environmental impact of building materials and educational materials related to the impact of construction on the natural environment. According to the UN Global Status Report 2020 for buildings and structures, the construction industry in the world consumes 35% of energy, which is associated with high CO2 emissions [2]. We define the negative ecological footprint as GWP (Global Warming Potential). In the construction industry, we can define two types of carbon footprint. The first of them is related to the extraction and production of building materials, while the second is related to the operation of facilities. A fundamental change is needed in the way of proceeding and thinking about architecture in order to achieve climate neutrality in the construction market. It is necessary at every stage of the construction process. The approach to the design, production of building materials, implementation of investments and the use of energy sources must be redefined [3–9]. As some regulations on the energy consumption of the construction process already exist in our legislation, it can be expected that these steps will result in the emergence of carbon footprint reference thresholds, and the need to check the calculations whether a given implementation falls within certain thresholds. This obligation is likely to fall on architects. When preparing for these steps, we should at every stage of the design path and the implementation of the task reflect on the impact of our activities on the natural environment. In addition, preparations should also include expanding our knowledge in this area, as well as preparing to take on new challenges related to determining the impact of our activities on the environment. The research carried out for the purposes of this article is based on certificates related to running a self-propelled design practice in which specific objects are implemented. The work shows a representative part of the entire implementation, which results in conclusions that will be implemented in further professional practice. Changes should be treated as necessary, for the continued existence better late than later, this is the slogan that should guide us in further project activity [10].

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2 Research Method The presented method is based mainly on the study of literature and a comparison of two specific cases of the method of erecting buildings in wooden construction. Two completed objects in a wooden structure were selected for the analysis. The first erected using CNC technology (Cross laminated WOOD) the second in wood frame construction. The analysis was based on the Life Cycle Assessment (LCA), which relays on the assessment of the entire life cycle of the building or its main elements [11]. The PN-EN 15978:2012 standard defines the individual phases of the analysis. Phase I (A1–A3) – in this process, CO2 emissions from the moment of extraction, processing and production of construction products are analyzed. Phase II (A4 and A5) – takes into account the process of erecting objects, taking into account the incorporation of individual materials and their transport from the facility to the construction site as well as indirect storage. Phase III (B1–B7) – this is the service life of the facility, until the end of the service cycle. In the process of use, aspects such as cooling, electricity consumption, heating, water supply, and waste disposal are taken into account. Phase IV decommissioning (C1–C4) – which begins when the building ends its useful life, ceases to be used and can be demolished. Phase V of External Influence (D) – demolition is treated as a process resulting in materials that can be a source of recycling. The emission analysis was carried out for the first phase (A1–A3) in terms of two types of wooden structures. The subject of the research are original projects erected in years 2020–2022 in the field of professional practical activity. The objects were carried out in parallel, which allowed for the simultaneous analysis of both cases. The time in which both objects were completed was not included into the in analysis, however, the author notes that the shorter period of implementation of the object with the help of massive wood technology was a big advantage of this solution. 2.1 Frame Construction The first example studied is a wood frame structure. Residential building commissioned by a private investor. It was designed in a lightweight frame construction due to economic and time conditions. The rapid implementation of the facility significantly affected the possibility of launching the function provided for the building and thus increased the profitability of the intention implemented by this investment project. The facility was built within 8 months of obtaining the permission. During the task, foundations were erected, a prefabricated structure was assembled, and the object was finished both inside and outside. CO2 emission tests were carried out in module A1–A3 in the first phase in accordance with PN-EN 15978:2012 (Figs. 1 and 2).

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Fig. 1. a) Arrangement of structural walls with window openings. b) Wall layers, frame structure. Author: Kamila Wilk, source: own archive.

Fig. 2. a) View from the assembly of the frame structure. b) View of the implementation of the frame structure. Photo: Kamila Wilk, source: own archive. Table 1. GWP module A1–A3. Material

Impact/m3 kg CO2 eq/m3

Volume [m3 ] Area [m2 ] Thickness [mm] Results [kg CO2 eq]

Paint

2,851.0

0.2

Gypsum board

169.6

200.00

1

5.95

570.2 1,009.1

Constr.tim

−680.0

22.75

MDF

−669.0

0.59

198.29

3

−398.0

40

200.00

200

2,816.0

0.20

200.00

1

Stone wool PP roof..mem Total

70.4 271.5

120

−15,470.0

54.3 −11,418.4

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2.2 Mass Timber Structure The second example covered by the research is a massive wood construction. The building with a residential function was designed for the needs of an individual recipient. The function has been adapted to the user’s requirements. The investment process was carried out at the beginning of March 2020 in the initial phase of pandemic restrictions. The investor wanted to minimize the implementation time and adapt the form and function of the building to his preferences. Thanks to the preparation of technological and installation openings and aperture needed for the implementation of window joinery during prefabrication, it was possible to order materials in advance and build them immediately after the assembly of the house. The facility was completed within four months of obtaining the building permit. This enabled quick implementation [12, 13]. The GWP module A1–A3 was used for the calculations, the A4–A5 phase was not analyzed due to the comparable mounting method in terms of emissions [kg CO2 eq/m3 ]. After making the calculations, the CO2 emissions during the production of materials necessary to erect the structure of the object were demonstrated in Tables 1. and 2..

Fig. 3. a) Arrangement of structural walls with window openings. b) Wall layers, mass timber structure. Author: Kamila Wilk, source: own archive.

Fig. 4. a) View from the assembly of the structure of massive wood. b) View of the completed building. Photo: Kamila Wilk, source: own archive.

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Table 2. GWP module A1–A3. Material

Impact/m3 kg CO2 eq/m3

Volume [m3 ]

CLT

−664.0

Stone wool

70.4

Construction timber

−680.0

Total

Area [m2 ]

Thickness [mm]

Results [kg CO2 eq]

30

100

−19,920.0

15

200

1,056.0

120

−3,400.0 −22,264.0

3 Results The result of the analysis is to maintain the assumption. The mass timber wood construction is a structure with lower CO2 emissions than frame structure. It does not have to be machined with finishing materials (Figs. 3 and 4). An additional advantage of the implementation of a building object in this construction is the acceleration of the erecting process [14–17]. It is also worth noting that for the production of mass timber wood it is possible to use diseased trees and windbreaks, which are not possible to use in the production of frame construction. The use of wood that occurs locally reduces transport costs and additionally inscribes the object in the environment using the context to create the project. According to the author’s assumption, the study confirmed the observations related to the erection of both objects, and indicated the more ecologically efficient one. It should be noted, however, that it would be necessary to examine the remaining phases of the building’s life cycle. This would allow for a broader comparison between the two technologies.

4 Conclusions A lower carbon footprint generates lower costs. Architects and constructors should think about this aspect during the implementation of concepts and construction projects. Public awareness of emissions is growing. Massive wood structures perfectly fit into the ideas of sustainable construction. It replaces other construction materials due to low CO2 emissions and convenience of implementation. Solid wood components are manufactured off-site according to precise specifications, which reduces waste, speeds up construction and reduces labor costs; this also results in improved energy efficiency, as a controlled production process means that a better building partition can be built with fewer air gaps and connections generating energy losses [18, 19]. Prefabrication does not limit the creativity of architects, it allows us to implement bold design ideas, using a beautiful material such as wood. As an organic element of interior finishing is a natural material for man, which causes a very good reception of the object erected in this technology by the user [20]. It should be noted that it would be necessary to analyze the determinants related to CO2 emissions

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when considering the entire life cycle of a building. Each of the stage of material production as well as other phases such as exploitation, demolition, or recycling. Particularly important for the issue related to sustainable construction is social awareness.

