Civil Engineering and Disaster Prevention : Proceedings of the 4th International Conference on Civil, Architecture and Disaster Prevention and Control (CADPC 2023), Suzhou, China, 24-26 March 2023 [1 ed.] 9781032546186, 9781032546209, 9781003425823
Civil Engineering and Disaster Prevention focuses on the research of civil engineering, architecture and disaster preven
118 24 693MB
English Pages xv; 741 [758] Year 2023
Cover
Title Page
Copyright Page
Table of Contents
Preface
Committee Member
Civil seismic detection and structural reinforcement
Impact of subway station upper span construction on existing
railway tunnel structure
1 Introduction
2 Construction Details
2.1 Project description
2.2 Geological condition
3 Numerical-Analytical Methods
3.1 Material parameters
3.2 3-Dimensional model analysis
4 Results and Discussion
5 Conclusions
Conflict of Interest Statement
References
The safety of adjacent road structures in a tunnel construction project
1 Introduction
2 Project Overview
3 Numerical Simulation Analysis of Project Construction
3.1 Constitutive model
3.2 Calculation of unit type and model parameters
3.3 Calculation profile analysis
4 Conclusions
References
Experimental and finite element parameter analysis of modular assembled composite shear wall considering corner structure optimization
1 Introduction
2 Test Survey
2.1 Specimen design
2.2 Test loading and measuring point arrangement
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3 Experiment Results And Analyses
3.1 Specimen failure characteristics and crack development
3.2 Specimen failure characteristics and crack development
3.3 Stiffness degradation
4 Numerical Model Establishment and Verification
5 Parameter Influence Analysis
5.1 Numerical model establishment and verification
5.2 Analysis of calculation results
5.2.1 Influence of the length-to-span ratio.
5.2.2 Influence of rib thickness ratio.
5.2.3 Influence of thickness ratio.
6 Conclusions
References
Engineering application of the tunnel disease rapid detection system
1 Introduction
2 Tunnel Detection Vehicle Function Test
3 Engineering Application of Tunnel Disease Rapid Detection System
3.1 Comparison of lining diseases
3.2 Comparison of interior decoration diseases
4 Conclusions
References
Deep learning-based recognition method of ground penetrating radar images for cracks inside pavement structures
1 Introduction
2 GPR Detection and Data Processing
2.1 GPR detection of pavement structures
2.2 Data processing of GPR images
3 Object Detection Model
3.1 Data processing of GPR images
3.2 Evaluation index of the YOLOv5 model
4 Results and Discussion
5 Conclusions
References
Shaking table test of frame structure considering bidirectional earthquake
1 Introduction
2 Model Design
2.1 Similarity relationship
2.2 Model design drawing
3 Test Scheme
3.1 Seismic wave selection
3.2 Test condition
4 Analysis of test Results
4.1 Dynamic characteristic analysis
4.2 Acceleration response analysis
4.3 Displacement response analysis
4.4 Inter-storey shear force
4.5 Residual deformation
5 Conclusions
References
Establishment method of evaluation index for the impact of chloride ion erosion on bearing capacity of reinforced concrete structure of tidal sluice
1 Introduction
2 Damage Mechanism of Reinforced Concrete Structure by Chloride Ion Erosion
2.1 Mechanism of concrete corrosion by chloride ions and model selection
2.2 Corrosion mechanism and corrosion rate estimation of steel bars in chloride ion environment
3 Establishment of Bearing Capacity Index of Concrete Structure Corroded by Chloride Ions
3.1 Influence of steel bar corrosion on bond stress between steel bar and concrete
3.2 Establishment of an evaluation index for structures affected by chloride ion erosion
4 Conclusions
References
Evaluation of technical status of concrete beam bridge based on machine learning
1 Introduction
2 Ensemble Learning Algorithms
2.1 XGBoost
2.2 LightGBM
2.3 CATBoost
3 Modeling and Analysis
3.1 Bridge status database
3.2 Bridge technical condition classification model
3.3 Model evaluation
4 Conclusions
References
Reinforcement effect of the anti-slide pile on structural slope plane based on point safety factor method
1 Introduction
2 Computing Method
2.1 Calculation model
2.2 Calculation parameters
2.3 Computing method
3 Results &x00026; Discussion
3.1 Safety factor analysis of main structural surface points under natural conditions
3.2 Analysis of safety factor of main structural surface points under reinforcement condition
4 Conclusions
References
Experimental study on the influence of stirrup on axial compression performance of full-scale concrete square column
1 Introduction
2 Test Overview
2.1 Specimen design and production
2.2 Mechanical properties of specimen materials
2.2.1 Reinforcement
2.2.2 Concrete
2.3 Measurement scheme
3 Description of Failure Process and Morphological Characteristics of the Specimen Test
4 Test Results and Analysis of Specimens
4.1 Ultimate bearing capacity of specimens
4.2 Stress-strain curves of concrete in the constrained core area
4.3 Analysis of influencing factors of concrete stress-strain relationship in the confined core area
5 Conclusion
References
The method of monitoring the health status of buildings based on Beidou
1 Introduction
2 Communication Network Selection
3 Sensor Layout Principles
4 Monitoring Program Layout
5 Design of Building Security Monitoring Indicators
5.1 Tilt monitoring
5.2 Settlement monitoring
6 Conclusion and Outlook
References
Based on Beidou navigation satellite system bridge variation monitoring analysis and research
1 Introduction
2 BDS Working Principle
3 Monitoring Scheme
3.1 Communication network selection
3.2 Layout location selection
4 Determination Method of Key Indicators
4.1 Tilt determination method
4.2 Measuring method of deflection
4.3 Vibration measurement method
5 Conclusions
References
Numerical simulation of reinforced concrete beam four-point bending test based on dual-particle Peridynamics
1 Introduction
2 PD Theory Introduction
2.1 Basic idea and basic model construction
2.2 Model construction
3 Benchmark Problem
4 RC Four-Point Bend Simulation
4.1 Model basic information introduction
4.2 Explanation of the model running process
4.3 Image analysis of model results
4.4 Conclusion
5 Conclusion
5.1 Strengths of the results in this paper
5.2 The problems that still exist
References
Discussion of foundation improvement methods for thick rock-filled gravel based on Wudangshan Airport, Shiyan, China
1 Introduction
2 Parameters and Models
3 Analysis of the Simulation Results
3.1 Maximum allowable post-work settlement
3.2 Numerical simulation results
3.3 The law of post-work settlement variation
3.4 Variation in the height of thick rock-filled gravel
4 Discussion of Foundation Improvement Methods
5 Conclusions
References
Effects of length, shape, and dosage of steel fiber on mechanical properties of steel fiber reinforced concrete
1 Introduction
2 Influence of Steel Fiber Length on Mechanical Properties of Concrete
2.1 Experimental materials and methods
2.2 Experimental result and analysis
2.2.1 Compressive strength
2.2.2 Flexural strength
2.3 Conclusion
3 Influence of Steel Fiber Shape on Mechanical Properties of Concrete
3.1 Experimental materials and methods
3.2 Experimental result and analysis
3.2.1 Compressive strength
3.2.2 Flexural strength
3.3 Conclusion
4 Influence of Steel Fiber Dosage on Mechanical Properties of Concrete
4.1 Experimental materials and methods
4.2 Experimental result and analysis
4.2.1 Compressive strength
4.2.2 Flexural strength
4.3 Conclusions
5 Discussion
6 Conclusion
References
Experimental study on the effect of fly ash proportion on the mechanical properties of sand concrete made by waste ultra-fine sand
1 Instructions
2 Materials and Methods
2.1 Sand
2.2 Cement
2.3 Fly ash
2.4 Water reducing agent
2.5 Ultra-fine sand concrete preparation
2.6 Mechanical property of sand concrete
3 Result and Discussion
3.1 Compressive strength
3.2 Splitting tensile strength
3.3 Immersion compressive strength
4 Summary
References
Deformation analysis of enclosure structure affected by foundation pit excavation
1 Introduction
2 Engineering Summary
2.1 Engineering background
2.2 Foundation pit excavation and support plan
3 Numerical Modeling
3.1 Modeling
3.2 Calculation of numerical model
3.3 Model validation
4 Analysis of Results
4.1 Effect of different support materials and numbers on the horizontal displacement of the enclosure structure
4.2 Influence of envelope parameters on the horizontal displacement of the maintenance structure
5 Conclusions
References
Key technology research on the establishment of the prefabricated component library of assembled structure based on BIM
1 Introduction
2 Prefabricated Structural Design Based on Bim
2.1 Prefabricated structural design process based on BIM
2.2 Creation procedure of prefabricated components information
3 Research on the Creation of the Prefabricated Component Library of Assembled Structure
3.1 Classification and coding of the stored prefabricated components
3.2 Creation of prefabricated components information
3.3 Preliminary establishment of prefabricated component library of assembled structure based on BIM
4 Research on Establishing the Bim Model for Prefabricated Shear Walls
5 Conclusions
References
Sensor fault classification for bridge SHM using LSTM-based with 1D-CNN feature extraction
1 Introduction
2 Sensor Fault Classification Related Theory
2.1 Feature extraction of one-dimensional convolutional neural network method
2.2 Fault classification by LSTM neural network method
2.3 Establishment of sensor fault classification model
3 Numerical Example Verification
3.1 Data collection
3.2 Model performance verification
4 Conclusions
References
Demolition technology of long-span concrete box girder in the upper span closed frame channel
1 General Instructions
2 Project Overview
3 Temporary Support Design
3.1 Beam segmentation
3.2 Temporary support structure
4 Analysis of the Cutting Step Sequence of the Beam
4.1 Design of the cutting step sequence of the beam
4.2 Numerical calculation results
5 Support Structure Design of the Lifting Equipment
5.1 Equipment support structure
5.2 Verification and calculation of stability
6 Conclusions
References
Design and application of the three-dimensional model for seepage deformation of diversion tunnel
1 Introduction
2 Key technologies
2.1 BIM + GIS
2.2 Key technologies for establishing dimensional parametric models
3 System overall design
3.1 Architecture design
3.2 System data organization
4 System function realization
5 Conclusion
References
Seismic response analysis of large-span structure considering rotational ground motion under uneven settlement sites
1 Introduction
2 Build a Structural Model
3 Modal Analysis
4 Selection of Ground Vibrations and Acquisition of Rotational Components
4.1 Selection of ground vibrations
4.2 Acquisition of rotational components
4 Setting of Uneven Settlement Conditions
5 Structural seismic response analysis
5.1 Structural displacement response
5.2 Structural torsional response
6 Conclusions
References
Numerical simulation of mechanical properties of fiber-reinforced recycled concrete beam-to-column joints under monotonic loading
1 Introduction
2 Establishment of Numerical Model
2.1 Numerical model
2.2 Constitutive relationship
3 Result Analysis
3.1 Load-displacement curves
3.2 Characteristic load
3.3 Shear deformation of the core area
4 Conclusions
References
Prediction and experimental research on fatigue life of corroded prestressed concrete beams
1 Introduction
2 Fatigue Life Prediction Method for Corroded PC Beams
2.1 Fatigue crack propagation model of corroded steel strand
2.2 Assessment of corrosion fatigue bond degradation
2.2.1 Fatigue bond stress
2.2.2 Calculation method considering residual strain of reinforcement under cyclic loading
2.3 Fatigue damage of concrete
2.4 Calculation process of fatigue life prediction for corroded PC beams
3 Experimental Program and Model Validation
3.1 Specimen details
3.2 Loading procedure
3.3 Concrete crack development
3.4 Failure modes and fatigue life
3.5 Model validation
4 Conclusions
References
Size effect of fly ash geopolymer concrete at different temperatures
1 Introduction
2 Test
2.1 Specimen model
2.2 Raw materials and mix ratio
2.3 Preparation and curing of test pieces
2.4 Test device and test method
3 Test Result
4 Size Effect of Fly Ash Geopolymer Concrete
5 Conclusions
References
The key problems of arch supported roof structural system design of a sports building in Xiaogan
1 Introduction
2 Structural System
3 Key Problems in Structural Design
3.1 Arch foot horizontal thrust
3.2 Super long plane temperature effect
3.3 Out-of-plane stability of Vierendeel truss arch
3.4 Out-of-plane stability of roof plane truss
4 Conclusions
References
An explainable two-stage data-driven approach for risk modelling in tunnel construction
1 Introduction
2 Development of a Hybrid ARIMA-DNN Model
3 Case Study
3.1 Project overview
3.2 Construction of the hybrid ARIMA-DNN model
3.2.1 Anomalies detecting monitoring data
3.2.2 Hyperparameters setting
3.3 Prediction results
3.3.1 Prediction accuracy
3.3.2 Prediction explainability
4 Conclusions
References
Experimental study on the collapse resistance of precast monolithic reinforced concrete beam-column subassembly with additional connecting steel bars
1 Introduction
2 Experimental Program
2.1 Specimen design
2.2 Materials
2.3 Test set-up
3 Test Results
3.1 Compressive arch action stage
3.2 Catenary action stage
3.3 Load-displacement curves
4 Conclusions
References
Experiment on the fatigue behavior of steel bridge deck pavement structures paved by high-content hybrid steel fiber reinforced self-compacting concrete
1 Introduction
2 Confirmation of the Most Unfavorable Loading Mode and Position
2.1 The most critical loading position
2.1.1 Model geometry parameters and model establishment
2.1.2 Model building and meshing
2.1.3 Model calculation results
2.2 Experimental loading mode and method
2.2.1 The method for loading the specimen
2.2.2 Selection of the specimen
2.2.3 The size of the specimen
2.3 Materials and proportioning
2.4 Experimental loading
2.4.1 Loading force and frequency
2.4.2 Loading times and damage guidelines
2.4.3 Experimental equipment
3 Results and Analysis
3.1 Fatigue damage patterns
3.1.1 Self-compacting concrete specimens without the addition of steel fibers
3.1.2 Self-compacting concrete specimens with single steel fibers
3.1.3 Hybrid steel fiber reinforced self-compacting concrete specimens
4 Discussion
4.1 A comparison of fatigue damage cracks
4.2 A comparison of fatigue life
5 Conclusion
Conflicts of Interest
References
Structural performance design analysis of a high-rise office building beyond the limit
1 Introduction
2 Project Overview
2.1 Basic information
2.2 Structural design
3 Structure Overrun and Seismic Performance Target
3.1 Identification of overrun
3.2 Determination of performance objectives
4 Structural Calculation and Analysis
4.1 Elastic analysis under frequent earthquake
4.1.1 Mode decomposition response spectrum method
4.1.2 Elastic time-history analysis method
4.2 Floor stress analysis under seismic fortification
4.3 Dynamic elastoplasticity under large earthquake
4.3.1 Plastic hinge state of frame beam and column
4.3.2 Plastic hinge state of shear wall
5 Main Seismic Strengthening Measures
6 Conclusion
References
Study on stress intensity factor of steel wire with double surface cracks
1 Introduction
2 Crack Propagation Analysis Software FRANC3D
3 Calculation Model of Double Crack Steel Wire
4 Stress Intensity Factor Analysis of Steel Wire Double Crack
4.1 Concept of interaction between double cracks
4.2 Calculation results of single crack stress intensity factor
4.3 Stress intensity factor calculation results of parallel double cracks
4.4 Stress intensity factor calculation results of collinear double cracks
5 Conclusion
References
Intelligent building and equipment installation technology
Utilization of recycled construction waste filler in urban greenway
1 General Instructions
2 Proposing the Concept of Sponge City Selecting
3 Improvement of Greenway Pavement Structure
3.1 Permeable slow lane
3.2 Slow traffic road with construction waste recycling materials as permeable base
4 Application Technology of Recycled Material Permeable Base
4.1 Grading selection
4.2 Grading correction considering secondary crushing of construction waste
4.3 Trial mixing test of recycled permeable base material
4.4 Strength test of recycled permeable base material
4.5 Drainage performance test of recycled pervious concrete base material
5 Discussion
5.1 Technical performance of recycled permeable concrete base
5.2 Suggestions for improving the drainage capacity of recycled permeable concrete base
6 Conclusions
References
Surveying and mapping ancient buildings with 3D laser scanning technology
1 Introduction
2 Overview of 3d Laser Scanning Technology
2.1 Principle of three-dimensional laser scanning technology
2.2 Technical process
3 Case Analysis
3.1 Project overview
3.2 Field data collection
3.3 Point cloud data processing
3.4 3D modeling
3.5 Texture mapping
3.6 Accuracy verification
3.7 Thematic map drawing
4 Conclusion
References
Building Integrated Photovoltaic (BIPV): Applications and development
1 Introduction
2 The Design Principle of BIPV
2.1 Mathematical model
2.2 Simulation results
2.3 Optimization measures
3 BIPV Project Examples
4 Conclusions
References
Design method of municipal water supply and drainage pipeline based on BIM technology
1 Introduction
2 Design of Dynamic Translation Model Based on BIM
3 Simulation Test Analysis
3.1 Simulation environment
3.2 Comparison of optimization effects
3.3 Pipe collision detection
3.4 Pipeline optimization
4 Conclusions
References
Lateral stability analysis of single-column pier bridge based on geometric nonlinearity
1 Introduction
2 Calculation of Lateral Stability of Bridge with Geometric Nonlinearity
3 Experimental Test
3.1 Test platform
3.2 Experimental test results
4 Conclusions
References
Design and application of cable crane for Zhangjiajie Grand Canyon Glass Bridge
1 Introduction
2 Key and Difficult Points in the Design and Construction of Cable Crane
3 Key Technology of Cable Crane Design
3.1 Overall design
3.2 Cable system design
3.3 Tower design
3.4 Tower foundation and anchor design
4 Application of Cable Crane
4.1 Tower and foundation construction
4.2 Cable installation
4.3 Installation of crown blocks and lifting points
4.4 Test lifting inspection of cable crane
5 Conclusions
References
A comparison study on different suction bucket pre-piling templates for offshore wind jacket foundation
1 Introduction
2 Marine and Soil Data
3 Numerical Model Setup and Load Case
3.1 PPTP model
3.2 PPT model
3.3 Suction bucket model
3.4 Loading conditions and load cases
4 Result and Discussion
4.1 Structure strength analysis
4.2 Suction bucket analysis
5 Engineering Applications
6 Conclusions
References
Numerical simulation study on ground subsidence caused by cavity under buried PCCP socket
1 Introduction
2 Finite Element Analysis Model
2.1 PCCP model size
2.2 Boundary constraints and interlayer relationships
2.3 Constitutive relations and model materials
3 Result Analysis
3.1 Influence of cavity dimensions and soil thickness on pavement stability
3.2 Distribution of ground subsidence
3.3 Distribution law of pavement surface settlement displacement curve
3.4 Effect of the cavity on the PCCP pipeline body
4 Conclusions
References
The application of BIM technology in urban utility tunnel
1 Introduction
2 Significance of the Development of Urban Utility Tunnel
3 Application of Bim Technology in the Design Stage of an Urban Utility Tunnel
3.1 Line selection
3.2 Collaborative design
3.3 3D visualization technology
4 Application of Bim Technology in the Construction Stage of an Urban Utility Tunnel
4.1 Assembled pipe gallery
4.2 Construction quality control
4.3 Construction progress and cost control
5 Application of Bim Technology in the Operation and Maintenance Stage of the Urban Utility Tunnel
5.1 Daily patrol management
5.2 Environmental monitoring and alarm
6 Conclusions
References
Experimental study on indoor model of water vapor humidification remodeling unsaturated loess under negative pressure conditions
1 Introduction
2 Model Test Overview
2.1 Test equipment
2.2 Test soil sample
2.3 Experimental protocol
2.4 Sensor arrangement
3 Analysis of Experimental Results
3.1 Negative pressure transfer analysis
3.2 Analysis of water separation field results
3.3 Analysis of temperature field results
4 Conclusions
References
Construction technology of retaining structures for deep foundation pit with H-shaped composite steel anchor piles-precast slab walls by compaction grouting method
1 Background
2 Engineering Project Application: A Construction Method of the Retaining Structure for Deep Foundation Pits With H-Shaped Composite Steel Anchor Piles- Precast Slab Walls
2.1 Design of the retaining structure for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls
2.2 Construction process
2.3 Technical essentials
3 Conclusion
References
Construction method and application of Dynamic Ultimate-Strength Control Technology (DUSCT) for ultra-shallow buried excavation
1 Introduction
2 Dynamic Design and Regulation
3 Ultimate-Strength Control Technology of Ultra-Shallow Buried Excavation
4 Ultra-Shallow Buried Excavation Technology
5 Dynamic Monitoring Technology
6 Engineering Test and Application
6.1 Project overview
6.2 Engineering geology and gydrology
6.3 Excavation and construction process
6.4 Monitoring and measurement of construction process
6.5 Measuring results
6.5.1 Monitoring and analysis of land subsidence
6.5.2 Analysis of soil pressure measurement results
6.5.3 Analysis of stress measurement results of reinforcement
7 Conclusions
References
Optimization and safety evaluation of dismantlement scheme for point-supported glass curtain wall in airport terminal buildings
1 Introduction
2 Project Overview
2.1 Component information
2.2 Demolition scheme
2.3 Monitoring scheme
3 Model Building and Verification
3.1 Element simulation
3.2 Element subdivision and division
3.3 Determination of boundary conditions
3.4 Model validation
4 Optimal Design of Existing Curtain Wall Removal Scheme
4.1 Partial removal scheme of the glass panel
4.2 Displacement monitoring analysis of the main truss
4.3 Support rod displacement monitoring analysis
4.4 Internal force monitoring and analysis of support rod
5 Security Verification before and after Renovation of the Curtain Wall Structure
5.1 Engineering design parameters
5.2 Safety assessment of curtain wall structural members
5.2.1 The strength and deflection checking of the glass panel
5.2.2 The deflection checking of the main truss
5.2.3 The deflection checking of the support rod
6 Conclusions
References
5D building information modeling status based on bibliometrics
1 Introduction
2 Methodology
2.1 Database choice
2.2 Filtering and controlled criteria
2.3 Article classification based on content
3 Bibliometric Analysis
4 Content Analysis
5 Conclusions
References
Hydraulic fracturing test on in-situ stress and its distribution regularities of a flood drainage construction engineering site
1 Introduction
2 Engineering Geology
3 Principle of the Hydraulic Fracturing Method
4 Field Test
4.1 Test parameters
4.2 Test procedure
5 Data Analysis of Test Results
5.1 Principal stress in different test sections
5.2 Direction of principal stress
5.3 The trend of the stress distribution
5.4 Lateral pressure coefficient of in-situ stress
5.5 Characteristics of in-situ stress
5.6 In-situ tensile strength of rock
6 Conclusions and Suggestions
References
Cable forces and optimization of construction process on composite girder of cable-stayed bridge
1 Introduction
2 The Xiangsizhou Bridge
3 Finite Element Model
4 Determination of Reasonable Completion State
5 Force Analysis During the Construction Phase
5.1 Construction process optimization
5.2 Reasonable construction state
5.2.1 Structural internal force analysis during the construction phase
5.2.2 Structural internal force analysis in the bridge completion stage
6 Conclusions
References
The design and optimization of the building electromechanical system based on BIM technology
1 Introduction
2 Research Background
3 Research Overview
3.1 Development status
3.2 BIM concept
4 Bim Technology in the Main Application of Building Mechanical and Electrical Engineering
5 Design and Optimization of Building Electromechanical System Based on Bim Technology
6 Conclusions
References
Application of Internet of Things and BIM technology in building intelligent operation and maintenance system
1 Introduction
2 Overview of Internet of Things Technology and Bim Technology
2.1 Internet of Things technology
2.2 BIM technology
3 Module Design of Intelligent Building Operation and Maintenance System
3.1 Visual display system
3.2 Equipment monitoring system
3.3 Safety management system
3.4 Energy consumption management system
4 Development Measures of Internet of Things Technology in Intelligent Operation and Maintenance
5 Conclusion
References
Horizontal top displacement change patterns of the pile foundation in slope under freeze-thaw cycles
1 General Instructions
2 Engineering Background of the Pile Foundation on Permafrost Slope
3 Mechanical Characteristics of Pile Foundation in the Slop in Permafrost Areas
4 Finite Element Modeling and Calculation
5 Boundary Conditions and Initial Conditions
6 Analysis of Numerical Simulation Results of the Pile on the Slope Soil Foundation
7 Conclusions and Prospects
References
Experimental study on deformation characteristics of cement soft soil under the influence of multiple factors
1 Introduction
2 Experimental Study Scheme
2.1 Test material
2.2 Factors considered in the experimental study and their composition
3 Comparative Analysis of Compression Test Results
3.1 Effect of cement dosage and curing age on deformation of cement soft soil
3.2 Effect of moisture content and age of maintenance on the deformation of soft cement soils
4 Conclusion
References
Application of composite foundation with vibro-crushed stone column in saturated liquefied sand ground treatment of a gas-fired power station in Myanmar
1 General Instructions
2 Project Overview
3 Comparative Selection of Foundation Treatment Scheme
4 Design of Composite Foundation with Vibro-Crushed Stone Column
5 Evaluation of Foundation Treatment Effect
5.1 The static load test
5.2 Standard penetration test
5.3 Dynamic penetration test
6 Conclusions
References
Application of oblique photography technology in airport site selection
1 Introduction
2 Oblique Photography Fieldwork
2.1 Site overview
2.2 Equipment selection
2.3 Route planning
3 3D Modeling of the Site to be Selected
3.1 Original photo pre-processing
3.2 POS data solution and export
3.3 Oblique photographic modeling
4 Accuracy Analysis and Visual Display of 3D Model
4.1 Precision analysis
4.2 3D model display and measurement
5 Conclusions
References
Satellite positioning and inertial navigation based firefighters positioning system
1 Introduction
2 Materials and Methods
2.1 Materials
2.2 Methodology
2.3 GPS/BeiDou module and IMU data fusion
3 Results and Analysis
4 Conclusions
References
Moisture accumulation in wall thermal insulation layer under the action of coupled heat and moisture during heating period in cold regions
1 Introduction
2 Coupled Heat and Moisture Transfer Model
2.1 Moisture control equation
2.2 Heat control equation
2.3 Initial conditions and boundary conditions
3 Model Calculation and Result Analysis
3.1 Typical insulated wall structure and thermal parameters
3.2 Moisture accumulation in a typical wall with external thermal insulation
3.3 Influence of installation of air layer
3.4 Influence of installation of vapor barrier
3.5 Influence of setting of finish coat
4 Conclusions
References
Bridge strain based on real bridge and indoor experiment
1 Introduction
2 Field Experiment
2.1 Experimental site
2.2 Sensor layout
2.3 Loading method
2.4 Experimental loading principles
2.5 Data preprocessing
2.6 Experimental results
3 Indoor Experiment
4 Conclusions
References
Design of test platform for early fire detection of power transmission and transformation equipment
1 Introduction
2 Fire Characteristics of Power Transmission and Transformation Equipment
3 Aspirated Smoke Detection Principle
3.1 The detection principle of the cloud chamber
3.2 Time-sharing circulating sampling fire location device
4 Fire Monitoring System Platform for Power Transmission and Transformation Equipment
4.1 Sampling pipe layout
4.1.1 Layout of sampling pipe of the switch cabinet
4.1.2 Layout of sampling pipe of cable trench
4.1.3 Layout of sampling pipe of the switch cabinet
4.2 Transmission network layout
4.3 Fire monitoring and early warning system
5 Fire Monitoring System Platform Test
6 Conclusions
References
Landscape conservation planning based on forest fire prevention: A case study of Lushan Forest Park in Sichuan Province
1 Introduction
2 Methodology
2.1 Research area
2.2 Data source
2.3 Experimental procedure
2.4 Spatial analysis model
3 Results and Discussion
3.1 Weight analysis and single factor evaluation
3.2 Comprehensive evaluation of ecological sensitivity
3.3 Ecological conservation planning of Lushan Forest Park
3.3.1 Ecological Conservation Zone (ECZ)
3.3.2 Restricted Development Zone (RDZ)
3.3.3 Moderate Development Zone (MDZ)
3.3.4 Tourism Development Zone (TDZ)
4 Conclusions
References
Acoustic design comparison for one auditorium indoor environment: A case study of bell-shaped University hall in China
1 Instructions
2 A Case Study
2.1 Description of the Auditorium
2.2 Design requirement
3 Materials and Methods
3.1 Optimization parameters
3.2 Objectives of sound quality evaluation
3.3 Software simulation
4 Results and Discussion
5 Simulation Results Analysis
6 Conclusions
References
Intelligent disaster prediction and prevention technology
Intelligent disaster prediction and prevention technology
1 Project Overview
2 Model Construction and Zoning Methodology
2.1 Delineation of zoning units
2.2 Flood risk analysis methods
2.3 Flood risk zoning
2.4 Flood prevention and control zoning
3 Flood Model Construction
3.1 Research data
3.2 Model range
3.3 Boundary conditions
3.4 Model construction
3.5 Model parameter calibration and validation
4 Analysis of Zoning Results
4.1 Analysis of flood risk elements results
4.2 Flood risk zoning analysis results
4.3 Analysis results of flood control zoning
5 Conclusion
References
Factors affecting thermal insulation performance of aluminum honeycomb: A numerical simulation study
1 Introduction
2 Method
3 Result and Discussion
3.1 Different honeycomb cell sizes
3.2 Different honeycomb core thickness
4 Conclusions
References
Analysis of Geo-electric observation anomaly before the 2021 MS 6.4 Yangbi earthquake, Yunnan, China
1 Introduction
2 Observation Background
3 Data Analysis
3.1 Analysis of earth resistivity data
3.2 Geoelectric field data analysis
4 Conclusion and Discussion
References
Influence of a bridge on flood control downstream of a sharp bend of a mountain river
1 Introduction
2 Study Area, Mothod and Data
2.1 Study area
2.2 Method
2.3 Data
3 Results
4 Conclusions
References
Engineering geological survey and treatment plan of landslide in Yangbi County
1 Introduction
2 Purpose and Method of Investigation
2.1 Geological survey and mapping
2.2 Drilling method
2.3 Sampling
2.4 In situ tests
2.5 Laboratory test
3 Engineering Geological Conditions and Hydrogeological Conditions
3.1 Regional geological structure
3.2 Hydrogeological conditions of slippery slope area
3.2.1 The surface water
3.2.2 Groundwater
4 Geological Characteristics of Landslide Engineering
4.1 Landslide form
4.2 Landslide structure
4.2.1 The sliding body structure
4.2.2 Sliding band (surface) features
4.2.3 Slide bed structure
4.3 Type and scale of the landslide
4.4 Cause analysis of landslide
4.4.1 Geological factors
4.4.2 The foot of the slope forms the air surface
4.4.3 The slope is in an unstable state
4.5 Landslide stability evaluation
4.6 Engineering geological evaluation of landslide
5 Project Disposal Proposal
6 Conclusions
References
Early warning and supervision of foundation pit excavation construction of buried sewage treatment plant
1 Introduction
2 Hardware Design
2.1 Early warning platform
2.2 Management platform
3 Software Design
4 Example Analysis
4.1 Project overview
4.2 Monitoring frequency and period
4.3 Monitoring alarm value
5 Conclusions
References
Application of comprehensive geophysical prospecting technology in a Rrban geological survey
1 Introduction
2 Geological and Geophysical Characteristics of the Survey Area
3 Interpretation of Geophysical Detection Results
3.1 Concealed fault detection
3.2 Underground cavity detection
4 Conclusions
References
Fire scenario deduction and simulation in a three-story island subway station
1 Introduction
2 Theoretical Basis
2.1 DBN model
2.2 Theoretical model
3 Subway Fire Accident Scenario Inference
3.1 Mechanism and disposal process of subway fire disaster
3.2 Situational element determination
3.3 Scenario state probability calculation
4 Accident Simulation Analysis Based on Scenario Construction
4.1 Accident scenario construction of typical scenarios
4.2 Analysis of numerical simulation results
5 Conclusions
References
Centrifuge modelling optimization of deformation control in using isolation piles for existing tunnel adjacent to excavations by numerical simulation
1 Introduction
2 Design of the Centrifuge Model Test
2.1 Description of the test
2.2 Layout of test materials and components
2.3 Sensor arrangement
3 Numerical Simulation
3.1 Numerical model
3.2 Calculation result
4 Optimization of the Test Design
4.1 Selection of test materials
4.2 Excavation simulation
4.3 Optimization of measurement
5 Conclusions
References
Risk assessment of mountain torrent disaster in Nanping City based on GIS and random forest
1 Introduction
2 Overview of the Study Area and Data Sources
2.1 Overview of the study area
2.2 Data sources and processing
2.2.1 Index selection and processing
2.2.2 Sample data
3 Model and Method
3.1 Random forest model
3.2 Index screening
3.2.1 Multicollinearity analysis
3.2.2 Index importance measure
3.3 Model building and evaluation process
3.4 Model testing
4 Results and Discussion
4.1 Index importance analysis
4.2 Risk assessment results
4.2.1 Risk assessment accuracy evaluation
4.2.2 Risk zoning statistics
5 Conclusions
References
Application and error analysis of free station method based on the electronic total station in soft rock tunnels
1 Introduction
2 Layout of Measuring Points and Manufacture of Laser Target
3 Measurement Principle
4 Error Analysis
4.1 Measurement model
4.2 Evaluation of uncertainty component
5 Conclusions
References
The performance variation of different cross-sectional road surface types
1 Introduction
2 Method
3 Test Results and Discussions
4 Conclusions
References
The influence of warming-humidifying and reclaiming wasteland on hail disasters on both sides of the Central Tianshan Mountains
1 Introduction
2 Experimental Design
2.1 Overview of the study area
2.2 Data
2.3 Methods
2.3.1 Construction of disaster index
2.3.2 The contribution rate of influencing factors
3 Results
3.1 Spatial variability of hail disasters
3.2 Monthly variability of hail disasters
3.3 Annual variability of hail disasters
4 Discussion
5 Conclusions
References
Study on the scheme of lowering the flood control section
1 Introduction
2 Existing Problems
3 Case Analysis
3.1 Research background
3.2 Research purposes and ideas
3.3 Research analysis
3.3.1 Basin characteristics
3.3.2 Design storm
3.3.3 Design flood
3.4 Flood discharge capacity of downstream river
3.5 Recommended scheme for downward movement of the control section
3.6 Flood level review
4 Conclusions
References
The effect of traffic blockage on fire smoke spreading and safety escape in highway tunnel
1 Introduction
2 Methodology
2.1 Simulation software
2.2 Model settings for fire smoke spread simulations
2.2.1 Tunnel and fire source model
2.2.2 Vehicle-blockage conditions
2.3 Model settings for personnel escape simulations
2.3.1 Hypothesis
2.3.2 Settings of traffic characteristics
2.3.3 Setting of personnel characteristics
2.3.4 Vehicle and personnel distribution
2.3.5 Case settings and boundary conditions for evacuation simulation
3 Results and Discussion
3.1 Influence of blocked vehicles on smoke diffusion
3.1.1 Influence of blocking area and blocking ratio on critical wind speed
3.1.2 Influence of height of blocked vehicles on vault temperature
3.2 Analysis of personnel escape influencing factors in the event of traffic blockage
3.2.1 Influence of stair slope on personnel escape
3.2.2 Influence of resting platform on personnel escape
3.2.3 Influence of stair spacing on personnel escape
4 Conclusion
References
Simulation of sunshine temperature field of steel box girder under marine environment
1 Introduction
2 Measurement and Results of Solar Irradiation Temperature on Steel Box Girder
2.1 Temperature measurement
2.2 Measurement result
2.2.1 Atmospheric temperature
2.2.2 Measuring point temperature
3 Sunshine Temperature Field Simulation of Steel Box Girder
3.1 Structural model parameters
3.2 Physical and thermal parameters of the material
3.3 Finite element model of sunshine temperature field
4 Analysis of Temperature Field on Steel Box Girder Under Solar Radiation
4.1 Temperature field distribution law in different periods
4.2 Top plate temperature change rule
4.3 Temperature variation rule of the bottom plate
5 Conclusions
References
Comparison and analysis of three common ground treatment methods for the treatment effect of airport thick fill foundation
1 Preface
2 Theoretical Analysis
3 Parameter Setting and Model Establishment
4 Simulation Results
5 Conclusions
References
Online optimization for enhanced Tunneling Boring Machine (TBM) attitude control during the tunneling process
1 Introduction
2 Related Studies
3 Methodology
3.1 Virtual model construction of TBM thrust system
3.2 Online genetic optimization for PID parameters
4 Case Study
4.1 Case background
4.2 Virtual thrust system verification
4.3 Result analysis
5 Discussions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
6 Conclusions
References
Research and application of a mountain flood forecasting and early warning system based on Xinanjiang model
1 Introduction
2 System Introduction
2.1 Monitoring system
2.2 Forecasting system
2.2.1 Evapotranspiration calculation
2.2.2 Runoff production calculation
2.2.3 Water source calculation
2.2.4 Confluence calculation
2.3 Early warning system
2.3.1 Forecasting and early warning
2.3.2 Monitoring and early warning
3 System Application Effect
4 Conclusions
References
Microscopic investigation methods for the integration of old and new asphalt and prospects for application in foamed warm-mix asphalt recycling mixture
1 Introduction
2 Microscopic Test Methods of Old-New Asphalt Integration
2.1 Fourier Transform Infrared spectroscopy (FTIR)
2.2 Gel Permeation Chromatography (GPC)
2.3 Atomic Force Microscopy (AFM)
2.4 Scanning Electron Microscopy (SEM) and Fluorescence Microscopy (FM)
3 Microscopic Test Applications of Old-New Asphalt Integration
3.1 Sample preparation method
3.2 Application of Fourier Infrared spectroscopy (FTIR)
3.2.1 Evaluation of the degree of integration of old-new asphalt by the aging index
3.2.2 Evaluation of the degree of integration of old-new asphalt by adding tracer
3.3 Application of Gel Permeation Chromatography (GPC)
3.3.1 Qualitative evaluation of old and new asphalt integration
3.3.2 Quantitative evaluation of old and new asphalt integration
3.4 Application of Atomic Force Microscopy (AFM)
3.5 Application of Scanning Electron Microscopy (SEM) and Fluorescence Microscopy (FM)
3.5.1 Application of Scanning Electron Microscopy (SEM)
3.5.2 Application of Fluorescence Microscopy (FM)
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4 Microscopic Research Prospect on Asphalt Integration in Foamed Warm-Mix Recycling Mixture
4.1 Microscopic test sample preparation method
4.1.1 Sample preparation using uncompacted asphalt mixture
4.1.2 Sample preparation using compacted asphalt mixture
4.2 Evaluation of the asphalt integration by adding tracers
4.3 Evaluation of the asphalt integration without adding tracers
5 Conclusions
References
Intelligent diagnosis and evaluation method of continuous beam bridge based on knowledge graph and management data
1 Introduction
2 Methods
2.1 Knowledge graph of bridge health
2.2 Construction of bridge intelligent diagnosis and evaluation model
2.2.1 TF-IDF
2.2.2 NBM.
2.2.3 SVM
2.2.4 RF
3 Results
4 Discussion
5 Conclusions
References
The number of independent variables in vibration frequency-based parametric identification of bridge cables
1 Introduction
2 Cable Vibration Equation and its Frequency Equation
3 Number of Independent Variables of the Cable Vibration Frequency Equation
4 Number of Independent Variables of the Cable Finite Element Frequency Equation
5 Number of Independent Variables in Dynamic Stiffness Frequency Equation
6 Example Verification
7 Conclusions
Declaration of Competing Interest
References
Ultimate bearing capacity and failure mode of weathered sandstone foundation considering edge load effect
1 Instruction
2 Field Test Overview
2.1 Site engineering geology and test overview
2.2 Analysis of field test results
2.2.1 1 Analysis of ultimate bearing capacity of foundation
2.2.2 Analysis of foundation deformation and failure mode
3 Numerical Simulation Analysis
3.1 Model establishment and parameter selection
3.2 Analysis of stress and deformation of foundation
3.2.1 Foundation stress analysis
3.2.2 Foundation deformation analysis
4 Conclusions
References
Double-loop dynamic prediction method for remaining life of RC bridges based on Bayesian probability box
1 Introduction
2 Analysis of Epistemic Uncertainty Based on Bayesian Probability Box
2.1 Corrosion model of reinforcement
2.2 Pitting spatial variability
2.3 Bayesian probability box
3 Target Reliability Index Analysis Based on Personal Safety
3.1 Impact of design reference period on target reliability
3.2 Personal safety criteria
3.3 Target reliability calculation
4 Case analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.1 Reliability analysis
4.2 Analysis of epistemic uncertainty parameters
4.3 Prediction of bridge remaining life
5 Conclusions
References
Seismic vulnerability curve fitting of power grid facilities based on Wenchuan earthquake
1 Introduction
2 Seismic Vulnerability of Substations
2.1 Calculation method of seismic vulnerability of substations
2.2 Construction of calculation formula for seismic damage index of substations
2.3 Classification of seismic damage level of substations
2.4 Seismic vulnerability curve fitting of the substation
3 Seismic Vulnerability of Electric Transmission Lines
3.1 Calculation method of seismic vulnerability of electric transmission lines
3.2 Calculation of seismic damage index of electric transmission lines
3.3 Classification of seismic damage levels of electric transmission lines
3.4 Seismic vulnerability curve fitting of electric transmission lines
4 Conclusions
Data Availability
Conflicts of Interest
References
Fire simulation at the end of cable tunnel based on high voltage cable burning experiment
1 Introduction
2 Cable burning test
2.1 Experimental instruments and samples
2.2 Analysis of experimental results
2.2.1 Density
2.2.2 Heat release
2.2.3 Quality loss
2.2.4 Flue gas production
2.2.5 Other Gases and toxicity indicators
3 Numerical Simulation
3.1 Model building
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.2 Grid division
3.3 Boundary conditions
4 Simulation Results and Analysis
4.1 Fire development and smoke dispersion
4.2 Visibility
4.3 Temperature change
4.4 CO generation
5 Post-Disaster Smoke Evacuation
6 Conclusion
References
Tensile properties of glass fiber reinforced polymer bars wrapped with carbon fiber reinforced polymer sheet
1 Introduction
2 Test Design
2.1 Test materials
2.2 Experimental apparatus and loading system
3 Test Results
3.1 Failure modes
3.2 Analysis of text data
4 Conclusions
References
Centrifuge model test study of deformation characteristics of roadbed with geotextiles at different zones
1 Introduction
2 Centrifuge Model Tests
2.1 Equipment
2.2 Models
2.3 Test schemes
3 Observation
3.1 Settlement response
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
3.2 Displacement development
4 Deformation of the Roadbed
4.1 Deformation distribution
4.2 Behaviors of the roadbed
5 Reinforcement Characteristics of Geotextiles
6 Conclusion
References
Vulnerability evaluation of geological disaster based on improved fuzzy comprehensive evaluation method
1 Introduction
2 Project Profile
3 Vulnerability Evaluation Method
4 Vulnerability Analysis of Landslide Hazards
4.1 Selection of evaluation factors
4.2 Evaluation index classification standard
4.3 Vulnerability calculation results
5 Conclusions
References
Geological disaster risk warning model under different rainfall conditions in Xindu district
1 Introduction
2 Project Overview
3 Rainfall Type Analysis
4 Mechanism of the Action of Rainfall on Geological Disasters
5 Determination of Rainfall Thresholds
6 Xindu District Early Warning Model
7 Conclusion
References
Author index
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CIVIL ENGINEERING AND DISASTER PREVENTION
Civil Engineering and Disaster Prevention focuses on the research of civil engineering, architecture and disaster prevention and control. These proceedings gather the most cuttingedge research and achievements, aiming to provide scholars and engineers with valuable research direction and engineering solutions. Subjects covered in the proceedings include: l l l l l
Civil Engineering Engineering Structure Architectural Materials Disaster Prevention and Control Building Electrical Engineering
The works of these proceedings aim to promote the development of civil engineering and environment engineering. Thereby, fostering scientific information interchange between scholars from the top universities, research centers and high-tech enterprises working all around the world.
Taylor & Francis Taylor & Francis Group http://taylorandfrancis.com
PROCEEDINGS OF THE 4TH INTERNATIONAL CONFERENCE ON CIVIL, ARCHITECTURE AND DISASTER PREVENTION AND CONTROL (CADPC 2023), SUZHOU, CHINA, 24–26 MARCH 2023
Civil Engineering and Disaster Prevention Edited by
Abhijit Mohanrao Zende Dr. Daulatrao Aher college of Engineering, India
Xin Ren Nanjing Tech University, China
Qingfei Gao Harbin Institute of Technology, China
First published 2023 by CRC Press/Balkema 4 Park Square, Milton Park, Abingdon, Oxon, OX14 4RN and by CRC Press/Balkema 2385 NW Executive Center Drive, Suite 320, Boca Raton FL 33431 CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business 2024 selection and editorial matter Abhijit Mohanrao Zende, Xin Ren & Qingfei Gao; individual chapters, the contributors The right of Abhijit Mohanrao Zende, Xin Ren & Qingfei Gao to be identified as the author of the editorial material, and of the authors for their individual chapters, has been asserted in accordance with sections 77 and 78 of the Copyright, Designs and Patents Act 1988. All rights reserved. No part of this book may be reprinted or reproduced or utilised in any form or by any electronic, mechanical, or other means, now known or hereafter invented, including photocopying and recording, or in any information storage or retrieval system, without permission in writing from the publishers. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. British Library Cataloguing-in-Publication Data A catalogue record for this book is available from the British Library Library of Congress Cataloging-in-Publication Data A catalog record has been requested for this book ISBN: 978-1-032-54618-6 (hbk) ISBN: 978-1-032-54620-9 (pbk) ISBN: 978-1-003-42582-3 (ebk) DOI: 10.1201/9781003425823 Typeset in Times New Roman by MPS Limited, Chennai, India
Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Editor(s), ISBN 978-1-032-54618-6
Table of Contents xiii xv
Preface Committee Member
Civil seismic detection and structural reinforcement Impact of subway station upper span construction on existing railway tunnel structure Silei Li, Chunfu Huang, Yu Yang & Hai Tian The safety of adjacent road structures in a tunnel construction project Limei Wang, Ting Hu, Yan Zhang & Xiaoya Huang Experimental and finite element parameter analysis of modular assembled composite shear wall considering corner structure optimization Fuchen Wu, Zhulin Nie, Jihua Mao, Wei Chen, Dayang Wang, Ye Yang, Chuanglian Luo & Rongxin Guo Engineering application of the tunnel disease rapid detection system Chao Wang Deep learning-based recognition method of ground penetrating radar images for cracks inside pavement structures Qian Liu Shaking table test of frame structure considering bidirectional earthquake Hongmei Ren, Gongsheng Peng, Fuwen Zhang & Hongwei Si
3 9
14
29
36 42
Establishment method of evaluation index for the impact of chloride ion erosion on bearing capacity of reinforced concrete structure of tidal sluice Bin Yan, Lei Chen, Jiabao Song & Chengfa Deng
51
Evaluation of technical status of concrete beam bridge based on machine learning Chengyu Li
56
Reinforcement effect of the anti-slide pile on structural slope plane based on point safety factor method Xun Zhang, Yixin Chen & Yi Kuang
63
Experimental study on the influence of stirrup on axial compression performance of full-scale concrete square column Yunda Shao
70
The method of monitoring the health status of buildings based on Beidou Zhuang Zhu, Shaolin Hu & Ye Ke Based on Beidou navigation satellite system bridge variation monitoring analysis and research Jie Zhang, Shaolin Hu & Ye Ke
v
77
84
Numerical simulation of reinforced concrete beam four-point bending test based on dual-particle Peridynamics Yuanze Xu & Zili Dai
91
Discussion of foundation improvement methods for thick rock-filled gravel based on Wudangshan Airport, Shiyan, China Bin Yan & Zhiheng Shang
102
Effects of length, shape, and dosage of steel fiber on mechanical properties of steel fiber reinforced concrete Yuyang Wu & Tianyu Shao
109
Experimental study on the effect of fly ash proportion on the mechanical properties of sand concrete made by waste ultra-fine sand Yixuan Wang, Zijun Tang, Aohui Tu & Chaohua Jiang
118
Deformation analysis of enclosure structure affected by foundation pit excavation Jianghao Guo, Zheng Yang, Yike Dang & Chunting Lu
123
Key technology research on the establishment of the prefabricated component library of assembled structure based on BIM Xue Wang, Lu Zhao, Shangang Wang, Miao Zhang, Guojiao Wen, Xiaohong Gao & Wen Liu
130
Sensor fault classification for bridge SHM using LSTM-based with 1D-CNN feature extraction Yufei Guo
142
Demolition technology of long-span concrete box girder in the upper span closed frame channel Yanyang Li, Yanhui Cao & Jinhua Ye
148
Design and application of the three-dimensional model for seepage deformation of diversion tunnel Zhijing Xu, Peng Zhao, An Dong & Yuhuan Gao
156
Seismic response analysis of large-span structure considering rotational ground motion under uneven settlement sites Haolin Han, Hao Zhang & Hongnan Li
162
Numerical simulation of mechanical properties of fiber-reinforced recycled concrete beam-to-column joints under monotonic loading Tianbei Kang, Ye Yuan, Jinghai Zhou, Yibo Liang, Jingtong Qu & Liwei Jin
169
Prediction and experimental research on fatigue life of corroded prestressed concrete beams Lizhao Dai & Jingjin Liu
175
Size effect of fly ash geopolymer concrete at different temperatures Shuangxing Wang The key problems of arch supported roof structural system design of a sports building in Xiaogan Yifeng Wu, Han Ji, Jun Dong, Hongsheng Li, Wei Wang & Zhifang Li
vi
185
191
An explainable two-stage data-driven approach for risk modelling in tunnel construction Fenghua Liu, Wenli Liu, Yangyang Chen & Yafei Qi
201
Experimental study on the collapse resistance of precast monolithic reinforced concrete beam-column subassembly with additional connecting steel bars Yihua Zeng & Yanpeng Shen
209
Experiment on the fatigue behavior of steel bridge deck pavement structures paved by high-content hybrid steel fiber reinforced self-compacting concrete Hua Zou, Yu Pang, Long Feng, Jiyun Zhang, Jiansong Liu, Chengyang Wang, Qingguo Yang, Ying Li & Xuefeng He Structural performance design analysis of a high-rise office building beyond the limit Xiaofang Cao, Ling Huan & Erhong Hu Study on stress intensity factor of steel wire with double surface cracks Hongsheng Xu & Ying Yang
216
231 239
Intelligent building and equipment installation technology Utilization of recycled construction waste filler in urban greenway Peichen Cai, Xuesong Mao, Xiaoyong Lai, Qian Wu, Xiang Li & Xiaojun Shi Surveying and mapping ancient buildings with 3D laser scanning technology Dongyang Huang Building Integrated Photovoltaic (BIPV): Applications and development Shilong Jia, Yinchu Ma & Jinghui Gao
251
258 264
Design method of municipal water supply and drainage pipeline based on BIM technology Yuhong Gan & Haibin Yu
270
Lateral stability analysis of single-column pier bridge based on geometric nonlinearity Wenxue Wang & Jianlei Ma
276
Design and application of cable crane for Zhangjiajie Grand Canyon Glass Bridge Qiang Yi, Junlong Zhou, Xiaomin Liu, Chuan Yan, Wenbin Geng & Penglin Xie
282
A comparison study on different suction bucket pre-piling templates for offshore wind jacket foundation Zhenya Tian, Rongsheng Zhang, Ronghua Zhu, Jiezhan Chen, Hanqiu Liu, Meiyang Zhang & Xiang Sun Numerical simulation study on ground subsidence caused by cavity under buried PCCP socket Lin Cheng, Pengsheng Pan, Yuheng Zhang, Zengguang Xu & Yue Jiang The application of BIM technology in urban utility tunnel Chuanhua Xu & Tiantian Ma
vii
289
298 310
Experimental study on indoor model of water vapor humidification remodeling unsaturated loess under negative pressure conditions Yan-feng Zhang & Qiong Xia
316
Construction technology of retaining structures for deep foundation pit with H-shaped composite steel anchor piles-precast slab walls by compaction grouting method Luheng Gao
327
Construction method and application of Dynamic Ultimate-Strength Control Technology (DUSCT) for ultra-shallow buried excavation Zhao Zhang, Fei Liu & Huanqiu Li
334
Optimization and safety evaluation of dismantlement scheme for point-supported glass curtain wall in airport terminal buildings Jian Li, Jian Hong, Shiyao Liu, Yuzai Zhou, Guangbo Wang & Chengxiang Xu
343
5D building information modeling status based on bibliometrics Hui Sun, Terh Jing Khoo, Jiao Wang & Maoying Wang
358
Hydraulic fracturing test on in-situ stress and its distribution regularities of a flood drainage construction engineering site Ping Lu, Jinliang Zhang, Guoxiang Guo, Qi Nie, Yong Cheng & Yiming Wen
366
Cable forces and optimization of construction process on composite girder of cable-stayed bridge Ersen Huang, Hongjun Ke, Zhuoyi Chen & Huanhuan Hu
373
The design and optimization of the building electromechanical system based on BIM technology Huize Wu
383
Application of Internet of Things and BIM technology in building intelligent operation and maintenance system Fengyi Han, Dejun Kong, Kaixiang Wang, Fei Du & Zhihan Zhu
388
Horizontal top displacement change patterns of the pile foundation in slope under freeze-thaw cycles Gaokai Lu & Chunxiang Guo
394
Experimental study on deformation characteristics of cement soft soil under the influence of multiple factors Yujie Zhang, Hongjun Liu & Zijie Liang
400
Application of composite foundation with vibro-crushed stone column in saturated liquefied sand ground treatment of a gas-fired power station in Myanmar Zheng Zheng, Wei Zhu, Honebo Li, Yafeng Lou & Yang Zhang
407
Application of oblique photography technology in airport site selection Zizhu Zhang, Lan Cheng, Xinpu Feng, Xiao Wang, Guoliang Zhai, Ke Tang & Shiman Sun
413
Satellite positioning and inertial navigation based firefighters positioning system Xiqing Liu, Ansong Feng, Chonglin Gu & Guozhan Wang
420
viii
Moisture accumulation in wall thermal insulation layer under the action of coupled heat and moisture during heating period in cold regions Weihua Zheng, Yuan Su, Yun Gao, Chao Song, Caihong Qi, Zehua Feng & Wenfei Zhao Bridge strain based on real bridge and indoor experiment Xiaofan Feng, Yu Tang, Lu Peng, Bin Li & Lin Bai
426
436
Design of test platform for early fire detection of power transmission and transformation equipment Guo-qiang Liu, Guo-chun Li, Ya-nan Hao & Zhi-peng Zhao
444
Landscape conservation planning based on forest fire prevention: A case study of Lushan Forest Park in Sichuan Province Meixin Qiu & Zigang Yao
453
Analysis on the widening reconstruction of existing bridge based on grillage analysis theory Bing Li & Jinhui Chen
460
Acoustic design comparison for one auditorium indoor environment: A case study of bell-shaped University hall in China Ziqing Tang, Zhengguang Li & Meijun Jin
470
Intelligent disaster prediction and prevention technology Comprehensive risk analysis method for flood and prevention regionalization in Luzhou urban area Hao Shen, Daling Cao, Shu Liu, Furen Jiang & Hongtao Wan
481
Factors affecting thermal insulation performance of aluminum honeycomb: A numerical simulation study Junhao Gao, Rongnan Yuan & Shouxiang Lu
494
Analysis of Geo-electric observation anomaly before the 2021 MS 6.4 Yangbi earthquake, Yunnan, China Jie Miao, Ye Fan, Bing Han, Xiaona Dong & Lei Xv
500
Influence of a bridge on flood control downstream of a sharp bend of a mountain river Hua Ge, Rui Long & Chunyan Deng
508
Engineering geological survey and treatment plan of landslide in Yangbi County Jianghua Liu
514
Early warning and supervision of foundation pit excavation construction of buried sewage treatment plant Jiajun Shen
520
Application of comprehensive geophysical prospecting technology in a Rrban geological survey Bin Xu & Guo-feng Yin
526
Fire scenario deduction and simulation in a three-story island subway station Xin Zhang, Zhanyou Sa, Nan Li, Jingbo Wu & Yongliang Yang
ix
531
Centrifuge modelling optimization of deformation control in using isolation piles for existing tunnel adjacent to excavations by numerical simulation Bingyi Wang, Yiming Du, Yu Diao & Xiangyu Zhao Risk assessment of mountain torrent disaster in Nanping City based on GIS and random forest Xuting Ma, Ronghua Liu, Xiaolei Zhang, Guiyi Zheng, Wenbo Wu, Jingyu Yue & Shufen Chen Application and error analysis of free station method based on the electronic total station in soft rock tunnels Yuanding Xing, Yongxiang Shi, Yingcheng Peng & Shiqian Song The performance variation of different cross-sectional road surface types Xufeng Li The influence of warming-humidifying and reclaiming wasteland on hail disasters on both sides of the Central Tianshan Mountains Yuan Liu, Xi Wang, Ngai Cheong & Guohong Zheng Study on the scheme of lowering the flood control section Ting Zhang & Junbo Yao The effect of traffic blockage on fire smoke spreading and safety escape in highway tunnel Zubin Ai, Zhensheng Cao, Shengjun Hou, Chenchen Jiang & Jianbin Zang Simulation of sunshine temperature field of steel box girder under marine environment Bin Zhou, Linghua Zeng, Dong Zhang, Yong Zhang, Yuewei Yang, Zhaoting Liu & Xiangfan Cao
540
550
561 568
573 582
588
602
Comparison and analysis of three common ground treatment methods for the treatment effect of airport thick fill foundation Bin Yan, Jie Ma & Yecheng Li
611
Online optimization for enhanced Tunneling Boring Machine (TBM) attitude control during the tunneling process Kunyu Wang & Limao Zhang
620
Research and application of a mountain flood forecasting and early warning system based on Xinanjiang model Hongri Zheng, Shun Yu & Yong Luan
629
Microscopic investigation methods for the integration of old and new asphalt and prospects for application in foamed warm-mix asphalt recycling mixture Kai-wen Lei & Yong-chun Qin
635
Intelligent diagnosis and evaluation method of continuous beam bridge based on knowledge graph and management data Qiao Guo, Guodong Shen & Zhu Liang
646
The number of independent variables in vibration frequency-based parametric identification of bridge cables Libo Meng, Jingbo Liao, Gang Liu, Cong Luo, Xiaoping Qin, Guangwu Tang, Wenbing Chen & Banfu Yan
x
653
Ultimate bearing capacity and failure mode of weathered sandstone foundation considering edge load effect Guan-hua Zhao, Yan-jie Zhang, Jian-dong Li & Zhao Long
666
Double-loop dynamic prediction method for remaining life of RC bridges based on Bayesian probability box Lizhao Dai & Peng Liu
679
Seismic vulnerability curve fitting of power grid facilities based on Wenchuan earthquake Rushan Liu, Zhenhui Wang, Lupeng Yan, Deyuan Tian & Meng Wu
689
Fire simulation at the end of cable tunnel based on high voltage cable burning experiment Shuhan Wang, Yongkang Zhang, Ling Wang, Qi Yuan, Chenbin Wu, Cheng Wang & Jianbin Zang
701
Tensile properties of glass fiber reinforced polymer bars wrapped with carbon fiber reinforced polymer sheet Linjie Li
713
Centrifuge model test study of deformation characteristics of roadbed with geotextiles at different zones Minwei Deng, Tianyi Chen, Zhenhua Liao & Taiping Mu
719
Vulnerability evaluation of geological disaster based on improved fuzzy comprehensive evaluation method Fuji Gu, Ronghui Shen, Di An, Yupu Wu & Yugang Li
727
Geological disaster risk warning model under different rainfall conditions in Xindu district Yupu Wu, Guoqin Yan, Yu Liu, Jie Zhao & Lifeng Wang
733
Author index
739
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Preface The International Conference on Civil, Architecture and Disaster Prevention and Control originates from 2020 and has attracted various promising researchers who are developing new directions in the field of civil, architecture and disaster prevention and control. The 2023 4th International Conference on Civil, Architecture and Disaster Prevention and Control (CADPC 2023), held via virtual form in Suzhou, China from March 24th to 26th, 2023, is meant to unite researchers, experts, professors and academic staff as well as a wide range of participants of professional community from China and other countries concerned. Under the promising unquantifiable damage brought by different hazards, how to promote disaster risk reduction of human life, environment, infrastructure and major projects through the development of science and technology has become the most important, popular and challenging subject in engineering discipline. And the promotion of architectural-related technologies and technologies for disaster forecasting, prevention and control are among the most critical part. Therefore, CADPC 2023, with more than 120 participants, was held to further enhance construction technologies for predicting and resisting disasters. Highly eventful program of the Conference, the list of well-established organizationsparticipants and devoted engagement of many colleagues throughout all the stages of Conference preparation instill confidence in practical importance of mutual initiative. The framework of the Conference comprises keynote speeches, oral presentations, and academic investigation, discussing open and significant problems facing construction technologies, proposing innovative ideas and approaches to solution of the problems, and considering the new possibilities of application and development of cutting-edge architecture technology. The Conference has become one of the venues for demonstrating scientific achievements of international class by leading scientists and young scholars and would facilitate the strengthening of academic cooperation, meanwhile the Conference proceedings would be expressed in the form of international scientific publishing. Various topics of the papers, gone through a vigorous peer review process, are covered in the proceedings, including: Civil Engineering Surveying, New Construction Technology, Engineering Monitoring and Testing Technology, Smart Disaster Prevention, Application of Big Data in Disaster Rescue, etc. On behalf of the Conference Organizing Committee, we would like to thank the referees for their efficient and thoughtful actions. We are grateful to the members of the Technical Program Committee for their efforts in making and shaping the program for CADPC 2023. Particularly, we acknowledge the publishing support from the members of CRC Press / Balkema – Taylor & Francis Group. We hope that the future CADPC will be as successful and stimulating, as indicated with the contributions presented in this volume. The Committee of CADPC 2023
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Editor(s), ISBN 978-1-032-54618-6
Committee Member Conference Chair Prof. Abhijit Mohanrao Zende, AGTI’S, Dr. Daulatrao Aher College of Engineering, India Technical Program Committee Prof. Xin Ren, Nanjing Tech University, China Prof. Dr. Edyta Plebankiewicz, Cracow University of Technology, Poland Prof. Fadi Hage Chehade, Lebanese University, Lebanon Prof. Volkan Kahya, Karadeniz Technical University, Turkey Prof. Wenliang Lu, Beijing Jiaotong University, China Prof. Tarun Kumar Lohani, Arba Minch University, Ethiopia Assoc. Prof. Jun Xie, Hebei University of Architecture, China Assis. Prof. Abdollah Tabaroei, Eshragh Institute of Higher Education, Iran Organizing Committee Assoc. Prof. Qingfei Gao, Harbin Institute of Technology, China Senior Engineer Feng Xu, China Merchants Chongqing Communications Research & Design Institute Co., Ltd., China Assoc. Prof. Houssam Khelalfa, Selinus University of Science and Literature (SUSL), Italy Asst. Prof. Shamshad Alam, Jazan University, Kingdom of Saudi Arabia Assoc. Prof. Young-Jin Cha, University of Manitoba, Canada Lecturer Gao Lv, Xi’an Shiyou University, China Publication Chair Assoc. Prof. Qingfei Gao, Harbin Institute of Technology, China
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Civil seismic detection and structural reinforcement
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Impact of subway station upper span construction on existing railway tunnel structure Silei Li, Chunfu Huang, Yu Yang & Hai Tian* School of Civil and Architecture Engineering, Nanchang Institute of Technology, Nanchang, China
ABSTRACT: In recent years, the rapid development of subway and underground comprehensive pipe corridor has led to frequent encounters with the problem of the close connection between underground structures and existing subway section structures during engineering construction. This paper uses the finite element analysis software ABAQUS to analyze the impact of the construction process of the Sanya Bay Metro Station on the Huofengshan tunnel structure. The aim is to verify and analyze the influence of the construction process of the Sanya Bay Metro Station on the Huofengshan tunnel structure. The calculation results of the two structural models show that the construction of the upper span of the subway station affects the stress field and displacement field of the tunnel surrounding rock, as well as the stress changes in the Huofeng Mountain tunnel lining, which are less than the strength value of the lining concrete. The tunnel surrounding rock and lining structure will experience additional stress and deformation, including changes in the lining’s additional tensile stress amplitude and the impact of tunnel vault unloading. Station operation reloading leads to the redistribution of stress in the tunnel surrounding, as well as changes in the tunnel lining structure and track structure morphology, especially for the tunnel operation deepening impact. The construction of the upper span of the subway station affects the structural safety of the existing tunnel, and the safety factor of the secondary lining structure is reduced to a certain extent, providing a reference for similar projects.
1 INTRODUCTION With the rapid development of urbanization in China, subways have become the preferred solution for alleviating traffic problems in cities. As the construction of subway lines increases and existing buildings along the perimeter of the line come into play, there are more and more cases of tunneling projects spanning over and under existing building structures. During the construction process, the foundation pit and underground engineering can cause deformation and changes in the force state of neighboring buildings and structures, affecting their structural safety. For current subway deep foundation pit projects, analyzing the impact of deformation on adjacent buildings is more important than analyzing the pit itself (Wan 2001). Numerical analysis calculations provide more consistent results with field measurements, and the use of finite element analysis can provide a reference for the design and construction of the pit (Marta 2001). Studies have been conducted on the interaction of two vertically intersecting mining method construction tunnels in soft soils, including the laws of ground stress redistribution, ground settlement, and tunnel deformation (Hamid Chakeri 2010). Comparative analyses of open and concealed excavation construction schemes for tunnels over existing railroad tunnels have also been carried out (Liu 2012; Wan 2013). The *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-1
3
displacement variation laws caused by shield tunnels close up over existing mining method construction tunnels have been analyzed (Zan 2016). Simulation and analysis of the effect of pit construction on the displacement changes of the surrounding subway structure have also been conducted (Ma 2018; Zhai 2019; Zeng 2018). The construction of the upper tunnel leads to higher settlements and bending moments compared to the construction of the lower tunnel (F. Hage Chehade 2008). Model tests and numerical analysis were performed for the behavior of the near-surface ground in the tunnel crossing area and the behavior of the existing tunnel located above the newly excavated tunnel. The results show that the upper tunnel interferes with the stress flow generated by the longitudinal arch effect caused by the lower tunnel excavation. The overlap zone between the upper and lower tunnels is a zone of vertical relaxation, and there is a tendency for the vertical tunnel stresses around the overlap zone to deform and distribute stresses during tunnel excavation (Byun 2006). Based on examples of the effects of pit excavation on adjacent tunnels and a comparison of numerical analysis with actual monitoring results, the trend of tunnel deformation was evaluated, pointing out that tunnels with greater stiffness will be subjected to greater bending moments (Sharma 1972). Currently, there is a limited amount of research on the construction and operation of subway stations over existing tunnels. Building subway stations presents a complex threedimensional structural deformation problem. This study focuses on the new Sanya Bay subway station, which is built over the Huofeng Mountain tunnel. The study models and analyzes the perimeter rock and stress changes during construction, and proposes preventive measures. The findings provide a valuable reference and guidance for future projects. 2 CONSTRUCTION DETAILS 2.1
Project description
Sanya Bay Station is an underground island crossing station that is 11m deep and has a total length of 353m. It was constructed using the open excavation method. The station intersects with the Huofeng Mountain Tunnel of the Shanghai to Chengdu (also known as ‘Hu-Rong’) Railway, with a corresponding mileage of K22 + 172.451K22 + 229.551. The minimum net distance between the station floor and the Huofeng Mountain Tunnel is 14.73 m, while the minimum net distance between the pile bottom of the station foundation pit and the Huofeng Mountain Tunnel is 11.73 m. The Sanya Bay Station, which is proposed to be built, will span the Huofengshan Tunnel of Hu-Rong Railway at an oblique angle. The inner rail top elevation of the Huofengshan Tunnel is approximately 222.83 m, while the tunnel top elevation is about 231.73 m, resulting in a height difference of approximately 46m from the surface. The composite lining will be reinforced with grade III surrounding rock, with the initial support sprayed with C25 concrete having a thickness of 23 cm. The secondary lining will be 40 cm thick and made of C30 concrete. 2.2
Geological condition
The station is situated on a denuded tectonic hill that has been significantly impacted by human activities. The terrain is relatively gentle, with a slope angle of approximately 2 and a ground elevation ranging from 274–278 m. The maximum height difference is around 4 m. The lithology of the surrounding rock in the section of Huofeng Mountain Tunnel (mileage HLZD2K21 + 400 HLZD2K21 + 600) on Sanya Bay Station across Hu-Rong Railway is composed of sandy mudstone and sandstone. The rock joints are more developed at this site, and the sandy mudstone and sandstone have a medium-thick laminated structure with a general combination between layers. According to the acoustic test data, the compression wave velocity of the surrounding rock in the Huofeng Mountain tunnel is 3.43.8 (km/s), which classifies it as Class III surrounding rock. 4
3 NUMERICAL-ANALYTICAL METHODS 3.1
Material parameters
The formation of mechanical parameters and concrete parameters are shown in Tables 1 and 2. Table 1.
Physical and mechanical parameters of geotechnical materials.
Concrete grade
Density (kg/m3)
Elastic modulus (MPa)
Poisson’s ratio
Tensile strength (MPa)
Compressive strength (MPa)
C25 C30
2200 2500
23 29.5
0.2 0.2
28 38.3
17 20
Table 2. Physical and mechanical parameters of concrete and soil materials.
3.2
Category
Density (kg/m3)
Elastic modulus (MPa)
Poisson’s ratio
Friction angle ( )
Plain fill Sandstone Sandy mudstone 1 Sandy mudstone 2
1950 2500 2560 2560
500 4512 1648 1668
0.3 0.13 0.38 0.35
28 38.3 31.4 32.2
3-Dimensional model analysis
In this calculation, the large-scale general finite element analysis software ABAQUS is utilized for numerical simulation analysis. Firstly, the displacement and stress calculation results of the surrounding rock and secondary lining during the construction of the Sanya Bay Station of Chongqing Rail Transit Line 10 are analyzed to determine the impact of the station’s construction on the structural safety of the Huofengshan Tunnel of the Hu-Rong Railway. Then, the structural analysis of the Huofengshan Tunnel is conducted based on the changes in displacement and stress of the surrounding rock when the Sanya Bay Station crosses the tunnel. The Mohr-Coulomb model is used for the overlying rock and surrounding rock, while an elastic model is used for the tunnel lining. Generally, structures outside 3 to 5 times the diameter of the tunnel on both sides have minimal impact on the tunnel, as shown in Figure 1(a). Therefore, the calculation range is taken as the spanning section with a width of 120m along the longitudinal direction of the station, combined with the characteristics of the excavation of the foundation pit of the Sanya Bay Station. The distance between the ground
Figure 1.
The finite-element model of earth-tunnel structure.
5
level and the bottom of the station foundation pit and the vault of the tunnel structure is 31.27 m and 46 m, respectively, as shown in Figure 1(b).
4 RESULTS AND DISCUSSION Figure 2 illustrates the variation of the horizontal stress component, vertical stress component, maximum principal stress, and minimum principal stress of the surrounding rock before and after the construction of the Sanya Bay Station. From the perspective of the distribution changes of surrounding rock stress components and the maximum and minimum principal stresses before and after the construction of the Sanya Bay Station, it is observed that the excavation of the foundation pit of the station and the application of the load of the Sanya Bay Station redistributed the stress of the rock mass. The vertical stress and the minimum principal stress of the surrounding rock before and after the construction of the Sanya Bay Station changed significantly. After the construction and operation of the Sanya Bay Station, the load balance of the overlying structure partially recovered the stress value of the surrounding rock compared to that before the excavation of the foundation pit.
Figure 2.
Influence of station construction on tunnel surrounding rock stress.
Figure 3.
Horizontal stress change curve of surrounding rock observation point.
Figure 3 illustrates the distribution of vertical displacement of the surrounding rock water after the excavation of the Sanya Bay Station foundation pit and the completion of the station’s main construction. Figure 4 shows that significant displacement of the surrounding rock occurred near the foundation pit of Sanya Bay Station after its construction. The inverted arch rose by approximately 8.4 mm compared to its initial state after excavation and unloading, while the station 6
floor rebounded by approximately 2.1 mm after construction and operation. Additionally, the vault was uplifted by approximately 1.4 mm after excavation and unloading, and the station floor settlement after construction and operation was 4.9 mm compared to its initial state. Figure 4 displays the displacement distribution of the Huofengshan Tunnel lining structure before and after the excavation of the Sanya Bay Station, along with the change curve of the maximum and minimum principal stress of the lining structure.
Figure 4.
Influence of station construction on tunnel lining structure.
The excavation of the Sanya Bay Station foundation pit resulted in the displacement of the Huofengshan Tunnel lining structure, with the maximum vertical displacement occurring at the inverted arch of the lining structure. After the excavation of the foundation pit, the vault subsided by approximately 0.14 mm, and the inverted arch was uplifted by 4.56 mm. After the station’s construction and operation, the vault displacement was approximately 4.5 mm lower than its initial state, and the inverted arch was uplifted by approximately 3.5 mm. Before and after the excavation of the Sanya Bay Station and during the operation stage of the station, the tunnel lining vault was under pressure, and the side walls were under tension. After the excavation of the station foundation pit, the stress changed significantly, and the value was smaller than the compressive and tensile strength values of the lining concrete of the Huofengshan tunnel. Based on the above analysis, the excavation and operation process of the Sanya Bay Station foundation pit has a significant impact on the Huofengshan tunnel, and the safety factor of the tunnel lining is lower than before excavation.
5 CONCLUSIONS This paper focuses on the upper span construction of the Sanya Bay Metro Station on the Huofengshan Tunnel of the Hu-Rong Railway. Through finite element analysis of the rockstratum structure model, the following conclusions were drawn: 7
(1) The excavation of Sanya Bay Station has a significant impact on the unloading of the tunnel vault. The reloading of station operations results in the redistribution of stress around the tunnel, affecting the lining structure and track shape, particularly the tunnel’s operation. These factors should be considered in the station’s construction measures. (2) Numerical calculations of the stratum structure model show that the stress and displacement fields of the surrounding rock of the Huofengshan tunnel changed after the construction of the Sanya Bay Station. The maximum settlement of the vault of the Huofengshan Tunnel occurred in the excavation condition of the foundation pit in Sanya Bay, which was about 4.5 mm. (3) The construction of the Sanya Bay Metro Station causes additional stress and deformation to the surrounding rock and lining structure of the Huofengshan Tunnel, reducing the tunnel structure’s safety. The additional tensile stress of the tunnel lining varies greatly, posing a risk to the operation of the Huofengshan Tunnel of the Hu-Rong Railway, which should be controlled to ensure safety.
CONFLICT OF INTEREST STATEMENT The authors declare no conflicts of interest.
ACKNOWLEDGMENTS This work received financial support from the Jiangxi Province Department of Education Science and Technology Research Project (No. GJJ211907), the National Science Foundation of China (No. 52068053), and the Jiangxi Province 2021 Graduate Innovation Special Fund Project (YC2021-S819).
REFERENCES Byun. G. W, Kim D G, Lee S D. (2006) Behavior of the Ground in Rectangularly Crossed Area Due to Tunnel Excavation Under the Existing Tunnel. Tunneling and Underground Space Technology 21(1):1–6. F. Hage Chehade, et al. Numerical Analysis of the Interaction Between Twin-tunnels: Influence of the Relative Position and Construction Procedure. Tunneling and Underground Space Technology, 2008, 23(2):210–214. Hamid Chakeri, Rohola Hasanpour, Mehmet Ali Hindistan, B Unver. (2010) Analysis of Interaction Between Tunnels in Soft Ground by 3D Numerical Modeling. Bulletin of Engineering Geology and the Environment, 70 (3): 439–448. Liu M.G. (2012) Analysis and Research of a New Tunnel Over-passing an Existing Railway Tunnel with Short Distance. Modern Tunnelling Technology, 49 (5): 79–84 (in Chinese). Marta D. (2001) Tunnel Complex Unloaded by a Deep Excavation. Computers and Geotechnics, 28: 469–493. Ma Z.L. (2018) Fluid-solid Coupling Numerical Simulation of Influences of Foundation Excavation on Adjacent Tunnel[J]. Construction Technology, 47 (19):85–88 (in Chinese). Sharma J S. (1972) Effect of Large Excavation on Deformation of Adjacent MRT Tunnels. Tunneling and Underground Space Technology, (7):11–23. Wan J H. (2001) Research on Foundation Excavation’s Effect to Adjacent Subway Stations and Intervals Track. Beijing: China University of Geosciences (in Chinese). Wan F., Tan Z.S., Chen Y. (2013) Research on Construction Program of New Tunnel Passing above Existing Tunnel at a Short Distance. Journal of Beijing Jiaotong University, 37 (1): 40–45 (in Chinese). Zan Z.H. (2016) Influence of Up-crossing Shield Tunnel Construction on Deformation of Adjacent Existing Tunnels. Journal of Wuhan Institute of Technology, 38 (1): 61–67 (in Chinese). Zhai C., Chen T. (2019) Influence of Foundation Pit Construction on Itself and Existing Subway. Journal of Water Resources and Architectural Engineering, 17(1):174–178 (in Chinese). Zeng K, Cai H.B., Shi L. Analysis on Influence of Foundation Pit Construction on Neighboring Existing Metro Shield Tunneling[J]. Low Temperature Architecture Technology, 2018, 40(1):117–121 (in Chinese).
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The safety of adjacent road structures in a tunnel construction project Limei Wang, Ting Hu, Yan Zhang & Xiaoya Huang* Chongqing College of Architecture and Technology, Chongqing, China
ABSTRACT: The deformation of the adjacent road structure during the construction of the proposed tunnel project and after it is completed and put into use is analyzed through the threedimensional and two-dimensional numerical simulation calculation of the typical section, and the safety of the adjacent road structure is evaluated. The analysis results show that the excavation and construction of the adjacent road slope only cause stress changes in its local scope and have little impact on the stress of the follow-up ramp tunnel implementation scope. It can be considered that the excavation of the adjacent road slope does not affect the surrounding rock conditions of the later ramp tunnel implementation. After the excavation and construction of the side slope adjacent to the road are completed, the tunnel can meet its stability under the support conditions during the subsequent tunnel excavation. The deformation caused by the subsequent tunnel construction is only in the local area around the tunnel and has little impact on the deformation of the retaining wall. Therefore, the ramp tunnel construction is safe and has little impact on the adjacent road retaining wall, meeting the requirements.
1 INTRODUCTION In recent years, with the rapid progress of urbanization in China, the number of urban migrants has been increasing, and the traffic volume is also increasing rapidly. To relieve the pressure of urban travel, the road network system must develop to the underground space, and the working conditions of tunnel engineering increasingly appear in urban traffic construction (Fu 2020). There is no mature experience to follow to ensure that the tunnel construction reduces to an acceptable range for existing buildings (Li 2015). The finite element analysis software MIDASGTS is used to establish a physical model to simulate the construction process of a pipe jacking tunnel. The model is calculated to obtain the results. It is verified whether the model is applicable by comparing the simulation results with the measured data (Li 2015). To study the surface deformation law of the “CC construction method” construction, relying on the actual project by sorting out and analyzing the actual monitoring results of the surface deformation during the construction process, the linear regression analysis method is used to fit the data (Li 2019). Taking an interval shield tunnel of Suzhou Metro Line 5 as the research object and based on the tunneling parameters during construction and the measured monitoring data of the tunnel surface, the influence of geological conditions, bunker pressure, advance speed, and other factors on the surface deformation in the shield tunneling project is analyzed (Lou 2019). The underground excavation section of the entrance and exit section of Changchun Metro Line 1 pass under the ring expressway. The tunnel has a large span, shallow burial depth, great construction difficulty, and high requirements for ground deformation control. Therefore, the large pipe curtain advance support, double side heading method excavation, and ground deformation monitoring schemes adopted in the construction of shallow tunnels *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-2
9
were studied, emphasizing the characteristics of ground deformation caused by on-site tunnel excavation (Cui 2022). The siltstone stratum adopts the construction method of full face excavation and the reinforcement method of combining the self-advancing grouting anchor bolt and the advance support of small grouting tremie, which can effectively control the settlement and deformation while speeding up the construction progress and can be used for reference by similar projects (Fan 2020). The shield construction mainly caused soil displacement inside the threesided enclosure area, with almost no soil movement outside. The roof and floor of the underground road above the tunnel axis had a large additional deformation, and the transverse duck egg deformation of the tunnel lining ring under the three-sided enclosure was weaker than that in the semi-infinite space (Yang 2018). Based on the Aobaogou Tunnel of Bazhun Railway, FLAC3D software, and the Mohr-Coulomb elastoplastic model, numerical simulation is carried out on the influence of staged excavation of shallow buried tunnels under heavily loaded roads and heavy loading of roads on the deformation during excavation. Heavy-loaded coal vehicles influence the ground and arch crown settlements by nearly one-third. In contrast, the influence on the deformation of the arch bottom is very small (Wu 2014). After the tunnel excavation, the road structure above the tunnel produced obvious longitudinal tensile stress. When the maximum settlement is close to 3 cm, the maximum tensile stress generated by the surface and base course is close to the allowable tensile stress. The long-term existence of the settlement and tensile stress, together with the repeated action of vehicles, increases the possibility of damage to the road structure (Sun 2011). This paper mainly analyzes the deformation of the approach road structure during the construction of the proposed project and after it is completed and put into use. It also evaluates the safety of the approach road structure through the three-dimensional and twodimensional numerical calculations of typical sections. 2 PROJECT OVERVIEW The proposed tunnel is located in the underpass structure of the ramp. The design range is K0 + 159.300 to K0 + 652.754, with a total length of 493.454 m. The underpass structure mainly includes an open cut section and concealed cut section. The adjacent road is an urban expressway, a two-way six-lane road. The design speed is 100 km/h, the road width is about 27 m, and asphalt pavement is used. This is shown in Figure 1. The tunnel project crosses the adjacent road section vertically. The anchor piles are arranged parallel on the south side of the adjacent road. The distance between the pile edge and the adjacent road’s outer edge is about 16.3 m. The vertical distance between the adjacent road surface and the proposed tunnel is 6.5 m. The distance between the pile edge and the outer edge of the existing road is about 13.5 m. The vertical distance between the adjacent road pavement and the proposed tunnel bottom is 4.8 m.
Figure 1.
Topographic photo of the south slope of the project.