References 1. Williamson, T.J., Williamson, T., Radford, A., Bennetts, H.: Understanding Sustainable Architecture. Taylor & Francis, Milton Park (2003) 2. United Nations 2020 Global Status Raport For Buildings and Constructions. https://globalabc. org/resources/publications/2020-global-status-report-buildings-and-construction. Accessed 31 May 2022 3. ZabalzaBribián, I., ValeroCapilla, A., ArandaUsón, A.: Life cycle assessment of building materials: comparative analysis of energy and environmental impacts and evaluation of the eco-efficiency improvement potential. Build Environ. 46(5), 1133–1140 (2011) 4. Pajchrowski, G., Noskowiak, A., Lewandowska, A., Strykowski, W.: Wood as a building material in the light of environmental assessment of full life cycle of four buildings. Constr. Build. Mater. 52, 428–436 (2014). https://doi.org/10.1016/j.conbuildmat.2013.11.066 5. Petersen, A.K., Solberg, B.: Environmental and economic impacts of substitution between wood products and alternative materials: a review of micro-level analyses from Norway and Sweden. For. Policy Econ. 7(3), 249–259 (2005) 6. Gustavsson, L., Sathre, R.: Variability in energy and carbon dioxide balances of wood and concrete building materials. Build. Environ. 41, 940–951 (2006) 7. Upton, B., Miner, R., Spinney, M., Heath, L.S.: The greenhouse gas and energy impacts of using wood instead of alternatives in residential construction in the United States. Biomass Bioenergy 32, 1–10 (2008) 8. Štefko, J., Múˇcka, M.: Environmentální vlastnosti dˇrevˇených stavebních konstrukcí In.: Sborník z konference Dˇrevostavby, Volynˇe, pp. 213–214 (2008) 9. Lindblad F.: Market structure and economic status for firms producing wooden single-family houses in Sweden. Licentiate Thesis no. 47/2016, Faculty of Technology, Linnæus University, Växjö, Sweden, 122 (2016) 10. Matová, H., Kaputa, V.: Attitudes of active and upcoming architects towards wood: the case study in Slovakia. Acta Facultatis Xylologiae Zvolen 60(2), 197–207 (2018) 11. PN-EN 15978:2012. Sustainable construction works – Assessment of the environmental performance of buildings – Method of calculation 12. Lindblad, F.A.: Comparative study of the environmental impact from transportation of prefabricated building elements using wood or concrete. Int. J. Eng. Technol. 11(3), 154–161 (2019) 13. Alazzaz, F., Whyte, A.: Linking employee empowerment with productivity in off-site construction. Eng. Constr. Archit. Manag. 22(1), 21–37 (2014) 14. PN-EN 13986. Wood-based panels for use in construction – Properties, conformity assessment and marking 15. Shahnewaz, M., Tannert, T., Alam, M.S., Popovski, M.: Capacity-based design for platformframed cross-laminated timber buildings. In: Structures Congress; ASCE, Denver, CO, USA, pp. 430–433 (2017) 16. Silva, C.V., Branco, J.M., Lourenço, P.B.: A Project Contribution to the Development of Sustainable Multi-Storey Timber Buildings. Universidade do Minho, Braga, Portugal. Portugal SB13 - Contribution of Sustainable Building to Meet EU 20-20-20 Targets, pp. 379–386 (2013)

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17. Popovski, M., Gavric, I.: Performance of a 2-Story CLT House Subjected to Lateral Loads. J. Struct. Eng. 142(4), E4015006 (2016) 18. Ehrhart, T., Brandner, R., Schickhofer, G., Frangi, A.: Rolling shear properties of some European timber species with focus on cross laminated timber (CLT): test configuration and parameter study. In: Timber Scientific Publishing, pp. 61–76 (2015) 19. Buchanan, A., Deam, B., Fragiacomo, M., Pampanin, S., Palermo, A.: Multi-storey prestressed timber buildings in New Zealand. Struct. Eng. Int. 18(2), 166–173 (2008). https:// doi.org/10.2749/101686608784218635 20. Steinhardt, N.: Liao: an architectural tradition in the making. Artibus Asiae 54(1/2), 5–39 (1994)

Rural Building in Opole Silesia During the Period of the Frederician Colonization Dariusz Bajno1(B) , Marcin Fiutak2 , and Maciej Niedostatkiewicz3 1 Faculty of Civil and Environmental Engineering and Architecture, Department of Building

Structures, Bydgoszcz University of Science and Technology, Bydgoszcz, Poland [email protected] 2 Przedsi˛ebiorstwo Projektowo-Usługowo-Handlowe „MARDOM” Marcin Fiutak, Byczy´nska 27, 46-200 Kluczbork, Poland 3 Faculty of Civil and Environmental Engineering, Department of Engineering Construction, Gda´nsk University of Technology, Gda´nsk, Poland

Abstract. The aim of the article is to present the rural construction in the Opole countryside (Poland) during the “Frederician colonization”. Showing socioeconomic areas of Opole villages, differing from each other, the differences of which resulted from the environmental conditions influencing the processes of enfranchisement and industrialization. These processes at the turn of the 17th and 19th centuries influenced the image of the Opole and Upper Silesian villages as well as the development of skeleton and brick construction. By gradually analyzing individual regions of Opole Silesia, one can notice large-scale changes in rural construction, differences in the population density and multiculturalism of the inhabitants of the Opole region, showing the complex settlement process, the effects of which can be seen in the Opole countryside to the present day. The article will discuss the process of “Frederician colonization” and its impact on the change of architecture in the region, as well as the accompanying construction and conservation techniques. A look at regional construction from the turn of the 17th and 19th centuries brings closer the cultural heritage of the region’s architecture, allows you to deepen your knowledge of architectural forms and techniques of erecting objects of a bygone era, thus enabling the protection of the historical substance of rural regions of Opole Silesia. The article will also discuss the degree of preservation of the historic substance described above along with the forecast of their durability in the further period of operation. Keywords: Frederician colonization · Opole village · Frederick II · Colonization processes · Half-timbered construction · Colonial style

1 Introduction The architecture of the Opole rural areas (in the Opole Voivodeship, Poland) until the turn of the 18th and 19th centuries was created based on the natural climatic and geographical conditions in this region. The intense afforestation occurring here satisfied the raw material needs of cheap and simple log construction and section with timbered, © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 234–241, 2023. https://doi.org/10.1007/978-3-031-26879-3_19

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based on easy carpentry processing of the building, which corresponded to the requirements of the local climate and also the logistic and technical possibilities of the then level of construction art. This style of construction had a centuries-old tradition in Silesia and was used here practically from the 10th century. This homogeneous development of a folk character in Opole Silesia changed as a result of the settlement processes that lasted from the second half of the 18th century to the mid-19th century when frame and brick construction developed. The beginning of this process dates back to the “Seven Year’s War” when there was a rapid increase in settlement. In 1773, Frederick II (King of Prussia) issued a declaration supporting the settlement process in Silesia [2]. Its purpose was to bring new settlements to the territory of Silesia, necessary for the launch and development of factory production [8]. New agricultural settlements, forestry, craft, and mining settlements began to emerge (Fig. 6). It is estimated that in the first period of colonization, about 14–15 thousand colonists came to Silesia, who created about 230 colony settlements [3]. Thanks to these intensified colonization processes, combined with the subsequent development of heavy industry, Silesia became one of the most industrialized lands in Germany and Central and Eastern Europe in the mid-nineteenth century. All this had a significant impact on the character and appearance of the then Silesian village. In the 17th century, in Upper Silesia, almost no other construction than the coronary one was known. It was only the Prussian colonization that introduced new building techniques and regulations as well as fire protection regulations concerning fire safety in forested rural settlements that were innovative at that time [2]. Ultimately, it was slowed down and even stopped only by the Napoleonic Wars [6]. For historical reasons, German-language literature was used when writing the paper. It is significant for the period of colonization which was taking to consideration [9].

2 Materials, Methods, Scope, and Aims of the Paper The aim of paper is characteristics of the described buildings, therefore own research was carried out, consisting in an inventory of the buildings and adequate literature studies [8]. The subject of the analysis are the objects, the location of which is shown in Fig. 1. In the newly created rural settlements, simple and economical frame construction began to be used, commonly known and used in other parts of Europe. It was different from the traditional one due to its features, which were determined by the nature and economic situation of a given region. The main advantage of this construction was the easy availability of the raw material, slight (not noticeable) deformation of the building structure (the main reason of which was wood swelling, a small amount of wood material (very little waste) and, above all, the simplicity of the structure and the high rate of erection of buildings (speed of filling the skeleton of the walls) [4]. The so-called “Prussian” province style in rural construction differed significantly from the technology of traditional timber-frame construction, used in other European countries at the turn of the 18th and 19th centuries [6, 23], and it was characterized by an extraordinary simplicity of form. It was also a time of rapid changes in the region. Until the mid-19th century, 471 province were established in Silesia, including as many as 225 in the Opole region [9]. As a result of changes also in the construction industry, the image of the

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Opole village changed significantly [24–26]. This process lasted until the outbreak of World War II, which should already be considered the end of half-timbered construction in this region. The above-mentioned simple construction processes of the “Frederician colonization” construction and the types of materials used (susceptible to environmental conditions and biological degradation) caused rapid technical wear and tear of these historic buildings [1]. Currently, there is possible to admire only single buildings that witness the culture and technology of a bygone era, such as the Church of the Blessed Virgin Mary in Radomierowice (Figs. 2 and 3), the Saint Jack’s Church in Wierzbica Górna (Fig. 5a), the bell towers of the church in Ko´scierzyce (Fig. 5b), and the housing estates, for example, in Radawa (Fig. 6) [8].

Fig. 1. Localization of the considered buildings: a) location of the Opole Voivodeship on the map of Poland, b) location of the considered objects within the Opole Voivodeship: 1 – Ko´scierzyce, 2 – Radomierowice, 3 – Ligota Turawska [30].

Fig. 2. Visualization and frame model of the church building in Radomierowice (from Marcin Fiutak’s archive).

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Fig. 3. The current appearance of the building with the visible deformation of the roof and the frame load-bearing structure infested by fungi and insects (from Marcin Fiutak’s archive).

The frame-timbered construction of the colonization period did not withstand the test [30] of time due to the lack of care for this cultural heritage on the part of the then authorities, as well as the lack of conservation and repair techniques for wooden structures (Fig. 3). Archival materials relating to the half-timbered folk architecture in Opole Silesia are very sparse, and it has not been properly documented (Fig. 4) [6, 27].