10
3 NUMERICAL SIMULATION ANALYSIS OF PROJECT CONSTRUCTION MIDAS/GTS, a large-scale general finite element software, is selected as the calculation platform in this assessment and analysis. MIDAS/GTS is a program directly developed for the functions required for structural analysis in geotechnical tunnels. It combines a general finite element program and geotechnical and tunnel expertise. Its new operation interface and 3D analysis function provide a powerful solution for geotechnical and tunnel engineers. 3.1
Constitutive model
The M-C model is adopted for rock and soil materials in this calculation. M-C constitutive model is the most commonly used model to simulate geotechnical materials. This model contains two failure criteria of shear and tension models, which correspond to different flow rules, respectively, and is very suitable for simulating the stress-strain characteristics of geotechnical materials. The elastic constitutive equation is adopted for the interval tunnel’s lining structure, retaining pile, and anchor bolt. 3.2
Calculation of unit type and model parameters
Quadrilateral plane strain solid elements simulate rock, soil, and tunnel structure. The element has four nodes, and each node has two degrees of freedom, which can simulate lateral deformation, vertical deformation, and relative rotation between any point. The track structure’s deformation law can be simulated using this element. The retaining pile is also simulated by a quadrilateral plane strain solid element. In the plane strain calculation model, the retaining structure is defaulted to be a structural entity with infinite longitudinal extension. The principle of equivalent stiffness must be adopted, as shown in the following formula so that the stiffness of the retaining structure is consistent with the actual situation. E w Iw ¼ E p Ip where Ew and Iw are the elastic modulus and inertia moment of the enclosure structure in the calculation. Ep and Ip are the retaining pile’s elastic modulus and inertia moment. The parameters used in this calculation are selected based on the latest survey results and the principle of considering the most unfavorable conditions. However, it should be noted that the mechanical properties of the surrounding rock around the tunnel may be reduced due to the impact of excavation blasting during the track construction. The deformation and strength parameters are reduced by 80% during the calculation and analysis. Table 1 parameters are used for calculation and analysis.
Table 1.
Project
Values of analysis parameters.
Material
Geotechnical Filling Moderately weathered sandy mudstone Tunnel C35 structure
Calculation model
Elastic modulus kPa
Mohr Coulomb
5 103 20 1.59 106 25.3
Elastic
3.15 107 25
11
Bulk density kN/m3
Cohesion kPa
Friction angle ( )
4 350
28 29
\
\
3.3
Calculation profile analysis
In this calculation, the model is simplified as follows: (1) The slope excavation does not consider step-by-step. It only considers one-step excavation, which makes the calculation result larger; (2) The tunnel is excavated in full sections at one time, and the calculation result is too large due to the 3D solid model. 3D model calculation size: 80 100 80 m. The numerical calculation model is shown in Figure 2.
Figure 2.
Numerical calculation model.
The calculation results are shown in Figure 3. It can be seen from the cloud diagram of surrounding rock deformation at the end of tunnel excavation that the maximum lateral deformation of the tunnel is about 1.2 mm at the end of tunnel excavation, which occurs at the arch waist of the upper tunnel after the completion of tunnel excavation, and is roughly evenly distributed along both sides of the tunnel. The maximum vertical deformation of the tunnel is 6.1 mm, which occurs near the arch crown of the lower tunnel, and the convergence and settlement are less than the specified limits. Therefore, it can be judged that the tunnel itself is stable during tunnel excavation.
Figure 3.
Cloud chart of surrounding rock deformation upon completion of tunnel excavation.
Figure 4 shows the surface subsidence deformation diagram. It can be seen from the figure that the maximum surface settlement of the adjacent road is 0.97 mm, and the maximum surface settlement of the widening ramp of the adjacent road is 0.20 mm, which shows that the effect on the surface settlement deformation of the adjacent road after the completion of tunnel excavation is very small. 12
Figure 4.
Surface subsidence deformation diagram.
4 CONCLUSIONS The finite element software MIDAS/GTS establishes the three-dimensional tunnel construction excavation and completion model. The tunnel and adjacent road surface deformation are analyzed through numerical simulation calculation. The following conclusions can be drawn: (1) The excavation and construction of the side slope adjacent to the road only cause stress changes within its local scope, which has little impact on the stress of the subsequent ramp tunnel implementation scope. It can be considered that the excavation of the side slope adjacent to the road does not affect the surrounding rock conditions of the later ramp tunnel implementation; (2) After the excavation and construction of the side slope adjacent to the road is completed, the tunnel can meet its stability under the support conditions during the subsequent tunnel excavation; (3) The deformation caused by the subsequent tunnel construction is only in the local area around the tunnel and has little impact on the deformation of the retaining wall; Therefore, the ramp tunnel construction itself is safe and has little impact on the adjacent road retaining wall, meeting the requirements.
REFERENCES Cui Guangzhen. Construction Technology of Shallow Buried Tunnel with Large Cross Section Undercrossing Expressway [J]. Building Construction. 2022, 44 (01): 118–120. Fan Changjie. Analysis of Road Deformation Caused by Tunnel Undercrossing Expressway Construction [J]. Highway. 2020, 65 (03): 299–305. Fu Lilei. Research on Deformation Control Technology of Super Shallow Buried Soft Rock Large Section Tunnel Passing Under Existing Traffic Subgrade [D]. China Academy of Railway Sciences. 2020. 12. Li Yufu. Research on the Influence of Surface Deformation of Tunnel Construction on Existing Buildings and Countermeasures [J]. Fujian Architecture. 2015, (08): 82–84. Li Dongfeng. Finite Element Analysis of Ground Settlement Caused by pipe Jacking Tunnel Construction [D]. Anhui University of Technology. August 2015. Li Peng, Li Yang, Gao Yi, Yu Shaohui, Li Yingfei. Analysis and Research on Ground Deformation Law of Pipe Jacking Tunnel Construction Based on “CC Construction Method”. Tunnel Construction (English and Chinese). 2019, 39 (11): 1838–1847. Lou Zhangyin. Control of Surface Deformation Caused by Shield Tunneling in Soft and Water Rich Strata [J]. Transportation Science and Technology and Economy. 2019, 21 (02): 61–67. Sun Jian, Song Hongwei, Wang Tianchun Study on the Influence of Tunnel Underpass on Road Structure’s Stress and Settlement Feformation [J] Journal of Water Resources and Building Engineering. 2011, 9 (05): 18–23. Wu Dongpeng, Yang Xin’an, Wu Chong. Research on Deformation Law and Control of Shallow Buried Soft Tunnel Under Heavy Load Road [J]. Journal of East China Jiaotong University. 2014, 31 (03): 23–28. Yang Xiao, Li Mingyu. Study on the Construction Deformation Law of the Shield Tunnel Overlapping Underpass. Journal of Hebei University of Engineering (Natural Science Edition). 2018, 35 (03): 49–53.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Experimental and finite element parameter analysis of modular assembled composite shear wall considering corner structure optimization Fuchen Wu Guangzhou Guangjian Construction Engineering Testing Center Co., Ltd., Guangzhou, China Guangdong Engineering Technology Research Center of Building Health Monitoring and Safety Early Warning, Guangzhou, China
Zhulin Nie* & Jihua Mao Guangzhou Guangjian Construction Engineering Testing Center Co., Ltd., Guangzhou, China School of Civil Engineering, Guangzhou University, Guangzhou, China Guangdong Engineering Technology Research Center of Building Health Monitoring and Safety Early Warning, Guangzhou, China
Wei Chen Guangzhou Guangjian Construction Engineering Testing Center Co., Ltd., Guangzhou, China Guangdong Engineering Technology Research Center of Building Health Monitoring and Safety Early Warning, Guangzhou, China
Dayang Wang & Ye Yang School of Civil Engineering, Guangzhou University, Guangzhou, China
Chuanglian Luo & Rongxin Guo Guangzhou Guangjian Construction Engineering Testing Center Co., Ltd., Guangzhou, China Guangdong Engineering Technology Research Center of Building Health Monitoring and Safety Early Warning, Guangzhou, China
ABSTRACT: In this study, three corner structure optimization schemes were designed for the shear wall of the modular steel-concrete combination, including modified chamfer (MACSW-J), modified chamfer stiffened (MACSW-Z), and modified rounded stiffened (MACSW-Y). The quasi-static tests of the three optimized shear walls and ordinary shear walls (MACSW) were compared and analyzed, including the determination of the damage degree and mechanical performance indicators of the specimens. Based on the collected test results, a numerical model is established. The influence of parameters is studied from the aspects of different length-span ratios, rib thickness ratios and plate thickness ratios. The results show that after the corner structure is optimized, the damage development of the specimen is significantly reduced, and the MACSW-Y is remarkable. Compared with the MACSW specimen, the ultimate bearing capacity of MACSW-Y increased by 34.19%, and the initial stiffness increased by 30.31%. Parameter analysis shows that the optimal values of the length-span ratio, rib thickness ratio and plate thickness ratio are 0.2, 1.8 and 3.0, respectively. The research reveals that the designed corner structure optimization scheme effectively delays the premature severe local damage at the corner of the shear wall, which leads to the loss of its bearing capacity, thereby ensuring the continuation of the overall mechanical properties of the specimen material.
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-3
1 INTRODUCTION In recent years, the sharp rise in energy carbon emissions and labour costs has been the focus and hot issue of global attention. According to the 2019 carbon emission data published in the China Building Energy Consumption Research Report (2021) released by the China Building Energy Efficiency Association, the construction industry accounts for about 40% of the total carbon emissions at the global level and 50.59% at the national level. It is worth noting that carbon emissions during the production of building materials and construction account for 53.74% of the total carbon emissions of the construction industry (China Building Energy Consumption Research Report 2020); in addition, the proportion of construction workers over the age of 40 reaches 50.6%, and young people no longer choose to be construction workers. According to the comprehensive data, the transformation and industrialization upgrading of the construction industry is the trend of the times. However, most of the prefabricated buildings adopt more complex “wet” connections, and most of them are transverse members and secondary structures, so it is difficult to ensure the speed and quality of construction at the same time. The vertical assembly system still has a great development prospect. Prefabricated steel plate composite shear wall is a widely used vertical structural member in prefabricated buildings strongly recommended by the state (Technical Specification for Steel Plate Shear Wall 2016). Many scholars at home and abroad have verified through a large number of studies that the steel plate composite shear wall has the excellent horizontal bearing capacity, ductility and energy dissipation performance, which is suitable for vertical prefabricated building structure systems (Astaneh 2001; Dey 2016; Guo 2011). Steel plate shear walls are prone to out-of-plane buckling deformation. To optimize the mechanical behaviour of the steel plate shear wall, Professor Astaneh proposed adding cast-inplace concrete on both sides of the steel plate to limit the out-of-plane buckling deformation of the steel plate. Compared with the conventional steel plate shear wall, the advantage of the built-in steel plate-encased concrete shear wall is that the lateral stiffness is larger. The disadvantage is that the concrete is easily crushed when the layer displacement angle is large (Zhao 2004, 2006). Guo et al. improved the method by setting a gap between the shear wall and the beam column, which greatly improved the energy dissipation and ductility of the composite shear wall (Guo 2011). Based on the pseudo-static test results of the anti-buckling steel plate composite shear wall, Gao et al. carried out a large number of numerical analyses. The results show that, on the one hand, the precast concrete slab has a considerable restraining effect on the embedded steel plate, and the overall performance of the member is improved. On the other hand, the steel plate embedded in the corner of the member will yield locally in the early stage of loading. The researchers suggest that the local reinforcement of corners should be further studied in the follow-up study of anti-buckling steel plate shear walls (Gao 2011). The tension band formed by the out-of-plane buckling of the steel plate improves the lateral bearing capacity of the structure (Hosseinzadeh 2014), but under the horizontal load, the columns with edge constraints are subjected to a great load. Therefore, to prevent the side column of the anti-buckling steel plate shear wall from yielding prematurely, it is necessary to greatly improve the column stiffness (Jahanpour 2015). Xue et al. explored the performance of anti-buckling steel plate shear walls connected on both sides through continuous tests and numerical analysis. The results show that the length-span ratio, rib thickness ratio and plate thickness ratio affect the seismic performance of buckling steel plate shear walls (Xue 1994). In conclusion, the seismic performance of the prefabricated composite shear wall is superior, but it has the disadvantages of complex working procedures and low assembly efficiency. For example, steel bars and concrete should be tied and poured at the construction site. In addition, due to the characteristic of the aspect ratio, the corner of a shear wall under stress is prone to serious local damage in the early loading process, and the failure form is bending compression failure. This leads to the premature degradation and failure of the overall bearing capacity of the members due to the lack of corner strength, which cannot give full play to the overall material mechanical properties of the members. There is still no suitable solution to this problem in the existing 15
research. Given this, a new type of modular assembled steel-concrete composite shear wall considering corner structure optimization is proposed in this study. To explore the mechanical properties, damage failure mode and failure mechanism of the new modular assembled steel-concrete composite shear wall, three simple and effective optimization schemes are designed and made according to the characteristics that the corner is most prone to bending and compression failure. Four groups of shear wall specimens with a scale ratio of 1:3 were designed and made, and the experimental study on mechanical properties was carried out under low cyclic loading. Then, the numerical model is established by ABAQUS finite element software and compared with the experimental results to verify the accuracy of the numerical model. Finally, the parameter optimization is carried out by setting different length-span ratios, rib thickness ratios and plate thickness ratios. This paper deeply explores the performance improvement effect of the specimen under different optimization schemes to find out the best corner optimization improvement scheme and provide a reference for the engineering design of the same type of prefabricated shear wall. 2 TEST SURVEY 2.1
Specimen design
Four sets of specimens with a scale ratio of 1:3 were made to explore the mechanical properties of the corner-optimized shear wall. The structure of the specimens and the corner structure are shown in Figure 1. The test piece is prefabricated in the processing plant, with steel plates inside and concrete wrapped outside. The size of the test piece is 1300 mm 600 mm 100 mm, the thickness of the built-in steel plate is 3 mm, and the other material models are Q235 steel, C30 concrete, and HPB-300 steel bar. The four groups of modular assembled steel-concrete composite shear wall specimens are common type (MACSW), improved knuckle type (MACSW-J), improved knucklestiffened type (MACSW-Z) and improved rounded-stiffened type (MACSW-Y). Table 1.
Components size information.
Specimen
Optimizing Types
L/mm
Lengthspan ratio a
MACSW MACSW -J MACSW -Z MACSW -Y
Common Improved knuckle Improved knuckle-stiffened Improved roundedstiffened
– 100 100 100
– 1/6 1/6 1/6
Ta/mm
Rib thickness ratio b Tb/mm
Thickness ratio l
– – 6 6
– – 2 2
– 3 3 3
– 9 9 9
Note: The reinforced length of the corner structure is L, the thickness of the stiffener at the corner is Ta, and the thickness of the reinforced steel plate at the corner is Tb.
Figure 1.
The structure of the specimens and the corner structure.
The key parameters of the corner structure of the specimen were designed using dimensionless parameters, and the values of the dimensionless parameters of the four specimens are shown in 16
Table 1. The length-span ratio a was the ratio of the corner reinforcement length to the span of the specimen. The rib thickness ratio b was the ratio of the thickness of the corner stiffener to the thickness of the built-in steel plate. The plate thickness ratio l was the ratio of the thickness of the corner reinforced steel plate to the thickness of the built-in steel plate. The numerical analysis results of (Chen 2018) clarified that the value range of the ratio of length-span ratio, rib thickness ratio and plate thickness ratio of the modular steel-concrete composite shear wall should not be greater than 0.3, 1.4 and 3.0, respectively. Construction processing and other factors, the values of the three parameters are 0.17, 2.0 and 3.0 in turn. Compared with the optimal value in the literature, when the length-span ratio is 0.17, the initial stiffness of the specimen differs by 3.95%, and the energy dissipation differs by 7.8%. When the rib thickness ratio is 2.0, the initial stiffness and energy dissipation of the specimen differs by 8.12% and 5.85%, respectively. The calculation result shows that the difference between the test piece parameter value and the optimal value is small. The maximum is only 8.00%, indicating that the key parameters of the test piece are reasonable to exert the best mechanical properties of the test piece. 2.2
Test loading and measuring point arrangement
Figures 2(a) and (b) show the loading device of the specimen. The MTS-244.41-100 t hydraulic actuator with maximum thrust up to 1500 kN is adopted. One end of the actuator was fixed with the reaction frame, and the other end was connected to the top beam of the specimen through a splint screw. The high-strength bolt connected the wall of the specimen with the bottom beam, and the horizontal movement of the bottom beam was constrained by the reaction beam at both ends, and the vertical motion of the bottom beam was restrained by the compression beam. This study stipulated that the positive direction (+) is pushed, and the negative direction (-) is pulled back. The test is carried out in two forms: force loading and displacement loading, as shown in Figure 2(c). The test specimen bears the vertical load, so considering the bearing capacity and safety of the specimen loading device, the axial compression ratio of the test specimen is set to 0.15, which represents the vertical load of 36t. The initial loading is controlled by the load to check whether there are errors in each connecting device. After the force loading confirms that the device is normal, displacement control loading is carried out. According to JGJ/T101-2015 “Code for Seismic Test of buildings” (Code for Seismic Test of Buildings 2015), the loading ends when the interstory displacement angle is 1 37, or when the horizontal shear force decreases to 85% of the peak shear force. The arrangement scheme of the shear wall steel bar strain and displacement meter is shown in Figure 3. The bending deformation was measured by DW5 and DW6, the shear deformation was measured by DW3 and DW4, the lateral displacement of the shear wall was monitored by d1-d4, and the vertical displacement of the bottom beam was monitored by d5. The monitoring and observation points of the traditional data acquisition system were limited, such as the lack of fine process monitoring, such as full-field deformation, damage cracking, strain development, and so on. Therefore, in this test process, a three-dimensional digital correlation (Digital Image Correlation, DIC) data acquisition system (ARAMISSRX three-dimensional camera system) is used to monitor the strain and crack development of the specimen in real-time during the whole loading process, as shown in Figure 2(b). Figure 4 shows the calibration and adaptation process of the DIC system before testing.
Figure 2.
The structure of the specimens and the corner structure.
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Figure 3. Arrangement scheme of strain gauge and displacement meter.
Figure 4.
Test of DIC.
3 EXPERIMENT RESULTS AND ANALYSES 3.1
Specimen failure characteristics and crack development
The failure characteristics of specimen MACSW are shown in Figures 5(a) and 7. Cyclic loading starts when the displacement angle is + 1/400, and two oblique cracks with a width of about 0.08 mm gradually appear at the lower right of the concrete slab. When the displacement angle is + 1/100 rad, there is a huge crack sound of concrete for the first time, which results in a significant convex phenomenon caused by further extrusion, accompanied by horizontal cracks of long 290 mm and wide 0.3 mm at the bottom of the concrete slab. When the displacement angle is 1/50 rad, after the first forward loading, the bottom crack runs through the wall, and the width expands to 4 mm. When the displacement angle is + 1/37 rad, after starting the forward loading, the oblique cracks interweave with each other, the upper right side of the concrete slab appears the drum bending phenomenon, and the lower right corner is seriously squeezed off, which determines that the shear wall is damaged and stops loading. The failure characteristics of specimen MACSW-J are shown in Figure 5(b) and Figure 7. When the displacement angle was +1/100 rad, there were many oblique cracks in the concrete wall, but no horizontal cracks. When the displacement angle was +1/50 rad, when the first positive load was applied, the concrete made a loud tearing sound, and the cracks at the bottom of the wall spread along the horizontal direction. The crack length was about 290 mm, and the width was about 3 mm. During the third forward loading, the buckling deformation of the outer steel plate is up to 5 mm. The displacement angle is + 1/37 rad. After the first forward loading, the tearing damage of concrete and the buckling deformation of the steel plate is more serious. The shedding displacement of the concrete surface reaches 15 mm, the buckling deformation of the outer steel plate on the right reaches 15 mm, and the specimen is seriously damaged.
Figure 5.
Failure mode diagram of the specimen under different loading displacement angles.
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Figure 5.
(Continued)
Figure 6.
Main strain nephogram of test piece from the perspective of DIC.
Figure 7.
Development and width of hand-painted cracks in concrete.
The failure characteristics of specimen MACSW-Z are shown in Figure 5(c) and Figure 7. Displacement angle +1/100 rad, after the first cyclic loading, the concrete at the corner of the wall is torn along the direction of the stiffener, and the tear width is 0.75 mm. Displacement angle +1/50 rad, after the first positive loading, there is a huge tearing sound in the concrete, and the cracks at the bottom spread along the horizontal direction. The crack length is about 235 mm, the width is about 2 mm, and the concrete tearing width along the direction of the stiffener reaches 4 mm. The displacement angle is +1/37rad. After the first forward loading, the outsourcing steel plate gradually buckled and deformed, and the concrete surface shed outward and displaced up to 23 mm. The buckling deformation of the right outsourcing steel plate reached 8 mm, and the specimen was destroyed. The failure characteristics of specimen MACSW-Y are shown in Figure 5(d) and Figure 7. Displacement angle +1/100 rad, after the first cyclic loading, the corner concrete is torn along the direction of the stiffener, and the tear width is 0.40 mm. Displacement angle +1/50rad, after the first positive loading, there was a huge tearing sound in the concrete, and the concrete cracks at 19
the bottom developed along the horizontal direction. The crack length was about 225 mm, and the width was about 0.70 mm. The concrete tearing width along the direction of the stiffener reaches 3.50 mm, and the concrete on the right-side wall is not peeled off due to extrusion. The displacement angle is +1/37 rad. After the first forward loading, the concrete surface falls off, and the displacement reaches 8 mm. The damage of the specimen is relatively light, so continue to load. After the first negative loading, buckling deformation of 2 mm gradually appeared on the left outer cladding steel plate. After the second forward loading, the front of the specimen was severely damaged, and the right outer cladding steel plate began to buckle 5 mm. Figure 5 shows that the four specimens begin to load to the ultimate loading state, and the high-strength bolts connecting the wall and the bottom beam are intact, indicating that this connection is safe and reliable. The bolt installation and disassembly process are simple and reflect the connection performance of the components in the real structure. Figure 6–Figure 8 show that cracks gradually occur in MACSW, MACSW-J, MACSW-Z and MACSW-Y4 specimens at the displacement angle 1/400rad, and the maximum cracks are 0.08 mm, 0.08 mm, 0.05 mm and 0.05 mm, respectively. When the displacement angle is loaded to + 1/50 rad, the maximum cracks of the specimen are 4 mm, 3 mm, 2 mm and 1.5 mm. Compared with the specimen MACSW, the crack width of the improved corner structure specimen is reduced by 62.5%. DIC measurements showed microcracks that were difficult to detect with the naked eye. Especially when the displacement angle is very small, DIC monitors the crack development of each loading process, which is helpful in exploring the dynamic development process of specimen damage. The results show that the improved corner specimen significantly reduces the damage development of the specimen, and the MACSW-Y of the specimen is the best.
Figure 8.
3.2
Maximum crack width of concrete.
Specimen failure characteristics and crack development
Figure 8 shows the measured horizontal load-load displacement curve and cumulative energy dissipation curve of each specimen. The results show that the hysteretic curves of the four shear wall specimens are full, indicating that the energy dissipation capacity of this kind of shear wall is excellent, and the concrete slabs, embedded steel plates, connectors and studs work smoothly. The hysteresis loop of specimen MACSW is small, and the pinch occurs in the later stage of loading. The hysteresis loop of the other three groups of corner optimization specimens did not produce a pinch phenomenon. At the initial stage of loading, the bearing capacity of the four specimens increases linearly with the increase of the displacement angle. With the aggravation of the damage of members, the bearing capacity increases slowly. The curve decreases after the specimen reaches the maximum bearing capacity. Table 2 shows that in the loading stage of R = + 1/37rad, the mechanical properties of MACSW-J, MACSW-Z, and MACSW-Y are 7.22%, 13.10% and 34.19% higher than that of 20
Table 2.
Displacement and bearing capacity of the test piece at the representative stage. Yield stage
Peak stage
Test piece
Displacement of yield/mm
Load of yield/ kN
Displacement of yield/mm
Load of yield /kN
R = + 1/50 rad Load/kN
R = + 1/37 rad Load /kN m
MACSW MACSW-J MACSW-Z MACSW-Y
16.20 16.42(1.40%) 16.80(3.69%) 16.84(3.97%)
276.84 304.51(9.99%) 306.55(10.73%) 310.42(12.13%)
28.55 28.58(0.10%) 28.65(0.36%) 38.85(36.09%)
328.85 359.61(9.35%) 370.98(12.81%) 382.87(16.43%)
328.85 359.61(9.35%) 370.98(12.81%) 375.67(14.24%)
285.32 305.93(7.22%) 322.71(13.10%) 382.87(34.19%)
1.76 1.74 1.70 2.28
Note: The values in brackets are the growth rates of the three improved specimens relative to the MACSW specimens; Displacement ductility coefficient = ultimate displacement/yield displacement.
Figure 9.
Hysteresis curve and cumulative energy consumption curve of the test piece.
MACSW, respectively. The corner structure is optimized to delay the yield displacement and peak displacement of the specimen to a certain extent, and the bearing capacity is improved, among which the effect of MACSW-Y is the best. In Figure 8(e), the cumulative energy consumption of the specimen is positively correlated with the loading displacement angle. The change of cumulative energy consumption in the early stage of loading is similar, with the maximum value of the MACSW-Y specimen and the smallest value of the MACSW specimen in the later stage of loading. In addition, the ductility of MACSW, MACSW-J and MACSW-Z are similar, and the displacement ductility coefficients (m) are 1.76,1.74,1.70, respectively. However, the displacement ductility coefficient of specimen MACSW-Y is 2.28, which increases significantly. 3.3
Stiffness degradation
According to the Code for Seismic Test methods of buildings, the stiffness variation characteristics of specimens are expressed by Secant stiffness (Zhou 2014). As shown in Figure 10, among the four specimens, the initial test stiffness of the MACSW specimen is the smallest, indicating that the optimization scheme of the corner structure is beneficial to improve the initial stiffness of the shear wall. And among the three optimization schemes, the corner treatment method of stiffening fillet reinforcement is the most beneficial to improve the initial stiffness of steel-concrete composite shear wall specimens. With the increase of the displacement angle, the Secant stiffness of the specimen MACSW-Y is the highest, and the stiffness 21
degradation is the lowest, which means that the mechanical stability of the shear wall with stiffened fillet reinforcement is better. This corner structure optimization scheme further avoids the premature degradation and failure caused by the serious local damage at the corner of the shear wall. It makes full use of the mechanical properties of the specimen materials.
Figure 10.
Secant stiffness and stiffness degradation ratio.
4 NUMERICAL MODEL ESTABLISHMENT AND VERIFICATION The finite element model of the specimen is established by ABAQUS software. The stress triaxiality steel damage constitutive model is selected for the steel constitutive model (Zhou 2014). The constitutive model of steel concrete adopts the concrete damage plastic model, which comes with ABAQUS software. The three-dimensional solid element (C3D8R) of the eight-node reduction integral format is selected for each component in the finite element model. The material type is the same as the test. The MACSW finite element model of the specimen and the MACSW-Y meshing of the specimen are shown in Figure 11. Combined with the test device, the bottom beam of the finite element model is consolidated, and the top beam constrains the out-of-plane translational degree of freedom (U3) and the rotational degree of freedom of the loading direction (UR1 and UR2). Penalty friction contact is adopted in the tangential direction of the boundary between the embedded steel plate and the outer concrete plate, “hard contact” is used in the normal direction, and “Tie” constraints are used between the stud and the embedded steel plate, and between the built-in steel plate and the surrounding steel connectors. The first buckling mode is used to simulate the initial test defects of the steel plate. In this study, the defect amplitude is 1/1000 of the long edge of the embedded steel plate.
Figure 11.
Finite element model.
Figure 12.
Comparison of buckling of MACSW-Y.
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Figure 13. Table 3.
Comparative analysis of hysteresis curve and skeleton curve. Comparison of ultimate bearing capacity. Result of test/kN
Result of numerical/kN
Error (%)
Test piece
Push
Pull
Push
Pull
Push
Pull
MACSW MACSW-J MACSW-Z MACSW-Y
328.9 359.6 371.0 375.7
338.0 380.0 369.7 378.8
334.2 360.4 368.3 375.3
339.3 372.8 379.4 387.2
1.61 0.22 0.72 0.10
0.37 1.91 2.63 2.22
The hysteretic curves of MACSW and MACSW-J and the skeleton curves of MACSW-Z and MACSW-Y are compared in Figure 13. The data in Table 3 show the ultimate bearing capacity of four specimens tested and numerically simulated. Combined with Figure 13 and Table 3, the numerical simulation results of the four specimens are consistent with the test results, with a maximum error of 1.61% for forward loading and 2.63% for negative loading. The hysteretic curve of the numerical analysis is fuller, mainly because the gap between the steel connectors in the test reduces the lateral stiffness to a certain extent. The numerical analysis assumes that the material properties and boundary contact in the model are ideal, which is different from the experiment, so the result of the numerical analysis is slightly larger than that of the experiment. From the overall point of view in Figure 12, the accuracy of the numerical model meets the engineering requirements and accurately reflects the real mechanical behaviour of the specimen.
5 PARAMETER INFLUENCE ANALYSIS 5.1
Numerical model establishment and verification
The analysis of the test results shows that the mechanical properties of the specimen MACSW-Y are the best. Therefore, the influence of parameters on the stiffening fillet improved module assembled steel-concrete composite shear wall (MACSW-Y) is analyzed. 23
To explore the effects of different length-span ratios a, rib thickness ratio b and plate thickness ratio l on the energy dissipation capacity and bearing capacity of MACSW-Y, and the optimal range of values. The calculation parameters of the model are shown in Table 4, and the influence analysis of 3 key parameters is carried out under 18 working conditions. Table 4.
Parameters of model calculation.
Case
The length-span ratio a
The rib thickness ratio b
The plate thickness ratio l
14 511 1518
0.1, 0.2, 0.3, 0.4 0.2 0.2
2 1.0, 1.2, 1.4, 1.6, 1.8, 2.0, 2.2 1.8
3 3 1, 1.5, 2, 2.5, 3, 3.5, 4
5.2
Analysis of calculation results
5.2.1 Influence of the length-to-span ratio. Comprehensive Figures 14–17 and Table 5 show that the maximum areas of concrete cracks, steel plate stress and damage are concentrated in the corner of the shear wall. With the increase of the length-span ratio of the shear wall, the initial stiffness, ductility, ultimate bearing capacity and energy dissipation capacity increase at first and then decrease. For example, the corresponding energy dissipation capacity of the four operating conditions is 59.98 kN m, 72.44 kN m, 64.41 kN m, and 59.16 kN m. The damage factor quantifies the damage development of concrete. In Figure 15, the compression damage factor of concrete in the yield stage is 0.950, 0.913, 0.922 and 0.927, respectively, and the damage factor decreases at first and then increases with the change in length-span ratio. With the increase of the length-span ratio, the stress of the steel plate decreases gradually, and so does the damage factor in the ultimate bearing capacity stage of the embedded steel plate. Considering the actual performance-to-price ratio of the corner structure, it is suggested that the best lengthspan ratio of MACSW-Y is 0.2 (condition 2).
Figure 14.
Curve of load-displacement.
Figure 16.
Stress nephogram of steel plate/kN.
Figure 15.
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Dissipation of energy.
Figure 17. Table 5.
Analysis of concrete damage. Performance index of shear wall.
Initial stiffness Case /(kN/mm)
Yield point
Ultimate bearing capacity
Load/ Displacement kN angle dy/%
Load/ Displacement kN angle dm/%
Ductility coefficient m
Steel plate damage factor at limit stage
1 2 3 4
325.57 347.43 344.26 343.58
387.85 403.57 393.36 392.16
1.63 2.24 1.98 1.98
0.48 0.45 0.44 0.41
32.67 33.12 33.05 30.79
1.07 1.21 1.13 1.13
1.75 2.24 1.98 1.98
Note: The initial stiffness is determined by the elastic stage of the skeleton curve, and the yield point load is obtained by interpolation.
5.2.2 Influence of rib thickness ratio. Table 6 shows that with the increase in rib thickness ratio, the restraint effect of stiffening ribs on the corner reinforced steel plate is getting stronger and stronger, and the overall performance of the member is getting better and better. However, with the continuous increase of rib thickness ratio, the yield-bearing capacity, initial stiffness and ultimate bearing capacity of members are affected by other structures, and the increasing effect of stiffeners on the mechanical properties of specimens becomes smaller. In Figure 18, the damage factors of the seven working conditions of the embedded steel plate in the ultimate bearing capacity stage are 0.50, 0.49, 0.47, 0.46, 0.46, 0.51 and 0.57, respectively, which means that the damage factor of the embedded steel plate decreases gradually with the increase of rib thickness ratio. However, when the rib thickness ratio continues to increase to 2.0 or 2.2, the damage factor increases in the reverse direction. Considering the economic factors, it is suggested that the ratio of rib thickness should be 1.8 (working condition 9).
Figure 18.
Nephogram of steel plate damage at ultimate bearing stage /kN.
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Table 6.
Performance index of shear wall. Yield point
Ultimate bearing capacity
Initial stiffness Case /(kN/mm)
Load/ kN
Displacement angle dy/%
Load/ kN
Displacement angle dm/%
Ductility coefficient m
Steel plate damage factor at limit stage
5 6 7 8 9 10 11
332.07 338.83 341.74 343.29 345.10 345.43 346.12
1.21 1.23 1.23 1.23 1.24 1.24 1.26
396.72 405.04 409.38 410.58 411.11 411.21 411.42
1.96 1.97 1.99 1.99 1.99 1.98 1.99
1.61 1.61 1.62 1.62 1.61 1.60 1.60
0.50 0.49 0.47 0.46 0.46 0.51 0.57
32.93 33.10 33.12 33.19 33.32 33.34 33.36
5.2.3 Influence of thickness ratio. In Figure 19 and Table 7, with the increase in plate thickness ratio, the initial stiffness, yield load and ultimate bearing capacity increase gradually, but the increase slows down. With the increase of plate thickness ratio in Figure 20, the energy consumption of the specimens loaded to 1/50 rad under seven working conditions from 12 to 18 is 67.88 kN m, 71.15 kN m, 72.24 kN m, 73.36 kN m, 73.49 kN m, 73.54 kN m and 73.88 kN m, respectively. In the ultimate bearing capacity stage, the damage factors of embedded steel plates under seven working conditions are 0.62, 0.60, 0.59, 0.57, 0.56, 0.55, 0.53 and 0.57, respectively. It is revealed that the damage factor of the embedded steel plate decreases gradually with the increase of the thickness ratio. When the thickness ratio is more than 3.5, the damage factor increases gradually and the trend changes. In Figure 21, The damage factor decreases with the increase in plate thickness ratio. According to the comprehensive analysis results, the recommended thickness ratio is 3. Table 7.
Performance index of shear wall. Yield point
Ultimate bearing capacity
Initial stiffness Case /(kN/mm)
Load/ kN
Displacement angle dy/%
Load/ kN
Displacement angle dm/%
Ductility coefficient m
Steel plate damage factor at limit stage
12 13 14 15 16 17 18
345.76 351.22 350.85 354.95 361.21 365.23 365.58
1.24 1.26 1.29 1.35 1.42 1.48 1.49
395.99 400.80 405.35 409.44 409.65 410.66 410.90
1.97 1.98 1.98 1.99 1.96 1.99 1.99
1.58 1.57 1.53 1.47 1.38 1.34 1.34
0.62 0.60 0.59 0.57 0.56 0.55 0.53
33.90 34.19 34.69 35.12 33.36 34.48 34.88
Figure 19.
Curve of load-displacement.
Figure 20.
26
Dissipation of energy.
Figure 21.
Analysis of concrete damage.
6 CONCLUSIONS In this paper, three corner structure optimization schemes for the new modular assembled steel-concrete composite shear wall are designed, and experimental research and parameter influence analysis is carried out. The conclusions are as follows: l
l
l
l
The damage of the specimen MACSW is the earliest and the most serious during the test loading. The improved corner specimen reduces the damage development of the specimen, among which the optimization scheme of MACSW-Y is the best, followed by MACSW-Z and MACSW-J. The DIC equipment equipped in the test is helpful in measuring the micro-cracks, which are difficult to find and avoids the disadvantages of traditional manual observation of concrete damage. In particular, when the in-situ displacement angle is very small, the crack development of each loading process is monitored, and the damage development process of the specimen is accurately reflected. The hysteretic energy consumption performance of three kinds of corner optimized shear walls is full, while the common shear walls appear to pinch phenomenon in the later stage of loading. Compared with specimen MACSW, the ultimate bearing capacity of specimens MACSW-J, MACSW-Z, and MACSW-Y increased by 7.22%, 13.1% and 16.43%, and the initial stiffness increased by 5.47%, 18.14% and 30.31%. The results show that the corner structure optimization scheme avoids serious local damage at the corner of the shear wall prematurely and then ensures that the mechanical properties of the whole material are fully utilized. The hysteretic curve and skeleton curve of the numerical model are consistent with the experimental results, and the maximum error of ultimate bearing capacity is only 2.63%. The results of parameter influence analysis show that the initial stiffness, yield load and ultimate load of energy dissipation and damping ratio increase at first and then decrease with the increase of length-span ratio. With the increase of rib thickness ratio and plate thickness ratio, the initial stiffness, yield load and ultimate load of energy dissipation and damping ratio increase at first, but the increase decreases gradually. Considering the optimization of the corner structure, the recommended values of the length-span ratio, ribthickness ratio and plate-thickness ratio of the modular assembled steel-concrete composite shear wall are 0.2, 1.8 and 3.0.
ACKNOWLEDGEMENTS The research grants that funded this work presented herein were No. 51778162 and No. 51878191 by the National Natural Science Foundation of China and No. 2022-K19574922 by the Science Technology Innovation Program of the Department of Housing and 27
Urban-Rural Development of Guangdong Province, and No. GJGJBG [2022]056 by the Science and technology projects of Guangzhou Guangjian Construction Engineering Testing Center Co., Ltd. of China.
REFERENCES Astaneh-Asl A, Ph. D. Seismic Behavior and Design of Steel Shear Walls[C]. SEOANC Seminar, 2001. China Building Energy Consumption Research Report 2020 [J] Building Energy Efficiency (Chinese and English), 2021, 49 (02): 1–6. Chen H S Improved Analysis of the Aseismic Structure of New Modular Fabricated Steel-concrete Composite Shear Wall [D] Guangzhou: Guangzhou University, 2018. Code for Seismic Test of Buildings: JGJ/T 101-2015 [S] Beijing: China Construction Industry Press, 2015. Dey S and Bhowmick A K. Seismic Performance of Composite Plate Shear Walls[J]. Structures, 2016:59–72. Guo L H, Li R, Rong Q and Zhang S. Cyclic Behaviour of SPSW and CSPSW in the Composite Frame[J]. Thin-Walled Structures, 2011, 51. Guo Y L, Zhou M and Dong Q L. Experimental Study on Three Types of Steel Plate Shear Wall Structures [J] Journal of Building Structures, 2011, 32 (1): 17–29. Gao H, Sun F F and Li G Q. Finite Element Analysis of Composite Steel Plate Shear Wall Connected on Both Sides [J] Building Structure, 2011, 41 (S1): 1112–1114. Hosseinzadeh S A A, Tehranizadeh M. The Wall–frame Interaction Effect in Steel Plate Shear Wall Systems [J]. Journal of Constructional Steel Research, 2014, 98(9):88–99. Jahanpour A, Moharrami H. Evaluation of the Behaviour of the Secondary Columns in Semi-supported Steel Shear Walls[J]. Thin-Walled Structures, 2015, 93:94–101. Technical Specification for Steel Plate Shear Wall: JGJ/T 380-2015 [M] China Construction Industry Press, 2016. Xue M and Lu L W. Interaction of Infilled Steel Shear Wall Panels With Surrounding Frame Members[J]. The Voice of Australian Steel, 1994: 339–354. Zhao Q H, Astaneh-Asl A. Cyclic Behaviour of Traditional and Innovative Composite Shear Wall[J]. Journal of Structural Engineering (ASCE), 2004: 271–283. Zhao Q H, Astaneh-Asl A. Seismic Behaviour of Steel and Compose Shear Wall Systems and Application of Smart Structures Technology[C]. Proceeding of US-Korea Workshop on Smart Structures Technology for Steel Structures, Korea, 2006. Zhou T H, Li W C, Guan Y and Bai L. Damage Analysis of Steel frame Under Cyclic Loading Based on Stress Triaxiality [J] Engineering Mechanics, 2014, 31 (07): 146–155.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Engineering application of the tunnel disease rapid detection system Chao Wang* Liaoning Transportation Planning and Design Institute Co., Ltd, Liaoning Datong Highway Engineering Co., Ltd., Shenyang
ABSTRACT: Highway tunnels will suffer from diseases in operation. The traditional detection methods mainly focus on manual detection, which is inefficient and subjective. In this paper, the tunnel disease fast detection system is applied. First, the function of the tunnel detection vehicle is tested and applied to the actual detection project. The detection results are compared with those of traditional methods. The results show that the tunnel disease rapid detection system has the advantages of high automation, high detection speed and accuracy, and objective and true results. It provides a reliable scheme for the scientific and rapid detection of tunnel diseases.