Fig. 4. Fragments of archival documentation – a rural cottage in a frame construction: a) elevation [2], b) beams with their markings [9], c) room arrangement [2].

The remaining small number of buildings from the period of “Frederician colonization” (Figs. 1, 2, 3, 5, and 6) no longer reflects the proper character and image of the diversity of architectural forms used in this region, but nevertheless, it should be recognized that folk construction framework, initiated by German settlers, had a great influence on the appearance and character of the then and contemporary Opole village [2].

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Fig. 5. View of the church building: a) in Wierzbica Górna, b) in Ko´scierzyce (from Marcin Fiutak’s archive).

3 Development of Frame Construction and Brief Description of Frame Structures Used in Opole Silesia In Opole Silesia, i.e. the region of southern Poland of mainly lowland character, a form referring to the post structure was used, the characteristic feature of which was the system of vertical columns constituting the main frame of the building. It was a kind of frame structures in which the ground beam, cap, columns with transoms, and stiffening struts formed the structural and skeletal system of vertical partitions of buildings. Their foundations were mainly made of oak wood whose task was to transfer the building’s own weight as well as operational and climatic loads to the foundation. For joining and stiffening wooden structures, notches and tenons, as well as a wooden frame reinforced with braces were used. The corner elements of the structure significantly influenced the deformation of walls and roof trusses [4]. The weak nodes in the skeletal structures were the corner joints, including the foundations (the latter due to corrosion, i.e. the brown distribution of wood and considerable pressure on the beams in the direction across the grain). Figure 6 shows historic residential buildings that are properly maintained and kept in good condition [5]. The corner columns of the frame structures often had a much larger cross-section than the middle columns. In frame structures used in Opole region construction, struts were placed only in corner fields, unlike structures in other parts of Europe where they were also used in fields adjacent to the corner fields and in the middle fields of external walls. The outer posts were divided with transoms. All transoms in the structure were connected with mullions and struts, they also replaced traditional lintels and window sills (Fig. 3). Each wall of the frame structure was ended with caps. The tacks were tied together at the corners like ground beams. Ceiling beams and elements of the roof structure were supported on the gratings. All these construction forms presented a large variety of construction solutions, from simple and primitive to more complex and additionally decorated. It was wrapped with straw and covered with a mixture of chaff and clay. The whole was mash smooth and then whitewashed. Later frames construction were filled with raw or burnt brick on lime mortar. This resulted in a mixed development including

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Fig. 6. Well-preserved and maintained residential buildings modelled on the colony villages of the housing estate in the southern Poland – still used for their intended purpose (from Marcin Fiutak’s archive)

the technology of log structures (log, butt), frame, brick, and half-timbered structures [28, 29]. As a consequence of this, mixed constructions were created, such as transom and brick, transom and timber, log and brick, and half-timbered and brick [5].

4 Discussion Until now, in Opole Silesia, few original witnesses of frame construction from the period of “Frederician colonization” have survived. Currently, we can admire only single objects from that period and also notice a clear influence of the consequences of urban solutions in the villages of the colony. The small, remaining number of buildings does not provide the proper spectrum, which proves the diversity of construction and architectural forms. However, it can be observed that the frame construction and urban solutions initiated by German settlers influenced the character of the Opole village. The low resistance of the building material and the simplicity of execution have undergone the destructive effects of weather and parasitic influences, disasters and fires. The tendency to remove monuments of construction in favor of brick construction has also caused such a small number of monuments of wooden construction in the landscape of the contemporary Opole village. The goal of each generation should be to preserve the cultural values of the past periods, including architecture and techniques used in construction.

5 Summary It often happens that buildings in a high stage of technical degradation are left to themselves, which is more and more often the reason for their disappearance from our landscape. Applying the principle that it is better to build anew than to save a damaged building substance, we tacitly consent to the disappearance of such “traditions”, without even having documented solutions used in the construction industry at that time. Such an infamous example is the disappearing technical culture of “Frederician colonization” construction. The distinctive sacral buildings are kept in relatively good technical condition, while the low homestead buildings fall into oblivion and annihilation. None

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of the newer solutions used in Opole Silesia is inspired by “Frederician colonization” architecture. There is a lack of awareness of the preservation or only documentation of a small region of southern Poland for subsequent generations of the past period in the local construction industry. This article draws attention to the disappearing fragment of history and at the same time inspires the authors to conduct research documenting those not very technically advanced solutions used in rural construction, which in a moment may completely disappear from our surroundings or be modernized [1, 7]. For future generations, the knowledge of the applied techniques of solutions for architectural and construction details, which are not present in modern construction today, should be preserved. In the past, they mainly determined the comfort and safety of use [10–22].

References 1. Gładyszowa, M.: Górno´sl˛askie budownictwo drewniane. Wrocław (1978) 2. Gładyszowa, M.: Górno´sl˛askie budownictwo ludowe. Wrocław (1978) 3. Chrzanowski, T.: udownictwo wiejskie Ziemi Opolskiej. In: Ziemia, vol. II, pp. 123–139 (1966) 4. Mielnicki, S.: Ustroje budowlane, Spółdzielnia Wydawnicza „Meta”, Katowice (1949) 5. Kopkowicz, F.: Ciesielstwo Polskie, Arkady, Warszawa (1958) 6. Czerwi´nski, T.: Budownictwo ludowe w Polsce, Wydawnictwo MUZA SA, Warszawa (2006) 7. Zabierzowski, A.: Praktyczne budownictwo wiejskie. Reprint from 1956, Volumina, Radom (2015) 8. Queries and authors’ own research 9. Schlenger, H.: Landlicher Sidlungen in Schlesien. Beitrage zur Morphologie der schlesischen Kulturlandschaf, Wrocław (1930) ´ aska Opolskiego: od s´redniowiecza do ko´nca w. 10. Chrzanowski, T., Kornecki, M.: Sztuka Sl˛ XIX, Wydawnictwo Literackie, Kraków (1974) ´ asku, Ossolineum, Wrocław (1975) 11. Matuszczak, J.: Ko´scioły drewniane na Sl˛ 12. Trocka, E., Wiatrzyk, S.: Konstrukcje s´cian w budownictwie regionalnym. Przegl˛ad Budowlany, 1 (1984) 13. Salmonowicz, S.: Prusy – Dzieje Pa´nstwa i Społecze´nstwa, Wydawnictwo Pozna´nskie, Wrocław (2002) 14. Zimmermann, K.: Fryderyk Wielki i jego kolonizacja rolna na ziemiach polskich, vol. 1–2, Drukarnia s´w. Wojciecha, Pozna´n (1915) 15. Juros, J.T.: W dolinie Małej Panwi. Historia fryderycja´nskiej osady hutniczej Ozimek/Malapane od 1754 do 1945 roku, Stowarzyszenie Dolina Małej Panwi, Ozimek (2010) ´ aski w Opolu, Opole 16. Golachowski, S.: Studia nad miastami i wsiami s´l˛askimi. Instytut Sl˛ (1969) 17. Adamiec, F.: Przemiany społeczno-ekonomiczne na wsi opolskiej w XIX w. i I połowie XX w. Wczoraj Dzisiaj Jutro. 2, 12–26 (1970) 18. Biesiekierski, T.: Studium architektury regionalnej wsi Dolno´sl˛askiego Pasma Sudetów. Politechnika Wrocławska, Wrocław (1979) 19. Gil, E.: Z bada´n Muzeum Wsi Opolskiej nad drewnianym budownictwem ludowym Opolszczyzny. Opolski Rocznik Muzealny, VIII (1985) ´ asku Opolskim. 20. Trocka-Leszczy´nska, E.: Architektura wiejskich ko´sciołów drewnianych na Sl˛ Architectus 2(8), 43–53 (2000) 21. Zin, W.: Typy i formy w polskiej architekturze drewnianej, Kraków (1956)

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22. Chilczuk, M.: Osadnictwo wiejskie. Metody bada´n koncentracji zabudowy i kształtów wsi. PWN, Warszawa (1975) 23. Adamczyk, K., Brol, J., Malczyk, A.: Wady podparcia drewnianej konstrukcji dachu w drewnianym ko´sciele. In: XXVI Konferencja Naukowo-Techniczna Awarie Budowlane – 2013, Szczecin – Mi˛edzyzdroje, pp. 143–150 (2013) 24. Białkiewicz, A.: Zagro˙zenia obiektów sakralnych. In: XXX Konferencja NaukowoTechniczna Awarie Budowlane – 2022, Szczecin – Mi˛edzyzdroje, pp. 21–32 (2022) 25. Błaszczy´nski, T., Oleksiejuk, H., Firlei, E., Błaszczy´nski, M.: Wielostopniowy monitoring i zabezpieczenie budynków pod ochron˛a konserwatorska przed awari˛a lub katastrof˛a. In: XXV Konferencja Naukowo-Techniczna Awarie Budowlane – 2011, Szczecin – Mi˛edzyzdroje, pp. 395–402 (2011) 26. Hejducki Z., Sawicki M.: Wybrane zagadnienia remontów pokry´c dachowych. In: VII Konferencja Naukowo-Techniczna Problemy Remontowe w Budownictwie Ogólnym, WocławSzklarska Por˛eba, pp. 87–94 (1996) 27. Nowogo´nska B.: Przyczyny stanu przed awaryjnego XVII wiecznego ko´scioła drewnianego. In: XXIV Konferencja Naukowo-Techniczna Awarie Budowlane – 2009, Szczecin – Mi˛edzyzdroje, pp. 217–225 (2009) 28. Tajchman J., Jurecki, A.: Historia technik budowlanych. Fundamenty, rusztowania, mury, wi˛ez´ by, sklepienia. PWN, Warszawa (2020) 29. Trochonowicz, M.: Wilgo´c w obiektach budowlanych. Problematyka bada´n wilgotn´sciowych. Budownictwo i Architektura 7(2), 131–144 (2010) 30. https://pl.wikivoyage.org/wiki/Wojew%C3%B3dztwo_opolskie. Accessed 25 Nov 2022

Other Issues: Wind Engineering, Innovative Solutions, Reconstruction, and Organization of Building Site

Wind-Based Form-Finding of a Tensegrity Pavilion Lenka Kabošová1(B)

, Tomáš Baroš2 , Stanislav Kmeˇt3 and Dušan Katunský2

, Eva Kormaníková3

,

1 Technical University of Košice, CRIC – Center for Research and Innovation in Construction,

Park Komenského 10, 042 00 Košice, Slovakia [email protected] 2 Civil Engineering Faculty, Institute of Architectural Engineering, Technical University of Košice, Vysokoškolská 4, 042 00 Košice, Slovakia 3 Civil Engineering Faculty, Institute of Structural Engineering, Technical University of Košice, Vysokoškolská 4, 042 00 Košice, Slovakia

Abstract. The subject of this paper is a nature-driven digital architectural formfinding of a small-scale pavilion made of bi-pyramidal tensegrities covered in a textile membrane. Inspired by nature, we propose to create architecture as an organism that evolves based on ambient influences. Within the form-finding process, digital software, such as Grasshopper for parametric designing, plug-ins such as Karamba 3D for structural analysis, and Computational Fluid Dynamics (CFD) simulation plug-in Butterfly for Grasshopper are employed. The wind force and the resulting wind surface pressure influence the specific spatial configuration of the pavilion structure and are the main form-influencing factor affecting the attachment of new layers of tensegrities. The structural performance is tested recursively, with the influence of the 8.5 and 26 m/s wind velocity, the corresponding overall deflection of the pavilion’s structure, and the shape optimization in one design loop. Only the best-performing option is subsequently analyzed in the 26 m/s wind employing CFD simulation. The proposed design process is presented through a case study in Košice, specifically on the site of the Technical University campus. Keywords: Tensegrity · Form-finding · Weather data · Grasshopper · CFRP

1 Introduction 1.1 Digital Design in the Climate Change Era The hardware and software continual evolvement has enabled expanding the search for innovative approaches and unconventional solutions in architectural design. Platforms like Grasshopper for Rhino allow parametrically designing architecture and encompassing other disciplines, including structural engineering, mathematical optimization, or environmental simulations. Nowadays, when changing climate is not just an abstract concept [1–3], a multidisciplinary architectural design approach during the first design steps is pivotal and leads to sustainable architectural designs [4–7]. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 245–257, 2023. https://doi.org/10.1007/978-3-031-26879-3_20

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The reciprocal relation of the wind and architectural shape and principles of leveraging the wind to improve the local microclimate is still an unexplored design territory. If the wind flow and its effects on architecture are investigated early in the conceptual design phase, environment-driven design choices result [8–11]. Entwining factors from multiple disciplines, such as architecture, structural engineering, meteorology, and others, into the architectural design process during the first design phases, the weather- and structural-related performance can be predicted, and adjustments made until the desired performance is reached. To mention just few demonstrations, by leveraging architectural and urban planning and digital tools, pedestrian wind comfort can be achieved [12–14], natural ventilation can be supported [8, 15, 16], or optimum sun exposure attained [17–19]. 1.2 Generative Design and Optimization Ideally, the digital design approach should be time-saving, leading to fast, informed, and sustainable design decisions [20]. Considering multiple alternatives with various structural or environmental performances demands implementing optimization methods and genetic algorithms into the architectural design [21–23]. Associating the approach with the principles of natural selection involves, in architectural design too, a gradual evolution towards the best-performing solution(s) under specific ambient conditions. Evolutionary techniques help systematically explore many possible solutions and thus lead the process to improved designs that meet stated design goals. Novel shape configurations, otherwise overlooked, can be generated [24]. The new architectural shape grammar can originate from this approach, expanding the design space. Nevertheless, design optimization (a formulated problem with an expected convergence) should be distinguished from the design exploration (an assumption about the system evolution without any proof of convergence) [25]. 1.3 Muscles/Bones: Design of Tensegrity Structures In nature, physical stresses act on living organisms, which, as an answer, have developed strategies to reduce the impact. When we link this idea to technical systems, the motivation is to design lightweight, flexible, yet strong structures that can withstand higher acting forces. Tensegrities have the potential to change their shape configuration as a response to acting load [25, 26] while reconfiguring the equilibrium of forces. Within such a system, struts must always be under compression like bones in organisms; cables under tension are like muscles connecting and controlling the bones. Few architectural applications of this innovative structural system include experiments with kinetic tensegrity facades, responsive to daylight and solar exposure [27], or regulating interior ventilation [28]. 1.4 CFD in the Tensegrity Structures Design The function of tensegrity structures as windbreaks was tested through digital Computational Fluid Dynamics (CFD) simulations as well as physical prototyping [29, 30]. Employing the tensegrity structure as a blockage in a narrow gap between two buildings demonstrated that such structures could act as windbreaks and enhance pedestrian wind

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comfort by interfering with the Venturi effect and dissipating the wind flow. Moreover, the design process was tangled with the optimization of the spatial configuration of the tensegrity cluster [29]. Tensegrities covered with textile membranes used as building envelopes could, through the passive shape change in the wind flow creating a “dimpled” building surface, favorably impact the wind comfort around buildings [30]. 1.5 The Pavilion An idea of a lightweight pavilion consisting of bi-pyramidal tensegrity units and a textile membrane as an envelope is explored in this paper. The vision is to design the pavilion on the Technical University of Košice campus, protecting the pedestrians from the wind gusts (Fig. 1). The pavilion is designed by employing progressive software in the form-finding process.

Fig. 1. The location of the tensegrity-membrane pavilion within the Technical University of Košice campus.

2 Design Method The design process is purely digital, leveraging the capabilities of the digital techniques. There are six steps within the proposed weather-driven design process: 1) Obtaining weather data. 2) Digital geometry and material definition of one tensegrity unit. 3) Attachment of tensegrities based on pre-defined rules. 4) Application of loads (tensegrity system). 5) Structural analysis (tensegrity system). 6) CFD simulations (membranes). 2.1 Obtaining Weather Data The design process is nature-driven. In the digital environment, the pavilion responds to the weather conditions while its shape is just forming. As the first step, weather data are acquired from the EnergyPlus web database [31]. The wind rose for Košice, generated through Ladybug for Grasshopper, shows strong winds, which are 33.46% of the time northerly. In the simulations, this wind direction with the wind speeds of

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8.5 m/s represents the median value of the standard wind conditions, whereas 26 m/s is the fundamental value of the basic wind velocity for this part of Slovakia (taken from the National Annex of the European standard, Actions on structures, Part 1–4: general actions, Wind actions) [32]. 2.2 Digital Geometry and Material Definition of One Tensegrity Unit One bi-pyramidal tensegrity unit, made of CFRP (Carbon-Fiber-Reinforced-Polymer) struts and steel tendons, is designed through the particle-spring solver Kangaroo for Grasshopper, through which the equilibrium state is found for the specified geometry and the material parameters, employing dynamic relaxation. The square CFRP frame of the bi-pyramidal tensegrity has a 0.5 m-long arm, and the CFRP middle rod can have a length ranging from 0.25 to 0.5 m. The equilibrium state in a geometrically non-linear static analysis is expressed by [25]: KT · u = F

(1)

KT = KE + KG

(2)

where K T is a tangential stiffness matrix, u is a vector of nodal displacements, F is a vector of nodal loading forces, and K E and K G are elastic and geometric stiffness matrix respectively (all in global coordinates). The material properties of struts and tendons are listed in Table 1. Figure 2 depicts the bi-pyramidal tensegrity unit. Table 1. Material properties as defined in Kangaroo and Karamba 3D. CFRP struts (Orthotropic material)

Steel tendons (Isotropic material)

E 1 [GPa]

142

210

E2 [GPa]

10.3

–

G12 = G31 = G32 [GPa]

7.2

80.76

ν [-]

0.27

–

γ [g/cm3 ]

1.58

7.85

f T1 = f C1 [MPa]

± 2 280

± 235

f T2 = f C2 [MPa]

± 57

–

t 12 [MPa]

71

–

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Fig. 2. One bi-pyramidal tensegrity unit. The frame measures 0.5 m × 0.5 m.