1 INTRODUCTION At present, the number of tunnels is increasing continuously. In the process of tunnel operation, the lining structure cannot avoid the existence of cracks, deformation, breakage, falling block, and leakage phenomenon, which may seriously threaten the safety, stability and durability of the lining structure. So, it is often necessary to detect and maintain the tunnel lining structure (Hong 2020; Yao 2014). The traditional detection methods mainly rely on manual detection, which is heavy in workload, time, and labour cost. At the same time, there may be misjudgment, or lead to the disease not being discovered in time, causing tunnel safety accidents. Therefore, a rapid and accurate intelligent identification tunnel disease detection system is of great significance for tunnel disease detection (Yang 2018). Nishikawa et al. proposed an automatic image processing method based on genetic programming, which is mainly used to detect cracks in concrete structure surface images (Nishikawa 2012). Tsai et al. proposed a semi-automatic crack detection algorithm based on the minimum path (Tsai 2013). Huang et al. extracted features using a full convolution network (FCN). A new image recognition algorithm for semantic segmentation of cracks and leakage defects in metro shield tunnels is proposed (Hang 2017). Xue et al. established a tunnel lining feature image classification model based on the GoogLeNet network structure of the improved Inception module (Xue 2018). Byunghyun et al. used the Mask R-CNN algorithm to detect cracks. Fractures are extracted and then quantified using morphological operations (Byunghyun 2019). Ren et al. proposed an improved neural network for deep full coiler according to the crack characteristics (Ren 2020). In this paper, the tunnel disease rapid detection system is applied. Firstly, the function of the tunnel detection vehicle is tested and applied in the actual detection project. The detection results are compared with the traditional detection methods. The results can provide scientific references for the rapid detection of tunnel disease. *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-4
29
2 TUNNEL DETECTION VEHICLE FUNCTION TEST Mathematically, the definition of image matching is: if an image is represented by a twodimensional sequence, the gray value at the point in one image is I1 ðx; yÞ, and the other is I2 ðx; yÞ, then the matching relationship between the image I1 and I2 is represented by the Formula (1.1): I2 ðx; yÞ ¼ gðI1 ðf ðx; yÞÞÞ
(1)
where I1 is the reference image, and I2 is the image to be matched; f is a 2D geometric transformation function; g represents a 1D grayscale transformation function. It can be seen from Formula (1.1) that image matching specifically includes two meanings. The solution of the function reflects the matching in geometric space; The solution of the function reflects the matching of gray levels between corresponding pixels. Therefore, under the premise of considering the distortion, through the matching between images, we should get the best spatial transformation relationship and gray transformation relationship between images so that the image features can get the most accurate matching. In general, the gray transformation relationship is unnecessary, so the key is finding the spatial geometric transformation relationship. Ignore the gray transformation relationship, and the previous relationship can be simplified as Formula (1.2): I2 ðx; yÞ ¼ I1 ðf ðx; yÞÞ
(2)
In Formula (1.2), if the two-dimensional function can be expressed by the product of two one-dimensional functions, that is: f ðx; yÞ ¼ f1 ðxÞ f2 ðyÞ
(3)
In the process of matching, the storage capacity will be significantly reduced, and the computational efficiency will be improved. The function test of tunnel detection vehicles mainly focuses on the application of tunnel comprehensive detection system in the actual highway tunnel detection. Due to certain differences in specific application scenarios in the actual application of tunnel detection vehicles, the results collected by the system will also be very different. The key contents to be tested include the image acquisition effect of different tunnel inner surfaces, the influence of different tunnel sizes on the image acquisition effect, and the ability of the system to continuously collect image data under the condition of a long-distance tunnel. According to the actual situation of the tunnel in Liaoning Province, the Dayu tunnel (up) in the Benxi City of Liaoning Province is selected as the test environment of the tunnel detection vehicle. The tunnel type is highway long tunnel, the route number is G1113, the centre pile number is K154 + 791, and the inner surface diameter is 11 m.
Figure 1.
The basic situation of the Dayu Tunnel.
30
During the detection process, the tunnel detection vehicle runs at a speed of 65km/h in the single lane of the tunnel while carrying out the detection work. When the detection vehicle runs in the right lane, the detection device collects the right half of the tunnel image information. When the detection vehicle runs in the left lane, the detection device collects the tunnel lining image information. Finally, the graph acquisition results of 6 line-scanning system components are shown in Figure 2.
Figure 2.
Test results of Dayu Tunnel.
This tunnel detection test is mainly to observe the detection effect of the tunnel detection vehicle on the cement surface in the actual klyrev detection. From the image acquisition results, the reflection effect of the cement surface is good, and the pictures taken are generally clear. The crack and other disease information can be clearly seen. In addition, from the shooting results of the six cameras, the image information from the vault to the side wall is clear. Therefore, it is no problem for tunnel detection vehicles to carry out detection under such tunnel conditions.
3 ENGINEERING APPLICATION OF TUNNEL DISEASE RAPID DETECTION SYSTEM The Jinshan Tunnel (ascending) was selected as the background, and the results of the tunnel disease rapid detection system were compared with those of traditional methods. The data was obtained by the traditional method of manual and device detection mode. 3.1
Comparison of lining diseases
The Jinshan Tunnel (ascending) was regularly inspected in 2018 and 2019, followed by the traditional method in 2018 and the tunnel disease rapid detection system in 2019. Compared 31
with the results of regular testing in 2018, 3 new lining diseases were found in Jinshan Tunnel (ascending) in 2019, which were as follows: (1) Newly added lining disease: there is a circumferential crack in the left arched waist of No. 007 board, 2.02 m away from the end of the board, with a length of 1.46 m and a width of 0.20 mm.
Figure 3.
New lining disease 1.
(2) Newly added lining disease: in the right arched waist of No. 007 plate, 1.13 m away from the end of the plate, 5.23 m away from the centre line of the vault, longitudinal cracks, 1.89 m in length and 0.3 mm in width.
Figure 4.
New lining disease 2.
(3) Newly added lining disease: in the vault position of No. 062 plate, 4.56 m away from the end of the plate, 1.18 m away from the centre line of the vault, longitudinal crack, length of 2.5 m, width of 0.4 mm.
Figure 5.
New lining disease 3.
As shown in Figures 3–5, the length and width information of tunnel lining diseases can be directly seen, and the tunnel disease information detected by the tunnel disease rapid detection system can be relied on. On-site manual inspection of disease information was carried out. It can be seen from the previous detection experience that the newly discovered diseases of tunnel lining in 2019 are likely to be non-new diseases, but missing and undiscovered diseases in previous years. 32
3.2
Comparison of interior decoration diseases
Compared with the results of regular testing in 2018, there were nine newly added interior decoration diseases, part of which were as follows: (1) Internal decoration disease: Inboard No. 089, it is 3.08 m away from the end of the board and 4.87 m away from the centre line of the vault, and the fire protection coating on the left arched waist fell off, with an area of 1.35 1.10 m2 (repaired in 2019).
Figure 6.
Interior decoration disease 1.
(2) Internal decoration disease: Inboard No. 090, it is 6.19 m away from the centre line of the vault, and the fireproof coating on the left arched waist fell off, covering an area of 1.10 1.18 m2 (repaired in 2019).
Figure 7.
Inner decoration disease 2.
(3) Interior decoration disease: In No. 129 plate, it is 1.64 m away from the end of the plate and 4.03 m away from the centre line of the vault, and the left arch waist fireproof coating fell off, with an area of 1.11 1.41 m2 (repaired in 2019).
Figure 8.
Interior decoration disease 3.
As shown in Figures 6–8, the area information and shape of the disease installed in the tunnel can be directly seen. The tunnel disease information detected by the tunnel disease rapid detection system is reliable. On-site manual inspection of disease information was 33
carried out. It can be seen from the previous detection experience that the newly discovered diseases in tunnel decoration in 2019 are likely to be non-new diseases, but missing and undiscovered diseases in previous years. The tunnel disease rapid detection system can collect the tunnel disease and profile information continuously, dynamically and completely, which can be used for regular tunnel inspection. In the background of automatic data analysis, statistical tunnel lining cracks, spalling, water leakage, exposed tendons, and other apparent diseases of geometric parameters and conditions. The following table shows the technical indexes of the tunnel disease rapid detection system. Table 1.
Technical indicators of tunnel disease rapid detection system.
Serial number
Project
Index
1
Detection index
2 3 4 5 6 7 8
Camera resolution Detection accuracy (tunnel lining) Detection speed Contour scanning accuracy Positioning accuracy Detection accuracy Inspection operator
Cracks, water seepage and other apparent diseases 4096*1 0.2 mm 080 km/h 6 mm 90% 23 people
If only the traditional manual method is used to detect the disease of a highway tunnel, the tunnel needs to be closed, and it takes six people three days to complete the 1 km tunnel [9-10]. Due to the great influence of human factors, there are problems such as low efficiency, poor accuracy, and high cost of lane sealing. The tunnel comprehensive detection system requires no road closure, only one driver and one operator, and the speed can reach 60–80 km/h. Moreover, the internal processing efficiency is high. Each person can independently process 24 km of tunnel data every day and issue a complete detection report. The following table is a comparison table of tunnel diseases detected by traditional manual detection and tunnel comprehensive detection systems. Table 2.
Comparison between traditional manual detection and tunnel disease rapid detection system.
Project
Traditional manual detection
Detection efficiency Field inspection Cost analysis
56 m/ person/day 860 yuan/m Rearrange the site pictures, manually count the location of the disease, and calculate the length or area of the disease
Industry data
Large human influence, subjective factors, high requirements for testing personnel technical experience It is difficult to ensure the safety of testing personnel
Treatment efficiency
Tunnel disease rapid detection system 60–80 km/h Fuel costs, high-speed costs The system can automatically read the length and location of the disease, and each person can process about 1km of tunnel data by himself every day Objective and true results
High security and easy operation
Field collection efficiency: Taking the 2 km tunnel as an example, the detection time of a single tunnel by the traditional method is 4–16 hours. However, it can be shortened to 10 minutes by using the tunnel disease rapid detection system. Internal data processing 34
efficiency: it takes 48 hours for traditional manual detection to sort out data and make reports, while it takes about 24 hours for the tunnel disease rapid detection system to check data, which is 5080% higher than the traditional method.
4 CONCLUSIONS This paper proposes a rapid detection system for tunnel diseases. The optimized image processing method is adopted in the system, which can detect the apparent diseases of tunnel lining efficiently and accurately and make disease statistics through supporting software. Tunnel integrated system has many advantages, such as high automation, fast detection speed, high precision, objective and true results and full coverage of the information. Through practical engineering application, it is found that the tunnel disease rapid detection system can greatly reduce the labour intensity of the inspectors and improve the efficiency of tunnel detection.
REFERENCES Byunghyun K, Soojin C. Image-based Concrete Crack Assessment Using the Mask and Region-based Convolutional Neural Network[J]. Structural Control and Health Monitoring, 2019, 26(8): e2381.1– e2381.15. Hang H, Sun Y, Xue Y, et al. Inspection Equipment Study for Subway Tunnel Defects by Grey-scale Image Processing[J]. Advanced engineering informatics, 2017, 32(Apr.): 188–201. Hong K, Feng H. Development Trends and Views of Highway Tunnels in China Over the Past Decade[J]. China Journal Highway Transportation, 2020, 33(12): 62–76. Nishikawa T, Yoshida J, Sugiyama T, et al. Concrete Crack Detection by Multiple Sequential Image Filtering [J]. Computer-Aided Civil and Infrastructure Engineering, 2012, 27(1): 29–47. Ren Y, Huang J, Hong Z, et al. Image-based Concrete Crack Detection in Tunnels Using Deep Fully Convolutional Networks[J]. Construction and Building Materials, 2020, 234: 117367. Tsai Y J, Kaul V, Yezzi A. Automating the Crack Map Detection Process for Machine Operated Crack Sealer [J]. Automation in Construction, 2013, 31(May.): 10–18. Xue Y, Li Y. A Fast Detection Method Via Region-based Fully Convolutional Neural Networks for Shield Tunnel Lining Defects[J]. Computer-Aided Civil and Infrastructure Engineering, 2018, 33(8):638–654. Yao Y, Tung S T E, Glisic B. Crack Detection and Characterization Techniques-An Overview[J]. Structural Control & Health Monitoring, 2014, 21(12): 1387–1413. Yang J, Liu X, Liu X, et al. Review of Rapid Test Vehicles for Highway Tunnel Structure[J]. Journal of East China Jiaotong University, 2018, 35(4): 30–38.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Deep learning-based recognition method of ground penetrating radar images for cracks inside pavement structures Qian Liu* The College of Electrical Engineering, Zhejiang University of Water Resources and Electric Power, Hangzhou, China
ABSTRACT: To detect and identify internal cracks in asphalt pavement intelligently and precisely, MALA ground-penetrating radar (GPR) and image processing steps like the Fourier transform are used. Images are collected with automatic data adjustment and high-pass filtering to eliminate interference and unnecessary noise. As a result, a total of 2346 GPR detection images were intercepted and labeled. Then, GPR crack images were located and recognized using you only look once (YOLO)v5 multi-scale feature fusion detection model. Experimental results show that the YOLOv5 model has an 85.2% detection accuracy and 85.1% F1 score, and the F1 score and mAP were improved by 4.3% and 4.4% after the improvement. The proposed model could meet the maintenance requirements of asphalt pavement.
1 INTRODUCTION To improve the performance of a road structure, it is necessary to evaluate the surface and interior distresses (Liu et al. 2023; Liu et al. 2023; Liu et al. 2023), among which the most common and difficult internal diseases to detect are cracks (Shangguan et al. 2014; Chen et al. 2022; Liu et al. 2022). In the past, distress detection is mainly based on manual methods (Liu 2019; Liu et al. 2019; Liu et al. 2021), which has such shortcomings as insufficient accuracy (Liu, Gu et al. 2022), low efficiency and high cost (Liu et al. 2021). Recently, GPR is gradually used to detect road structures because of its non-destructive (Long et al. 2023), anti-interference, fast continuous, and safe detection features. A large number of images generated by GPR rapid detection lack timeliness and effectiveness (Liu et al. 2022). So, the above shortcomings seriously restrict the popularization and application of this technology (Mardeni et al. 2010). First of all, because of data acquisition speed, noise, and other reasons (Liu et al. 2022), the error in GPR detection is caused (Leng & Al-Qadi 2014; Liu et al. 2021). To cope with this problem, it is necessary to study the pre-processing and feature analysis of the detection data (Shangguan et al. 2014). However, the artificial image processing method is very tedious, and the GPR image evaluation is fuzzy (Kang et al. 2019). These shortcomings lead to the inability to carry out effective detection and accurate maintenance of road structure distress (Liu et al. 2023). After that, artificial intelligence (AI) (Liu et al. 2021; Liu et al. 2023) and computer vision (CV) (Liu et al. 2022) make it possible to use deep learning to realize automatic detection of GPR images (Li et al. 2021; Solla et al. 2021). At present, the combination of artificial neural networks and deep learning algorithms has made some exploration and breakthroughs in GPR image processing (Tong et al. 2018). There are two detection methods— two-stage and one-stage. The two-stage method mainly includes R-CNN, Fast R-CNN, and Faster R-CNN (Wang et al. 2022). The implementation of the algorithm is divided into two processes: coarse *Corresponding Author: [email protected]
36
DOI: 10.1201/9781003425823-5
positioning and fine classification. The single-phase detection frameworks mainly include YOLO and SSD (Liu et al. 2022). This method does not need to carry on the region candidate operation and directly obtains the types and position from input images so that the localization problem of object boundary is transformed into a regression problem. The detection model represented by YOLOv3 has been widely used in transportation infrastructures (Liu et al. 2023), such as the detection of potholes and cracks in the road surface (Liu et al. 2022; Wang et al. 2022). The above series models are widely used in image recognition (Pang et al. 2020), such as medical image analysis (Raza & Singh 2021), road distress detection (Kim et al. 2018), and so on. However, the effectiveness of these models for GPR images is not discussed. At the same time, due to the relatively expensive GPR equipment and difficult data acquisition, it lacks a large dataset for GPR images, which also restricts the development of object detection algorithms in GPR image recognition of pavement distress. The YOLOv5 detection model was selected to detect the features of internal cracks in GPR images to realize the research on the recognition effect of the target detection model on GPR images. In addition, its accuracy and reliability are further improved by selecting the best image enhancement strategy. Finally, a control experiment was conducted on the mainstream target detection models, and combined with the evaluation index results. The model performance was analysed, guiding the intelligent detection and precise maintenance of internal road distresses.
2 GPR DETECTION AND DATA PROCESSING 2.1
GPR detection of pavement structures
By receiving, processing, and analysing reflected electromagnetic waves, GPR can determine the internal condition of pavement structures based on the difference in the dielectric constant of different media. Antenna arrays and stepping frequencies are key characteristics of GPR pavement detection technology, taking into account the depth of detection and the resolution required. Based on this, GPR equipment adopts GX750 (RTS&HDR) equipment produced by MALA, Sweden, which is equipped with the third-generation digital radar host ProEx. Integrated with the host and antenna, the detection system becomes faster and the signal-to-noise ratio is higher. The antenna frequency of 750 MHz can meet distress detection needs within the range of the road structure layer. Detailed detection information is shown in Table 1. Table 1.
Technical parameters of the MALA GX750.
Index
Frequency (MHz)
Time window (ns)
Signal noise ratio (DB)
Working temperature ( C)
Detection speed (km/h)
Values
900
50
> 64
30–60
40–80
2.2
Data processing of GPR images
According to the imaging results in previous studies, due to the heterogeneous discrete characteristics of the structural materials inside the road, there will be various noises in the detection image of ground-penetrating radar, which results in less observable information and blurred images. Therefore, when the detection image is obtained, it needs to be filtered. Reflexw software is used to effectively filter the original ground-penetrating radar image by removing DC drift, static excising, gain, removing horizontal and high and low-frequency interference signals. And transverse smoothing, noise, clutter, and other interference can be removed. The radar image before and after processing is shown in Figure 1. From Figure 1(b), internal cracks in the GPR images are characterized by an obvious hyperbola. 37
Figure 1.
Data processing diagram of GPR images.
3 OBJECT DETECTION MODEL 3.1
Data processing of GPR images
YOLO series is a single-step object detection framework, which directly implements classification probability regression and bounding frame coordinate regression on the input image to realize object detection. Since YOLOv3, the pooled layer and the fully connected layer have been abandoned, and Darknet53 has been used instead of Softmax for multi-scale prediction of target characteristics. YOLOv4 is an improvement of YOLOv3. This model consists of CSPDarknet53 backbone, spatial pyramid pooling, and PANet structures. At the same time, various data enhancement technologies are used to improve detection performance. mAP value increased by 10% compared with the YOLOv3 model. In 2020, Ultralytics released YOLOv5, which has the same performance as YOLOv4 but is faster in reasoning speed, and its model framework is more convenient for engineering deployment. YOLOv5 uses C3Darknet to extract rich information from GPR images, and PANet as Neck to aggregate features. The model detection layer is the same as that of YOLOv3, and the activation function of the convolutional module adopts the SiLU function. In addition, YOLOv5 set depth-multiple and width-multiple parameters to adjust the backbone network’s depth and width and divided four magnitude models: s, m, l, and x. In our study, the YOLOv5s model was used to detect GPR crack images. Its structure is presented in Figure 2. As e result, 2346 GPR images were used for model detection. A 6:2:2 ratio was used to divide the data set into training, testing, and verification.
Figure 2.
3.2
Data processing diagram of GPR images.
Evaluation index of the YOLOv5 model
The mean average precision (mAP) is used as an important index to measure the detection accuracy in target detection tasks. AP is the value of the area surrounded by the precisionrecall curves and axes. Precision (P) and Recall (R) are defined as Equations (1) and (2): Precision ¼ Recall ¼
CP CP þ NP
CP CP þ NF
38
(1) (2)
F1score ¼ mAP ¼
2PR PþR
(3)
PðRÞdR
(4)
ð1 0
where CP represents the number classified as cracks correctly, NP represents the number classified as cracks incorrectly, and NF denotes the number classified as non-cracks incorrectly. F1 score is a comprehensive evaluation index of the object detection model.
4 RESULTS AND DISCUSSION Table 2 summarizes the results of evaluation indexes of these experimental and control models: P, R, F1 score, and mAP. In most cases, evaluation indicators have been improved in some way after enhancements and improvements. To investigate the degree of improvement brought by image enhancement to the two target detection models from qualitative and quantitative perspectives, the YOLO series, SSD, and Fast R-CNN models were used as the experimental groups. In the YOLOv5 model, the P, R, F1 score, and mAP indexes were 84.4%, 85.7%, 85.2%, and 85.1%, respectively. As for the other groups, the model experiment results of YOLOv3 and YOLOv4 were close to each other. The F1 score and mAP of these two models were all about 76%, which indicates the YOLOv5 model is significant in the detection effect of the YOLO series version. The detection accuracy of the SSD model was only about 70%, indicating that it is not suitable for crack image detection from GPR images. In the Faster R-CNN model. Besides, the F1 score and mAP were all 80.8%, indicating that the two-stage model had more effective performance than the one-stage detection models. However, the improved YOLOv5 model exceeded the Faster R-CNN model, with the F1 score and mAP increased by 4.3% and 4.4%, respectively. Table 2.
Detection results of the experimental models. Indexes
Models
P
R
mAP
F1 score
YOLOv5 YOLOv4 YOLOv3 SSD Faster R-CNN
0.846 0.748 0.744 0.712 0.811
0.857 0.781 0.776 0.724 0.805
0.852 0.766 0.758 0.717 0.808
0.851 0.764 0.763 0.718 0.808
5 CONCLUSIONS The improved YOLOv5 model was performed to detect internal cracks from GPR images. Faster R-CNN, SSD, and other YOLO series models were also used to compare detection effects. The model performance of these models was further compared and analysed. The model’s robustness was enhanced and the detection accuracy was increased. The proposed YOLOv5 model achieved 85.2% detection accuracy and 85.1% F1 score, and the F1 score and mAP improved by 4.3% and 4.4% after the improvement. This paper further verified the superiority of the two-stage detection models in the detection accuracy, and the effect achieved can meet the practical needs of pavement maintenance engineering. 39
ACKNOWLEDGMENTS This work was financially supported by the Science and Technology Plan Project of the Department of Water Resources of Zhejiang Province (RC2165).
REFERENCES Chen, Y., X. Gu, Z. Liu and J. Liang (2022). “A Fast Inference Vision Transformer for Automatic Pavement Image Classification and Its Visual Interpretation Method.” Remote Sensing 14(8): 1877. Kang, M.-S., N. Kim, S. Im, J. J. Lee and Y.-K. An (2019). “3D GPR Image-based UcNet for Enhancing Underground Cavity Detectability.” Remote Sensing 11: 2545. Kim, N., K. Kim, Y.-K. An, H.-J. Lee and J. J. Lee (2018). “Deep Learning-based Underground Object Detection for Urban Road Pavement.” International Journal of Pavement Engineering 21: 1–13. Leng, Z. and I. Al-Qadi (2014). “An Innovative Method for Measuring Pavement Dielectric Constant Using the Extended CMP Method with Two Air-coupled GPR Systems.” NDT & E International 66. Li, S., X. Gu, X. Xu, D. Xu, T. Zhang, Z. Liu and Q. Dong (2021). “Detection of Concealed Cracks from Ground Penetrating Radar Images Based on Deep Learning Algorithm.” Construction and Building Materials 273: 121949. Liu, Z. (2019). Research of Airport Pavement Management System Based on BIM and WebGL. Liu, Z., Y. Chen, X. Gu, J. K. W. Yeoh and Q. Zhang (2022). “Visibility Classification and Influencingfactors Analysis of Airport: A Deep Learning Approach.” Atmospheric Environment 278: 119085. Liu, Z., X. Gu, Y. Chen and Y. Chen (2021). System Architecture and Key Technologies for the Whole Life Cycle of Smart Road. Liu, Z., X. Gu and Q. Dong (2019). Multi-Scale 3D Display of the Internal Quality of the Pavement Based on BIM. Liu, Z., X. Gu, Q. Dong, S. Tu and S. Li (2021). “3D Visualization of Airport Pavement Quality Based on BIM and WebGL Integration.” Journal of Transportation Engineering Part B-Pavements 147(3). Liu, Z., X. Gu, X. Dong, B. Cui and D. Hu (2023). “Mechanism and Performance of Graphene Modified Asphalt: An Experimental Approach Combined with Molecular Dynamic Simulations.” Case Studies in Construction Materials 18: e01749. Liu, Z., X. Gu and R. Hong (2023). “Fire Protection and Evacuation Analysis in Underground Interchange Tunnels by Integrating BIM and Numerical Simulation.” Fire 6(4): 139. Liu, Z., X. Gu and H. Ren (2023). “Rutting Prediction of Asphalt Pavement with Semi-Rigid Base: Numerical Modeling on Laboratory to Accelerated Pavement Testing.” Construction and Building Materials 375: 130903. Liu, Z., X. Gu, H. Ren, X. Wang and Q. Dong (2022). “Three-dimensional Finite Element Analysis for Structural Parameters of Asphalt Pavement: A Combined Laboratory and Field Accelerated Testing Approach.” Case Studies in Construction Materials 17: e01221. Liu, Z., X. Gu, H. Ren, Z. Zhou, X. Wang and S. Tang (2022). “Analysis of the Dynamic Responses of Asphalt Pavement Based on Full-scale Accelerated Testing and Finite Element Simulation.” Construction and Building Materials 325: 126429. Liu, Z., X. Gu and L. Wang (2021). Research on Information Management of Airport Pavement Quality Based on BIM and GIS Integration. Green and Intelligent Technologies for Sustainable and Smart Asphalt Pavements, CRC Press: 432–436. Liu, Z., X. Gu, C. Wu, H. Ren, Z. Zhou and S. Tang (2022). “Studies on the Validity of Strain Sensors for Pavement Monitoring: A Case Study for a Fiber Bragg Grating Sensor and Resistive Sensor.” Construction and Building Materials 321: 126085. Liu, Z., X. Gu, W. Wu, X. Zou, Q. Dong and L. Wang (2022). “GPR-based Detection of Internal Cracks in Asphalt Pavement: A Combination Method of DeepAugment Data and Object Detection.” Measurement 197: 111281. Liu, Z., X. Gu, H. Yang, L. Wang, Y. Chen and D. Wang (2022). “Novel YOLOv3 Model With Structure and Hyperparameter Optimization for Detection of Pavement Concealed Cracks in GPR Images.” IEEE Transactions on Intelligent Transportation Systems 23(11): 22258–22268. Liu, Z., L. Sun, X. Gu, X. Wang, Q. Dong, Z. Zhou and J. Tang (2023). “Characteristics, Mechanisms, and Environmental LCA of WMA Containing Sasobit: An Analysis Perspective Combing Viscositytemperature Regression and Interface Bonding Strength” Journal of Cleaner Production: 136255.
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Liu, Z., S. Wang, X. Gu, Z. Li, Q. Dong and B. Cui (2022). “Application of a Novel EWMA-f Chart on Quality Control in Asphalt Mixtures Production.” Construction and Building Materials 323: 126264. Liu, Z., W. Wu, X. gu, S. Li, L. Wang and T. Zhang (2021). “Application of Combining YOLO Models and 3D GPR Images in Road Detection and Maintenance.” Remote Sensing 6: 1081. Liu, Z., J. K. W. Yeoh, X. Gu, Q. Dong, Y. Chen, W. Wu, L. Wang and D. Wang (2023). “Automatic Pixellevel Detection of Vertical Cracks in Asphalt Pavement Based on GPR Investigation and Improved Mask R-CNN.” Automation in Construction 146: 104689. Long, J., Q. Luo, Z. Liu and Z. Zhu (2023). Road Distress Detection and Maintenance Evaluation Based on Ground Penetrating Radar. Advances in Civil Function Structure and Industrial Architecture, CRC Press: 474–481. Mardeni, R., r. s. a. raja abdullah and H. Shafri (2010). “Road Pavement Density Analysis Using a New NonDestructive Ground Penetrating Radar System.” Progress In Electromagnetics Research B 21: 399–417. Pang, L., H. Liu, Y. Chen and J. Miao (2020). “Real-time Concealed Object Detection from Passive Millimeter Wave Images Based on the YOLOv3 Algorithm.” Sensors 20: 1678. Raza, K. and N. Singh (2021). “A Tour of Unsupervised Deep Learning for Medical Image Analysis.” Current Medical Imaging Reviews 17. Shangguan, P., I. Al-Qadi, A. Coenen and S. Zhao (2014). “Algorithm Development for the Application of Ground-penetrating Radar on Asphalt Pavement Compaction Monitoring.” International Journal of Pavement Engineering 17: 1–12. Solla, M., V. Perez-Gracia and S. Fontul (2021). “A Review of GPR Application on Transport Infrastructures: Troubleshooting and Best Practices.” Remote Sensing 13: 672. Tong, Z., G. Jie and H. Zhang (2018). “Innovative Method for Recognizing Subgrade Defects Based on a Convolutional Neural Network.” Construction and Building Materials 169: 69–82. Wang, D., Z. Liu, X. gu, W. Wu, Y. Chen and L. Wang (2022). “Automatic Detection of Pothole Distress in Asphalt Pavement Using Improved Convolutional Neural Networks.” Remote Sensing 14: 3892. Wang, L., X. Gu, Z. Liu, W. Wu and D. Wang (2022). “Automatic Detection of Asphalt Pavement Thickness: A Method Combining GPR Images and Improved Canny Algorithm.” Measurement 196: 111248.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Shaking table test of frame structure considering bidirectional earthquake Hongmei Ren School of Digital Construction, Shanghai Urban Construction Vocational College, Shanghai, China BIM Engineering Center of Anhui Province, Anhui Jianzhu University, Hefei, China
Gongsheng Peng Yihai Kerry Arawana Holdings Co., Ltd., Shanghai, China
Fuwen Zhang Shanghai Research Institute of Building Sciences Co., Ltd., Shanghai, China
Hongwei Si* School of Architectural and Environmental Art, Shanghai Urban Construction Vocational College, Shanghai, China
ABSTRACT: To study the seismic performance of frame structures under bidirectional earthquake, a four-story frame structure model was designed according to the geometric scale of 1:3, and shaking table tests were carried out on the model. The dynamic characteristics of the model, such as natural vibration period and damping ratio, were identified. The displacement and acceleration responses of the model under different seismic wave inputs at different seismic level stages were recorded, and the seismic response differences between the transverse frame and the longitudinal frame were analyzed. The results provide a basis and reference for further research on the seismic performance of frame structures under bidirectional earthquakes.
1 INTRODUCTION As a kind of sudden natural disaster, earthquakes are mainly reflected in the damage or collapse of buildings caused by the earthquake and the casualties and huge economic losses caused by flood, fire, and other secondary disasters, which makes the seismic design research of engineering structures become the focus (Liu 2008). The seismic response analysis and design methods of building structures have been developed from the static stage, response spectrum stage, dynamic stage, and performance-based seismic design theory stage to resilient structures. The shaking table test is one of the important means to simulate the seismic performance of building structures under laboratory conditions. It can timely present the ground motion and dynamic effect on buildings under earthquake and effectively evaluate the seismic performance of the whole building structure (Wang 2010). Experts and scholars have researched shaking table test methods from theoretical and experimental perspectives (Huang 2012; Lu 2017; Shen 2010; Shang 2019; Tian 2008; Zhou 2012, 2019, 2022; Zhou & Lu 2012). In this paper, the seismic performance of reinforced concrete frame structures under bidirectional earthquake will be studied from model similarity relationship, model design, test conditions, selection of seismic excitation, the input of seismic excitation, and analysis of test results. *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-6
2 MODEL DESIGN 2.1
Similarity relationship
In this test, model design, manufacture, and seismic excitation input of the model are carried out by the similarity theory. The model and the prototype must be geometrically similar in size and maintain a certain proportion. The model and the prototype must be similar in material or have some similar relationship. The geometric scale ratio of the model selected in this experiment is 1:3, the elastic modulus is 1:1, and the other similarity coefficients are shown in Table 1. Table 1.
Structural similarity coefficient of the model.
Physical property
Physical parameter
Geometric property Length Material properties Strain Equivalent modulus of elasticity Equivalent stress Mass density Mass Load performance Concentration Force Line load Surface load Bending moment Dynamic Period Frequency performance Speed Acceleration
2.2
Expression
Similarity Dimension constant
Sl Se ¼ Ss =SE SE
L / FL2
1:3 1:1 1:1
Ss ¼ SE Sr ¼ SE =ðSl Sa Þ Sm ¼ Sr Sl3 Ss Sl2 Ss Sl Ss SM ¼ Ss Sl3 ST ¼ ðSE =Sr Þ0:5 Sl Sw ¼ ðSE =Sr Þ0:5 Sl1 Sv ¼ ðSE =Sr Þ0:5 Sa ¼ SE =ðSl Sr Þ
FL2 FT2L4 FT2L1 F FL1 FL2 FL T T1 L T1 L T2
1:1 1.8:1 1:15 1:9 1:3 1:1 1:27 0.447:1 2.236:1 0.707:1 1.667:1
Model design drawing
The structure of the test model is 4 layers, the storey height is 1.0 m, the total height is 4.0 m, the design strength of the concrete is C25, and the reinforcement is HPB300. The structure plan and elevation are shown in Figures 1–3.
Figure 1.
Layout plan (horizontal X, vertical Y).
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Figure 2.
Lateral frame elevation.
Figure 3.
Elevation of longitudinal frame.
3 TEST SCHEME 3.1
Seismic wave selection
According to the requirements of seismic fortification (GB 50011-2010, 2016), El Centro seismic wave is selected as the shaking table input excitation in this test. The time history curve and response spectrum after the adjustment of peak acceleration (0.1 g) are shown in Figure 4. When the excitation is input on the shaking table, the peak acceleration of the El Centro seismic wave should be adjusted according to the similarity coefficient of acceleration = 1.667 and the corresponding relationship between each seismic level and the designed basic seismic acceleration value. For example, under the basic intensity of 7 degrees (0.1 g), the peak acceleration of the input seismic wave on the shaking table should be 0.1 g 1.667 = 0.167 g. In addition, the seismic wave holding time should be compressed to 0.447 times the original record holding time according to the time similarity coefficient = 0.447.
Figure 4.
3.2
El centro wave time history curve and response spectrum.
Test condition
The simulated seismic tests were carried out on the model structure in the order of 7 degrees frequent, 7 degrees basic, 7 degrees rare, 8 degrees rare, 8.5 degrees rare, 9 degrees rare, 10 degrees basic, and 10 degrees rare under the test loading conditions. Before and after the input of different level seismic waves, the model was scanned with white noise to test the characteristic dynamic parameters of the structure, such as natural vibration frequency, vibration mode, and damping ratio. The test conditions are shown in Table 2. 44
Table 2.
Test condition table.
Serial number
Condition number
1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28
SWEEP1 F7-EL-XY F7-EL-YX SWEEP2 B7-EL-XY B7-EL-YX SWEEP3 R7-EL-XY R7-EL-YX SWEEP4 R7.5-EL-XY R7.5-EL-YX SWEEP5 R8-EL-XY R8-EL-YX SWEEP6 R8.5-EL-XY R8.5-EL-YX SWEEP7 R9-EL-XY R9-EL-YX SWEEP8 B10-EL-XY B10-EL-YX SWEEP9 R10-EL-XY R10-EL-YX SWEEP10
Intensity
Seismic excitation
1st white noise 7 degrees frequent 2nd white noise 7 degrees basic 3rd white noise 7 degrees rare 4th white noise 7.5 degrees rare 5th white noise 8 degrees rare 6th white noise 8.5 degrees rare 7th white noise 9 degrees rare 8th white noise 10 degrees basic 9th white noise 10 degrees rare 10th white noise
Input value in the X direction (gal)
Input value in the Y direction (gal)
50 58.3 49.6 50 166.7 141.7 50 366.7 311.7 50 516.8 439.3 50 666.8 566.8 50 850 722.5 50 1033.3 878.3 50 1200 1020 50 1500 1275 50
50 49.6 58.3 50 141.7 166.7 50 311.7 366.7 50 439.3 516.8 50 566.8 666.8 50 722.5 850 50 878.3 1033.3 50 1020 1200 50 1275 1500 50
4 ANALYSIS OF TEST RESULTS 4.1
Dynamic characteristic analysis
In this test, white noise scanning was carried out on the model before and after the working conditions at each stage. The natural vibration frequency, period, and damping ratio calculated under the working conditions with white noise are shown in Table 3. The frequency reduction rate is the percentage reduction of the structure’s natural vibration frequency about the structure’s initial state after different peak acceleration test conditions. Table 3.
Dynamic characteristics of the model under different white noise conditions. First-order translation in the X direction
First-order translation in the Y direction
White noise condition
Frequency Period Damping /Hz /s ratio
Frequency reduction rate
Frequency Period Damping /Hz /s ratio
Frequency reduction rate
Initial state 7 degrees frequent 7 degrees basic 7 degrees rare 7.5 degrees rare 8 degrees rare 8.5 degrees rare 9 degrees rare 10 degrees basic
6.75 6.75 5.63 3.75 3.00 2.88 1.75 1.38 1.25
– 0.00% 16.59% 44.44% 55.56% 57.33% 74.07% 79.56% 81.48%
7.13 7.13 5.88 4.00 3.13 3.00 1.88 1.38 1.13
– 0.00% 17.53% 43.90% 56.10% 57.92% 73.63% 80.65% 84.15%
0.15 0.15 0.18 0.27 0.33 0.35 0.57 0.72 0.80
1.96% 1.43% 2.09% 4.03% 9.50% 12.4% 8.26% 11.02% 21.04%
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0.14 0.14 0.17 0.25 0.32 0.33 0.53 0.72 0.88
1.94% 1.98% 3.62% 6.04% 12.00% 4.83% 10.21% 14.73% 16.58%
As seen from Table 3, no damage or failure occurs in the model, and the natural vibration frequency of the structure does not change after 7 degrees seismic test conditions. After the 7 degrees basic seismic test condition, although no obvious cracks appear in the model, the structure is damaged. The natural vibration frequencies in the X direction and Y direction decrease by 16.59% and 17.53%, respectively, compared with the initial state, and the damping ratio of the structure also increases. After the rare earthquake test condition of 7 degrees, the model appeared to have obvious cracks, and the structure was damaged. The structure’s natural vibration frequency reduction rate was the largest, and the X direction and Y direction decreased by 44.44% and 43.90%, respectively, compared with the initial state. After the basic seismic test condition of 10 degrees, the structure was seriously damaged. The natural vibration frequency in the X and Y directions decreased to 81.48% and 84.15%, respectively, compared with the initial state. The damping ratio in the X and Y direction reached 21.04% and 16.58%, respectively. 4.2
Acceleration response analysis
The acceleration response of the structure is related to the spectrum characteristics of seismic waves, the natural vibration period of the structure, and the damping ratio of the structure, which is an important parameter of the dynamic response of the structure. The maximum acceleration measured by the acceleration sensor placed on the model base is the reference standard. The maximum acceleration of each model structure layer divided by the foundation base’s maximum acceleration is the acceleration amplification coefficient of each model layer under this working condition. Figure 5 shows the comparison of structural acceleration amplification coefficients of the model under different working conditions.
Figure 5.
Comparison of floor acceleration amplification coefficients.
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As shown in Figure 5, the structural acceleration amplification coefficient shows an obvious trend of increasing with the rise of floors in all working conditions. In addition, with the increase of the input seismic wave acceleration peak, the acceleration amplification coefficient of each floor shows a decreasing trend. This is because, with the increase of the input acceleration peak, the structural beam-column joints fail to produce a plastic hinge, reducing the structural acceleration response. After 9 degrees rare occurrence and 10 degrees basic seismic test conditions, severe damage occurs to the components. The structure’s acceleration response decreases significantly with the increase of the floor and presents characteristics similar to the whole translation of the seismic isolation structure system. 4.3
Displacement response analysis
The interlayer displacement angle is an important sign to measure the damage degree of structure and an important basis for evaluating the damage degree of structure from the macro level. According to the model’s measured displacement time history data under each working condition, the displacement time history between each layer can be obtained by subtracting the displacement time history between two adjacent layers. The maximum displacement angle between each layer can be obtained by dividing the maximum displacement time history between the two layers by the layer’s height. Figure 6 compares displacement angles between model layers under different working conditions.
Figure 6.
Comparison of displacement angle between layers.
As can be seen from Figure 6, the structure has a similar pattern of inter-story deformation in the X and Y directions. The inter-story displacement angle firstly increases and then decreases with the increase of the floor. The weak layer of the structure is the bottom two 47
layers, and the inter-story deformation of the structure shows an overall trend of increasing with the increase of the input acceleration peak. Under the working condition of 7 degrees earthquake, the maximum displacement angle between the model’s layers in the X and Y direction is 1/1250 and 1/1000, respectively, much smaller than the standard value of 1/550. With the increase of the input acceleration peak value, the structure is damaged, and the interlayer displacement angle of the structure gradually increases. Under the 10 degrees basic seismic test condition, the interlayer displacement angle of the model in the X direction and the Y direction reaches 1/13.7 and 1/16.7. 4.4
Inter-storey shear force
The shear force between each layer of the model can be calculated according to the absolute acceleration response of measuring points at each layer. To facilitate calculation, the mass between the model floors is concentrated at the floors. Figure 7 shows the shear force distribution between model layers under different working conditions.
Figure 7.
Shear force distribution diagram between layers.
It can be seen from Figure 7 that the inter-story shear force of the structure decreases with the increase of the floor. With the increase of the peak acceleration under the test condition, the interstory shear of the structure on the whole increases first and then decreases. Under the condition of 48
7-degree frequent seismic test, the inter-story shear force of the structure is the minimum. With the increase of the peak value of loading acceleration, the interlayer shear reaches the maximum value when the earthquake test is carried out at 8.5 degrees. With the increase of the peak value of loading acceleration, the interlayer shear of the structure showed a decreasing trend under subsequent test conditions. This is caused by structural damage. Under the 9 degree rare and 10 degree basic seismic test conditions, the structural damage has been very serious. The natural vibration frequency of the structure has attenuated by more than 80%. 4.5
Residual deformation
The model does not have the ability of structural self-centering, so the structural damage and residual deformation will gradually accumulate with the test. The residual deformation of the structure in the X and Y directions is shown in Figure 8.
Figure 8.
Distribution of residual displacement.
As shown in Figure 9, the residual deformation of each layer of the model increases with the increase of the peak value of the input acceleration, and the residual deformation of the structure in the X direction is smaller than that in the Y direction. Under the rare earthquake of no more than 8.5 degrees test conditions, the residual displacement of the structure increases slowly. The maximum residual deformation in the X direction is 7.3 mm, and in the Y direction is 4.5 mm. In the case of 9 degrees rare seismic test, the residual deformation of the structure increases sharply, and the maximum value of the residual deformation in the Y direction is 28.1 mm. In the case of 10 degrees basic seismic test, the residual deformation continues to increase, and the maximum value of the residual deformation in the Y direction is 53.7 mm. Table 4 shows the peak value of the residual interlayer displacement angle in each model layer. Table 4.
Peak value of residual interlayer displacement angle.
X direction Y direction
1F
2F
3F
4F
1/62.16 1/53.45
1/57.84 1/55.26
1/115.98 1/63.29
1/1006.55 1/903.46
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5 CONCLUSIONS The model’s dynamic characteristics, failure mode, residual deformation, and other seismic performance indexes are studied through the shaking table test of the frame structure model considering bidirectional earthquakes. The main conclusions are as follows: 1. With the increase of seismic action, the natural vibration frequency of the structure in the X and Y directions decreases continuously, and the floor amplitude and damage increase continuously, eventually leading to the formation of the weak layer of the structure. 2. The structure’s acceleration amplification coefficient increases with the floor’s increase. In addition, with the increase of the input seismic wave acceleration peak, the acceleration amplification coefficient of each floor shows a decreasing trend. 3. With the increase of floor height, the inter-story displacement angle of the structure increases first and then decreases in both the X and Y directions. The inter-story deformation of the structure increases with the increase of the input acceleration peak. 4. The inter-story shear force of the structure decreases with the increase in floor height. With the increase of the peak acceleration under the test condition, the inter-storey shear force of the structure presents a changing trend of increasing first and then decreasing. 5. The residual deformation of each layer increases with the input acceleration peak value increase, and the residual deformation in the X direction is less than that in the Y direction. 6. The current research on structural seismic performance mainly focuses on unidirectional earthquakes, so it is suggested to carry out more in-depth research on bidirectional earthquakes from the depth and breadth of the research content. ACKNOWLEDGMENTS This work was financially supported by the BIM Engineering Center of Anhui Province (AHBIM2021KF02) and Shanghai Urban Construction Vocational College (cjky202231). REFERENCES Huang Sining, Guo Xun, Zhang Minzheng. (2012) Study of Design Method and Similitude for Small-scale Reinforced Concrete Structural Models. China Civil Engineering Journal., 45(7): 31–38. Liu Jianping. (2008) Shaking Table Model Test of Reinforced Concrete Frames. Hebei University of Engineering. Lu Xilin, Quan Liumeng, Jiang Huanjun. (2017) Research Trend of Earthquake Resilient Structures Seen from 16WCEE. Earthquake Engineering and Engineering Dynamics., 37(3): 1–9. Shang Shouping, Zhang Bin, Xiao Yifu, et al. (2019) Shaking Table Test of Base Isolation Low-rise Frame Structure. Earthquake Engineering and Engineering Dynamics., 39(04): 30–40. Shen Dejian, Lu Xilin. (2010) Experimental Study on the Mechanical Property of Microconcrete in Model Test. China Civil Engineering Journal., 43(10): 14–21. Tian Ye, Lu Xilin, Zhao Bin. (2008) Shaking Table Experimental Study on a Precast Concrete Frame. Structural Engineers., 24(1): 66–71. Wang Xin, Yan Zhenguo. (2010) Shaking Table Model Test and Seismic Evaluation of Reinforced Concrete Frames. Architectural and Structural Design., (02): 37–40. Zhou Ya, Feng Jian, Gai Jianguo, et al. (2022) Shaking Table Test on a Frame Structure Model of SCOPE System. Journal of Building Structures., 43(7): 100–110. Zhou Ying, Wu Hao, Gu Anqi. (2019) Earthquake Engineering: from Earthquake Resistance Energy Dissipation, and Isolation, to Resilience. Engineering Mechanics., 36(6): 1–12. Zhou Ying, Zhang Cuiqiang. (2012) Study on the Method of Ground Motion Selection and Input Sequence in the Shaking Table Test. Structural Engineers., 28(6): 60–65. Zhou Ying, Lu Xilin. (2012) Method and Technology for Shaking Table Model Test of Building Structures [M]. Beijing, Science Press. (2016) GB 50011-2010, Code for Seismic Design of Buildings [S]. 2016 ed. Beijing, China Architecture & Building Press.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Establishment method of evaluation index for the impact of chloride ion erosion on bearing capacity of reinforced concrete structure of tidal sluice Bin Yan* Zhejiang Institute of Hydraulics & Estuary (Zhejiang Institute of Marine Planning and Design), Hangzhou, Zhejiang Province, China Zhejiang Provincial Key Laboratory of Hydraulic Disaster Prevention and Mitigation, Hangzhou, Zhejiang Province, China
Lei Chen* Zhejiang Xianju Xia’an Reservoir Development Co., Ltd, Taizhou, Zhejiang Province, China
Jiabao Song* Hangzhou Dingchuan Information Technology Co., Ltd, Hangzhou, Zhejiang Province, China
Chengfa Deng* Zhejiang Institute of Hydraulics & Estuary (Zhejiang Institute of Marine Planning and Design), Hangzhou, Zhejiang Province, China Zhejiang Provincial Key Laboratory of Hydraulic Disaster Prevention and Mitigation, Hangzhou, Zhejiang Province, China
ABSTRACT: The tide sluice is significantly affected by chloride ion erosion, and the corrosion of steel bars leads to a decrease in the structure bearing capacity and endangers structure safety. Based on the analysis of the corrosion mechanism of steel bars caused by chloride ion intrusion into reinforced concrete, and the influence of steel bar corrosion on the bond stress between steel bars and concrete, this paper proposes a method for establishing a quantitative index related to chloride ion erosion in the safety evaluation of tidal sluice, that is, the relationship between chloride ion erosion and structure bearing capacity is established through finite element calculations, and the corrosion rate upper limit of tidal sluice components and the corresponding bearing capacity index value of the actual measured value of chloride ion can be obtained accordingly, which provides the necessary data basis for the comprehensive safety evaluation of tide sluice.
1 INTRODUCTION The tidal sluice is a large-scale tidal sluice project for comprehensive utilization of flood discharge, tidal protection, freshwater storage, and shipping. It plays an important role in regulating the water level of the river and preventing the backflow of seawater. The salt spray environment of the tidal sluice is different from the inland river sluice. The reinforced concrete structure will gradually age in the salt spray environment, and the bearing capacity of the concrete components will decrease accordingly. Chloride ion erosion related indicators. Based on the explaining of the mechanism of chloride ion intrusion and steel bar corrosion, this paper will study the analysis method of the impact of chloride ion erosion on the *Corresponding Authors: [email protected], [email protected], [email protected] and [email protected] DOI: 10.1201/9781003425823-7
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bearing capacity of reinforced concrete structures, establish the relationship between chloride ion erosion and the change of component bearing capacity, and formulate quantitative indicators related to chloride ion erosion.
2 DAMAGE MECHANISM OF REINFORCED CONCRETE STRUCTURE BY CHLORIDE ION EROSION 2.1
Mechanism of concrete corrosion by chloride ions and model selection
There are various ways for chloride ions to invade concrete, such as surface absorption, infiltration with water pressure, and chloride ion diffusion. Studies have shown that chloride ions are mainly transported to concrete by diffusion. There is a concentration gradient between the concrete surface exposed to the chloride salt environment and the internal pore solution, and then chloride ion diffusion occurs. The concentration gradient between the concrete surface concentration and the pore solution is the driving force for chloride ion diffusion. The diffusion process of chloride ions in concrete is complex and has many influencing factors. Scholars at home and abroad have done extensive research on this (Aloflso & Andrade 2000; Wang 2010). A chloride ion erosion model for concrete under saturated or unsaturated conditions has been established, which can meet the needs of different projects. The detection quantity of chloride salt erosion in the tidal sluice is mainly the penetration depth and chloride ion content. The chloride ion diffusion depth and concentration distribution in concrete can be simulated using the measured data combined with the diffusion model. Because Fick’s second law is in good agreement with the measured data, and it is a more effective model for simulating the diffusion of chloride ions. This model can be used to estimate the chloride ion concentration inside concrete when there is a lack of measured data. The analytical solution of the chloride ion concentration of the model is as follows: x C ðx; tÞ ¼ Ci þ ðCS Ci Þ 1 erf pffiffiffiffiffiffi (1) 2 Dt Cðx; t ¼ 0Þ ¼ Ci 0 < x < 1 Cðx ¼ 0; tÞ ¼ Cs 0 < t < 1
(2)
where D is the diffusion coefficient, x is the concrete depth, t is the diffusion time, C is the chloride ion concentration of the pore solution at x, and Cs and D are parameters related to factors such as temperature, relative humidity, and hydration degree. 2.2
Corrosion mechanism and corrosion rate estimation of steel bars in chloride ion environment
Chloride ions are very active and have a strong penetrating power. Even if the concrete is not carbonated, chloride ions can reach the surface of the steel bar from the pore solution. As the diffusion continues, chloride ions will accumulate on the passivation film on the steel surface, and the acidification of chloride ions will reduce the pH value of the steel surface to 4 or even lower (Hong 1999). Regarding the problem of steel bar corrosion, domestic and foreign scholars have done a lot of research and proposed a steel bar corrosion model, and whether the protective layer is cracked can be divided into two categories: – Before the protective layer cracks, representative models include the Kim model (Vu & Stewart 2000) and the Youping Liu model (Liu & Weyers 1998), etc. – After the protective layer cracks, the representative models include the Li model (Li 2004), etc. 52
In this paper, the following method is used to analyze the influence of chloride ion corrosion on the corrosion of steel bars. This method can be used to estimate the steel bar corrosion of the concrete members of the tidal sluice that have been in service and obtain the average corrosion rate of the steel bars of the sampling members, providing data for the quantitative indicators of structural safety evaluation (Wang 2007). The average annual corrosion rate of steel bars after the protective layer is cracked is: lcl ¼ 11:6 i 103 3034 0:005r þ ln mcl T þ 273 3 10 c Ms1 ¼ Ms0 þ ðMs Ms0 Þ 1 erf pffiffiffiffiffiffiffiffiffi 2 Dtcr
ln i ¼ 8:617 þ 0:618 ln Ms1
(3) (4)
(5)
where lcl is the average annual corrosion rate of steel bars (mm/a), i is the corrosion current density of steel bars (mA/cm2), r is the concrete resistivity (kWcm), which can be taken according to the actual measured value, T is the steel bar temperature ( C), mcl is the local environmental coefficient, Ms1 is the measured chloride ion concentration on the steel surface (kg/m3), Ms is the measured concrete surface chloride ion concentration (kg/m3), Ms0 is the concrete preparation time Incorporation of chloride ions (kg/m3), D is the diffusion coefficient of chloride ions, c is the thickness of the concrete protective layer, and tcr is the time for rust swelling of reinforced concrete members. The average annual corrosion rate of steel bars after the protective layer is cracked is: lcl1 ¼ ð4:5 26lcl Þ lcl lcl1 1:8lcl (6) lcl1 ¼ 1:8lcl lcl1 < 1:8lcl where lcl is the average annual corrosion rate of steel bars before the concrete cover cracks, and lcl1 is the annual average corrosion rate of steel bars after the concrete cover cracks.
3 ESTABLISHMENT OF BEARING CAPACITY INDEX OF CONCRETE STRUCTURE CORRODED BY CHLORIDE IONS 3.1
Influence of steel bar corrosion on bond stress between steel bar and concrete
In the simulation analysis of the bearing capacity of the main stressed structure of the tidal sluice, the bond-slip constitutive relationship between the steel bar and the concrete must consider the corrosion factor of the steel bar to make the calculation and analysis of the reinforced concrete structure corroded by chloride ions more reasonable. The bond-slip constitutive relationship between steel bars and concrete is the relationship between the local bond stress t at each point within the bond range between steel bars and concrete and the corresponding relative slippage s. This paper intends to use the empirical formula proposed by Kang (1996) as the bond-slip constitutive relation of finite element simulation. The bond strength degradation coefficient adopts the bond performance degradation model proposed by Yuan et al. (2001). The bond-slip constitutive relation is: rffiffiffiffiffiffiffiffiffi fc 2 4 2 6 4 tðsÞ ¼ 5:3 10 s þ 2:52 10 s 5:47 10 s (7) 40:7 where t represents the bonding stress (N/mm2), s represents the relative slippage (mm), and fc represents the concrete axial compressive strength (MPa). 53
The formula of the bond strength degradation coefficient is:
c Fu ¼ 1:0 10:544 1:586 hs d
(8)
where Fu represents the degradation coefficient of bond strength, c is the thickness of the concrete cover, d is the diameter of the steel bar, and hs is the average corrosion rate of the steel bar. 3.2
Establishment of an evaluation index for structures affected by chloride ion erosion
The main factors that affect the safety of tide sluice are working bridges, traffic bridges, sluice piers, etc., and the more serious ones are working bridge girders, traffic bridge girders and other components, especially the bending area of the beams (Niu et al. 2002). This paper takes the girder of the working bridge as the research object, analyzes the bearing capacity of the girder under various corrosion rates, and takes the corresponding applied load when the mid-span deflection reaches the maximum allowable value as the ultimate load of the corresponding corrosion rate. In this paper, the concrete damaged plasticity model is used to simulate and analyze the influence of steel bar corrosion on the bearing capacity of components. The ultimate load data samples under different corrosion rates are obtained through the finite element analysis, and the relationship between the ultimate load and the corrosion rate is obtained by fitting. From this relationship, steel bar corrosion can be obtained when the design value of the opening and closing force is the ultimate load. It is considered that the corrosion rate is the maximum allowable corrosion rate of the working bridge. If the corrosion rate of steel bars reaches this value, it is considered that there is a relatively large risk of hidden danger, and measures must be taken to eliminate the hidden danger and make it safe to serve. Taking the girder of a tidal sluice working bridge as an example, a finite element model is established to calculate the ultimate bearing capacity corresponding to the corrosion rate of each steel bar, and the relationship between the corrosion rate and the ultimate load of the tidal sluice working bridge girder can be obtained from the fitting result data: F ¼ 0:135h4 þ 2:249h3 13:228h2 þ 28:631h þ 50:970
(9)
where F is the ultimate load of the corresponding corrosion rate (kN), and h indicates the corrosion rate of steel bars (measured or estimated value). Incorporate the impact of chloride ions on the bearing capacity into the safety evaluation index system of the tidal sluice. We take the ultimate bearing capacity that takes into account the erosion of chloride ions into the comprehensive evaluation system, and quantify and standardize all indicators when conducting comprehensive safety evaluation on the tidal sluice. In this paper, the ratio of the bearing capacity F of the corroded girder to the bearing capacity F0 of the uncorroded girder is used as the standardization method, and the remaining ratio of the bearing capacity (F/F0) is proposed as the impact index of chloride ion erosion.
4 CONCLUSIONS The main conclusions of this paper are as follows: – On the basis of analyzing the mechanism of chloride ion erosion, models with good applicability, such as chloride ion diffusion, estimation of steel corrosion rate and bond stress degradation between steel and concrete were selected. 54
– A method of using the finite element method to establish the relationship between the corrosion rate of steel bars and the bearing capacity of tidal sluice components was proposed, and the residual rate of girder bearing capacity after chloride ion erosion was proposed as an evaluation index, which was incorporated into the safety evaluation index system of the tidal sluice. – The influence relationship of steel bar corrosion on the bond constitutive model involved in this paper is determined by experts and scholars based on the summary of experiments. In the future, it is necessary to improve the indicators related to chloride ion corrosion according to the latest mature research results to make it more reasonable.
ACKNOWLEDGMENT This study was supported by the science and technology plan project of the Zhejiang Provincial Department of Water Resources of China (No. RB2214) and the President Fund of Zhejiang Institute of Hydraulics & Estuary of China (No. ZIHE21Y003).
REFERENCES Aloflso, C., Andrade, C. (2000) Chloride Threshold Values to Depassivate Reinforcing Bars Embedded in a Standardized OPC Mortar. Cement and Concrete Research, 30(7): 1047–1055. https://doi.org/10.1016/ S0008-8846(00)00265-9. Hong, N.F. (1999) Corrosion and Protective Technology of Rebar in Concrete(3) – rebar Corrosion by Chloride Salt. Industrial Construction, 1999(10): 60–63. https://doi.org/10.3321/j.issn:1000-8993.1999. 10.017. Kang, Q.L. (1996) RC Finite Element Analysis. China Water&Power Press, Beijing. http://www.waterpub. com.cn. Liu, Y.P., Weyers, R.E. (1998) Modeling the Time to Corrosion Cracking in Chloride Contaminated Reinforced Concrete Structures. ACI Mater J, 95(6): 675–681. https://doi.org/10.14359/410. Li, C.Q. (2004) Reliability-based Service Life Prediction of Corrosion-affected Concrete Structures. Structure Engineer, 130(10):1570–1577. https://doi.org/10.1061/(ASCE)0733-9445(2004)130:10(1570). Niu, D.T., Lu, M., Wang, Q.L. (2002) Research on the Calculation Method of the Flexural Capacity of the Corroded Reinforced Concrete Normal Section. Building Structure, 32(10): 14–17. https://doi.org/ 10.19701/j.jzjg.2002.10.004. Vu, K.A.T., Stewart, M.G. (2000) Structural Reliability of Concrete Bridges Including Improved ChlorideInduced Corrosion Models. Structural Safety, 22(4): 313–333. https://doi.org/10.1016/s0167-4730(00) 00018-7. Wang, C.K. (2010) Peak Value Distribution of Surface Chloride Concentration and Convection Depth of Concrete. Bulletin of The Chinese Ceramic Society, 179(02): 262–267. https://doi.org/10.16552/j.cnki. issn1001-1625.2010.02.041. Wang, Q.L. (2007) Standard for Durability Assessment of Concrete Structures. China Architecture Publishing & Media Co., Ltd, Beijing. http://www.cabp.com.cn. Yuan, Y.S., Jia, F.P., Cai, Y. (2001) The Structural Behavior Deterioration Model for Corroded Reinforced Concrete Beams. China Civil Engineering Journal, 34(3): 47–52. https://doi.org/10.3321/j.issn:1000131X.2001.03.009.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Evaluation of technical status of concrete beam bridge based on machine learning Chengyu Li* Research Institute of Highway, Ministry of Transport, Beijing, China
ABSTRACT: This work offers an intelligent evaluation approach for the technical status of Concrete Beam Bridges to diminish further the influence of artificial factor weights on the overall technical bridge status grade assessment in the highway bridge evaluation specification. Based on artificial intelligence, a disease status database is created using 350 pieces of detection data from short and medium-span concrete bridges of a Beijing-based bridge detection company, based on the technical status specification of highway bridges and the index system of technical status. Three ensemble learning algorithms–XGBoost, LightGBM, and CATBoost–are utilized to build a mapping model of bridge disease characteristics and illness status, using the crack disease of the upper structure as an example. The ideal parameters are identified by plotting the learning curve and using 5-fold cross-validation, and the ensemble learning algorithm’s evaluation performance is contrasted. The outcomes demonstrate that the LightGBM algorithm can be more effectively utilized to assess the technical state of bridges in this region. This method can utilize the data value accumulated in the detection report and further reduce the grading deduction flaws of the highway bridge evaluation specifications and the impact of artificial determinant weight on the grading evaluation of bridge technical status.
1 INTRODUCTION The term “bridge technical status evaluation” refers to a thorough assessment of the bridge’s technical condition by periodic inspection data and pertinent specifications to determine the bridge’s technical status level and provide a foundation for future maintenance. The “Highway Bridge Maintenance Specification” (JTG 5120-2021), published by the Ministry of Transportation, is one of China’s Bridge Assessment standards and has been in effect since November 1, 21. At the same time, the original “Code for Maintenance of Highway Bridges and Culvers” (JTG H11-2004) has been eliminated. The “Standards for Technical Condition Evaluation of Highway Bridges” (JTG/T H21-2011), as it is now known, contains specific guidelines for the evaluation procedures and benchmarks for the technical status of bridges. The Assessment Standard examines the technical state of the entire bridge hierarchically and according to the bridge’s structural system, components, and components. Combined with a single control index that is immediately determined to be five types of bridges, it also assesses the technical state of the entire bridge. According to the survey, it is discovered that there are notable disparities in the findings of the inspection and appraisal of the same bridge because of the diverse knowledge backgrounds and practical experience of bridge inspectors. A challenging issue in the research of bridge technical conditions is the inconsistent evaluation results of highway bridge conditions. Relevant academics, both *Corresponding Author: [email protected]
56
DOI: 10.1201/9781003425823-8
domestically and internationally, have proposed numerous solutions to the issues that plague the assessment of bridge technical conditions. The techniques are largely separated into two groups. The first step is to assess the bridge’s technical state using fuzzy math and hierarchical analytic techniques. Jiang et al. evaluated the stone arch bridge using the analytical hierarchy process and developed a new illness deduction and overlay formula considering the component weight and rate of degradation (Jiang 2015). I adjusted the weight of the components of the concrete beam bridge in the specifications based on the analytic hierarchy process. I put forward a set of systems for the status evaluation of the beam bridge (Li 2014). The disadvantage of hierarchical analysis, fuzzy mathematics, and other methods is that they all need to determine the weight of the factors that affect the bridge status, and the subjective factors strongly influence the evaluation results. The second is the bridge assessment using machine learning and large data analysis methods. Miao et al. used the maintenance and inspection data of 3368 bridges and realized the prediction of three bridge status levels through an artificial neural network and sensitivity analysis method (Miao 2021). Gong evaluates the technical condition of simple-supported beam bridges based on a support vector machine, trains and tests the support vector machine with expert samples, and obtains a classifier that can judge the scale of the supported beam bridges according to the disease characteristics (Gong 2019). Machine learning is more advantageous than traditional evaluation methods, which can not only make full use of the data value accumulated in the detection reports but also avoid calculating Bridges’ overall technical condition score based on cumbersome formulas and weights in the detection specifications. This paper uses crack disease of the upper structure as an example, uses three machine learning algorithms to learn the benefits and drawbacks of machine learning algorithms in applying the technical status of bridges, establishes the mapping model of bridge status evaluation values, and provides the best model by comparison, all based on 350 assessment reports of the technical status of small and medium span concrete bridges from a bridge detection company in Beijing. As an illustration, the model is used to forecast the technical state of bridges.
2 ENSEMBLE LEARNING ALGORITHMS Instead of using a single machine learning method, ensemble learning builds numerous models from the data and then combines them to create a more accurate model. Weak learners are the models that are involved in ensemble learning. Examples of common ensemble learning implementations include the Boosting and Bagging frameworks. The representative model for bagging is the random forest, and the representative model for boosting is XGBoost. 2.1
XGBoost
XGBoost is an enhancement to the gradient lifting algorithm (Chen 2016). To reduce the complexity of the model, avoid overfitting, and boost the model’s generalizability, a regularization item is introduced to the loss function. The following are some of this algorithm’s benefits: 1. It pre-orders the data, utilizes a parallel approach to the feature granularity, drastically decreases computation, and then selects the feature with the highest gain to divide. Due to the quantity and quality of bridge data, calculation efficiency has increased. 2. It eliminates the potential for misclassification of many cracks and lessens over-fitting. This algorithm’s drawback is that it requires a high-performance server and takes up a lot of storage space. 2.2
LightGBM
LightGBM is a framework for GBDT algorithm implementation that facilitates effective parallel training (Qi 2017). The following are some of this algorithm’s benefits: 1. When 57
using the histogram technique, it is not only unnecessary to retain the results of the presorting, but it is also possible to save only the values of the discretized features, which reduces memory use and processing costs. 2. Using the optimized feature parallel and data parallel methods to speed up the calculation, a parallel voting strategy can also be used when the amount of data is very large. This eliminates a lot of pointless computations and significantly reduces the time complexity. It also provides a quick iteration basis for the data model of bridges, allowing for continuous machine optimization. The algorithm has the following drawbacks: 1. The algorithm is iterative, each iteration will adjust the weight of the prediction results of the previous iteration, and the model bias will gradually reduce, so it is easy to be sensitive to some noise easily generated by the bridge characteristics. 2. A deep decision tree may grow, and a maximum depth limit needs to be added in advance to ensure high efficiency and prevent over-fitting. 2.3
CATBoost
CATBoost is a GBDT framework with fewer parameters, support for categorical variables, and good accuracy based on a symmetric decision tree as a basic learner (Dorogush 2018). Handling category characteristics effectively and fairly is the key problem to be solved. These are some of this algorithm’s benefits: 1. It can deal with category and numerical characteristics, simplifying the steps of turning bridge category characteristics into numerical features. 2. It eliminates the requirement for multiple hyperparametric adjustments, decreases over-fitting, and increases the robustness of the bridge model. The algorithm has the following drawbacks: 1. The processing of class-type features consumes a significant amount of memory and processing time, and the computing server’s performance is quite high. 2. Different bridge models can be obtained by weighted voting since different random number settings will have some impact on the results of the model prediction. 3 MODELING AND ANALYSIS 3.1
Bridge status database
All data sources (such as monitoring and assessment reports, traffic volume reports, etc.) can only augment this study’s regular inspection reports of 350 medium and small-span bridges. First, since the test report’s Word document served as the original data source, extracting the information using natural language processing was necessary. Next, the basic information about the bridge and the scoring table for the superstructure, substructure, and deck system had to be extracted. Finally, the information had to be classified according to the evaluation component classification and disease classification before the design load, bridge age, and other characteristic indicators could be extracted as features. When dealing with the issue of each feature having a significantly distinct value range, dimensional normalization and data standardization are frequently used. This offers two benefits: Scaling data from various ranges to 0–1 can help the model be more accurate. The second step is to convert dimensional attributes into unified attributes and eliminate dimensional effects. At the same time, not all attributes sufficiently contribute to the label, so it is necessary to filter the database attributes, keeping the attributes related to the overall technical status of the bridge and removing the attributes that appear to be unrelated, such as the bridge number, name, etc., to finally get a simplified database. The superstructure crack disease is taken as an example and “Standards for Technical Condition Evaluation of Highway Bridges” (JTG/T H21-2011) is used. The component and component scores in the conventional bridge inspection report are the training feature vector. The technical information about the bridge’s structural elements includes the beam type, the location of cracks on the surface, the surface of the crack, the crack direction, the crack length, the crack width, the pantothenic phenomenon, the number of cracks, design load, the bridge age, and the bridge width. 58
3.2
Bridge technical condition classification model
The training and test sets are created by randomly dividing the bridge status database’s data. The test set evaluates the discriminatory power of the learned model parameters to new samples. In contrast, the training set is used to learn potential data distribution. This study uses 80% of the data as a training set and 20% as a test set. Three popular ensemble learning algorithms—XGBoost, LightGBM, and CATBoost—are contrasted in this work. Cross-validation, learning curves, or grid searches are typically used to determine a model’s ideal parameters for the greatest evaluation performance possible. They are using cross-validation to monitor model stability. It randomly and evenly splits samples into k parts, uses one as the test set, the remaining k-1 as the training set, and then takes the mean of the accuracy of the k tests as the final accuracy. It is possible to see whether the model is under-fitting or over-fitting by drawing the learning curve. The regularization penalty factor of XGBoost L2 is determined by continually debugging with step 0.01 after repeated debugging and narrowing the search interval within the (0, 5) interval. The model performed best when = 0.55, as shown in Figure 1. The L2 regularized penalty coefficient for the LightGBM and CATBoost models’ best score can also be found. The classifier in the training set is replaced with the combined parameters of XGBoost, LightGBM, and CATBoost, and 5-fold cross-validation is utilized. As can be seen in the figure, it is discovered that neither the training set accuracy nor the validation set accuracy produces the best outcomes. The training set sample size is too small, which is the cause. The accuracy of the training set and test set will get closer and closer throughout cross-validation if the training samples keep growing. Eventually, convergence and accuracy will reach the ideal value.
Figure 1.
3.3
XGBoost L2 regularized penalty coefficient.
Model evaluation
The technical status of the bridge sample data is classified and forecasted in the test set using cross-validation, learning curves, and grid searches to determine the appropriate parameters for each algorithm model. This work employs eight widely used machine learning model assessment metrics to assess the performance of the display classification model: accuracy, recall rate, F1 score, KS value, confusion matrix, ROC curve and AUC value, and feature importance. Using cracks as an example, statistical analysis of 350 small and medium-span bridges reveals that the primary goal is to categorize diseases of Types 2 and 3. In contrast, diseases of Types 1 and 4 are insufficient or insufficiently prevalent to support the categorization. The test set had 56 Type 3 diseases and 248 Type 2 diseases. This is a problem of two classes. Table 1 compares the assessment indices of training duration, accuracy, precision, regression rate, F1 score, AUC value, and KS value for the three ensemble learning algorithms, XGBoost, LightGBM, and CATBoost. 59
Table 1.
Evaluation indices of different algorithms.
Evaluation index
XGBoost
LightGBM
CATBoost
Training time Accuracy Precision Regression rate F1 score AUC value KS value
0.073 s 98.68% 93.33% 100.00% 96.55% 99.19% 98.39%
0.022 s 98.68% 93.33% 100.00% 96.55% 99.19% 98.39%
3.990 s 98.36% 91.80% 100.00% 95.73% 98.99% 97.98%
Compared with LightGBM, the shortest total training time is 0.022 seconds, the accuracy of XGBoost and LightGBM is 98.68%, the accuracy is 93.33%, the F1 value is 96.55%, the AUC value is 99.19%, and the KS value is 98.39%, which is the highest of the three types of ensembles learning algorithms, and the recall rate is excellent. The confusion matrix for the test set is calculated to demonstrate the model’s effectiveness, as shown in Figure 2. The confusion matrix is a square matrix, with each column representing the predicted technical status level and each row representing the actual technical status level. Elements on the diagonal line of a matrix represent the proportion of disease correctly classified. In contrast, elements on the non-diagonal line of a matrix represent the proportion of diseases incorrectly classified. Therefore, the higher the diagonal value is, the higher the accuracy of the classifier is. Non-diagonal data is not found to be 0, indicating a correlation between the two and three diseases. The first reason is that the performance of
Figure 2.
Confusion matrix of different algorithms.
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the evaluation index or algorithm used in this paper needs to be better at the stage of bridge grade transition. Second, there are some limitations in evaluating the scale of components in the test report by standardizing graded deduction methods. Overall, the ensemble learning algorithm is more stable for predicting three kinds of crack diseases than the two kinds of diseases, reflecting the limitations of the evaluation indexes and the current highway bridge evaluation standards. Most of the indicators in the evaluation criteria are simple appearance information and non-destructive testing data, together with the subjectivity of the evaluation, which may lead to large deviations in the disease evaluation scale. XGBoost, LightGBM, and CATBoost offer a package of significant importance to return the importance of each feature following preprocessing-modeling-prediction-evaluation. To determine the value of each feature in the total model, they calculate the importance of each feature based on the weak classifier, add each feature’s contributions to each weak classifier, and then average them. The model can be better understood and comprehended by calculating the relevance of its features. The model is probably flawed if the model’s top attributes do not follow the evaluation logic. Each feature importance for each of the three ensemble learning algorithms is calculated in this study. The weights are then redistributed by the components under consideration, and the resulting feature importance is compared to that of the algorithm, as shown in Table 2. The crack length, width, and bridge width comprise the first three characteristics of the XGBoost algorithm. The first three characteristics of the LightGBM algorithm are the crack length, width, and bridge width. Design load, bridge width, and crack width are the top three attributes of the CATBoost algorithm. The crack’s length and width are considered in the quantitative description of the specification compared to the crack specification standard, as shown in the table. As a result, the XGBoost and LightGBM algorithms are closer to the quantitative criteria. Additionally, the XGBoost and LightGBM algorithms have no connection to the white analysis function, which scored the lowest of the three algorithms. Analysis may be necessary because it is challenging to discern between the 2 class and 3 class of pantothenic phenomenon. Table 2.
Features the importance of different algorithms.
Feature
XGBoost
LightGBM
CATBoost
The beam type Location of cracks on the surface The surface of the crack The crack direction The crack length The crack width Pantothenic phenomenon Number of cracks Design load The bridge age The bridge width
7.177033 6.937799 0.956938 8.61244 28.4689 18.66029 0 0.239234 5.263158 6.45933 17.22488
3.097345 6.415929 0.442478 9.292035 33.84956 20.13274 0 0.221239 3.982301 6.858407 15.70796
9.925929 3.147584 2.245694 9.517874 13.79393 18.25299 0.3763 0.195139 19.89539 3.89575 18.75342
4 CONCLUSIONS Based on the results and discussions presented above, the conclusions are obtained as below: 1. This paper creates a database of bridge status based on 350 technical status evaluation reports of medium and small-span concrete bridges from a bridge inspection company in Beijing. It confirms that the current bridge beam evaluation work is focused on using machine learning algorithms for disease prediction and evaluation of the technical status grade of the technical bridge status. 61
2. Three distinct ensemble learning methods are evaluated by comparing their accuracy, precision, recall, F1 score, KS value, and other indicators. The outcomes demonstrate that LightGBM performs at its peak. This does not imply, though, that LightGBM is always the best option. Finding the best method requires some knowledge and experimentation due to the impact of external influences on bridges in various places and the differences in engineers’ inspection and evaluation. 3. Using machine algorithms to evaluate and anticipate bridge technical conditions offers certain benefits but drawbacks: The number of training sets in this paper is too small, and it is necessary to continue collecting samples to converge the model to improve prediction accuracy. 1. When evaluating the model’s performance, it is discovered that improving and refining the evaluation indicators can continue improving accuracy. 2. The model is still overfitting when searching for the optimal parameters. 3. Currently, some diseases cannot be assessed or predicted due to a shortage of samples.
REFERENCES Chen T., Guestrin C (2016). XGBoost: A Scalable Tree Boosting System. Proceedings of the 22nd ACM SIGKDD International Conference on Knowledge Discovery and Data Mining. San Francisco California USA: ACM: 785–794. Dorogush, A. V., Ershov, V., & Gulin, A. (2018). Catboost: Gradient Boosting with Categorical Features Support. Gong H (2019). Technical Condition Evaluation of Simply Supported Bridges Based on Support Vector Machine Method. (Doctoral dissertation, Zhejiang University of Technology). JTG 5120-2021, Specifications for Maintenance of Highway Bridges and Culverts. JTG H11-2004, Code for Maintenance of Highway Bridges and Culvers. JTG/T H21-2011, Standards for Technical Condition Evaluation of Highway Bridges. Jiang, T.Y., Long, W (2015). A Method for Evaluating Technical Condition of Stone Arch Bridge Based on AHP. Journal of Highway and Transportation Research and Development, 32(9), 49–56. Li L.P (2014). The Application of the Variable Weight Synthetic Assessment Method for Condition Evaluation of Beam Bridge (Doctoral dissertation, Chang’an University). Miao P., Liu P (2021). Prediction-based Maintenance of Existing Bridges Using Neural Network and Sensitivity Analysis. Advances in Civil Engineering, 2021, 4598337. Qi M (2017). LightGBM: A Highly Efficient Gradient Boosting Decision Tree. Neural Information Processing Systems. Curran Associates Inc.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Reinforcement effect of the anti-slide pile on structural slope plane based on point safety factor method Xun Zhang Hangzhou Fuyang District Water Conservancy and Hydropower Engineering Quality and Safety Service Guarantee Center, Hangzhou Zhejiang, China
Yixin Chen & Yi Kuang* Hangzhou Fuyang District River Reservoir Management Center, Hangzhou Zhejiang, China
ABSTRACT: To study the application of the point safety coefficient in the reinforcement effect of structural slope planes, taking a certain slope as an example, based on FLAC3D numerical simulation, the point safety coefficient method was used to analyze the distribution pattern and changes of point safety coefficient of the main potential sliding structural planes of the slope before and after the reinforcement with anti-slip piles. The results showed that after the reinforcement with anti-slip piles, the point safety coefficient of the potential sliding structural planes was significantly improved, and the anti-slip piles significantly affected the stability of slope reinforcement. At the same time, from the distribution pattern of safety factors before and after reinforcement, it can be seen that the distribution of safety factors after reinforcement is smoother, indicating that anti-slip piles improve the overall stability of the slope. It provides an engineering case for the layout and design of anti-slip piles.
1 INTRODUCTION The anti-slide pile is a pile column passing through the potential sliding structural plane of the slope, which is used to support the sliding force of the sliding mass and stabilize the slope. It is suitable for shallow and medium-thick landslides. It is a common measure for slope anti-slide treatment. At present, the analysis methods for the effect of anti-slide piles on slope reinforcement are divided into two categories: the first is various numerical analysis methods based on strength reduction technology, and the second is various slice methods based on limit equilibrium theory. The research direction mainly focuses on the pile-soil interaction, the stability analysis of the slope strengthened by anti-slide piles, and the design of anti-slide piles. The stability analysis of slope strengthened by anti-slide piles, especially the influence of piles on the safety factor of slope stability, has attracted the interest of many researchers for many years. Dai et al. (2007) studied the application of the strength reduction method in the dynamic stability of complex rock slopes. Zhang et al. (2010) carried out reinforcement simulation and application of slope anti-slide pile based on FLAC3D structural unit. Lin (2020) studied the reasonable pile spacing of cantilever anti-slide piles. Chen et al. (2022) studied the influence of the anchorage length of the anti-slide pile on the sliding surface and the antislide capacity of the homogeneous slope. Ping et al. (2022) carried out impact analysis on the pile position and embedded depth of landslide anti-slide pile reinforcement in different evolution modes. Hou et al. (2022) carried out a sensitivity analysis of three-dimensional *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-9
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stability factors of slope strengthened by anti-slide piles. Liu et al. (2021), Ren et al. (2022), and Su et al. (2023) conducted a comparative study on the arching soil effect of anti-slide piles with different sections. But it has not yet obtained a universal and recognized method. This paper uses FLAC3D numerical simulation to study the reinforcement effect of antislide piles on the slope by analyzing the distribution law and change of the point safety factor of the main potential sliding structural plane of the slope before and after the anti-slide pile reinforcement, providing engineering cases and technical support for the arrangement and design of anti-slide piles.
2 COMPUTING METHOD 2.1
Calculation model
A slope is located in Ya’an City, Sichuan Province. The model is built with FLAC3D. The length of the model (X direction) is 720 m, the width (Y direction) is 690 m, and the maximum height (Z direction) is 230 m. The model is divided into 32340 hexahedral elements, with 35518 nodes. The slope is cut into three potential sliding blocks by three weak structural planes, sh1, sh2, sh3, and tensile cracks j11, j102, and fault f4. The anti-slide piles are arranged in three rows, with 57 anti-slide piles in the first row, arranged along the exposed part of sh1. The pile length is 40 m, the spacing is 6m, and the section size is 3.6 m 2.4 m; There are 53 anti-slide piles in the second row arranged along the exposed part of sh2. The pile length is 25 m, the spacing is 6 m, and the section size is 3.0 m 2.0 m; 47 anti-slide piles in the third row are arranged along the exposed part of sh3, with the pile length of 25 m, spacing of 6 m, and section size of 2.4 m 1.6 m. The calculation model and distribution of anti-slip piles are shown in Figure 1. The distribution of the main structural planes of the slope is shown in Figure 2. The corresponding relationship between potential sliding blocks and structural planes is shown in Table 1.
Figure 1.
FLAC3D slope and anti-slip pile calculation model.
Figure 2.
Schematic diagram of slope segmentation.
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Table 1.
Slope block table.
Block number
Corresponding structural plane
1# block 2# block
Cut by sh1, j11, tensile crack j102, and fault f4 Cut from weak structural plane sh2 with clay and debris, trailing edge tensile crack j11, tensile crack j102, and fault f4 Cut from weak structural plane sh3 with debris and clay, trailing edge tensile crack j11, tensile crack j102, and fault f4
3# block
2.2
Calculation parameters
According to experimental tests, the parameters of the anti-slip pile are elastic modulus 20 GPa, Poisson’s ratio 0.2, density 2400 kg/m3, f = 1.40, c = 2.25 MPa; The physical and mechanical parameters of rock mass, tensile cracks, and weak structural planes are shown in the table below: The physical and mechanical parameters of the rock mass are shown in Table 2. The physical and mechanical parameters of the main structural planes are shown in Table 3. Table 2.
Physical parameters of rock mass testing.
Rock mass area Slightly to unweathered rock mass on the slope Slope cracking and loosening area Slope ancient landslide accumulation area Slope tension and relaxation area
Table 3.
2.3
Poisson’s ratio
Internal friction angle ( )
Deformation Cohesion/ modulus/GPa MPa
0.3
70
11
0.3 0.3 0.3
35 31 39
3.0 2.5 3.0
1 0.15 0.1 0.30
Physical parameters of tensile crack and weak structural surface tests.