The tensegrity unit designed this way is modified by transformation components (rotation and movement) to create a ring chain. One row of the ring chain represents 35 units placed on a circumscribed circle of the polygon (octatriacontagon). 2.3 Attachment of Tensegrities Based on Pre-defined Rules In the Grasshopper script, other bi-pyramidal units are added to create a cylindrical pavilion structure, utilizing the parametric nature of Grasshopper (Fig. 3). The cylindrical structure measuring 2.9 m in height and 7.7 m in length is created by the first layer of 175 tensegrities, 5 rows of 35 tensegrities creating the circle. The formation of the second, third, fourth, and fifth rows is performed by a vector movement (representing the displacement of the lower corner of the rectangular frame to the opposite corner of the opposite tendon).

Fig. 3. The digitally-created tensegrity-membrane pavilion through the algorithmic design software Grasshopper.

Each first-layer tensegrity unit is coated with a textile membrane to form “cushions” while the membrane made of 5 segments spans through the interior of the semi-open pavilion.

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The second layer of tensegrities follows, connecting to the structure of the previous layer, based on the pre-defined rules: tensegrity units from the second layer are shifted into the gaps between the units of the first layer, rotated 45° around the middle rod, and pushed outwards from the center. Additional selected units in three alternatives represent the third layer. The resulting composition of the third layer units is determined to be the most suitable in terms of structural engineering load impact assessment. The third layer consists of membranecoated tensegrities, which are located: a) randomly, b) only on the windward side of the structure, and c) regularly (diagonally) on the cylindrical structure, connecting to the second layer. The shifted tensegrity units are rotated (–30°) along individual axes so that there are no collisions between the individual units (Figs. 4 and 5). The three options, each with a different placement of the third layer of tensegrities, will be examined further. The pavilion’s foundation is envisioned as 35 bored micro-piles 2.5 m underground. The lowest row of the first pavilion layer is anchored to the support in the bottom corner of each tensegrity unit.

Fig. 4. The layers of the tensegrity-membrane pavilion. The first layer is displayed in yellow, the second in blue, and the third in green.

The structure without membranes subsequently enters the process of structural engineering assessment, the outcome of which is the resulting overall displacement of the designed tensegrity system due to the considered load and, therefore, the best-performing option out of the three tested variants can be selected. 2.4 Application of Loads Through Karamba 3D (Tensegrity System) In the next step, employing Karamba 3D for Grasshopper, the pavilion’s tensegrity structure is subjected to the wind force. The material properties and characteristics are taken from Table 1. The pretension of steel tendons and gravitational force are applied as well. Two wind speeds (transformed to wind force) are used for loading the structure: a) 8.5 m/s, representing the average wind gusts in Košice, and b) 26 m/s, representing the wind speed used for designing the structures in this part of Slovakia, following the EU standard [32]. In Karamba 3D, the wind force is applied through the Point load, where the input is required as a vector load in kN. First, the Bernoulli equation is employed to convert the

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Fig. 5. The tensegrity-membrane pavilion. The “cushions” are placed at the first and the third layer of tensegrities. The inner membrane consists of 5 parts.

wind speed to the wind pressure in Pascal, depending on the air density and velocity. That is followed by the expression for calculating the windward side area of the cylindrical surface. Last, the resulting pressure acting on the surface is divided by the number of points loaded by the wind force. 2.5 Structural Analysis (Tensegrity System) The tensegrity-membrane pavilion consists of the tensegrity structure and textile membranes that create protection from the wind. Five segments of the inner textile membrane are anchored to the endpoints of the upper and bottom tensegrities of the first layer. The spatial membrane “cushions” are coating the first and third-layer tensegrities to create a semi-enclosed pavilion. The Kangaroo 2 physical engine is used to form-find these membranes. Through the parametric environment, the height of the tensegrities of the first layer can be varied, influencing the dimensions of other tensegrity layers. The permeability of the structure is thus adaptively variable. That, however, is not a subject of experimentation in this paper. The structural response of the tensegrity system, excluding membranes, is observed, employing Karamba 3D. The total deflection of the pavilion’s structure is inspected. In the analysis, the material (CFRP and steel) and geometric (cross-sections) characteristics of the input tensegrity elements, as well as loads (gravity, wind force, pretension) are defined through components. Subsequently, a finite element model from given entities is created, which is assessed by the Analyze component, which calculates the deflections of a given model using first-order theory.

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Three different options for the placement of the third-layer tensegrity components are evaluated through Karamba 3D: a) randomly placed, b) placed only on the windward side of the structure, and c) placed regularly (diagonally) on the cylindrical structure (Fig. 6). The numerical model in Karamba 3D is represented exclusively by beam elements. Each group of elements, ergo vertical rods, tensegrity units’ frames, and tendons, is defined as a beam structural element with different material and structural properties (Young’s modulus, cross-section geometry, etc.). One tensegrity unit has five struts and eight tendons. The pavilion is composed of a three-layer form. The first, second, and third layers have 875; 850; and 285 struts and 1,400; 1,360; and 456 tendons, respectively. Altogether, it is 3,216 cables and 2,010 struts for the Option c of the third-layer tensegrity (Fig. 6). In the future, the plan is to include the membrane as a shell element in the structural analysis.

Fig. 6. Three placement options for the third-layer tensegrities. The wind force is depicted with blue arrows. Membranes are displayed as well but are omitted from the structural analysis.

The displacements of the tensegrity structure under the acting wind force representing the 8.5 and 26 m/s winds are observed and described in Table 2. Table 2. Options a–c and the overall displacements of the tensegrity structure (without membrane) in the wind. Option

8.5 m/s wind max. Displacement d 1 [mm]

26 m/s wind max. Displacement d 2 [mm]

a) random

1.605

15.020

b) north-oriented

1.600

14.969

c) regular

1.532

14.338

The Karamba 3D analysis showed that, when placing the tensegrity elements of the last (third) layer, the position plays a role in the behavior of the tensegrity structure in the wind. The optimally-performing option, regarding the lowest overall displacement, is Option c. That might be due to the diagonal placement of the third-layer tensegrity units, which act as structural bracing.

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2.6 CFD Simulations (Membranes) Simultaneously, the effect of the shape on the wind surface pressure is investigated for the 8.5 m/s winds through CFD simulations in Butterfly for Grasshopper and subsequent post-processing through Paraview. The designed tensegrity structure is subtle, and the membranes covering the first and third tensegrity layer, along with the inner membrane of the pavilion, determine the wind flow. Therefore, only membrane geometries are examined in the wind. The effect of neighboring buildings is not considered in the simulations, however, the first-impact building in the northeast will presumably slightly affect the prevailing northerly winds by slowing them. The two-equation CFD turbulence model Realizable k-ε is used for the simulations, simple grading is employed for meshing the geometry, and the convergence criteria are designated as 1e−2 . For accelerating the convergence of the simulations, the relaxation factors are 0.5, 0.7, 0.3, and 0.7 for k (turbulent kinetic energy), U (velocity), p (pressure), and ε (dissipation rate of the turbulent kinetic energy). The simulations are decomposed into 12 processors to distribute the calculations, leading to a shorter run time. The simulations consist of 135,720; 187,860; and 187,860 cells for: 1) the analysis of the inner membranes, 2) the analysis of the first-layer membranes (“cushions”), and 3) the analysis of both, inner membranes and first-layer membranes together, respectively. Hexahedral meshing is used to mesh the case into finite volumes. The results of the CFD analysis of three cases are processed through Paraview to obtain the values of the wind surface pressure (Table 3), as well as colored images (Fig. 7). Table 3. Cases 1–3 tested in 8.5 m/s winds. Case

Static pressure ps [m2 ·s−2 ]

Kinematic pressure pk [Pa]

No. 1

12–35

15–43.75

No. 2

11–28

13.75–35

No. 3

12–27

15–33.75

Fig. 7. Cases 1–3, from left to right: 1) membranes, 2) “cushions” on the first layer of tensegrities, and 3) membranes plus the first-layer tensegrities “cushions”. The input wind speed is 8.5 m/s.

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Interestingly, no suction is present, and the values of the wind surface pressure are improved once the membranes are merged with small membranes (“cushions”). The system of layering the pavilion’s walls, as well as the spiky shape of the “cushions” leads to reducing the acting wind pressure. As the next step, the membrane structure is tested in the 26 m/s winds, representing the fundamental value of the basic wind velocity for Košice. Only Option c is tested, as Karamba 3D analysis showed it was performing the best in terms of structural integrity. Option c has the tensegrity units of the third layer ordered regularly along the cylinder’s surface (Fig. 6). The CFD simulation settings are the same as for the previous simulations. The wind tunnel consists of 197,253 hexahedral cells. The resulting wind surface pressure, as well as the wind flow pattern, are displayed in Fig. 8. Again, no wind suction on the surface is present, only the positive wind pressure values. The designed complex tensegrity-membrane structure provides a sheltered zone inside with a calm wind situation in the high wind speeds. It was demonstrated that the tensegrity structure could withstand the extreme wind speed of 26 m/s despite its lightweight appearance.