Structural plane
Thickness/ Internal friction m angle ( )
Deformation modulus/ GMPa
Poisson’s ratio
Cohesion/ MPa
f4 j102 j11 sh1 sh2 sh3
0.15 0.15 0.12 0.25 0.30 0.40
0.7 0.7 0.5 1.20 1.20 1.20
0.32 0.32 0.32 0.27 0.27 0.27
0.030 0.060 0.060 0.033 0.040 0.045
24 24 24 13 15 17
Computing method
The safety factor indicates the degree of slope stability against sliding and is the ratio of the sliding force to the sliding force. To study the stability of various rock and soil mass parts, some scholars have introduced the concept of the point safety factor. According to the Mohr-Coulomb criterion, the safety factor of point shear resistance is shown in Formula (1). K¼
f sn þ c tn
(1)
where sn and tn are the normal stress and shear stress on the stress slope, c and f are the friction coefficient and cohesion of geotechnical materials. It can be seen from Formula (1) that for a stress point, there is a point safety factor corresponding to each section. In practical application, it is necessary to take the extreme 65
value of the point safety factor and find the minimum point safety factor. The section with the minimum point safety factor is called the most dangerous. The minimum point safety factor can be calculated and derived, as shown in Formula (2). pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi 2 ðc f s1 Þðc f s3 Þ K¼ (2) ðs1 s2 Þ where s1 and s3 are the maximum and minimum principal stresses, respectively. The direction angle of the most dangerous section corresponding to the minimum point safety factor (the angle between the normal direction of the most dangerous section and the maximum principal stress) meets the conditions shown in Formula (3). pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi ðc f s1 Þ tan a ¼ pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi (3) ðc f s3 Þ
3 RESULTS & DISCUSSION 3.1
Safety factor analysis of main structural surface points under natural conditions
It can be seen from Figure 3 that under natural conditions, the point safety factor of structural plane sh1 ranges from 1.05 to 1.25, and the smaller point safety factor occurs in the area near structural plane f4 and the exposed part of structural plane sh1 near j102; It can be seen from Figure 4 that the point safety factor of structural plane sh2 ranges from 1.05 to 1.35, and its
Figure 3.
Safety factor at point sh1 of the structural plane under natural conditions.
Figure 4.
Safety factor at point sh2 of the structural plane under natural conditions.
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distribution is similar to that of structural plane sh1. The smaller point safety factor occurs in the area near structural plane f4 and the exposed part of structural plane sh1 near j102; It can be seen from Figure 5 that the point safety factor of structural plane sh3 ranges from 1.10 to 1.45, and the smaller point safety factor occurs in the area near structural plane f4.
Figure 5.
3.2
Safety factor at point sh3 of the structural plane under natural conditions.
Analysis of safety factor of main structural surface points under reinforcement condition
It can be seen from Figure 6 that after the anti-slide pile reinforcement, the point safety factor of structural plane sh1 ranges from 1.20 to 2.40. Compared with natural conditions, the safety factor has been significantly improved. At the first row of piles, the safety factor reaches the maximum value of 2.40. It can be seen from Figure 7 that the point safety factor of the structural plane sh2 is 1.20–3.20 after the anti-slide pile reinforcement. The safety factor has been significantly improved compared with the natural static condition. At the first row of piles, the safety factor reaches 3.00, and at the second row of anti-slide piles, the safety factor reaches the maximum value of 3.20. It can be seen from Figure 8 that after the anti-slide pile reinforcement, the point safety factor of the structural plane sh3 ranges from 1.20 to 3.40. The safety factor has been significantly improved compared with the natural static condition. At the first row of anti-slide piles, the safety factor reaches 2.60. At
Figure 6.
Safety factor of structural plane sh1 point under reinforcement condition.
67
Figure 7.
Safety factor of structural plane sh2 point under reinforcement condition.
Figure 8.
Safety factor of structural plane sh3 point under reinforcement condition.
the second row of anti-slide piles, the safety factor reaches 3.00, and at the third row of antislide piles, the safety factor reaches 3.40. It shows that the point safety factor of the potential sliding structural plane is significantly improved after the anti-slide pile reinforcement, and the anti-slide pile improves the stability of the slope. At the same time, from the distribution law of the point safety coefficient before and after reinforcement, the distribution of the point safety coefficient after reinforcement is gentler, which shows that the anti-slide pile improves the integrity of the slope.
4 CONCLUSIONS Taking a slope project as an example, based on FLAC3D numerical simulation, this paper uses the point safety factor method to analyze the change of the safety factor of the slope’s main potential sliding structural plane before and after the anti-slide pile reinforcement. The results show that after the anti-slide pile reinforcement, the point safety factor of the 68
potential sliding structural plane is significantly improved, and the anti-slide pile obviously affects the stability of the slope reinforcement. At the same time, from the distribution law of the safety factor of the points before and after reinforcement, the distribution of the safety factor of the points after reinforcement is gentler, indicating that the anti-slide pile improves the integrity of the slope.
REFERENCES Chen, J.F., Guo, X.P., Tian, D., Yu, S.B., (2022) Study on the Influence of Anti-slide Pile Anchorage Length on the Sliding Surface and Anti-slide Capacity of Homogeneous Slope [J]. Journal of Tongji University (Natural Science),50 (01): 42–49. Dai, M.L., Li, T.C., (2007) Safety Evaluation Analysis of Complex Rock Slope Dynamic Stability Based on Numerical Calculation of Strength Reduction Method [J]. Chinese Journal of Rock Mechanics and Engineering, (S1): 2749–2754. Hou, C.Q., Ding, Y., Sun, Z.B., Li, J.F., (2022) Sensitivity Analysis of Three-dimensional Stability Factors of Slope Strengthened by Anti-slide Piles [J]. Journal of Hefei University of Technology(Natural Science),45 (08): 1092–1099. Kuang, Y., Zhang, X., Yixin Chen, Y.X., et al. (2022) Analysis of Influence of Design Parameters of Anti Slide Pile on Slope Stability, 2022 8th InternationalConference on Hydraulic and Civil Engineering: Deep SpaceIntelligent Development and Utilization Forum (ICHCE):790–794. Lin, Z.H., (2020) Research on the Reasonable Pile Spacing of Cantilever Anti-slide Piles Arching Together Behind the Pile and at the Pile Side [J]. Journal of Water Resources and Architectural Engineering,18 (03): 175–181. Liu, K., Luo, J., Chen, Q., et al. (2021) Numerical Simulation Study on the Anti-slide Pile Reinforcement Effect of the Alpine Ski Fill Track of the Beijing Winter Olympic Games [J]. Safety and Environmental Engineering,28 (04): 48–56. Ping, S.Y., Xia, Y.Y., Chen, C., Yan, Y.F., Yi, C.C., (2022) Analysis of the Influence of Different Evolution Modes of Landslide Anti-slide Pile Reinforcement Pile Position and Embedded Depth [J]. Journal of Wuhan University of Technology,44 (07): 67–74. Pan, T., (2021) Dynamic Response of Slope Under Seismic Load and Stability Analysis of Slope Strengthened by Anti-slide Piles [D]. Guangzhou: Guangzhou University. Ren, X., Luo, L.J., Li, F.T., et al. (2022) Test on the Formation and Evolution Process of Passive Soil Arch in Front of Piles in the Embedded Section of Anti-slide Piles in the Loess Area [J]. China Journal of Highway and Transport,35 (11): 86–96. Su, P.D., Zhang, H.C., Qi, Z.K., et al. (2023) Comparative Study on Soil Arch Effect of Anti-slide Piles with Different Sections [J/OL]. Journal of Disaster Prevention and Mitigation Engineering: 1–8. Zhang, Y., Dai, M.L., Kuang, Y., et al. (2010) Simulation and Application of Slope Anti-slide Pile Reinforcement Based on FLAC3D Structural Unit [J]. Water Resources and Power, 28 (12):95–97.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Experimental study on the influence of stirrup on axial compression performance of full-scale concrete square column Yunda Shao* The Open University of China, Beijing, China
ABSTRACT: In order to study the influence of stirrup on the axial compression performance of full-size concrete square columns, the axial compression test of four full-size concrete square columns was completed, and the ultimate bearing capacity and axial deformation of the specimen were measured. The influence of stirrup ratio and stirrup strength on axial compression capacity and deformation capacity of the full-size concrete square column is studied. The test results show that the stirrup ratio directly affects the bearing capacity and deformation capacity of the specimen. The higher the stirrup ratio is, the higher the bearing capacity and deformation capacity of the specimen are. High-strength stirrup can delay the decline rate of the bearing capacity of the specimen and significantly improve the deformation capacity of the specimen.
1 INTRODUCTION The axial compression of concrete columns under stirrups is the simplest and most basic stress state. (Dai 1984; Du 2010; Hamim 1980) Therefore, the study of the mechanical properties and deformation capacity of concrete columns under stirrups is the basis for the study of the mechanical properties and deformation capacity of concrete columns under a composite stress state.(Hu 1993; Li 2014; Yang 2010) This paper has completed the axial compression test of four full-size concrete square columns restrained by stirrups, observed and recorded the failure process and failure form of the specimen, studied the axial compression bearing capacity and deformation capacity of the concrete square columns restrained by stirrups, and analyzed the influence of stirrup ratio and stirrup strength on the axial compression bearing capacity and deformation capacity of the concrete square columns by comparison.(Park 1982; Sheikh 1982)
2 TEST OVERVIEW 2.1
Specimen design and production
In this test, a total of 4 reinforced concrete column specimens with stirrup constraints were completed, among which 2 were equipped with ordinary stirrups and 2 with high-strength stirrups. The height of all specimens was 1800 mm, and the section size was 600 mm 600 mm. During the design of the specimen, the main influencing factors to be considered include stirrup strength, stirrup form, stirrup diameter, longitudinal arrangement, volume stirrup ratio, etc. (Si 2010; Zhou 2007) (1) Stirrup strength: HRB400 ordinary reinforcement and HSHB1000 high-strength reinforcement are mainly used. (2) Stirrup forms mainly include type A and type B, as shown in Figure 1. *Corresponding Author: [email protected]
70
DOI: 10.1201/9781003425823-10
Figure 1.
Shapes of hoop.
(3) Layout of longitudinal reinforcement. Under the condition that the ratio of longitudinal reinforcement is equal as far as possible, the number of longitudinal reinforcement and section size of longitudinal reinforcement are different due to different stirrup forms, as shown in Table 1. Table 1.
Parameters of specimens. Longitudinal reinforcement
Stirrup Diameter/ spacing Form d (mm) (mm)
Stirrup ratio/rv
Collar characteristic value/l
1.05% 1.01%
A B
10 10
100 100
1.16% 1.74%
0.22 0.33
1.05% 1.01%
A B
9 9
100 100
0.94% 1.41%
0.52 0.77
Specimen number
Reinforcement Quantity ratio/r
Na-100(10) Nb-100(10)
12 12 8 12 12 8
Ha-100(9) Hb-100(9)
2.2
20 16 14 20 16 14
Stirrup
Mechanical properties of specimen materials
2.2.1 Reinforcement The measured elastic modulus of steel Es, yield strength fy, ultimate strength fu and yield microstrain mey are listed in Table 2. Table 2.
Mechanical properties of reinforcement.
Reinforcement type
Diameter/d(mm)
Es/MPa
fy/MPa
fu/MPa
me y
HRB400 HSHB1000
10 9
174700 188700
511 1044
641 1092
2925 5532
2.2.2 Concrete Table 3. shows the mechanical property parameters of the concrete sample. Table 3.
Mechanical properties of concrete.
Concrete type
Modulus of elasticity/Ec (MPa)
Axial compressive strength/fc (MPa)
Compressive strength/fcu (MPa)
C40
32463
32.0
47.9
71
2.3
Measurement scheme
The test was carried out in the Key Laboratory of Seismic Engineering and Structural Diagnosis and Treatment of Ministry of Education, Beijing University of Technology. The axial loading equipment was a 40000 kN multifunctional electro-hydraulic servo testing machine. During the test loading, the press transfers the load to the ball hinge, and then from the ball hinge bottom plate to the top of the specimen, which approximates the uniform load. In order to prevent local compression failure at the column end, steel plate sleeves with a height of 300 mm were arranged at both ends of the specimen. The detailed test loading device and its schematic diagram are shown in Figure 2.
Figure 2. Loading device and its schematic diagram. 1-40000 kN jack; 2 – upper ball hinge; 3 – force sensor; 4 – trolley and internal lower ball hinge; 5 – the upper and lower ends of the specimen are covered with steel plates; 6 – specimen; 7- Displacement meter.
3 DESCRIPTION OF FAILURE PROCESS AND MORPHOLOGICAL CHARACTERISTICS OF THE SPECIMEN TEST The failure course of all specimens under axial load is basically the same. From the perspective of specimen shape, it is the process of crack development, failure of the concrete protective layer, buckling of longitudinal reinforcement, and yielding or buckling of the stirrup. At the beginning of the test, when the axial load increases to a certain value, the first vertical crack appears above the cylinder surface. As the axial load increases step by step, many small cracks appear on the cylinder surface, and the original vertical cracks widen and extend. As the load continued to increase, cracks appeared in the middle and corner of the specimen, the width and length of the original cracks increased and gradually connected, some longitudinal bars buckled, and the concrete surface began to drop slag and peel. The load continued to increase again, the concrete protective layer peeled off a large number, and the specimen reached the ultimate bearing capacity; Subsequently, the axial deformation increased and the bearing capacity of the specimen decreases. Finally, the stirrup yields or breaks, losing its function of confining the core concrete, and the specimen is damaged.
Figure 3.
Breakage forms of NA-100(10).
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Figure 3 shows the cylinder failure morphology of specimen NA-100(10) at three stages: corner cracking, protective layer shedding, and test end.
4 TEST RESULTS AND ANALYSIS OF SPECIMENS Based on the results of axial compression tests of four full-size concrete columns with reinforced stirrup confinement completed in this paper, the ultimate bearing capacity, axial deformation, stress-strain curve, peak strain, and energy dissipation of concrete in the core area with stirrup confinement are discussed below. Parameters such as stirrup strain were discussed, and the influence of stirrup strength on axial compression performance was analyzed. 4.1
Ultimate bearing capacity of specimens
According to the collation and summary of test data, the main parameters of axial load and axial deformation of each specimen are given in Table 4, where Fp represents the ultimate bearing capacity of the specimen, Nu represents the axial bearing capacity of the concrete column calculated according to the measured value of the material in accordance with the normative formula, and Fp/Nu represents the ratio of the ultimate axial force of the test to the calculated bearing capacity of the specimen. The larger the value is, the more obvious the improvement of the bearing capacity of the specimen under constraints is. Up represents the peak axial deformation of concrete in the stirrup-constrained core area, U85 represents the axial deformation when the axial load of concrete in the stirrup-constrained core area drops to 85% of the peak load, and U85/Up represents the ductility coefficient of the specimen, The larger the value, the better the ductility. Table 4.
Main test results of specimens.
Specimen number
Fp(kN)
Nu(kN)
Fp/Nu
Up(mm)
U85(mm)
U85/Up
NA-100(10) NB-100(10) HA-100(9) HB-100(9)
16870 20525 17062 21780
13201 12822 13201 12822
1.27 1.60 1.29 1.70
15.13 17.49 15.80 19.49
21.51 25.05 27.41 32.43
1.42 1.43 1.73 1.66
As can be seen from Table 4, the ultimate load of the four specimens is greater than the axial compressive capacity calculated according to the standard formula. Among them, the ultimate load of two specimens with strong constraints, HB-100(9) and NB-100(10), exceeded 150% of the calculated bearing capacity, indicating that the enhancement of stirrup ratio and stirrup strength is very beneficial to the improvement of bearing capacity of concrete columns. 4.2
Stress-strain curves of concrete in the constrained core area
The stress-strain curve of concrete in the constrained core area under axial load reflects the basic mechanical characteristics of concrete under the constrained action, which is an important basis for the study of its strength and deformation. It also has a great influence on the whole process analysis of the elastoplastic process of the concrete column under the constrained action and the study of the stress distribution of the section of the concrete column under the limit state. Therefore, this paper summarized and analyzed the data results measured by the test. Figure 4 shows the stress-strain relationship curve of concrete in the constrained core area of each specimen. 73
Figure 4.
4.3
Stress-strain curve of core concrete of constraints.
Analysis of influencing factors of concrete stress-strain relationship in the confined core area
1. Influence of the stirrup ratio Figure 5(a) shows the stress-strain contrast curves of concrete in the constrained core area of specimens NA-100(10) and NB-100(10). Figure 5(b) shows the stress-strain contrast curves of concrete in the constrained core area of specimens HA-100(9) and HB-100(9). It can be seen in Figure 5(a) and (b) that the ascending section of stress-strain curves of specimens NA-100(10) and NB-100(10) almost coincide. Similarly, the ascending section of stress-strain curves of specimens HA-100(9) and HB-100(9) almost coincide, indicating that the restraint effect of the stirrup is not obvious at the initial loading stage and has little influence on the axial bearing capacity and deformation capacity of specimens. As the load continues to increase, the two curves are compared in Figure 5(a) and (b)
Figure 5.
Stress-strain curves of core concrete with a different stirrup ratio.
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gradually separate, that is, the stirrup begins to play a restrictive role. In Figure 5(a), specimen NA-100(10) with a small stirrup ratio reached the peak load before specimen NB-100(10) with a large stirrup diameter, and its corresponding peak strain was also smaller. In Figure 5(b), the peak load and peak strain of specimen HA-100(9) with a small stirrup ratio section of the curve, the stress descending rate of specimen NA-100(10) in Figure 5(a) is significantly faster than that of specimen NB-100(10), and that of specimen HA-100(9) in Figure 5(b) is also significantly faster than that of specimen HB-100(9). In summary, the test results and analysis show that the larger the stirrup ratio is, the more obvious the stirrup restraint effect is, the greater the ultimate bearing capacity of the specimen is, and the better the deformation capacity is to a certain extent. 2. The influence of stirrup strength Figure 6 shows the stress-strain contrast curves of concrete in the constrained core area with high-strength stirrup and with ordinary stirrup. Figure 6(a) shows the stress-strain contrast curves of concrete in the constrained core area of specimens NA-100(10) and HA-100(9); Figure 6(b) shows the stress-strain correlation curves of specimens NB-100 (10) and HB-100(9) in the confined core area of concrete.
Figure 6.
Stress-strain curves of core concrete with different stirrup strength.
It can be seen from Figure 6(a) and (b) that the entire ascending section of the two stressstrain curves of specimen NA-100(10) and HA-100(9) almost coincide, with little difference in peak stress. The two stress-strain curves of specimen NB-100(10) and specimen HB-100(9) almost coincide in the whole ascending section, and the peak stress of specimen NB-100(10) is slightly larger. For the descending section of the curve, the high-strength stirrup specimens HA-100(10) and HB-100(9) were much gentler than the ordinary stirrup specimens NA-100 (10) and NB-100(10), respectively. In particular, the descending section of HB-100(9) was “convex”. By comparing the two groups of specimens, it can be fully demonstrated that ordinary stirrup specimens with the same stirrup form and stirrup spacing and high-strength stirrup specimens have almost the same restraint effect before reaching the stress peak. Highstrength stirrup has no obvious effect on the improvement of the bearing capacity of specimens, while its effect on the improvement of the deformation capacity of specimens is very significant. The main reason is that when the specimens reach the ultimate bearing capacity, The stress of the high-strength stirrup is far from reaching its yield stress. However, after the peak deformation, with the increase of concrete transverse deformation, the stirrup strain increases, and the restraint effect of a high-strength stirrup becomes apparent. In addition, considering that the stirrup ratio of high-strength stirrup specimens is about 81% of that of ordinary stirrup specimens, and the stirrup characteristic value is twice that of ordinary stirrup specimens, “stirrup characteristic value” cannot be used as a single index to evaluate the constraint effect of the high-strength stirrup. 75
5 CONCLUSION In this paper, the axial compression test of four full-size concrete columns with reinforced stirrup confinement was completed, and the test results were systematically analyzed, and the following conclusions were obtained: 1. The failure process of full-size concrete column specimens constrained by stirrup under axial load is strengthened, that is, the process of crack development, concrete protective layer failure, longitudinal reinforcement buckling, stirrup yielding, or straining. 2. Strengthening stirrup restraint can significantly improve the bearing capacity and deformation capacity of full-size concrete columns. The larger the stirrup ratio, the more obvious the restraint effect of the stirrup, the larger the ultimate bearing capacity of the specimen, and the better the deformation capacity. 3. Compared with ordinary stirrup, high-strength stirrup has no obvious effect on improving the bearing capacity of specimens, but has a significant effect on improving the deformation capacity of specimens. If the constraint stress of a high-strength stirrup does not reach the yield strength, the “stirrup characteristic value” cannot be used as a single index to evaluate the constraint effect of the high-strength stirrup.
REFERENCES Dai Ziqiang, Zhang Zuguang, Chu Pengnian (1984). Experimental Study on Strength and Deformation of Confined Concrete Columns [J]. Journal of Tianjin University, (4): 16–24. Du Xiu-li, Fu Jia, Zhang Jian-wei (2010). The Experimental Study on Size Effect of the Large-size Reinforced Concrete Column Under Axial Loading [J] China Civil Engineering Journal, 43 (Suppl 1):1–8. Hamim A. Sheikh, S.M (1980). Uzumeri Strength and Ductility of Tied Concrete Columns[J]. ASCE J. Struct. Div, vol.106, NO. st5, 1079–1102. Hu Haitao, Ye Zhiman (1993). Experimental Study on Strength and Deformation of High-strength Confined Concrete Under Axial Compression [J]. Journal of Qingdao Institute of Architecture and Engineering, 14 (1): 1–8. Li Zhen-bao, Song Jia, Du Xiu-li (2014). Experimental Study on Size Effect of Compressive Response of Concrete Confined by Square Stirrups. Journal of Beijing University of Technology, 40 (2): 223–230. Park, M.J.N. Priestley, W.D. Gill (1982). Ductility of Square-confined Concrete Columns[J], Journal of the Structural Division, ASCE, 108(ST4): 929–951. Sheikh, Uzumeri (1982). Analytical Model for Concrete Confinement in Tied Columns. ASCE. Si Bingjun, Hu Zhong, Sun Zhiguo (2010). Review on the Mechanical Properties of High-strength Concrete Columns Confined by High-strength Stirrups Under Axial Compression [J]. Concrete, (11): 39–42. Yang Kun, Shi Qingxuan, Wang Qiuwei (2009). Performance Analysis of High-Strength Stirrup Confined Concrete Under Axial Compression [J]. Journal of Xi’an University of Architecture and Technology, 41 (2): 161–167. Zhou Wenfeng, Lu Ying (2007). Literature Review of Confined Concrete [J]. Sichuan Building Science Research, 33 (3): 144–168.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
The method of monitoring the health status of buildings based on Beidou Zhuang Zhu* Jilin Institute of Chemical Technology, Jilin, China Guangdong University of Petrochemical Technology, Maoming, China
Shaolin Hu* & Ye Ke* Automation School, Guangdong University of Petrochemical Technology, Maoming, China
ABSTRACT: There are many parameters for evaluating building health and safety. This paper emphasizes two parameters, the settlement and the tilt related to the overall state of the building while briefly introducing the current situation and applications of building monitoring at home and abroad. Then it compares the characteristics of existing communication networks and selects the transmission method suitable for building monitoring scenarios. Based on the sensor layout principles, we design a layout scheme and metrics about building settlement and tilt parameters to evaluate the impact of various factors on building health and to improve the risk perception capability for building disasters. 1 INTRODUCTION In recent years, first-tier cities such as Beijing, Shanghai, Guangzhou and Shenzhen, Nanjing, Suzhou, and other new first-tier cities are increasing the height of building construction or expanding the urban plate. Coupled with the old buildings of the last century in disrepair, the state of the building and related safety parameters will change at any time. The frequency of accidents caused by building safety has seriously affected people’s lives, so the issue of building health and safety is gradually becoming a concern for all sectors of the community. In addition, natural disasters are frequent. On 12 May 2008, an 8.0 magnitude earthquake struck Wenchuan. On Dacheng Road, Ningbo, Zhejiang, a 5-storey residential house of 23 years old collapsed on 4 April 2014. A residential building collapsed due to tilting in Luohu, Shenzhen, on August 28, 2019. All of them put forward strict requirements for building safety state monitoring. Then, the research and development of building safety monitoring systems came into being on this demand. It is of great practical significance to improve the risk perception ability to build construction, to achieve real-time monitoring and timely response. Direct measurements are mostly used for the monitoring of building safety. We take the tilt monitoring parameters as an example. The use of inclinometers attached to the building’s perimeter can reach an accuracy of 0.001 . But monitoring each point is completed independently, leading to disadvantages that emphasize the locality and ignore the whole. Domestic building safety monitoring technology has been put into market use. Jiang (2014) designed a building health monitoring system based on IoT technology; Ye (2020) completed the design of an automatic building monitoring system through Lora technology. There is also much to learn from foreign building monitoring technology regarding how it is technically implemented. Martinez et al. (2020) used UAV clustering technology for common tall building hazard types in their tall building monitoring program, enhancing *Corresponding Authors: [email protected], [email protected] and [email protected] DOI: 10.1201/9781003425823-11
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identifying and assessing hazards in high-rise projects. Oh & Park (2022) used convolutional neural networks to monitor the long-term structural health of buildings, thereby building an urban safety network and achieving good results. In this paper, we introduce a Beidou-based monitoring method that uses attached measurements, which has the advantages of being mobile, low-cost, and non-damaging to buildings while being effective in monitoring the overall health status of buildings.
2 COMMUNICATION NETWORK SELECTION In building monitoring, it is necessary to consider the complexity of the environment in which the application object is located. For example, multiple wall obstacles and long distances will make information transmission more difficult. Therefore, using various wireless communication methods for data transmission is more appropriate. The traditional structure monitoring system mainly uses a wired sensor network for data transmission, but the wired sensor network has a large amount of wiring and high system cost. The wireless sensor network constructed by sensor nodes equipped with a microprocessor and wireless communication module has the advantages of low price, fast construction, simple maintenance, and strong expansibility (He 2012). At the same time, wireless communication can form a network between sensor nodes in a wireless self-organizing manner. New nodes can join at any time without affecting the original network. It has good scalability and is easy to install. Currently, the mainstream wireless communication technologies are GPRS, Wi-Fi, Zigbee, Bluetooth, UWB, and LPWAN (Low Power Wide Area Network), among which the main representatives of LPWAN are NB-loT and LoRa communication technologies. LPWAN features long-range, low power requirement, low power, low cost, etc. Compared with WiFi, Bluetooth, ZigBee, and other technologies, LPWAN can achieve ultra-longrange coverage for IoT at a low cost. In Table 1, five communication technologies, such as GPRS, are introduced with relevant parameters, while Table 2 introduces two Table 1. Comparison of the characteristics of several mainstream wireless communication technologies. Wireless Communication Technology
GPRS
Frequency Band Transmission distance Transmission rate
935–960 MHz 2.4 GHz 20 km 100 m 115 kbps 11 Mbps
Network Architecture Typical Applications
Star type Star type China Mobile Home Networking
Table 2.
Wi-Fi
Zigbee
Bluetooth
UWB
2.4 GHz 10–100 m 20–250 kbps Star type Sensor Networks
2.4 GHz 10-20 m 1 Mbps
3.1–10.6 GHz Under 10 m 400 Mbps
Star type Voice Transmission
Star type Indoor positioning
Comparison of the characteristics of NB-loT and LoRa technologies.
LPWAN
NB-loT
LoRa
Technical Features Network Deployment
Honeycomb Multiplexing with existing cellular base stations Carrier Authorized Bands Long Distance
1 x1;j ðtk Þ þ x3;j ðtk Þ x2;j ðtk Þ þ x4;j ðtk Þ 1X > > x j ðtk Þ ¼ þ xi;j ðtk Þ ¼ > > 2 2 2 4 i¼1 > > > > < 4 1 y1;j ðtk Þ þ y3;j ðtk Þ y2;j ðtk Þ þ y4;j ðtk Þ 1X (2) þ ¼ yi;j ðtk Þ ðj ¼ 0; 1; 2Þ yj ðtk Þ ¼ > 2 2 2 4 i¼1 > > > > 4 > 1 z1;j ðtk Þ þ z3;j ðtk Þ z2;j ðtk Þ þ z4;j ðtk Þ 1X > > > þ z ðt Þ ¼ zi;j ðtk Þ ¼ : j k 2 2 2 4 i¼1
Settlement is the vertical displacement that occurs in the Z-axis, but the settlement of the building is mostly uneven. The 12 sensors that have been placed on the perimeter of the building and the reference points outside the building are differenced in Z-axis quantities to produce 12 relative quantities to assess the uneven settlement of the building. In addition, excessive uneven settlement of the foundation can cause the building to tilt. To distinguish the two cases of settlement and tilting, the difference in Z-axis quantities between the three central coordinates 1-2, 2-3, and 1-3 is introduced. The values of these three central coordinates are recorded when the building is in a healthy state. If the difference
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in the central coordinate changes, the type of disaster occurring in the building is tilting, and if it remains unchanged, it is the settlement.
6 CONCLUSION AND OUTLOOK The development trend of urban building safety management is using sensor technology and Beidou navigation technology to monitor and warn about building health and avoid the risk of building collapse. The research on the monitoring method and parameter metrics design of the parameters related to the overall state of the building can make the monitoring effect more comprehensive. Building monitoring will expand the monitoring content from more perspectives, from point to surface, from part to whole. At the same time, combined with artificial intelligence, data mining, and other technologies, it is integrated into the remote cloud platform for development, constantly meeting the market’s needs and realizing and improving more functions.
ACKNOWLEDGEMENT Guangdong University of Petrochemical Technology High-level Talent Scientific Research Start-up Fund Project.
REFERENCES Fang Qian-qian & Hang Xiao-Jian. (2017). Study on Dangerous Building Health Monitoring System Based on Cloud Platform. Jiangsu Architecture (04), 57–60. He Pei. (2012). Research of Building Structure Monitoring System Based on Wireless Sensor Network (Master thesis, Jiangnan University). Jian Xia. (2022). Application Research of Railway Slope Monitoring Based on Beidou Positioning Technology. Automation and Instrumentation, (09)2022: 89–92. DOI: 10.14016/j.cnki.1001–9227.2022.09.089. Jiang Shuai. (2014). Research and Design of Building Health Monitoring System Based on Internet of Things Technology (Master thesis, Changan University). Jourdan, D. B., & de Weck, O. L. (2004, May). Layout Optimization for a Wireless Sensor Network Using a Multi-objective Genetic Algorithm. In 2004 IEEE 59th Vehicular Technology Conference. VTC 2004Spring (IEEE Cat. No. 04CH37514) (Vol. 5, pp. 2466–2470). IEEE. Martinez, J. G., Gheisari, M., & Alarcón, L. F. (2020). UAV Integration in Current Construction Safety Planning and Monitoring Processes: Case Study of a High-rise Building Construction Project in Chile. Journal of Management in Engineering, 36(3), 05020005. Oh, B. K., & Park, H. S. (2022). Urban Safety Network for Long-term Structural Health Monitoring of Buildings Using Convolutional Neural Network. Automation in Construction, 137, 104225. Wu, C. H., Su, W. H., & Ho, Y. W. (2010). A study on GPS GDOP Approximation Using Support-vector Machines. IEEE Transactions on Instrumentation and Measurement, 60(1), 137–145. Xiang Jianliang. (2019). Design of Building Environment Monitoring System Based on LoRa Wireless Sensor Network (master’s thesis, Nanjing University of Technology). DOI: 10.27241/d.cnki.gnjgu.2019.000525. Ye Yi-Ping. (2020). Design of Building Automatic Monitoring System Based on LoRa (master’s thesis, Shenzhen University). DOI: 10.27321/d.cnki.gszdu.2020.000834.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Based on Beidou navigation satellite system bridge variation monitoring analysis and research Jie Zhang* Guangdong University of Petrochemical Technology, Maoming, China and Jilin Institute of Chemical Technology, Jilin, China
Shaolin Hu* & Ye Ke* Automation School, Guangdong University of Petrochemical Technology, Maoming, China
ABSTRACT: Bridge monitoring plays an important role in ensuring the safe operation of bridges and improving the efficiency of bridge maintenance. Based on the theory of bridge health monitoring, combined with the structural characteristics of long-span continuous beam bridges, this paper studies the variation of the bridge. Aiming at the problem that the bridge tilt, deflection, vibration, and other indicators are not easy to measure, a method of long-term monitoring of bridges using multiple BDS devices is proposed. The bridge data was obtained by BDS device, and the index information of the bridge was obtained by further analysis. It provides some reference for effectively improving the level of bridge information monitoring and provides effective data support for the subsequent maintenance and repair of bridges.
1 INTRODUCTION Bridges occupy an important proportion of the transportation system and play an important role in ensuring smooth traffic and ensuring the stable development of the national economy (Ge et al. 2022). The bridge is accompanied by the increase of traffic pressure and the influence of natural factors such as temperature, humidity, earthquake, flood, and other natural factors during the use of the bridge, which increases the dynamic response of the bridge, causes the vibration of the bridge to increase, the damage of the bridge structure, and the strength of the bridge surface will also be reduced, resulting in a significant reduction in the actual service life of the bridge, and even lead to sudden collapse. Therefore, to ensure the safety of bridge traffic, it is necessary to effectively monitor the bridge’s state. With the gradual development of measurement technology, various measurement equipment is constantly updated, and the monitoring method of bridge variation increases. The conventional geodetic method generally uses theodolite, level, and other measuring equipment to obtain the deformation of the measured object by measuring angle, level, and other methods. Most of them collect data manually, and the workload is large. The physical sensor method mainly installs sensors such as strain gauges, displacement meters, and accelerometers to monitor the state of the bridge. It can only be used to monitor the local deformation of the bridge and is insufficient for the overall monitoring of the bridge. Commonly used vibration frequency detection tools, such as accelerometers, need to integrate and calculate vibration parameters, which cannot measure long-period quasi-static displacement (Nan et al. 2019), the commonly used tilt monitoring tools are inclinometers, *Corresponding Authors: [email protected], [email protected] and [email protected]
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DOI: 10.1201/9781003425823-12
which need to be embedded in the surface of the bridge and the measurement accuracy is not high. Therefore, researchers have developed a large number of monitoring equipment, such as robot total station (Panos et al. 2007), photogrammetry equipment (Jahanshahi et al. 2013), 3D laser scanner (Zhao et al. 2015), GNSS (Zhao et al. 2021), synthetic aperture radar (Liu et al. 2021), etc. to monitor structural deformation. Due to the increase in sight distance, the accuracy of photogrammetry and laser scanning methods will decrease. Synthetic aperture radar can only measure the deformation in the line of sight direction, and the deformation measurement in the vertical and line of sight direction is lacking. However, with the continuous development of satellite navigation technology, BDS has shown more and more unique advantages in bridge deformation monitoring. As early as 2000, the British scholar Roberts et al. (2000) placed four GPS receivers and a GPS / GLONASS receiver on the Humber bridge deck, placed a GPS receiver on a truck, and placed a reference receiver on land. The receiver collects data at a frequency of 5Hz and combines the finite element model to monitor the deflection deformation of the bridge, which is beneficial to calibrate the finite element model to optimize the bridge design. In 2020, Lin et al. (2020) designed and measured the network of the first-level GNSS control network of the Taizhou Bay Bridge and the connection project, and solved the longdistance cross-sea bridge and the construction of the connection project. The demand for the first-level control network verifies that GNSS technology is a reliable method for bridge monitoring, but there are still deficiencies in the calculation of specific monitoring indicators. In this paper, a single-hole bridge is taken as the object, and multiple BDS devices are used to monitor multiple points on the bridge. The coordinate data of the points are collected for related calculation and processing. Finally, the bridge is monitored from the indexes of bridge inclination, vibration, and deflection. Compared with the original method, the use of BDS equipment can monitor the external deformation of the bridge around the clock, with high measurement accuracy, high sampling frequency, high automation level, and the ability to provide coordinate measurement. Moreover, its continuous observation ability can grasp the running state of the bridge at all times. When the bridge changes abnormally, it can be found and processed in time, thus ensuring the safety of the bridge and providing reliable data for future maintenance work.
2 BDS WORKING PRINCIPLE The satellite positioning system is a technology that uses satellites to accurately locate something. It realizes the functions of navigation, positioning, and timing at any time and any point on the Earth as long as four satellites can be observed at the same time. The principle of satellite positioning is mostly a three-ball rendezvous measuring principle (Song et al. 2018). However, the Beidou positioning principle is that the ground monitoring station continuously monitors all visible satellites of the navigation receiver, generates pseudo-range and carrier observation information, and sends the original data to the ground control station after preprocessing. The ground master station verifies and evaluates the original data, calculates the satellite orbit and clock correction, generates the correction and other related parameter enhancement information according to the protocol, and transmits it to the high-orbit satellite by the uplink station. The high-orbit satellite broadcasts through the PPP-B2b (Xu et al. 2021) line, and the user can perform real-time precise point positioning after receiving the correction information. When BDS is used for bridge deformation monitoring, some means are needed to improve the measurement accuracy due to the large positioning error of equipment and other factors. The main technical method is RTK, and the commonly used RTK model is the double difference model (Jin et al. 2020). As shown in Figure 1, two receivers (a base station and a rover) are observing satellite data, and the base station transmits the received carrier phase signal (or carrier phase difference correction signal) through its transmitting station; the mobile station receives the radio signal of the base station through its receiving station while receiving the satellite signal. Based on these two signals, the solidification software on the 85
Figure 1.
The basic principle of RTK.
mobile station can realize differential calculations, to accurately determine the spatial relative position relationship between the reference station and the mobile station. Nonetheless, in this process, due to the influence of observation conditions, signal sources, etc., there will inevitably be some errors. 3 MONITORING SCHEME This time, a total of 11 BDS devices were used, of which 10 were used as monitoring stations and 1 as reference stations. The single point positioning accuracy error of a single universal GNSS receiver device is about 3 meters, the static accuracy error is about 5 mm, the RTK accuracy error is about 15 mm, and the wireless communication is rich, supporting 4 G, NBIoT, BT, LoRa and other communications. The data update frequency can be selected as 1 Hz, 2 Hz, 5 Hz, and 10 Hz. 3.1
Communication network selection
Most of the traditional embedded sensors use wired cables for data transmission, which has strong anti-interference, high stability, and certain confidentiality. However, wired communication is often greatly affected by the environment, with weak scalability and poor mobility. In reality, the bridge environment is complex and often passes by vehicles and pedestrians. The use of wired transmission may have an impact on traffic, which is not conducive to the development of this study. Therefore, wireless communication is used for data transmission. The first is the wireless communication network selection of the bridge monitoring site. 11 devices are used as monitoring stations on the bridge deck, and 1 device is used as a reference station on the land near the bridge. The distance between the monitoring station and the reference station is within 1 km. The LoRa mode is selected for communication, and the 4 G mode is selected for communication over 1 km, as shown in Figure 2. The wireless transmission also has good expansibility and adaptability. In the later stage, additional monitoring equipment can be added according to the need to access the communication network well. 3.2
Layout location selection
The deformation monitoring equipment is generally placed in the characteristic parts of the deformation body to observe the position change of the monitoring mark. Therefore, the selection of monitoring sites will reflect whether the trend of bridge changes over time can be effectively monitored. According to the general idea of bridge variation monitoring, aiming at the monitoring of bridge inclination, deflection, vibration index, and the environment of the 86
Figure 2.
Bridge monitoring communication network.
monitoring area, it is required that the point layout must meet the requirements of safety, feasibility, reasonable layout, prominent key points, non-destructive installation, and can meet the accuracy requirements, to facilitate long-term monitoring. Therefore, this paper chooses to symmetrically place monitoring points on the midpoint and sides of the bridge deck, onequarter of the two sides and one-eighth of the two sides, a total of 11 monitoring stations, as shown in Figure 3. The built-in power supply of the receiver can meet long-term monitoring. If the receiver is placed at the corresponding point, it will not damage the bridge structure to achieve non-destructive installation, and will not affect the reasonable layout of the traffic. Among them, one-quarter and one-eighth of the receivers are used to monitor the bridge tilt index; a quarter, an eighth, and a midpoint receiver are used to monitor the bridge deflection index; all receivers can be used to monitor bridge vibration indicators. The reference station is required to be established in a place where the foundation is stable. At the same time, it is also necessary to meet a location that is conducive to safe operation, easy to preserve for a long time, has no high-power wireless emission source within 200 meters around, and no large building occlusion. Therefore, the roof of a building near the bridge is selected.
Figure 3.
Navigation receiver placement point.
4 DETERMINATION METHOD OF KEY INDICATORS 4.1
Tilt determination method
Tilt refers to the ratio of the settlement difference in the tilt direction of the two ends of the foundation to its distance. In the bridge, the skew angle of the bridge generally refers to the angle between the bridge axis of the plate bridge and the vertical line of the supporting line. 87
This indicator uses four BDS receivers (A, B, C, D) to be placed at the corners at both ends of the bridge deck. The receiver uses the CGCS2000 coordinate system, and the coordinate origin is the mass center of the entire earth including the ocean and atmosphere. The spatial coordinates of the receiver at any time are Aðx1 ðt! k Þ; y! 1 ðtk Þ; z1 ðtk ÞÞ, Bðx2 ðtk Þ; y2 ðtk Þ; z2 ðtk ÞÞ, C ðx3 ðtk Þ; y3 ðtk Þ; z3 ðtk ÞÞ, Dðx4 ðtk Þ; y4 ðtk Þ; z4 ðtk ÞÞ, d1 , d2 using the position information of the four receivers, and are two vectors of the non-collinear bridge deck, as shown in Figure 4.
Figure 4.
Analysis of bridge inclination measurement.
( ! d1 ðtk Þ ¼ ðx1 ðtk Þ x3 ðtk Þ; y1 ðtk Þ y3 ðtk Þ; z1 ðtk Þ z3 ðtk ÞÞ ! d1 ðtk Þ ¼ ðx2 ðtk Þ x4 ðtk Þ; y2 ðtk Þ y4 ðtk Þ; z2 ðtk Þ z4 ðtk ÞÞ ! ! ! We use l to represent the normal vector of the surface of d1 , d2 : y1 ðtk Þ y3 ðtk Þ z1 ðtk Þ z3 ðtk Þ x1 ðtk Þ x3 ðtk Þ z1 ðtk Þ z3 ðtk Þ ! ; l ð tk Þ ¼ ; y2 ðtk Þ y4 ðtk Þ z2 ðtk Þ z4 ðtk Þ x2 ðtk Þ x4 ðtk Þ z2 ðtk Þ z4 ðtk Þ ! x 1 ð tk Þ x 3 ð tk Þ y 1 ð tk Þ y 3 ð tk Þ x 2 ð tk Þ x 4 ð tk Þ y 2 ð tk Þ y 4 ð tk Þ :
(1)
(2)
The inclination angle qðtk Þ of the bridge deck at a time tk is calculated by the following formula: 1 0 x1 ðtk Þ x3 ðtk Þ y1 ðtk Þ y3 ðtk Þ B x2 ðtk Þ x4 ðtk Þ y2 ðtk Þ y4 ðtk Þ C C: ! qðtk Þ ¼ arccos B (3) @ A l ðtk Þ
4.2
Measuring method of deflection
Bridge deflection refers to the linear displacement of the center of the bridge cross-section along the vertical direction of the axis when affected by factors such as the bridge’s load-bearing, bridge deck moving load and ambient temperature (Xie et al. 2021). The short-term deformation of the bridge is called deflection deformation. The short-term deformation can be restored to the state before the external force is applied when the external force disappears. Short-term deformation can be divided into static and dynamic. The static component is a slow change close to static caused by factors such as temperature, driving load, and continuous wind. The dynamic component is the small vibration generated by the bridge structure under the excitation of environmental factors such as earthquakes, wind, and driving load. This method is mainly used for the static deflection deformation of the bridge. The spatial coordinates of the bridge intermediate receiver at any tk time are E ðx5 ðtk Þ; y5 ðtk Þ; z5 ðtk ÞÞ;and the center coordinates of the plane formed by A, B, C, and D are M ðx6 ðtk Þ; y6 ðtk Þ; z6 ðtk ÞÞ, as shown in Figure 5. 88
Figure 5.
Analysis of the principle of bridge deflection measurement.
The deflection b at the time tk : x6 ðtk Þ ¼ 21 21 ðx1 ðtk Þ þ x3 ðtk ÞÞ þ 21 ðx2 ðtk Þ þ x4 ðtk ÞÞ
(4)
y6 ðtk Þ ¼ 21 21 ðy1 ðtk Þ þ y3 ðtk ÞÞ þ 21 ðy2 ðtk Þ þ y4 ðtk ÞÞ
(5)
z6 ðtk Þ ¼ 21 21 ðz1 ðtk Þ þ z3 ðtk ÞÞ þ 21 ðz2 ðtk Þ þ z4 ðtk ÞÞ
(6)
h i b ¼ ððx5 ðtk Þ x6 ðtk ÞÞ2 þ ððy5 ðtk Þ y6 ðtk ÞÞ2 þ ððz5 ðtk Þ z6 ðtk ÞÞ2 Þ :
(7)
4.3
Vibration measurement method
Bridge vibration is a kind of finite amplitude vibration, which generally does not lead to the instability of the bridge. However, due to the occurrence of vibration, the local fatigue failure of the bridge will occur, which seriously endangers the driving safety of the bridge deck. Through the monitoring of bridge vibration frequency, it can provide a reliable scientific basis for determining the running state of the bridge. Firstly, there is no load on the bridge at the time t0 to cause vibration, and the coordinate information of the navigation receiver at the time t0 is recorded. When the bridge vibrates at the time tk , the coordinates of the navigation receiver are recorded, and then the coordinates of the receiver at a time tk and the coordinates at a time t0 are subtracted to obtain the longitudinal, lateral, and vertical displacement Sx ðtk Þ, Sy ðtk Þ, Sz ðtk Þ of the bridge in unit time. Taking the vertical displacement Sz ðtk Þ as an example, the displacement data Sz ðtk Þ is Fourier transformed, and the time domain signal is transformed into the frequency domain signal to obtain the vibration spectrum of the bridge. The amplitude Hz ðpÞ and frequency Fz ðpÞ of the point are determined according to the value Xz ðpÞ of the spectrum. Finally, the frequency corresponding to the maximum value of Xz ðpÞ is taken as the vibration frequency of the bridge, where the sampling frequency is fs, the frequency domain number is p, the sampling length is T, and the number of sampling points is N. T1 X
Sz ðtk Þej2ppk=T
(8)
Hz ðpÞ ¼ jXz ðpÞj
2 N
(9)
Fz ðpÞ ¼ ðp 1Þ
fs N
(10)
Xz ðpÞ ¼ DFT ½Sz ðtk Þ ¼
k¼0
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5 CONCLUSIONS In order to solve the problem of bridge safety monitoring and ensure the safe and stable operation of the bridge, this paper uses BDS navigation receiver positioning technology with a wireless transmission network to achieve effective monitoring of bridge tilt, deflection, vibration, and other abnormal conditions without affecting the use of the bridge. At the same time, this method processes the bridge data through the collaborative calculation of multiple navigation receivers, which makes up for the deficiency that some sensors can only monitor the local changes of the bridge. When abnormal changes occur in the monitoring bridge, this method can detect hidden dangers in time, remind relevant personnel to take effective maintenance measures to avoid accidents and provide a certain reference value for other large-scale infrastructure change monitoring.
ACKNOWLEDGMENT Guangdong University of Petrochemical Technology High-level Talents Scientific Research Start-up Fund Project.
REFERENCES Ge Guolian, Liu Lei. (2022). Research on lnvestment Calculation and Construction of Rural Public Infrastructure in China under the Background of Rural Revitalization. Issues in Agricultural Economy (10),133–144. doi:10.13246/j.cnki.iae.2022.10.010. Jahanshahi, M. R., & Masri, S. F. (2013). A New Methodology for Non-contact Accurate Crack Width Measurement Through Photogrammetry for Automated Structural Safety Evaluation. Smart Materials and Structures, 22(3), 035019. Jin Jianjian, Gao Chengfa, Zhang Ruicheng, Wang Bo. (2020). Accuracy Analysis of Short Baseline Calculation for GPS and BDS2, BDS3 Fusion Data. Bulletin of Surveying and Mapping (03),83–86+95. doi:10.13474/ j.cnki.11-2246.2020.0083. Lin Yan, et al. (2020). The Design and Survey of ‘the First-level GNSS Control Network of Taizhou Bay Bridge and Wiring Project. Journal of Changchun Institute of Technology (Natural Sciences Edition) (02),50–53+68. Liu Zhiping, Luo Xiang, He Xiufeng. (2021). Vibration Detection of High-speed Railway Bridge Using Millimeter Wave Radar Measurement System. Journal of Tongji University (Natural Science)(04),561–568. Nan Shen, Liang Chen, Jingbin Liu... & Ruizhi Chen. (2019). A Review of Global Navigation Satellite System (GNSS)-Based Dynamic Monitoring Technologies for Structural Health Monitoring. Remote Sensing (9). doi:10.3390/rs11091001. Panos A. Psimoulis & Stathis C. Stiros. (2007). Measurement of Deflections and of Oscillation Frequencies of Engineering Structures Using Robotic Theodolites (RTS). Engineering Structures (12). doi:10.1016/j. engstruct.2007.09.006. Roberts, G. W., Dodson, A. H., Brown, C. J., Karunar, R., & Evans, A. (2000). Monitoring the Height Deflections of the Humber Bridge by GPS, GLONASS, and Finite Element Modelling. In Geodesy Beyond 2000: The Challenges of the First Decade IAG General Assembly Birmingham, July 19–30, 1999 (pp. 355– 360). Springer Berlin Heidelberg. Song Peishuai, Ma Jing, Ma Zhe, Zhang Shuyuan, Si Chaowei, Han Guowei... & Wang Xiaodong. (2018). Research and Development Status of Quantum Navigation Technology. Laser and Optoelectronics Progress (09),29–43. Wang, X., Zhao, Q., Xi, R., Li, C., & Li, G. (2021). Review of Bridge Structural Health Monitoring Based on GNSS: From Displacement Monitoring to Dynamic Characteristic Identification. IEEE Access, 9, 80043–80065. Xie Chong-hong, Zhang Yan-chang, Li Wen. (2021). Research on Bridge Structural Health Monitoring Based on Wireless Sensor Technology. Intelling Building and Smart City (11),172–173. doi:10.13655/j.cnki.ibci.2021.11.079. Xu Yangyin, Yang Yuanxi & Li Jinlong.(2021).Performance Evaluation of BDS-3 PPP-B2b Precise Point Positioning Service. GPS Solutions (4). doi:10.1007/S10291-021-01175-2. Xuefeng Zhao, Hao Liu, Yan Yu... & Jingping Ou. (2015). Bridge Displacement Monitoring Method Based on Laser Projection-Sensing Technology. Sensors (4). doi:10.3390/s150408444.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Numerical simulation of reinforced concrete beam four-point bending test based on dual-particle Peridynamics Yuanze Xu* Undergraduate Student at Shanghai University, Shanghai University, Shanghai, China
Zili Dai* Associate Professor, Department of Civil Engineering, Shanghai University, Shanghai, China
ABSTRACT: Peridynamics is a relatively new numerical calculation method to describe material properties, which can well avoid some discontinuity problems. This paper outlines the basic idea and calculation method of the Peridynamics method. Based on the idea of Peridynamics, the Peridynamics model and the calculation program of the four-point bending test of ordinary concrete beams and reinforced concrete beams are prepared on the computer with the C++ program, which reveals the crack extension process and final damage mode of concrete beams. The results show that the Peridynamics model can accurately and effectively solve the buckling problem of concrete beams.
1 INTRODUCTION Concrete is a composite material formed by mixing aggregates of varying particle sizes and cementing agents and then hardening. However, its engineering applications are limited because most forms of damage of plain concrete are brittle. Since Monier invented reinforced concrete in 1849, its properties have been greatly improved, and concrete engineering applications have been enriched. As it is difficult to control the location and size of the aggregate, it is particularly tough to use traditional mathematical methods to study the location, direction, and form of cracking during damage. Therefore, it is necessary to construct a reasonable model. Traditional finite element methods such as FEM and FDM (Farahani & Taghaddos 2020; Gerstle & Xie 1992; Ibrahimbegovic et al. 2010; Zou et al. 2021; Zárate et al. 2018) are macroscopic methods based on the theory of continuous medium mechanics. They need to preset the crack’s location and size and define its fracture criterion to determine the direction of crack expansion. After the preset crack expansion, they must re-divide the mesh with a specific mesh dependence. Even if the meshless method is proposed at a later stage to get rid of the mesh constraint, the bottleneck of computational accuracy and computational efficiency is still challenging to break through for complex three-dimensional problems and the intersection of multiple cracks, which is due to the need to solve the spatial differential equations to satisfy the continuity condition in the analysis process. However, the material is inevitably accompanied by the discontinuity feature when the damage produces cracks. Thus, the computational accuracy will be affected. Peridynamics (2000) has the advantages of both molecular dynamics and meshless methods and breaks the limitations of molecular dynamics in terms of computational scales while avoiding the singularity problems arising from the meshless method at the development of the *Corresponding Authors: [email protected] and [email protected] DOI: 10.1201/9781003425823-13
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rift. It has received much attention from scholars worldwide because of its high accuracy and efficiency in solving macroscopic and microscopic discontinuity problems. Peridynamics has yielded good research results in studying many materials and structures. Based on his proposed Peridynamics model, Gerstle et al. (2007) proposed a micropolar model and used its invisible solution to solve the concrete structural damage and cracking problem effectively; Yaghoobi & Chorzepa (2017) used the micropolar model and developed a model applicable to fiber-cemented materials. They developed a model applicable to fibercemented materials using a micropolar model and demonstrated the accuracy of the fracture analysis model; Zhao et al. (2020) applied a partially homogenized stochastic Peridynamics model to study the fracture mode of concrete after reinforcement corrosion; Tuniki (2012) developed a Peridynamics concrete constitutive model and verified that this principal structure model could be well applied to concrete axial force problems; Wu & Huang (2022) developed a plastic constitutive model in the framework of Peridynamics, which considered the inhomogeneity inside the concrete and depicted the complex dynamic mechanical response of concrete under high-speed impact; Xia et al. (2021) applied the micropolar model to the modeling of reinforced concrete and simulated good results. Most of the studies mentioned above are based on plain concrete itself, so only one particle of concrete particles is set, which has high requirements on the material properties of the research object itself and does not apply to the study of composite materials such as reinforced concrete. Therefore, based on the previous study, a PD model consisting of concrete and steel reinforcement as two different particle units are developed in this paper. Regarding the parameter settings in the literature (Dai et al. 2021; Gu et al. 2016; Lu 2019; Shen et al. 2012; Zhao et al. 2018), the computational program is compiled based on C++ language and simulated for four-point bending tests of plain concrete beams and reinforced concrete beams to compare the effect of adding steel particles on the damage of concrete and whether the simulation results match with the actual experimental results.
2 PD THEORY INTRODUCTION 2.1
Basic idea and basic model construction
Figure 1.
Material points and horizon.
Based on bond-based Peridynamics, we discretize the entire space into n material points with material physical properties, each of which can have different particle sizes. At any point in time t, each material point will generate an interaction force f of equal magnitude and opposite direction with any particle j within its field of approach. f is defined in Peridynamics as a function of the relative position vector and the relative displacement vector, satisfying both momentum conservation and angular momentum conservation so that that f can be
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expressed as: fðuðj; tÞ uði; tÞ; xðj; tÞ xði; tÞÞ After substituting u for uðj; tÞ uði; tÞ and x for xðj; tÞ xði; tÞ, we get: fðu; xÞ:
Figure 2.
The relative position and relative displacement between material points.
Also, according to Newton’s second law, we can obtain that: r
@ 2 uðx; tÞ ¼ @t2
ð fðu; xÞdVj þ bðx; tÞ Rx
In the above equation, r is the density of the material point material; b is the external body force vector; Rx is the domain of integration in the near field range. Mathematically the equation is written as: Ri ¼ fxðj; tÞ 2 R; kxðj; tÞ xði; tÞk dg For the bond-based Peridynamics model, it is essential to find paired force functions. For micro elastic materials, any pair of force functions is a pair of vectors parallel to u þ x, so we can conclude that: fðu; xÞ ¼ mcðdÞsðu; xÞ
uþx ku þ xk
In the above equation, cðdÞ represents the micro elastic modulus function, which can be derived from the strain energy equivalence principle and expressed by E and n as follows: cðdÞ ¼
3E pd3 ð1 nÞ
sðn; xÞ is expressed as the elongation of the particle to spacing: sðn; xÞ ¼
kn þ xk kxk kxk 93
where m is used as a judgment indicator to measure whether the bond is broken or not and can be given by the following equation: 1; s < s0 ; m¼ 0; other: When s > s0 , we usually consider that the bond between the particle pairs breaks, and this break is irreversible. After the break, we consider that there is no longer a force between the particle pairs, so we set m to 0. 2.2
Model construction
In this paper, we simulate the plain and reinforced concrete models. The model of plain concrete is simple and consists of only one kind of particle, while the reinforced concrete model requires specific requirements for particle size. We need to take the particle size as an approximation of the protective layer thickness to make the coordinates of the rebar particles coincide with the coordinates of the center of mass of the corresponding part of the rebar in the actual case. The model has two types of particles: reinforcing steel and concrete. Three types of particle pairs are concrete-concrete, concrete-rebar, and rebarrebar. It should be noted that the ultimate tensile strains of these three particle pairs are not the same. In this paper, according to the literature (Jiao & Chen 2022; Ministry of Housing and Urban-Rural Development of the People’s Republic of China 2015; Kilic 2008; Wang 2005), we take s0 of concrete-concrete particle pairs as 0.0001, s0 of steelrebar particle pairs as 0.01, and s0 of concrete-rebar particle pairs as 0.00035. Finally, structural damage is dynamic, but we will introduce artificial damping to approximate the quasi-static process loading for the model calculation needs. We refer to the literature to select artificial damping [18]. However, unlike what is described in the literature, there is no diagonal stiffness matrix for the reinforced concrete model, so we choose n times the modulus of elasticity as the artificial damping.
Figure 3.
Sketch of the model in this paper.
Figure 3 is a more straightforward example of the model in this paper, labeled with the three particle pairs used in the model. The particle pairs in the figure are simple, and the actual model is more complex than the figure in terms of positioning and formation of the particle pairs, requiring the pairing and classification of all particles in the horizon of a particular particle according to the theory in the basic model. In most cases, the size of the horizon cannot allow all its particles to be completely encompassed within the horizon. As shown in Figure 4, some of the edge particles will be partly inside and outside the horizon. Therefore, we need to apply a volume correction to these particles to obtain more accurate results in the calculation process, and the correction
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Figure 4.
Sketch of edge particles and internal particles.
formula is as follows. 8 < Vj ; kxk d 0:5Dx; Vj ¼ ½ðd þ 0:5Dx kxkÞ=DxVj ; d 0:5Dx kxk d : 0; other: In the above equation, x can represents the radius or edge length of the particle, which needs to be flexibly adjusted according to the needs of the actual model. To better distinguish the three particle pairs and their corresponding different parameters and to facilitate the calculation, we label the different particles in the program and apply three arrays to count the number of different particle pairs. The detailed program flow chart is as follows.
3 BENCHMARK PROBLEM Figure 6 shows the standard concrete test block strength test model. The side length of the concrete test block is 150 mm, the strength grade is C30, and the load Load is a slowly linear increasing force. Through a small quantity of restriction on the upper and lower surfaces, we simulate the actual situation of the standard concrete test Cyclo-hoop effect and use this to verify the feasibility of the model in this paper.
Figure 5.
Loading model sketch.
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Figure 6.
Model simulation results in this paper.
Figure 6 indicates the degree of concrete damage. The redder the color represents, the deeper the damage is. The model in this paper can reasonably simulate the Cyclo-hoop effect and the subsequent possible conical damage characteristics of the standard concrete specimens during the strength test.
Figure 7.
Stress-strain curves.
The stress-strain curve of the actual standard concrete block in Figure 7 can be well seen, and the model in this paper can reasonably simulate the stress-strain relationship during the loading process of the standard concrete block. The Pearson coefficient will be used to analyze Figure 7 quantitatively. For quantitative analysis of strains in the range of 0 to 0.002, 50 sets of data with different strains are taken out, and the strain difference between every two adjacent data sets is the same. Assuming that the vector X fx1 ; x2 ; x3 ; . . .; x50 g represents the stress values calculated in this paper, and the vector Y fy1 ; y2 ; y3 ; . . .; y50 g is the stress value computed using the FPC2D method, the Pearson correlation coefficient of these two vectors can be calculated as: n P
ðxi xÞðyi yÞ i¼1 CorrðX ; Y Þ ¼ sffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi : n n P P ðxi xÞ2 ðyi yÞ2 i¼1
i¼1
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The Pearson coefficient calculated by the above equation is 0.997, proving that the results obtained by the method in this paper and those obtained by using the PFC2D method are strongly correlated. Therefore, the model in this paper is feasible. In the following, the four-point bending condition of the concrete beam will be simulated by the model of this paper.
4 RC FOUR-POINT BEND SIMULATION 4.1
Model basic information introduction
A two-dimensional plain concrete and reinforced concrete supported beam model is shown in Figures 8 and 9. The beam has a clear span of 2.1 m and a height of 0.2 m. Concrete is assumed to be a homogeneous, homogeneous, and brittle elastic material. The reinforcement is arranged at the bottom of the beam with a protective layer thickness of 30 mm, material point spacing of 6 mm, and a total of 11550 material points. The boundary conditions are realized by constraining the displacement of the corresponding material points. The remaining parameters are shown in the following table.
Table 1.
Parameters of concrete particles.
Type of concrete
Modulus of elasticity
Poisson ratio
Density
C30
30 GPa
0.18
2500 Kg/m3
Table 2.
Parameters of rebar particles.
Type of rebar
Modulus of elasticity
Poisson ratio
Density
HRB400
200 GPa
0.33
7850 Kg/m3
Figure 8.
Schematic diagram of plain concrete model.
Figure 9.
Schematic diagram of reinforced concrete model.
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4.2
Explanation of the model running process
In setting the boundary conditions, we simulate the actual situation as much as possible and simplify the complexity of the model. The support points were set to three particles side by side without displacement or rotation during loading. The loading points were also set to three particles side by side without rotation during the loading process. Also, to simulate a quasi-static model loading process, N times the modulus of elasticity of concrete is introduced as artificial damping. N needs to be adjusted according to the model and load, an external force, to be as close as possible to the critical damping state.
Figure 10.
4.3
Four-point bending simulation of plain concrete beam.
Image analysis of model results
Figure 10 shows a schematic diagram of the crack expansion process during the damage of the plain concrete beam. It can be more obviously seen that firstly, the damage appears at the ends of the supports, then distributed cracks appear at the bottom of the beam. When the cracks expand to a certain degree, prominent brittle damage characteristics appear, i.e., the image appears to fracture. Because part of the concrete instantly loses the binding force after fracture, it makes the image appear to have prominent non-convergence characteristics. As the external force applied at the loading point becomes more significant, the time to appear the non-convergence feature becomes shorter and shorter.
Figure 11.
Four-point bending simulation of reinforced concrete beams.
Figure 11 shows the schematic diagram of the damage process of the reinforced concrete beam. The crack extension process is the same as a plain concrete beam. However, due to the
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addition of reinforcement units, it can be seen that after the distributed cracks are generated, there will be transverse damage between concrete and reinforcement units, which is caused by the material properties. Finally, the same shear-type cracks will be generated, extending from the loading to the support point. It is also worth noting that the addition of reinforcement to the concrete significantly improves the flexural properties of the concrete and changes the damaged form of the concrete, thus showing a transparent crack development process in the images of the same time in Figures 11(b) and (c). The overall damage time is also delayed compared to the plain concrete beam due to the improved tensile properties.
Figure 12.
Crack development process (Load = 50 kN).
Figure 13.
Crack development process (Load = 100 kN).
By comparing the images of Figures 12 and 13 at similar moments, it is evident that the concrete beams show different characteristics of damage forms under the action of different external forces. The bending crack in the middle section of the bottom of the beam appears first during the bending damage, and the crack gradually expands and slowly appears on the outside. In contrast, the shear damage appears first in the bottom of the beam against the outside at similar times, which indicates a fundamental change in the damage form.
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Figure 14.
Load-displacement curves.
For the analysis in Figure 14, the simulation results of the model in this paper can meet the requirements of the Chinese code well and outperform the finite element in terms of crack development simulation. It is worth noting that the model still shows some errors near the load of 40 kN, which should be due to the loss of load-bearing capacity of the concrete at the bottom of the beam after complete crack development. As mentioned in the previous paper, the model in this paper ignores the possibility of particle rotation, so it does not accurately simulate the reinforced concrete beam after using the reinforcement as the main load-bearing module, which needs subsequent improvement. 4.4
Conclusion
The four-point bending model PD simulations for plain and reinforced concrete, the crack development patterns, and damage forms agree with the field experimental results. 5 CONCLUSION 5.1
Strengths of the results in this paper
In this paper, the reinforcement and concrete are divided into two kinds of particles for consideration, good artificial damping is introduced to simulate the quasi-static process, and a model that can accurately reflect the damage form and crack development form of reinforced concrete is established. The following conclusions can be drawn from the simulations of the four-point bending of plain and reinforced concrete beams. Different damage characteristics can be well simulated for both models under the same pressurebearing conditions, and their damage characteristics are consistent with actual experiments. The model can reasonably simulate the reinforcement in reinforced concrete to strengthen the concrete material, change the form of damage, and accurately predict the location of cracks and subsequent development. 5.2
The problems that still exist
In this paper, the particle is treated as a material point that can only translate without taking into account the possible rotation of the particle, and the model of concrete is simplified to an isotropic homogeneous model. The model results may have some errors compared with the real results. 100
REFERENCES Code for Design of Concrete Structures: GB 50010-2010 (2015). Ministry of Housing and Urban-Rural Development of the People’s Republic of China. www.jianbiaoku.com/webarbs/book/209/2438396.shtml Dai, Z., Xie, J., Lu, Z., Qin, S., & Wang, L. (2021). Numerical Modeling on Crack Propagation Based on a Multi-Grid Bond-Based Dual-Horizon Peridynamics. Mathematics, 9(22), 2848. Farahani, A., & Taghaddos, H. (2020). Prediction of Service Life in Concrete Structures Based on Diffusion Model in a Marine Environment Using Mesh Free, FEM and FDM Approaches. Journal of Rehabilitation in Civil Engineering, 8(4), 1–14. Gerstle, W., Sau, N., & Silling, S. (2007). Peridynamic Modeling of Concrete Structures. Nuclear Engineering and Design, 237(12–13), 1250–1258. Gerstle, W. H., & Xie, M. (1992). FEM Modeling of Fictitious Crack Propagation in Concrete. Journal of Engineering Mechanics, 118(2), 416–434. Gu, X., Zhang, Q., & Huang, D. (2016). Peridynamics Used in Solving Penetration Problem of Concrete Slabs. Journal of Vibration and Shock (06), 52–58. doi: 10.13465/j.cnki.jvs.2016.06.009. Ibrahimbegovic, A., Boulkertous, A., Davenne, L., & Brancherie, D. (2010). Modelling of Reinforced‐concrete Structures Providing Crack‐spacing Based on X‐FEM, ED‐FEM and Novel Operator Split Solution Procedure. International Journal for Numerical Methods in Engineering, 83(4), 452–481. Jiao, Y., & Cheng, L. (2022). A New Anisotropic Plastic-Damage Model and its Numerical Implementation for Plain Concrete. Engineering Mechanics (08), 122–137. Kilic, B. (2008). Peridynamic Theory for Progressive Failure Prediction in Homogeneous and Heterogeneous Materials.https://doi.org/10.1016/j.compstruct.2009.02.015 Lu, Z. (2019). Numerical Simulations Research on the Damage and Fracture Behaviors of Concrete Based on Peridynamic Theory (Doctoral Dissertation, Huazhong University of Science and Technology). https://kns. cnki.net/KCMS/detail/detail.aspx?dbname=CDFDLAST2020&filename=1019923670.nh Shen, F., Zhang, Q., Huang, D., & Zhao, J. (2012). Peridynamics Modeling of Failure Process of Concrete Structure Subjected to Impact Loading. Engineering Mechanics (S1), 12–15. doi:CNKI:SUN: GCLX.0.2012-S1-005. Silling, S. A. (2000). Reformulation of Elasticity Theory for Discontinuities and Long-range Forces. Journal of the Mechanics and Physics of Solids, 48(1), 175–209. Tuniki, B. K. (2012). Peridynamic Constitutive Model for Concrete. www.researchgate.net/publication/ 277064005_Peridynamic_constitutive_model_for_concrete Wang, Y. (2005). Research on Some Issues of Bond Performance between Steel bar and Concrete (master’s thesis, Huazhong University of Science and Technology). https://kns.cnki.net/kcms2/article/abstract?v= TDFJfL_btVXpN0uc3_ IC01mZmOFN36too 99OJy8k1fYU7W9OzbRaQ9k3nek7GWSNFDXMlQyYli ZrL5fRJAA1KirCDEk07ore6dufvflh-vDGx_8DAG3dHQ==&uniplatform=NZKPT&language=CHS Wu, L., & Huang, D. (2022). Peridynamic Modeling and Simulations on Concrete Dynamic Failure and Penetration Subjected to Impact Loadings. Engineering Fracture Mechanics, 259, 108135. Xia, Y., Fan, C., Shen, F., & Qian, W. (2021). Simulating the Failure Process of Reinforced Concrete Structure by Using Peridynamics. Chinese Journal of Applied Mechanics (01), 143–149. Xie, B., & Wu, C. (2020). The “Hoop Effect” in Concrete Compressive Strength Testing. China Highways (17), 112–113. doi: 10.13468/j.cnki.chw.2020.17.037.\ Yaghoobi, A., & Chorzepa, M. G. (2017). Fracture Analysis of Fiber Reinforced Concrete Structures in the Micropolar Peridynamic Analysis Framework. Engineering Fracture Mechanics, 169, 238–250. Zárate, F., Cornejo, A., & Oñate, E. (2018). A Three-dimensional FEM–DEM Technique for Predicting the Evolution of Fracture in Geomaterials and Concrete. Computational Particle Mechanics, 5(3), 411–420. Zhao, C., Zhong, X., Liu, B., Zhang, T., & Shi, W. (2018). Numerical Simulation of the Failure Process of Reinforced Concrete Structures Based on the Rigid Body Spring Method. Journal of Mining Science and Technology (02), 129–138. doi: 10.19606/j.cnki.jmst.2018.02.004. Zhao, J., Chen, Z., Mehrmashhadi, J., & Bobaru, F. (2020). A Stochastic Multiscale Peridynamic Model for Corrosion-induced Fracture in Reinforced Concrete. Engineering Fracture Mechanics, 229, 106969. Zou, X., D’Antino, T., & Sneed, L. H. (2021). Investigation of the Bond Behavior of the Fiber Reinforced Composite-concrete Interface Using the Finite Difference Method (FDM). Composite Structures, 278, 114643.
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Discussion of foundation improvement methods for thick rock-filled gravel based on Wudangshan Airport, Shiyan, China Bin Yan* & Zhiheng Shang CAAC Central Southern Airport Design & Research Institute (Guangzhou) Co., Ltd., Guangzhou, China
ABSTRACT: To control the post-work settlement of the thick rock-filled gravel foundation of Wudangshan Airport to a reasonable level. In this paper, numerical simulation was used to investigate the law of post-work settlement under the conditions of maximum height of gravel layer and maximum filling height, respectively, to determine the required treatment depth. And then, based on analyzing the characteristics of existing means of deep foundation treatment, disposal suggestions for foundation treatment design were provided. 1 INTRODUCTION Wudangshan Airport is a high-fill airport in Shiyan, China, which is proposed to carry out a new expansion project shortly. Affected by the previous construction, a high height of gravel layer was filled in this proposed expansion area. The gravel is randomly filled without effective compaction, which can lead to severe settlement. Foundation treatment is difficult for reasons including: (1) Poor homogeneity. The composition is mainly excavation and blasting material with widely varying particle sizes (up to 1–2 m in diameter) and lack of adhesion. (2) Low compaction. Uneven size of pores and large diameter voids exist between particles due to random throwing without treatment. (3) High layer height. The landform before backfilling is undulating with obvious spatial changes in gravel height. The current application of mature treatment methods for gravel foundations mainly includes dynamic consolidation, composite foundation, etc. (Gong 2007). However, most existing methods are either limited in treatment depth or high cost, and the effectiveness of development for thick gravel is yet to be verified. Therefore, the foundation treatment method disposal proposal should be reasonably provided for the gravel layer exceeding the critical height to balance engineering applicability and economic practicality. To meet the post-work settlement design requirements of the flight area of the airport, the settlement of the site under the extreme conditions of the distribution of thick rock-filled gravel should be reasonably estimated. Therefore, in this research, we will use the numerical simulation method to estimate the settlement under two influencing factors of different load fill layer heights and gravel layer heights, respectively. The critical fill height is to be dealt with. Based on the simulations, the applicability of each foundation improvement method will be analyzed to provide a numerical reference for proposing a set of more applicable methods suitable for civil aviation. 2 PARAMETERS AND MODELS The numerical simulation software Abaqus was used to simulate the settlement of the thick rock-filled gravel foundation after loading the road fill to investigate the settlement law of *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-14
the soil without in-situ foundation improvement. The range of the height distribution of the rock-filled gravel was set between 0.4 and 69.3 m. The lower layers were strongly weathered and moderately weathered schist in order from top to bottom. The Mohr-Coulomb principal model was used for both foundation soil and road fill. Each soil layer’s physical and mechanical property indexes were set, as shown in Table 1. Table 1.
Physical and mechanical property indexes of each soil layer.
Soil (rock) name Fill Thick rock-filled gravel Strongly weathered schist Moderately weathered schist
Gravity density g/(kN/m3)
Deformation modulus E/MPa
Poisson’s ratio m
Vold ratio e
Cohesion c/kPa
Friction angle j
Permeability coefiicient k/(m/s)
23.0 18.8
30 5
0.25 0.25
0.2 0.45
15 5
28 19
0.0001 0.0002
26.9
44
0.15
–
70
25
–
26.0
1000
0.10
–
120
22
–
The modeling work was divided into two groups, A and B. On the one hand, the height of the thick rock-filled gravel layer was set as 70 m, and the heights of the fill were regulated in Group A. According to the design, the maximum height of apron fill can be up to 10 m, so a test was set up in the range of 0 m–10 m at every 2 m, as shown in Figure 1(a). On the other hand, the height of the apron fill was set to the maximum of 10 m, and the heights of rockfilled gravel were regulated in Group B. A test was set up in the 10 m–70 m every 10 m to investigate the critical rock-filled gravel layer height under the maximum apron fill layer height condition, as shown in Figure 1(b).
Figure 1.
Numerical simulation test conditions.
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To facilitate the setting of drainage conditions on the top surface of the apron fill layer, the gravity loaded on the apron surface layer was equated to a uniform pressure applied downward. We set the work period of the apron fill layer to 30 days. The settlement observation period can be divided into a 1-year construction and a 30-year operation. During the construction, considering the disturbance, a uniform pressure of 20 kPa is applied to the top surface of the gravel layer. During the operation, we consider the height of the apron surface layer to be 0.8 m and the gravity density to be 25 kN/m3. The pressure caused by the aircraft can be up to 30 kPa at the bottom of the apron surface layer (Tian 2019). Therefore, the maximum uniform pressure can be set to 50 kPa. The CPE4P unit was used for the model units of the thick rock-filled gravel layer and apron fill layer, while the CPE4 unit was used for the schist layers because the pore water pressure does not need to be considered. The apron loading period was set to 30 d, the construction period 360 d, and the post-work period to 10, 800 d (30 years) according to the designed service life. The horizontal displacement of left and right boundaries and the vertical displacement of the lower boundary of the model were restricted. The upper boundary of the apron fill layer was set to be fully drained, while the rest of the boundaries were not drained.
3 ANALYSIS OF THE SIMULATION RESULTS 3.1
Maximum allowable post-work settlement
When considering the post-work settlement index of the apron influence area, the runway settlement should be controlled during 0.2 m – 0.3 m while the taxiway and apron 0.3 m – 0.4 m, according to the relevant standard (MH/T 5027-2013 2013). A higher value is desirable for the high-fill foundation with poorly graded gravel. Therefore, the maximum allowable post-work settlement was 0.3 m at maximum. 3.2
Numerical simulation results
The settlement distributions at the end of the simulated operation period for Group A and B are shown in Figures 2(a) and (b), respectively. As seen from the figures, both will have a
Figure 2.
Settlement distribution under different conditions at the end of the simulated operation period.
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settlement at the end of the operation for the thick rock-filled gravel foundation without any reinforcement. The vertical displacement varies significantly in the thick rock-filled gravel layer but slightly in the strong and medium weathered schist or apron fill layer, indicating that the thick rock-filled gravel layer is significantly compressed. The vertical displacement of the upper layer changes more than the lowers, which leads to the conclusion that the settlement mainly occurs in the thick rock-filled gravel layer. 3.3
The law of post-work settlement variation
The vertical displacement at the midpoint of the top surface of the model in each group can be extracted separately from the end of the construction period to the end of the operation period for a period of 10, 800 days, which can be seen as the post-work settlement. The settlement of Group A is shown in Figure 3(a), which indicates the variation law of settlement with time under different apron fill layer height conditions, and (b) shows the variation law of post-work settlement with apron fill layer heights. From Figure 3(a), the settlements of different tests in Group A grow significantly within the first three days of applying the load. The growth rates develop fastest when the load is first applied. However, the new settlements are stable after three days. The post-work settlements corresponding to different fill heights develop similarly, so the settlements are nearly evenly distributed with the fill heights after stabilization. From Figure 3(b), the postwork settlements decrease with the increase in fill height. The maximum post-work settlement is 0.426 m for the no-fill model, and the minimum is 0.412 m for the 10 m-high-fill models. The post-work settlement reduction law is close to linear. The fitted curve of the relationship between the post-work settlement and the fill height is shown in Equation (1). y ¼ 0:4255 0:0013x; R2 ¼ 0:98874
(1)
Therefore, when the height of the thick rock-filled gravel layer is fixed at 70 m, the postwork settlement is approximately linearly and negatively correlated with the height of the fill layer but not significantly affected. After applying 50 kPa pressure during the operation period, the foundation settlement experiences a fast and then slow development, and stabilization is completed about 3 days after the operation starts. It indicates that the thick rockfilled gravel layer’s settlement is insignificant. It is presumed that the relatively large pore and drainage conditions at the upper boundary led to a fast consolidation rate.
Figure 3.
Post-work settlement development in Group A.
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The influence of fill height on foundation settlement is mainly reflected in the construction. The higher the height, the greater the settlement. However, the height could be more effective in reducing post-work settlement. According to Equation (1), the fill height must reach 96.6 m to meet the control index of post-work settlement of less than 0.3 m. Therefore, the post-work settlement cannot be effectively reduced by overload precompression because of the lack of implementation possibilities. 3.4
Variation in the height of thick rock-filled gravel
The settlement of Group B is shown in Figures 4(a) and (b). From Figure 4(a), the fastest settlement growth rates occur within three days after the load is applied. The higher the height of the thick rock-filled gravel layer, the greater the settlements before the beginning and after the end of the operation period. Settlement growth is not obvious when the height is less than 30 m, while the post-work settlement can be completed in 1 day. However, the settlement growth is relatively obvious when the height exceeds 60 m, and the post-work settlement needs to be completed in 3 days. Therefore, the settlement after the operation period begins to increase with the height of the thick rock-filled gravel layer, while the settlement time is also relatively longer. From Figure 4(b), the post-work settlement grows with the height of the thick rock-filled gravel layer. The post-work settlement reaches a minimum of 0.070 m when the height of the thick rock-filled gravel layer also reaches a minimum of 10 m. What’s more, the post-work settlement reaches the maximum of 0.412 m when the height also reaches the maximum of 70 m. The law of increasing post-work settlement is also close to linear, and the primary function fitting curve is shown in Equation (2). y ¼ 0:0122 þ 0:0057x; R2 ¼ 0:99882
(2)
Therefore, when the fill height is fixed at 10 m, the post-work settlement is approximately linearly and positively related to the height of the thick rock-filled gravel layer. According to Equation (2), when the height of the thick rock-filled gravel layer is higher than 50 m, the post-work settlement cannot meet the design index of less than or equal to 0.3 m. Therefore, it is necessary to consider foundation improvement measures to reduce post-work settlement. Since the height of the fill has only a slight effect, the growth of post-work settlement is mainly controlled by the thick rock-filled gravel layer. The higher the thick rock-filled gravel layer is with a high void ratio, the larger the space is available for compression. Therefore, the replacement method that reduces the gravel layer’s height or compaction treatment can effectively reduce the post-work settlement. Moreover, the compaction treatment method is relatively more economical and applicable.
Figure 4.
Post-work settlement development in Group B.
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4 DISCUSSION OF FOUNDATION IMPROVEMENT METHODS As a result, the improved characteristics of the thick rock-filled gravel of this project are as follows. (1) The main cause of settlement is proposed to be the high porosity of the gravel but not drainage difficulties. Therefore, compaction treatment methods should reduce the postwork settlement instead of overload precompression. (2) The post-work settlement of gravel is positively correlated with the layer height, so only the effective treatment of gravel below 20 m depth can control the post-work settlement within 0.3 m. Only shallow gravel can be treated by ordinary dynamic compaction, while deep gravel without effective treatment will still induce post-work settlement. The methods dealing with deep non-cohesive soil can be divided into the following kinds according to the reinforcement effect: (1) Ultra high-level energy dynamic compaction, which focuses on vertical compaction. (2) Vertical reinforcement composite foundation, which focuses on horizontal compaction. (3) Column hammer dynamic compaction or down-hole dynamic consolidation method (DDC pile method), which focuses on both vertical and horizontal compaction functions (Niu 2017). Ultra high-level energy dynamic compaction can reach more than 10, 000 kNm. At the same time, the reinforcement depth exceeds 10 m, which can also use the compaction energy to break the rock to improve the grain group gradation in addition to compaction (Yan et al. 2013). For example, 25, 000 kNm ultra high-level energy was used at Dalian Jinzhou Airport to treat thick rock-fill gravel foundations, indicating that the processing thickness can reach 17 m – 20 m (Xie 2017). However, ultra-high-level energy dynamic compaction lacks lateral crowding capacity. What’s more, the compaction surface must be fixed, which can only give priority to strengthening the shallow gravel layer (Gao et al. 2013). For airport projects, the huge compaction energy is unfavorable to the adjacent airport facilities. The vertical reinforced composite foundation method is a foundation improvement method that relatively meets the requirements of deep reinforcement, which can be divided into granular, flexible, and rigid pile composite foundations according to the kinds of material. Granular piles, such as vibratory immersed gravel piles, and flexible piles, such as cement mix piles. And rigid piles, such as concrete pipe piles, are all used in airport projects (Huang et al. 2018). Vibratory immersed gravel piles can play lateral squeeze and bite functions for granular piles. However, a vertical drainage channel with an expensive cost is not needed in Wudangshan Airport. For flexible piles, it is difficult to control the piling quality at the bottom with pile lengths over 20 m. Moreover, it is mainly used in soft-ground foundations (He 2002). For rigid piles, since low bearing capacity requirements are needed in apron foundation, the rigid pile has the disadvantages of waste bearing capacity and poor adhesion to gravel. Column hammer dynamic compaction and down-hole dynamic consolidation are compaction replacement methods that lower the ramming surface through ramming holes. Conversely, compaction is carried out by the rammer in the hole. On the other hand, replacement is realized by rock, soil, construction waste, and other fillers to squeeze the lateral pore space. Therefore, these methods can strengthen both the holding foundation and deep weak subgrade layers in the range of force layers. The differences between the two methods are as follows. Column hammer is used in the column hammer dynamic compaction method to ram directly on the surface of the foundation to form ramming holes. On the one hand, it causes a shallow depth of ramming holes and an effective reinforcement depth of 10 m – 15 m (Luo 2020). On the other hand, the impact on the surrounding buildings is only slightly less than that of the original dynamic compaction method. The rammer hole is formed by construction machinery in the down-hole dynamic consolidation method. The replacement function uses a spinning hammer to ram the filler and compact the rammed hole to form a bead-like pile. Down-hole dynamic consolidation method for gravel foundation can reach a depth of 40 m at a unit cost of only 88 ¥ per square meter (Chen et al. 2009). Moreover, deep compaction can reduce the impact on the surrounding environment, making it suitable for foundation improvement of renovation and expansion projects. 107
The engineering practicality and economic applicability should be taken into account in the selection of foundation improvement methods at the same time. They consider the environmental impact combined with the local construction conditions before choosing. Therefore, it is recommended to design a down-hole dynamic consolidation method for reinforcement tests and a composite foundation in some special areas.