Fig. 8. Left: wind surface pressure (ranging from 130–290 m2 /s2 , ergo 162.5–362.5 Pa). Right: wind flow displayed 1.75 m above the ground. The input wind speed is 26 m/s.

The plans for further research include investigating the structural behavior of the tensegrity structure coated with membranes, which will replace the tendons in Karamba 3D. Secondly, by incorporating artificial intelligence into the process, we could forecast the wind speed and direction and use them as inputs for the form-finding of the pavilion. This way, the structure could prepare for the expected weather (loading) impacts and respond accordingly through the appropriate spatial configuration and design.

3 Conclusions The paper intends to introduce a digital methodology for designing architecture considering the wind as a critical design parameter. The overall form of the semi-closed pavilion, located in Košice, Slovakia, is found digitally through multiple Grasshopper

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plug-ins. The proposed pavilion has the task of improving wind comfort for pedestrians in the locality and creating a space for exhibitions and performance events that can take place inside it. The proposed design process deals with structural performance-tested recursively, in Karamba 3D, with the incorporated influence of the weather. The result is the well-performing architectural form optimally reacting to the specific microclimate. Three alternatives to the composition of the individual layers of the tensegrity units set are studied, considering the best structural engineering assessment result. The tensegrity-membrane structure is tested as follows: First, the tensegrity structural system is analyzed in Karamba 3D to get the idea of the overall maximum displacement of the structure under loads. Second, the optimum alternative, regarding the structural response in the wind, is selected. The option with regularly placed third-layer tensegrities worked the best. Third, the membranes are used to coat the first and the third layer of tensegrities in this option, plus a 5-segment inner membrane is added. Wind surface pressure acquired through the CFD simulations (utilizing Butterfly) showed that the layering of the pavilion’s walls helps reduce the acting wind pressure. The designed pavilion efficiently withstands high wind speeds, despite its visual lightness, and provides wind protection also in the 26 m/s wind speed while generating no wind surface suction through its shape. Therefore, we can confirm that with this approach in design, we can experimentally and effectively verify the structural and architectural solutions for the tentative tensegrity-membrane forms and focus on their precise and informed manufacturing when implementing them into practice. Acknowledgements. The authors would like to acknowledge the financial support from the following funds: VEGA 1/0626/22, VEGA 1/0129/20, and VEGA 1/0374/19 (Grant Agency of the Slovak Republic).

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Energy-Efficient Reinforced Heating System Implemented as a Carbon Concrete Formwork Ralf Gliniorz(B) , Carolin Petzoldt, Kristin Mandel, and Sandra Gelbrich Chemnitz University of Technology, 09126 Chemnitz, SA, Germany [email protected]

Abstract. This paper reports the development of an energy-efficient reinforced heating system implemented as a carbon concrete formwork. At present, nearly 70% of the provided heating energy for private households is based on fossil resources. The aim of research and development is to exclusively use renewable energy, to avoid power stations und combustion plants. Therefore, innovative technologies are needed that enable the generation of heat from electric energy. To achieve this goal, carbon concrete heating elements (CCE) were developed by Chemnitz University of Technology (TUC). These elements consist of thin-walled textile reinforced concrete and can be used for curved forms and rational functionalities, combined with durability and economic efficiency. A carbon roving, meander structured and not impregnated, creates the heating energy. A roving, often used in composites, consists of endless parallel fibre filaments. For structural and mechanical reasons, the basis is a reinforcing glass fibre grid (GFG). The carbon fibre has a high electric conductivity and a negative temperature coefficient (NTC element), which enables fast heating up to the target temperature. Therefore, power consumption increases with the temperature. The practical implementation of the CCE involves with many challenges: The properties of the concrete have to be adjusted, options for wall fastening as well as energy input, steering, and performance of the system have to be developed. During production, the exact positioning of the integrated elements and the use of resource-friendly casting and forming technology must be ensured. Keywords: Carbon concrete heating · Renewable resources · Energy-efficiency

1 Introduction A promising opportunity to use renewable energies for thermal power generation are electric heating systems, like mobile electric radiators, night storage heating, electric floor heating, infrared heating, or natural stone heating systems. Especially natural stone heating systems have a high potential for energy efficient building, because of their high performance of up to 400 W (≙ 2000 W/m2 ) [1]. The temperature of a marble stone heating plate, with the dimensions 40 × 50 × 2 cm3 , can rise up to 90 °C after one hour, starting at 23 °C (Fig. 1). Over a long period, the maximum surface temperature is 94 °C under full performance. © The Author(s), under exclusive license to Springer Nature Switzerland AG 2023 Z. Zembaty et al. (Eds.): ECCE 2022, LNCE 322, pp. 258–272, 2023. https://doi.org/10.1007/978-3-031-26879-3_21

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Fig. 1. Natural stone heating by the company Granotech: a) front view, b) back view, c) heat-up curve, d) infrared picture during full power.

The marble heating consists of a metal heating element, which transforms electric energy into heat as a result of elastic bounces of electrons with other electrons, atoms, and photons. Thus, the resistance increases with the temperature and the performance falls. The complex production, like cutting from blocks, milling of slits, gluing the heating element, and the limited design of form and surface are disadvantageous. The new carbon concrete heating element (CCE) will follow up here. The mineral construction material concrete behaves similarly like marble, while an electro conductive carbon roving provides the heating power. These carbon roving is directly integrated during production and increase the effectivity of heating. The heating system of the future is free from fossil resources and only uses renewable energies. The aim is to achieve a significant contribution to the energy revolution by developing and realising an energy efficient integrated reinforced concrete heating system. A functional textile serves the reinforcement of concrete and creates the infrared emission. In addition, the development allows for the efficient production of a carbon heating. It is implemented in lightweight construction with high heating power (>1000 W/m2 ), low weight, and individual design.

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2 Materials and Methods 2.1 Components for Carbon Concrete Heating Elements The energy efficient reinforced carbon concrete heating systems consist of a textile reinforcement (alkali-resistant glass fibres) for higher strength, a carbon roving for creating the thermal power through an resistance heating, and a mineral matrix (fine-grained concrete). Textiles There are heating textiles on the market, e.g. by the company Gustav Gerster GmbH & Co. KG. They were developed for tempering polymer composites (Fig. 2) but are not useful for concrete matrix.

Fig. 2. Commercial carbon heating textiles on glass fibre textile from Gustav Gerster GmbH & Co. KG.

The carrier textile of product (Fig. 2a) is not useful for the combination with the concrete, because it has to be manually cut out of the textile structure before using. For that reason, the flexible semi-finished product cannot be freely positioned without tension and fixing. Furthermore, the mesh size of the other heating textile with yellow grid (Fig. 2b) is not big enough for concrete to flow through it smoothly. Hence, a carbon roving, fixed on a glass fibre grid (GFG), is not available on the marked now. So, an in-house development was needed, which has to ensure the reinforcement and the heating function. The reinforcement was implemented by glass fibre textiles and the heating function by carbon fibre roving. The company solidian GmbH has a GFG for plane elements: a rigid GRID Q87/87AAE-21 and for curved elements SITGrid200. Possible alternatives are GFG AR 184, AR 777 and AR 780 from Dr. Günther Kast GmbH & Co. Technische Gewebe, Spezial Fasererzeugnisse KG.

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Carbon Fibre Rovings A variety of carbon rovings were tested for the application of the heating wire. These are characterized by their high electrical conductivity and negative temperature coefficient (NTC). Therefore, a fast heating up to the target temperature is possible, because in contrast to conventional metallic heating wires, the resistance decreases with increasing temperature and the power consumption increases (Fig. 3).

Fig. 3. U-I characteristic of a carbon roving (black) compared to constant ohmic resistance behaviour (red).

The carbon rovings investigated in the project are listed in Table 1, including their characteristic resistance (Rc ). The resistance values were determined by using a multimeter. Table 1. Selected roving types and measurements. Roving type

Linear density [tex]

Spec. Resistance Rc [Ohm/m]

Toho Tenax HTA40 E13

400

72.1

Tenax

800

34

R + G Sigrafil C30 50k 3300 tex

3300

9.2

Zoltek 50K PX35 Charge 2

3710

9.1

Zoltek 50K PX35 Charge 1

3710

8.7

Bacuplast

3500

8

A carbon heater must be designed to provide the desired surface heating. Factors include the operating voltage, the maximum power of the elements, the heating surface,

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and the roving length. This results in the required specific resistance of the roving, which can be selected for production according to Table 1. Concrete For this study, Steinbeis Research Centre BetoTex developed a special concrete mixture for casting, as a self-compacting mixture matched to specific heating systems (Table 2). A special mixing procedure was processed, using an Eirich RO5T. Table 2. List of components for used concrete. Component CEM I 52,5 N grey Silica sand 0–2 mm Silica sand 0–4 mm Rock flour Defoamer Stabilizer PCE-Fluxing agent Water

Formwork The formwork is an essential part for the production of concrete elements. There are various possibilities to fill a formwork for concrete casting. Gravity is used the most. For a plain CCE, a standing formwork made of wood, acrylic as well as PU was used (Fig. 4). The surface design of the concrete elements correlates primarily with the structure or quality of the formwork, whereby the component surface corresponds exactly to the formwork surface. In addition to smooth element surfaces, structured component surfaces are also possible with structured PU moulded mats. Methods for Integration of the GFG with Heating Carbon Roving The GFG, which has been defined as the preferred variant, can be used in both plane and curved formwork. In the case of single curve, the GFG adapts to the given contour without any problems. For double-curved or 3D surfaces, a modification of the GFG is necessary [3]. For example, cuts or segmentations have to be made at certain points. The GFG can then only be used for these adapted contours. The integration depth of the GFG is decisive for the position of the heating structure. Various solutions were investigated, one GFG-layer was used. Besides fixing to the formwork by means of adhesives, screws and hold-downs to prevent floating, FriPOX spacers (HPF GmbH & Co. KG) were used (Fig. 5 [2]). These have been adapted to the specific requirements of the GFG and can be stacked on top of each other to enable the depth of integration at any point. The number of spacers depends on the integration depth, size, and stiffness of the GFG.

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Fig. 4. Formwork variant for plane carbon concrete elements: self-construction of an upright formwork made of wood with transparent side made of acrylic.

Fig. 5. Spacer FriPOX and GFG.

2.2 Sample Preparation The manufacturing process of the carbon heating elements is divided into the following work steps: 1) 2) 3) 4) 5) 6)

preparation of the functional textiles, preparation of formwork, integration of functional textile in formwork, casting technology, curing time, dry heating.

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An essential work step is the preparation of the functional textiles [5]. First, the GFG is cut to size and the carbon roving is applied to the reinforcement textile in a meandering pattern and fixed in place using hot glue. When laying the roving, a minimum distance of approx. 2 cm from each other and from the edge must be ensured in order to avoid voltage bridges and, thus, hotspots. Then the roving ends are contacted by means of crimping and the sensor is positioned in the middle of the GFG. The result is shown in Fig. 6.

Fig. 6. Prepared functional textile: GFG and carbon roving.

Furthermore, the integration elements (GFG, carbon roving, force introduction elements “FIE” [2–4], positioning aids, sensors, contacting and wire routing) are prepared accordingly and positioned in the formwork (Fig. 7). Depending on the design, the elements are produced lying or standing with the self-compacting concrete mix in the concrete casting process [8].

Fig. 7. Positioning of the components in the horizontal formwork.

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After a curing time of 1 to max. 3 days, the elements are demoulded and stored at 20 °C and at approx. 40% humidity for 14 days from the day of production. The CCE are then gently dried for at least another 14 days. For this purpose, the temperature is increased daily by 3 to 5 K in order to achieve gentle drying of the concrete matrix. If heating takes place too early or too quickly, water vapour bubbles form inside, which can lead to cracks or other damage of the components. 2.3 Test Set-Up Strength The 4-point bending tensile strength based on DIN EN 12390-5 was determined on the manufactured samples. The test is carried out on the Zwick/Roell Z250 with a load frame of 10 kN and samples measuring 500 × 106 × 20 mm3 (length × width × height). The support width (L) is 450 mm and the support distance (L/3) 150 mm (Fig. 8).

Fig. 8. Schematic representation of the 4-point bending test (force F, support width L, support radius R1 , impactor radius R2 ).

Heat-Up Process The CCEs are set up vertically at room temperature and heated up slowly with a gradual increase in power in order to avoid concrete damage [6] due to thermal stresses and a sudden escape of residual moisture (Fig. 9). A standard heat-up procedure is developed as a template for the control programme of the heating elements to avoid damage of the concrete [7]. The increase in temperature depends on the plate thickness (Fig. 9) and amounts to 3–5 K/d. The maximum plate temperature is 90 °C. In addition, this provided insights into the power design for the desired target temperatures and heating times.

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Fig. 9. CCE with grey concrete mix 100 cm × 60 cm × 3 cm.

3 Results and Discussion 3.1 Influence of the Fibre-reinforcement on Concrete Strength Figure 10 shows the distribution scheme of 4-point bending (DIN EN 12390-5) of differently orientated reinforced concrete samples.

GF/CF long GF/CF cross

carbon roving GFG concrete Fig. 10. Fibre distribution scheme of the 4-point bending test samples.

With regard to the 4-point bending strength fct of the panels, the GFG results in a strength increase of 57% compared to the unreinforced concrete. In the case of carbon

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Table 3. Results of the 4-point bending test DIN EN 12390–5. Sample

fct [MPa]

Concrete

Standard deviation

7,57

0,45

11,87

1,19

Concrete + GF/CF cross

8,14

1,34

Concrete + GF/CF long

10,97

0,70

Concrete + GF

roving, only a very slight increase in the maximum strength in the cross direction could be demonstrated compared to samples without GFG and carbon roving. The unimpregnated roving cannot absorb forces transverse to the fibre and acts as a failure point. However, the strength in the long direction of the carbon roving showed nearly the same compared to the GFG reinforcement. This can be attributed to the lack of impregnation of the roving and the resulting lack of bonding between fibre and matrix (see Figure 10 and Table 3). Thus, the carbon heating roving reduces the effectiveness of the GFG, which must be taken into account when designing the panels in terms of strength. Nevertheless, since the carbon heating plates are not exposed to large forces during use, the strengths achieved when using the carbon roving are still sufficient for practical use. 3.2 Heat-Up Properties The heating process of a CCE is shown as a thermographic image in Fig. 11. The heat signature of the individual heating roving is clearly visible on the surface.

Fig. 11. Thermographic image of the heating process of a CCE.

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Due to the low resistance of the carbon and the good heat transfer inside the component, surface temperatures of over 100 °C can be achieved within one hour with a power density of about 1 kW/m2 with the CCE. Figure 12 shows the temperature performance of the carbon heater in comparison to the marble heater, in each case at maximum power.

Fig. 12. Temperature curve of the surface of carbon heater at maximum output compared to the natural stone heater.

Compared to conventional natural stone heating with 2 kW/m2 , higher temperatures can be achieved in fewer time and with the half of the area output by the CCE. This clearly underlines the great potential of carbon heating in terms of energy efficiency and heating conduction. 3.3 Energy Consumption The results of the energy consumption measurement of the CCE are compared with those of a conventional marble heating panel (Fig. 13 and Fig. 14). The measurements show that the carbon heating plate reaches a higher maximum temperature about three times faster than the marble plate. This results in lower energy consumption during the heating phase. For heating up the CCE, the costs are 50% less compared to the marble plate. Due to the comparable heat storage capacity, the cooling and reheating times of the two heating plates are analogous at about 45 min each.

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Fig. 13. Temperature curve and energy consumption of a 50 × 40 × 2 cm3 conventional marble plate.

Fig. 14. Temperature curve and energy consumption of a 50 × 40 × 2 cm3 CCE plate.

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3.4 Reference Object The reference object of the carbon heating elements is designed as a column to show the potential of carbon heating as a design element with integrated heating in indoor and outdoor areas. These curved elements are made using a cardboard formwork and a polystyrene core (Fig. 15). When demoulding the concrete column, the cardboard formwork is first destroyed, removed, and recycled. The formwork of the polystyrene core is then removed by pulling it out at prepared positions in order to ensure that the core could be reused.

Fig. 15. Formwork variant for curved CCE: Tubbox formwork tube from Max Frank GmbH with polystyrene core.

Over a period of two weeks, the heating process described above is used to slowly heat at 5 K/d to the target temperature of 60 °C (limited contact temperature). This is done comfortably via the temperature control of Appelt & Appelt GbR, enclosed in the water-blasted and welded cover plate (Fig. 16). Semi-circular cavities are provided in the lower part of the column so that the room air can circulate through the interior and the cover plate. This creates convection heat in the room in addition to radiation heat.

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Fig. 16. Reference object “column” (left) and thermographic image of heated state (right).

4 Conclusions This scientific work showed the technical implementation of an energy-efficient reinforcement-integrated carbon concrete heating system, which uses a functional textile made of glass and carbon fibres to generate the IR radiation. The textile also serves as reinforcement and, thus, strengthens the load-bearing behaviour. Lasting functionality was comprehensively investigated through, both, the material behaviour and the operating characteristics of the system as a function of the ambient conditions and the heating regime. With 50% of the applied power, compared to marble heating, a higher temperature (95 °C) can be achieved in a shorter time ( BCWS) and lower than planned earned value (BCWP < BCWS) together with the recorded values of composite indexes (CPI, SPI < 1,000) signal a simultaneous delay and underestimation of works, additionally expressed as a variance from costs (CV < 0 00).

Fig. 3. Scenarios of monotonicity of the BCWP and CV for seven dates of the contractor leaving the site (interruption of construction works). Source: Own research and development, calculations in PLANISTA MAX.

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Figure 3 shows the actual progress in terms of time and costs of the works carried out on the construction site, juxtaposed with the accompanying financial losses. The most common reason for contractors to terminate construction works contracts is the fact that the works are underestimated at the stage of their estimation, or the diagnosis of delays in execution, preventing their timely completion. The contractor’s decision to leave the construction site must be preceded by an analysis of the necessary costs, generated by contractual penalties for failure to comply with the subject of the contract. Figure 3 summarizes the scenarios of monotonicity of the basic parameters of the EVM method – Budget Cost of Work Performed (BCWP) and cost variance (CV) – assuming further persistence of undesirable trends of delay and underestimation. These scenarios, in the number of seven, describe the course of further disruptions from the assumed schedule – in the case of continuation of works in the following months. The growing underestimation of the cost estimate of works, along with their delay, prompts an economic and financial analysis of the profitability of the entire project, compared with the pool of contractual penalties provided for in the construction contract. Figure 4 shows the course of the schedule performance index (SPI) and the cost performance index (CPI) for seven monthly contractor exit scenarios determined by increasing production losses.

Fig. 4. Scenarios of monotonicity of the SPI for seven dates of the contractor leaving the site (interruption of construction works). Source: Own research and development, calculations in PLANISTA MAX.

The analysis of Figs. 2, 3, and 4 gives the answer how to use the EVM to determine the economic consequences for the contractor of works in the event of their interruption (leaving the construction site).

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Along with the termination of works, the increase in the Budget Cost of Work Performed (BCWP), Actual Cost of Work Performed (ACWP) and Cost Variance (CV) is stopped – no production is carried out on the construction site. The value of Budget Cost of Work Scheduled (BCWS) continues to grow – according to the assumed schedule, to the final value of the project budget. The cost performance index (CPI) does not show changes in monotonicity, while the value of the schedule performance index (SPI) is constantly decreasing: depending on the adopted date of the contractor’s leaving the construction site, to a level in the range of 2.7–70.9% – on the planned date completion of the project. The decrease in SPI value in individual scenarios illustrates the actual pace of works implementation, decreasing over time in relation to the overall planned pace. The aforementioned range of values on the planned date of completion of works shows what percentage of the scope of works could be completed in a timely manner in this project, depending on the moment of its termination. The monotonicity of the SPI, dictated by the following variants of the course of works, also shows the highest partial budgets of the project: in the third and sixth month of construction (distances between the graphs). The results of the economic profitability analysis of the project are presented in Table 1. The forecast of the time and cost condition was prepared on the date of the planned completion of works: 30th September 2017 (EVM report). The final values of the SPI [−], and the corresponding values expressed in money (SPI [PLN]), referring to the planned budget (SPI [30th September 2017] × BCWS [30th September 2017]) has been assigned to this date. In order to analyze the growing financial losses of construction, the increasing cost variances (CV), the value of ACWP and the value of the BCWP index, describing the measure of the actual progress of works in the venture were also compared. The forecast for determining the underestimation and delays of works carried out according to the assumed schedule, together with contractual penalties, indicates that the analyzed project becomes unprofitable after the fourth month of works execution. During this period, the value of the losses incurred exceeds the costs of contractual penalties for failure to meet the subject of the contract (30/06/2017: |CV| > K(ZU)).). The comparison of the schedule performance index SPI [PLN] expressed in amounts with the Budget Cost of Work Performed BCWP [PLN] in individual months of works shows slight differences of 1–2%. This configuration, in its absolute values, should, however, be compared with the Budget Cost of Work Scheduled (BCWS): then an increasing delay in the schedule becomes visible, which in turn leads to additional costs in the form of penalties for delayed completion of the entire project (K(ZZ) = 20,106.85 PLN/day of delay). The description of increasing cost variances (CV) assumes a continuing trend of delays and underestimation of works carried out according to the schedule (no remedial actions on construction site). This parameter, in absolute value, represents the contractor’s financial losses in the event of a decision to continue the works.

921.206,88

736.965,50

BCWS [PLN]

3.236.540,32

Works delay: K(ZZ) = 0,1%/day x 20.106.851,97 = 20.106,85 PLN/day SPI = 0,750: delay 2,25 month (46 working days): 904.808,34 PLN

Breach of contract: K(ZU) = 15% x 20.2016.851,97 = 3.016.027,80 PLN

1.231.910,49 4.342.233.19

1.521.197,64 5.380.342,65

916.439,08

5.827.871,53 7.751.990,53

9.382.143,54

8.746.881,93 11.643.856,89

14.080.854,26

14.258.398,31 19.033.078,68

23.203.803,29

0.750

20.106.851,97

24.597.314,48

15.094.498,29

15.080.138,98

552.724,13

0.709 14.255.758,05

ACWP [PLN]

0.435 8.746.480,61

BCWP [PLN]

5.871.200,78

0.292

30.09 2017

−368.482,75 −604.758,56 −2.143.802,33 −3.554.272,01 −5.333.972,33 −8.945.404,98 −9.502.816,19

3.237.203,17

0.161

31.08 2017

CV [PLN]

924.915,19

0.046

31.07 2017

0.027

30.06 2017

542.885,00

31.05 2017

SPI [PLN]

30.04 2017

SPI [−]

31.03.2017

Date of EVM report: 30.09.2017 Contractor leaving the site (interruption of construction works) on

Table 1. Analysis of the profitability of the project in seven stages of its implementation EVM. Economic Conditions of Leaving the Construction Site 297

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3 Conclusions from Research and Analyzes The conducted economic analysis on the selected example of works execution, taking into account the effects of their interruption using the EVM method, gives important conclusions. The comparison of the results of the works progress report together with the costs of contractual penalties showed the unprofitability of the analyzed project after the fourth month of the works execution, in which the level of cost variances (CV, Fig. 3) exceeded the contractual amount for the contractor’s failure to perform the contract. The specified schedule performance index (SPI, Fig. 4) in the EVM shows the scenarios of stopping the works, depending on the moment of the decision to resign from their further performance. This index, presenting the actual pace of works implementation, also provides a forecast for future periods – with the final value of the project, describing the percentage of the project’s budget implemented on time. It is worth noting that a similar economic analysis of the works carried out could concern their profitability in the context of the processes to be performed, instead of monthly accounting periods. Then, the contractor’s decision-making would describe the times of completion of individual activities in the schedule. The presented earned value method (EVM), despite many disadvantages described in the literature, can be an effective tool for cost control and the actual efficiency of works. The presented approach may also be a solid contribution to increasing the decisionmaking efficiency of the contractor on the construction site.

References 1. Colin, J., Martens, A., Vanhoucke, M., Wauters, M.: A multivariate approach for top-down project control using earned value management. Decis. Support Syst. 79, 65–76 (2015) 2. Koskela, L.: Application of the new production philosophy to construction. Stanford University, CA, CIFE Technical Report 72, Stanford (1992) 3. PMBOK Guide (1st ed.): Practice for Earned value Management P.A. Newton Square: Project Management Institute, PMI (2005) 4. Lock, D.: Essentials of Project Management. Warsaw, pp. 92–95 (2009) 5. Marciniak, S.: Modern Management Instruments. Warsaw, pp. 11–12 (2008) 6. Marcinkowski, R., Krawczy´nska-Piechna, A.: Designing the Construction Implementation. Warsaw, pp. 310–316 (2019) 7. Marcinkowski, R.: Methods of Allocating the Contractor’s Resources in the Engineering and Construction. Warsaw, pp. 15–20, (2002) 8. Nowak, E.: Managing Costs and Effects. Wrocław, pp. 41–42 (2011) 9. Poło´nski, M., Komendarek, P.: Ongoing control of the costs of a construction building using the earned value method. Quant. Methods Econ. Res. XII(2), 279–290 (2011) 10. Przywara, D.: Time-cost analysis in the monitoring of construction schedule works. Doctoral dissertation, Opole University of Technology, Opole (2019) 11. Przywara, D., Rak, A.: Evaluation of time and cost variances using earned value method. Materiały Budowlane 6, 29–31 (2016) 12. De Koning, P., Vanhoucke, M.: Stability of earned value management: do project characteristics influence the stability moment of the cost and schedule performance index. J. Mod. Project Manag. 4(1), 8–25 (2016) 13. Smaga, E., Guzik, K.: Risk and Profitability of Financial and Material Investments. Cracow, pp. 101–107 (2013)

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14. Poło´nski, M.: Control of the financial advancement of an office building using the EVM in MS project. In: Proceedings of the Technical Conference “Problems of Preparation and Implementation of Construction Investments” (Conference monograph), pp. 101–110, Puławy (2008) 15. Willems, L., Vanhoucke, M.: Classification of articles and journals on project control and earned value management. Int. J. Project Manage. 33(7), 1610–1634 (2015) 16. Przywara, D., Rak, A.: Monitoring of time and cost variances of schedule using simple earned value method indicators. Appl. Sci. 11(4), 1357 (2021)