5 CONCLUSIONS 1. When the height of the thick rock-filled gravel layer reaches the maximum of 70 m, the post-work settlement exceeds the design requirement of 30 mm within 30 years of operation according to 50 kPa load under the design apron fill height, indicating that the thick rock-filled gravel layer must be compacted by in-situ foundation improvement to reduce post-work settlement. 2. When the apron fills height reaches the design maximum of 10 m, the height of the thick rock-filled gravel layer should keep less than 50 m to ensure that the post-work settlement is less than 30 mm, indicating that in-situ foundation improvement requires the least 20 m high rock-filled gravel be compacted to meet the design requirements. 3. Among the existing deep foundation improvement methods, ultra high-level energy dynamic compaction and column hammer dynamic compaction methods are simple to apply. Still, they need help to meet the design required depths and are environmentfriendly. Vertical reinforcement composite foundations can only have an effective application at a high cost. Down-hole dynamic consolidation method can simultaneously meet the engineering practicality and economic applicability requirements, but the actual application effect requires field tests to verify.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Effects of length, shape, and dosage of steel fiber on mechanical properties of steel fiber reinforced concrete Yuyang Wu* & Tianyu Shao Chang’an University, Xian, Shaanxi, China
ABSTRACT: In recent years, steel fiber reinforced concrete has developed rapidly at home and abroad and has excellent mechanical properties, which can be applied to many construction fields. Therefore, it is important to explore the factors affecting the strength of steel fiber reinforced concrete and find ways to enhance the strength of concrete for improving building stability and disaster prevention and damping. This study investigated the influence of the length, shape, and dosage of steel fiber on the mechanical properties of concrete. It cited relevant experiments to analyze and summarize the changes in concrete compressive strength and flexural strength under different fiber parameters. The results show that steel fiber with different lengths and shapes can improve the mechanical properties of concrete. In addition, proper fiber content can improve the function of steel fiber and then enhance the compressive strength, flexural strength, tensile strength, and impact resistance of concrete. In the future, these findings can provide a reference for building stability in existing engineering projects, enhance the ability to resist disaster damage and reduce casualties and economic losses in disasters.
1 INTRODUCTION Concrete materials are the most widely used engineering materials (Dong et al. 2021). In modern construction, the requirement for concrete materials is extremely high due to the construction of buildings, which is an essential part of the development of cities and society. However, because of the higher height and more complex structures for the buildings, ordinary concrete cannot meet the current requirements for the stability of the buildings. Defects like low ultimate tensile strength, weak deformation ability, and frequent cracking will cause many serious safety problems, such as collapse and falling. In addition, ordinary concrete materials are brittle, and climate change can easily influence their properties. To ensure the stability of the buildings, scientists have made great efforts to promote concrete capacities for decades. The fiber reinforced concrete (FRC) is one of the researchers’ achievements, which can present superior mechanical performances (Dong et al. 2021). Fibers do not change the chemical properties of various materials in concrete. Thus, they do not damage the durability of concrete. Adding fibers into concrete can not only effectively limit the expansion of cement-based micro-cracks, but also improve the strength of concrete, prolong its service life, and expand its application field (Brandt 2008). FRC can be divided into several types. Meanwhile, different kinds of FRC have different functions for reinforcing concrete. For example, synthetic fibers can enhance the tenacity and ductility to make the concrete possess deformation abilities. If high compressive, tensile, and bending resistance are preferred for a preparing building, taking steel fiber reinforced concrete (SFRC) as a construction material will be a nice choice (Shi et al. 2022). As for steel fiber concrete, because of its outstanding capacities, it exists in almost all building structures. *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-15
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This also leads to a new research direction on SFRC, exploring the material factors affecting its performance. In recent years, scientific studies have made great progress in this area. Han et al. (2019) pointed out that after adding steel fiber to concrete materials, the fracture energy of plain concrete will be significantly improved, and the reinforcement effect of steel fiber increases with fiber length, indicating that longer steel fiber in a certain range has a more effective reinforcement effect. In addition, Jin (2012) found that different steel fiber shapes will perform differently on compression. Meanwhile, the shape of steel fiber has a more significant impact on the flexural strength and fatigue strength of concrete than the compressive strength. Another factor that has been found in the fiber dosage is concrete. It has been proved that inserting an appropriate amount of steel fiber can improve the flexural strength of ultra-high-performance concrete. In contrast, excessive fibers will cause negative impacts on the strength of the concrete mixture. These findings will provide a series of guidance for project teams in future construction tasks. Based on the existing data, this essay will discuss in detail the influence of the length, shape, and dosage of steel fiber on the mechanical properties of concrete and explore the effect of the combination of different fiber factors on improving the bearing capacity of SFRC, to reduce the damage of severe climate change and earthquake and other natural disasters on the stability of buildings. 2 INFLUENCE OF STEEL FIBER LENGTH ON MECHANICAL PROPERTIES OF CONCRETE 2.1
Experimental materials and methods
Ying (2014) adopted 42.5 grade ordinary Portland cement, which contains 05 mm, 510 mm, and 1015 mm lime crushed stone as coarse aggregate 1:5:4 mixed and coarse sand with a fineness modulus of 2.8. Three hybrid steel fiber specimen groups were made using a certain material mix ratio. Then the compressive strength test was carried out on the NYL-1000 200 t pressure testing machine, and the flexural strength test was carried out on the WE-1000 hydraulic universal testing machine. 2.2 2.2.1
Experimental result and analysis Compressive strength
Figure 1.
Compressive strength of three types of steel fiber reinforced.
Through comparison from Figure 1, Ying (2014) found that only the volume dosage of ultrashort ultra-fine steel fiber in the concrete can exceed 6%, and the compressive strength peak
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of about 70 MPa when the volume dosage is about 5%, which is 165.8% of the same batch of plain concrete, greatly improving the compressive strength of the sample. Three types of steel fibers have a certain degree of strengthening effect on the compressive strength of concrete. When the volume dosage is more than 2%, the advantage of ultra-short and ultra-fine steel fiber to enhance compressive strength becomes increasingly obvious. 2.2.2
Flexural strength
Figure 2.
Flexural strength of three types of steel fiber-reinforced concrete.
Through comparison, it is found that only the volume dosage of ultra-short ultra-fine steel fiber in the concrete cannot exceed 3%. When the volume dosage is 4% to 5%, the flexural strength of hybrid fiber-reinforced cement concrete is increased by 124%, far greater than that of the other two samples. Hybrid steel fiber’s effect on increasing concrete’s flexural strength is far better than that of ordinary steel fiber and ultra-short and ultra-fine steel fiber at the same volume dosage (Ying 2014). 2.3
Conclusion
For steel fiber reinforced concrete, steel fiber with different lengths has different effects on SFRC. If steel fiber with a reasonable length range is selected, the performance of SFRC can be effectively improved. The research of Ying (2014) shows that the length of ordinary steel fiber in ordinary steel fiber concrete is large, and it takes work to mix evenly. However, the ultra-short ultra-fine steel fiber has a short length and a small aspect ratio. It is easier to form a disorderly distribution in the concrete with the same volume, and the effect of strengthening the compressive strength of concrete is more obvious. In addition, the fracture energy of plain concrete has been significantly improved after the addition of steel fiber. The reinforcement effect of steel fiber increases with fiber length, indicating that longer steel fiber in a certain range has a more effective reinforcement effect (Han et al. 2019). Within a certain range, with the increase of steel fiber length, the splitting strength and unstable fracture toughness of SFRC gradually increase. However, with the increase of fiber length again beyond a certain range, the corresponding mechanical properties will be affected and reduced. Han et al. (2019) show that the length of the steel fiber needs to be longer, making it easy to pull out, and the effect of strengthening mechanical properties needs to be more obvious, 111
reducing its effectiveness. However, we suppose the steel fiber should be shorter. In that case, it will tend to agglomerate during the mixing process, resulting in an uneven distribution of steel fiber and thus affecting its effect of strengthening toughness. When the length of the steel fiber exceeds its critical length, it exhibits tensile failure in the concrete fracture test, causing the concrete fracture to exhibit brittle failure, which is not conducive to early warning of steel fiber concrete fracture. The mechanical reinforcement effect is insignificant for steel fiber-reinforced concrete with a single length. When mixed steel fiber of varying lengths is used to strengthen concrete, it not only takes advantage of the benefits of shorter steel fiber for ease of construction but also fully utilizes the benefits of longer steel fiber for good reinforcement effect, which improves not only the utilization rate of steel fiber but also the strength of steel fiber concrete (Abbas et al. 2015).
3 INFLUENCE OF STEEL FIBER SHAPE ON MECHANICAL PROPERTIES OF CONCRETE 3.1
Experimental materials and methods
Jin (2012) selected four kinds of low-carbon steel fibers with different shapes: long straight, twisted, hook, and dumbbell. The length of the four rigid fibers is 30 mm, the equivalent diameter is 0.5 mm, and the dosage is 80 kg/m3 (Three variables are the same). Because steel fiber is easy to agglomerate, the fluidity of steel fiber reinforced concrete is poor and difficult to mix and tamp. Therefore, to ensure the uniformity of concrete, first, we use the mixer to dry mix the coarse aggregate and steel fiber (one minute), the second step is to add cement and fine aggregate and then continue to dry mix (one minute), and the third step is to add water with water reducer for mixing (three minutes). After the mold is installed and shaped, it shall be inserted evenly and placed on the vibration table for vibration and leveling. The sand ratio is determined to be 40%, and the design strength grade is C30, C40, and C50. We are carrying out 28-day compressive strength impact test and flexural strength impact test on test pieces. The concrete mix proportion is shown in Table 1. Table 1.
Concrete proportioning.
Cement
Water
Fly ash
Sand
Aggregate
Steel fiber
FDN
/kgm-3 298 357 397
/kgm-3 168 168 168
/kgm-3 53 63 70
/kgm-3 740 711 692
/kgm-3 1110 1067 1038
/kgm-3 80 80 80
/% 0.6 0.9 1.0
3.2
Experimental result and analysis
3.2.1 Compressive strength Figure 3 depicts the results. When the concrete strength is C30, the steel fiber-reinforced concrete mixed with four types of steel fiber has a compressive strength of about 40 MPa. With the improvement of the concrete strength grade, the strength of the four types of steel fiber reinforced concrete is also increasing steadily, with a growth rate of approximately the same, which shows that the shape of the steel fiber has little influence on the compressive strength of concrete, dumbbell steel fiber reinforced concrete has certain advantages in the compressive strength of low strength concrete, and long straight steel fiber reinforced concrete has a higher compressive strength in high strength concrete (Jin 2012).
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Figure 3.
Effect of steel fiber shape on the compressive strength of concrete.
3.2.2 Flexural strength Figure 4 shows the impact of four steel fiber shapes on concrete’s flexural strength. The histogram indicates that for C30 concrete with low strength grade, the flexural strength of dumbbell-shaped steel fiber concrete and twisted steel fiber concrete is higher than that of long straight steel fiber concrete and hook-shaped steel fiber concrete. With the increase in concrete strength, the flexural strength of four types of steel fiber concrete is also rising. For C50 concrete, the flexural strength of long straight steel fiber concrete is the lowest, and that of dumbbell steel fiber concrete is the highest, which is 15% higher than that of long straight steel fiber concrete (Jin 2012).
Figure 4.
3.3
Effect of steel fiber shape on the flexural strength of concrete.
Conclusion
In addition, the shape of steel fiber has a similar effect on the splitting tensile strength of concrete at 28 d age. Generally, the shape of steel fiber has a higher impact on the flexural strength and splitting tensile strength of concrete, while the impact on the compressive strength is lower. In the flexural and splitting tensile strength of concrete, the dumbbellshaped steel fiber C50 concrete has the best static performance, while the long straight steel
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fiber C50 concrete has the worst static performance. In the compressive strength of C50 concrete, the two are just the opposite. As a result, the bond strength between the dumbbell steel fiber and the matrix is the best, forming a large bond force. The hook steel fiber has a good embedding force. Both can make the steel fiber bear the stress as a whole and improve the mechanical properties of the steel fiber reinforced concrete. However, the distribution direction of the long straight steel fiber is horizontal to the tensile stress, the surface is the smoothest, and the bond strength between the steel fiber and the substrate is the smallest. Hence, it needs to be carefully selected in practical engineering (Jin 2012).
4 INFLUENCE OF STEEL FIBER DOSAGE ON MECHANICAL PROPERTIES OF CONCRETE 4.1
Experimental materials and methods
Sun (2021) uses cement, silica fume, sand, steel fiber, additives, and water as raw materials (the chemical components of cement and silica fume are shown in Table 2). Through proper mix design, the steel fiber volume is selected to be 0, 1%, 2%, 3%, and 4%, and the lengthdiameter ratio is 65, 80, and 90100, respectively. The shape of steel fiber is straight, hooked, and twisted. The specific mix design is shown in Table 3. Then we pour the cement, silica fume, and fine sand weighed according to the mix proportion into the horizontal concrete mixer (three minutes), add water and water reducing agent (five minutes), and finally, evenly add steel fiber. After 3 to 5 minutes, we test the slump of the test piece. After pouring, film removal, vibration, and curing are done (14 days). Each group of test pieces is three during the test, and the electronic universal testing machine is selected to test and take the average value. Table 2.
Chemical components of cement and silica fume. The ratio of different chemical components/%
Material
SiO2
Al2O3
Fe2O3
MgO
CaO
Cement Silica fume
20.8 95.5
6.12 0.38
4.73 0.29
1.21 0.66
65.70 0.61
Table 3.
Ultra-high-performance concrete mix ratio design. Steel fiber
Reference number
Cement (kgm-3)
Silica fume (kgm-3)-
Crushed Sand stones -3 (kgm ) (kgm-3)
1 2 3 4 5 6 7 8 9 10
880 920 967 1020 1069 967 967 967 967 967
140 150 150 150 150 150 150 150 150 150
725 698 675 640 610 675 675 675 675 675
1115 1055 1000 958 910 1000 1000 1000 1000 1000
Note: () is the amount of steel fiber, unit: kg/m3
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Water (kgm-3)
FDN (kgm-3)
Draw Dosage ratio
Shape
216 226 236 248 258 236 236 236 236 236
6 6 6 6 6 6 6 6 6 6
0(0) 1(78) 2(156) 3(234) 4(312) 2(156) 2(156) 2(156) 2(156) 2(156)
Long straight Long straight Long straight Long straight Long straight Long straight Long straight Long straight Twisted Hook
65 65 65 65 65 80 90 100 65 65
4.2
Experimental result and analysis
4.2.1 Compressive strength Through the analysis of Figure 5, Sun (2021) found the influence of the steel fiber dosage on the compressive strength of concrete. The compressive strength is also increasing with the gradual increase of the steel fiber content. When the steel fiber dosage is 0, the compressive strength is about 110 MPa, and when the steel fiber dosage reaches 4%, the compressive strength reaches 173.5 MPa, which is 57.7% higher than the former. However, the increase of compressive strength is decreasing, compared with 28.7% when the steel fiber dosage increases from 0% to 1% and only 3.8% when the steel fiber dosage increases from 3% to 4%. This shows that the increase of steel fiber content can enhance the friction between it and the mixture. It can form many fiber network structures inside the concrete, effectively controlling the generation and transmission of micro-cracks inside the concrete.
Figure 5.
Effect of steel fiber dosage on the compressive strength of ultra-high-performance concrete.
4.2.2 Flexural strength It can be seen from Figure 6 that when the steel fiber dosage is 0 and 1%, the flexural strength of concrete has no obvious change, which is in the range of 20 MPa to 25 MPa.
Figure 6.
Effect of steel fiber dosage on the flexural strength of ultra-high-performance concrete.
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When the dosage of steel fiber increases from 1% to 3%, the flexural strength of concrete increases significantly. When the dosage of steel fiber increases from 3% to 4%, the flexural strength of concrete decreases inversely. This shows that adding the appropriate amount of steel fiber can improve the flexural strength of concrete. Still, excessive steel fiber is prone to agglomeration, which will destroy the compactness of the mixture and then affect the flexural strength of concrete (Sun 2021). 4.3
Conclusions
The mix proportion of steel fiber concrete is determined according to the different steel fiber models and concrete grades, the construction process requirements, and other specific conditions. Niu et al. (2019) found that improving the steel fiber dosage can significantly improve the splitting tensile strength, bending strength, and impact resistance of concrete. In addition, reasonable fiber content can make the concrete structure more compact and reduce the generation and transmission of original holes and cracks. However, the amount of rigid fiber in large demand will also increase the difficulty of construction. The economic cost should also be considered to display the steel fiber’s performance fully displayed. Therefore, in the construction, the steel fiber dosage proportion should be optimized as much as possible.
5 DISCUSSION A construction project should consider several influencing factors, which can be primarily divided into internal and external factors. Gravity estimation should take the first place in thought for the internal factors. It directly affects the stability of the building through columns, which bear weights and loads from the floors above. It means that the columns should have a better capacity for compressive resistance. In addition, dynamic loads mainly come from people’s movement and are a significant part of internal factors. When designing a building, it is necessary to use high flexural resistance and compressive resistance concrete materials for beams and floors, then make assumption models on the population flow and formulate the maximum limitation of dynamic loads to control the volume of people. However, the abovementioned problems can be predicted in advance, and people have already mastered the methods and experience. Protection from external factors, usually unsteady and unpredictable, is an essential topic that modern constructions should consider. The climate can change in temperature and moisture, which may cause thermal deformation and frost-thaw cracking. It should also be noticed that in the location of earthquake stripes, the buildings should have a higher standard for the strength of the structure to achieve the goal of defending the natural disasters (Chao et al. 2021). To overcome these difficulties, the prime approach is to use SFRC as concrete material, especially paying attention to the related parameters of the fibers. Studies above have shown that for the fibers at the same length situation in the concrete mixture, if the length is longer and not at the “changing range”, the ability to improve compressive resistance and toughness will increase. In contrast, the ability will recede due to the obstacles of the stirring process when the length is beyond limitation. Furthermore, putting steel fibers with different lengths into the concrete will induce surprising results on reinforcing, which is better than just putting the same fibers. This information reminds the workers that accomplishing the stability purpose and achieving the goal of protecting buildings from natural disasters can be dealt with by changing the length of fibers in SFRC when taking it as construction material. Jin (2012) came up with different shapes of fibers and illustrated their fitness for different situations. It tells the workers that dumbbell-shaped steel fibers are the best choice if they are searching for great strength of cracking resistance and tensile resistance; if they are urging for better compressive resistance, the long straight shape should be taken more into 116
consideration. In addition, Sun’s research (2021) expounded that a suitable dosage of fibers will also influence the strength of SFRC. In real construction sites, it is difficult to command fibers’ content at which the strength ability is at the peak value. Commonly, excessive dosage of steel fibers results in collapse and falling accidents. The research gives workers a theoretic reference on the dosage aspect, which makes the construction task easier and safer. Finally, with toughness changes of SFRC, the columns can possess the extra deformation ability. Thus, they will not suffer earthquake breakage, which is crucial for disaster recovery.
6 CONCLUSION In conclusion, length, shape, and dosage can affect performance differently. Increasing length, compressive strength, and flexural strength will also be promoted for single-length fibers in a limited range. In contrast, varying lengths fibers can perform better than single lengths fibers in reinforced concrete. Secondly, different shapes of fibers will provide different improving characteristics for the concrete, and decision-makers can choose the most suitable type for their project to ensure the safety and stability of the buildings under high population flow and earthquake conditions. Finally, a moderate dosage will maximize the functions of steel fibers, especially for tensile strength, bending strength, and impact resistance. Based on these scientific discoveries, enhancing the stability of buildings gains a straightforward direction in the future. It should be highlighted that many unknown factors may influence the performance of SFRC. Revealing these connections is the task for scientists in their further research.
REFERENCES Brandt (2008) Fibre Reinforced Cement-based (FRC) Composites After Over 40 Years of Development in Building and Civil Engineering. Composite Structures., Volume 86, Issues 1–3, Pages 3–9. https://doi.org/ 10.1016/j.compstruct. Chao, S.-., Shamshiri, M., Liu, X., Palacios, G., Schultz, A.E. & Nojavan, A. (2021) Seismically Robust Ultra-high-performance Fiber-reinforced Concrete Columns. ACI Structural Journal., vol. 118, no. 2, pp. 17–32. https://go.exlibris.link/3QMspDss. Dong, F., Wang, H., Yu, J., Liu, K., Guo, Z., Duan, X. & Qiong, X. (2021) Effect of Freeze–thaw Cycling on Mechanical Properties of Polyethylene Fiber and Steel Fiber Reinforced Concrete. Construction & Building Materials., vol. 295, pp. 123427. https://go.exlibris.link/Jjh4gvCp. Hammood, R.A. & Mohsin, S.M.S. (2019) Effect Use of Steel Fiber on Mechanical Properties of Concrete Mixture. IOP Conference Series: Earth and Environmental Science., vol. 365, no. 1, pp. 12061. https://go. exlibris.link/GLXgT3pn. Juhong Han, Mengmeng Zhao, Jingyu Chen, Xiaofang Lan. (2019) Effects of Steel Fiber Length and Coarse Aggregate Maximum Size on Mechanical Properties of Steel Fiber Reinforced Concrete. Construction and Building Materials., vol. 209, pp. 577–591. https://doi.org/10.1016/j.conbuildmat.2019.03.086. Jin Yongtai (2012) Effect of Steel Fiber Shape on Mechanical Properties of Concrete. Cryogenic Building Technology., Issue 10 in 2012 (172 in total). 1001 - 6864 (2012) 10 - 0004 - 02. Kim, K., Yang, I. & Joh, C. (2018) Effects of Single and Hybrid Steel Fiber Lengths and Fiber Contents on the Mechanical Properties of High-Strength Fiber-Reinforced Concrete. Advances in Civil Engineering., vol. 2018, pp. 1–14. https://go.exlibris.link/wc72VL8B. Niu Longlong, Zhang Shiping, Wei Youxin (2019) Effect of Steel Fiber Content on Mechanical Properties of Concrete. China Concrete and Cement Products., 2019 No.3. pp. 211167. 10.19761/j.1000-4637.2019.03.051.04. Abbas, S., Soliman, A. M. and Nehdi, M. L. (2015) Exploring Mechanical and Durability Properties of Ultrahigh Performance Concrete Incorporating Various Steel Fiber Lengths and Dosages. Construction and Building Materials., vol. 75, pp. 429–441. https://doi.org/10.1016/j.conbuildmat.2014.11.017. Sun Yong (2021) Study on Influence of Steel Fiber on Construction and Mechanical Properties of Ultra-high Performance Concrete. Highway Engineering., Vol. 46, No. 1. 10. 19782 /j. cnki. 1674 - 0610. 2021. 01. 033. Ying Wenzong (2014) Mechanical Properties of Different Kinds of Steel Fiber Reinforced Concrete. Transportation Standardization., vol. 42, no. 21. 10.16503/j.cnki.2095-9931.2014.21.001.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Experimental study on the effect of fly ash proportion on the mechanical properties of sand concrete made by waste ultra-fine sand Yixuan Wang* Faculty of Medicine, Nursing and Health Sciences, Monash University, Melbourne, Victoria, Australia
Zijun Tang*, Aohui Tu* & Chaohua Jiang* College of Port, Coastal and Offshore Engineering, Nanjing, Jiangsu, China
ABSTRACT: Waste ultra-fine sand from waterway regulation is used as the main raw material to prepare sand concrete by vibrating forming. In this study, 7 d, 14 d and 28 d compressive strength, splitting tensile strength and immersion compressive strength are tested to determine the influence of fly ash content on the performance of sand concrete. The result shows that the best mixing ratio of this sand concrete is 66.8 % of waste ultra-fine sand,16.6 % of cement and 16.6 % of fly ash with a water/binder ratio of 0.38 and water reducing agent dosage of 0.45 %. The optimal 28d compressive strength, tensile strength and immersion compressive strength of the sand concrete can be achieved. For the areas with rich waste ultra-fine sand resources but lack ordinary sand materials for concrete, the study provides a technical approach for the utilization of waste ultra-fine sand from waterway regulation, having significant economic value and broad application prospects.
1 INSTRUCTIONS Study [1] on the sand concrete made by dune and river sand indicates that adding an appropriate amount of limestone filler can improve the rheological and mechanical properties of sand concrete. It can be seen that recent studies of sand concrete focused on the influence of the ratios of dune sand or river sand and limestone filler on the related properties at the initial stage of preparation and after hardening. However, few studies focus on waste ultra-fine sand. As a mineral blending to prepare concrete, fly ash has an “active effect”, “interfacial effect”, “micro-aggregate filling effect” and “water reducing effect”, which make the concrete exhibit many advantages [2]. With the forward pace of the waterway regulation in the Yangtze River, a huge amount of waste ultra-sand is produced by cutting slopes. It will cost a lot in the transportation and storage of waste ultra-sand and also be negative to the environment so that the development and utilization of waste ultra-fine sand have been increasingly concerned. Therefore, it is necessary to carry out studies on preparing sand concrete with waste ultra-fine sand from waterway regulation. Accordingly, the main target of this research is to investigate the feasibility of preparing sand concrete. In this way, it can provide a promising route to the resource utilization of waste ultra-fine sand in waterway regulation and the utilization rate can be improved. In this paper, the effects of fly ash content on the compressive strength, splitting tensile strength and immersion compressive strength of sand concrete specimens are the main focus of this study. *Corresponding Authors: [email protected], [email protected], [email protected] and [email protected]
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DOI: 10.1201/9781003425823-16
2 MATERIALS AND METHODS 2.1
Sand
The pH of the waste ultra-fine sand is weakly alkaline with a value of 7.67 whose moisture content is 5.04%. Before the tests, the sand is dried in a constant temperature oven. Table 1 presents the particle size distributions of waste ultra-fine sand, particle size of the sand below 0.075 mm only takes 1.29% mass proportion, which is less than 10%; The particles are mainly distributed in the range of 0.15–0.3 mm, and the content is as high as 78.55%. The fineness modulus of the waste ultra-fine sand is calculated as 0.82. Table 1.
Particle size distributions of waste ultra-fine sand.
Size (mm)
1.18
0.6
0.3
0.15
0.075
< 0.075
Percent (wt.%)
0.02
0.02
3.21
75.34
20.16
1.29
2.2
Cement
The OPC used for sample preparation is P.O.42.5 from Hailuo Company. 2.3
Fly ash
The Grade II fly ash density is 923 kg/m3, and the ratio of water demand is 104%. 2.4
Water reducing agent
The FDN water reducing agent is produced by Nanjing Hydraulic Research Institute. 2.5
Ultra-fine sand concrete preparation
In the case of a fixed water-binder ratio (0.38) and water reducing agent content (0.45%), the amount of waste ultra-fine sand and fly ash is changed to determine the proportion of sand concrete specimens. The sand concrete samples are prepared with a certain ratio of waste ultra-fine sand, cement, fly ash, etc., and water (Table 2). The mixture is stirred with water reducing agent by JJ-5 cement mortar mixer for 3 min, which is cast into a 70.7 mm 70.7 mm 70.7 mm triple joint mold. Then the mixture (including molds) is vibrated for 2 stages by HZJ-A concrete vibrator, each stage is for 1 min. After molding, the film is covered on the surface of the test mold, and specimens are demolded after 24 h. The cubic samples are cured to the specified age under standard conditions.
Table 2.
1 2 3 4 5
Proportion of sand concrete specimens (wt.%). Waste ultra-fine sand
Cement
Fly ash
83.1 75.3 66.8 59.8 48.9
16.9 16.5 16.6 16.1 17.0
0 8.2 16.6 24.1 34.0
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2.6
Mechanical property of sand concrete
According to the SL352-2006 Test code for hydraulic concrete, strength tests are carried out in SHT4350 microcomputer controlled electro-hydraulic servo universal testing machine.
3 RESULT AND DISCUSSION 3.1
Compressive strength
Figure 1 presents the results of compressive strength associated with different fly ash proportions. With the increasing fly ash in sand concrete, the 7 d, 14 d and 28 d compressive strength show a trend of rising first and then decreasing. When elevating the fly ash replacement from 0 to 16.6 %, with the increasing fly ash addition, a significant growth of compressive strength can be seen from the Figure. 7 d, 14 d and 28 d strength for sand concrete with 16.6 % fly ash exhibit the highest values, they are 13.23 MPa, 19.14 MPa and 22.56 MPa. At this time, the ratio of fly ash and cement content is 1:1. Compressive performance decreases with the increase of fly ash content in the range of 16.6 % - 34.0 %. Strength after 7 d, 14 d and 28 d of sand concrete specimens with 34.0 % drop down to 9.35 MPa, 14.56 MPa and 18.32 MPa, which are less than the case of no-fly ash added (they are 10.21 MPa, 16.35 MPa and 20.14 MPa).
Figure 1.
Compressive strength of sand concrete blended with fly ash.
During the pozzolanic reaction of fly ash, a hydrolysis layer exists around the fly ash microbeads with the formation of hydrating products. The surface of microbeads is continuously eroded by calcium ions through the hydrolysis layer. Meanwhile, the hydration products fill the hydrolysis layer. With the progress of the hydrated reaction, calcium silicate hydrate (C-S-H) gel and Ca(OH)2(CH) precipitate together to form a “double membrane layer”, which is closely combined with cement slurry. Therefore, fly ash can improve the degree of filling effect in sand concrete which reduces the pore size between the hydrating products (C-S-H, CH, etc.) [3], which is beneficial to optimizing the compressive strength of sand concrete. Besides, a small amount of C3A in cement can further stimulate the activity of fly ash, accelerate the hydration process of cement, and generate calcium carbo aluminate crystals [4]. This may be one of the reasons why the compressive strength of sand concrete increases rapidly in the early stage. 3.2
Splitting tensile strength
By adjusting the proportion of fly ash, the impacts of different fly ash and waste ultra-fine sand ratios on the splitting tensile strength of concrete specimens are shown in Figure 2. The 120
increase of 7 d, 14 d and 28 d splitting tensile strength can be observed and reach to 1.14 MPa, 1.74 MPa and 2.01 MPa when fly ash replacement goes from 0 % to 16.6 % (the ratio of fly ash and cement content is 1:1). When fly ash content exceeds 16.6%, splitting tensile strength of sand concrete specimens decrease with the increase of fly ash addition. When the ratio is up to 34.0 %, the splitting tensile strength of samples at 7 d, 14 d and 28 d reduce to 0.85 MPa, 1.32 MPa and 1.74 MPa respectively, which is higher than that at the corresponding age without fly ash blended (0.71 MPa, 1.28 MPa and 1.67 MPa). The fly ash particle size is small and particle strength is high relatively. Adequate fly ash can fill between the cement particles in the preparation stage of sand concrete, which enhances the “deflocculation” diffusion of cement particles, and increases the compactness of the pouring mixture so that densify the initial structure of concrete. At the early and late stages of hardening, fly ash can also play a physical and active filling role [3]. These effects can effectively reduce the pore volume and the size of coarse particles in sand concrete and improve the splitting tensile performance. However, fly ash particles are spherical and smooth, reducing the internal friction resistance and improving the sand concrete’s fluidity [1,2]. Excessive fly ash will gradation of cement - fly ash - sand mixing system and adversely affect the tensile strength of sand concrete. Therefore, it appears that the splitting tensile strength of sand concrete specimens increased first and then decreased with the increase of fly ash content. 3.3
Immersion compressive strength
Figure 3 shows the test results of immersion compressive strength. The variation trend of immersion compressive strength is similar to that without softening. After softening, the extreme compressive strength values at 7 d, 14 d and 28 d are 12.43 MPa, 17.80 MPa and 21.31 MPa (Fly ash proportion is 16.6%, and the ratio of fly ash and cement content is 1:1). Because of the hydration of cement, the strength of sand concrete increases at early hydration while the hydration products can promote the hydration effect of fly ash. After a certain period, the hydration of fly ash will accelerate and become the main source of strength increase. SiO32- and SO42- ions in sand concrete can replace CO32- ions to generate calcium silicate and ettringite respectively. Meanwhile, the presence of calcium carbonate in sand concrete will accelerate the hydration reaction of calcium silicate to effectively adjust the Ca/Si ratio in the hydrated calcium silicate. On the other hand, in the hydration process of calcium silicate, calcium carbonate can promote the hydration of cement and produce
Figure 2.
Splitting tensile strength of sand concrete blended with fly ash.
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Figure 3.
Immersion compressive strength of sand concrete blended with fly ash.
hydrated calcium carbo aluminate crystals [5]. The formation of calcium silicate, hydrated ettringite and hydrated calcium carbo aluminate will improve the immersion compressive strength of sand concrete.
4 SUMMARY Fly ash can improve the strength of sand concrete made by waste ultra-fine sand for waterway regulation significantly. According to this study, it is determined that the optimal ratio of sand concrete for waterway regulation is 66.8% sand, 16.6% cement, 16.6% fly ash, 0.38 water-binder ratio and 0.45% water reducing agent. 28 d compressive strength, splitting tensile strength and immersion compressive strength of sand concrete prepared in accordance with the optimal mix ratio reach to 22.56 MPa, 2.01 MPa and 21.31 MPa respectively, which are all the optimal values. For the areas with abundant waste ultra-fine sand resources and insufficient ordinary concrete sand and stone materials, the engineering construction of waterway regulation waste ultra-fine sand has significant economic value and broad application prospects.
REFERENCES [1]
[2] [3] [4]
[5]
Bédérina M, Khenfer M M, Dheilly R M, et al. Reuse of Local Sand: Effect of Limestone Filler Proportion on the Rheological and Mechanical Properties of Different Sand Concretes [J]. Cement and Concrete Research, 2005, 35 (6): 1172–1179. Huang Weirong, Guo Guixiang. Effect of Fly Ash Content on C30 Self-compacting Concrete [J]. Concrete, 2015 (4): 119–122. Pera J, Husson S, Guilhot B. Influence of Ground Limestone on Cement Hydratation [J]. Cem. Concr. Compos. 1999, 21: 99–105. Uchikawa H, Hamhara S. Influence of Microstructure on the Physical Properties of Concrete Prepared by Substituting Mineral Powder for Part of Fine Aggregate [J]. Cement and Concrete Research. 1996, 26 (1): 101–111. Zielinsha E. The Influence of Calcium Carbonate on the Hydration Process in Some Portland Cement Constituents (3CaO, Al2O3 and 4Ca, Al2FeO3) [J]. Pr. Inst. Technol. Organ. Prod. Bud. 1972, (3).
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Deformation analysis of enclosure structure affected by foundation pit excavation Jianghao Guo, Zheng Yang* & Yike Dang Xi’an Jiaotong University, China
Chunting Lu Installation Engineering Co., Ltd. Of CCSCEC 7th Division, China
ABSTRACT: Relying on the actual deep foundation excavation project, a threedimensional numerical model of the pit-support structure-building interaction was established by using the FLAC3D finite difference program. On the base of the model being verified to be correct, the effects of the support structure material, quantity as well as the thickness and insertion depth of the enclosure structure on the horizontal displacement of the enclosure structure were studied. The results show that the displacement of the enclosure tends to decrease with the increase of the number of supports, stiffness and thickness while the location where the maximum displacement of the enclosure occurs tends to develop downward with the increase of the insertion depth.
1 INTRODUCTION With the gradual acceleration of urban development and construction, the utilization of urban space is gradually increasing, and the excavation of foundation pits for new projects is inevitably close to existing buildings. However, the excavation of the foundation pit will destroy the original equilibrium state of the soil layer and cause the settlement of the surrounding soil layer and buildings. The main factor causing these problems is the horizontal displacement of the enclosure structure, so it is necessary to investigate the influence law of key parameters of foundation pit excavation on the displacement of the enclosure structure. Some scholars studied the settlement law of the enclosure structure and the soil behind the wall caused by the excavation of the foundation pit based on the field measurement data of the actual project (Ren et al. 2022; Tan et al. 2013; Yang et al. 2022), another part of scholars studied the influence law of different working conditions on the displacement of the enclosure structure, the soil behind the wall and the surrounding buildings during the excavation of the foundation pit using the hair method of numerical simulation (Wang et al. 2017; Wei et al. 2021; Zhu et al. 2013, 2021), and another part of scholars combined field measurement data and numerical simulation to study the influence law of the surrounding environment during the foundation pit excavation process (Han et al. 2018; Yu & Geng 2019; Ye et al. 2020). However, lots of studies have shown that the degree of influence on the surrounding environment will be different depending on the geographical conditions in which the pit is located, the excavation area of the pit and the support system used in excavation. Therefore, this study establishes a numerical model of the foundation pit and the *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-17
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neighboring buildings based on the actual deep foundation excavation project for refined analysis and explores the influence of the support material and quantity of the foundation pit as well as the thickness and insertion depth of the enclosure on the deformation law of the enclosure. It is expected to provide a reference for the design, construction and research of similar projects in the future. 2 ENGINEERING SUMMARY 2.1
Engineering background
This paper relies on the adjacent open-cut deep foundation pit project, and the building adjacent to the north of the pit is a historical building that needs protection. The length of the foundation pit is 144 m, the width is 20.5 m and the average excavation depth reaches 7.2 m, which is in the typical powdery soil, it is necessary to strictly control the impact of the foundation pit support deformation on the displacement of the surrounding soil and surrounding buildings. The foundation pit plan is shown in Figure 1, where D71, D72, D73, D74, D75 and D76 are the horizontal displacement monitoring points of the crown beam of the foundation pit.
Figure 1.
2.2
Foundation pit plane.
Foundation pit excavation and support plan
The average excavation depth of the foundation pit is 7.2 m, and each excavation depth is 1 m, excavating to the bottom of the foundation; the foundation pit support method adopts a rotary hole grouting pile and internal support method enclosure scheme, and the internal support includes concrete support and steel support. The diameter of the bored infill pile is 0.8 m, the spacing is 1.1 m, the depth is 14.4 m, and the arrangement becomes a temporary structure. After the excavation of the first layer is completed, the first concrete support is constructed with the size of 800 mm 800 mm, the spacing is about 9 m, and the distance from the ground is 0 m; after the excavation of the fourth layer is completed, the second steel support is constructed with the size of 609 mm in diameter and t = 16 mm in wall thickness, the spacing is about 4.5 m, and the distance from the ground is -3.7 m. The section of the foundation pit is shown in Figure 2 and soil information is shown in Table 1.
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Figure 2.
Section of foundation pit support structure and soil information.
Table 1.
Parameters of the soil strata.
No.
f ( )
c (kPa)
v
Es (MPa)
r (KN/m3)
1 2 3 4 5 6 7 8
5 18 19 16 28 18 32 17
5 10 11 26 0 14 0 19.5
0.35 0.31 0.3 0.34 0.25 0.3 0.25 0.31
5.00 8.25 11.89 5.12 16.50 14.06 25.00 9.10
1800 1910 1860 1940 1850 1930 1960 1960
Note: f = Internal friction angle; c = Cohesion;V = Poisson’s ratio; Es = Deformation modulus; r = Density of soil
3 NUMERICAL MODELING 3.1
Modeling
Since soil is a complex and loose material, the mechanical behavior is also complex, and under the action of external load soil, it will produce elastic deformation and permanent plastic deformation. The self-contained Moore Coulomb elastic-plastic model in Flac3D can well respond to the characteristics of soil (Liu & Zeng 2006), so this study uses this software for numerical simulation, the model schematic diagram is shown in Figure 3, and the upper boundary of the computational model is free, surrounded by normal constraints, and all constraints are on the bottom surface. To reduce the influence of the model boundary on the simulation results, the model length and width are taken to be about 5 times the excavation depth, and the depth direction of the model is taken to be 40 m. The size of the model is
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Figure 3.
Mesh of model.
established as 200 m 200 m 40 m. The foundation pit enclosure is considered to be a diaphragm wall in this model, and the elastic model is used, and its elastic modulus is considered according to C30 concrete. Because the equivalence to a diaphragm wall will improve the integrity of the enclosure structure, according to the previous paper, the strength reduction factor of the enclosure structure is 0.7 - 0.85 (Hu et al. 2022), and the modulus of elasticity is 80%, reducing to E = 24 GPa, and Poisson’s ratio is 0.2. The equivalent section flexural stiffness method is used to equate the enclosure pile to the continuous wall, and the calculation steps are as in Equation (1), and the thickness of the continuous wall is 0.6 m. ðD þ tÞh3 pD4 ¼ 12 64
(1)
where t is the pile center distance, h is the equivalent diaphragm wall thickness and D is the pile diameter. 3.2
Calculation of numerical model
The process of basic model analysis is as follows. (1) The model is established with the initial stress field generated. (2) All displacements are zeroed out and the enclosure unit is activated (3) monitoring recording points are set up, and the first layer of soil (the project adopts a layered excavation method, the thickness of each layer is 1 m) is excavated and the first concrete support is constructed. (4) When excavating to the fourth layer of soil, the second steel support will be constructed. (5) The bottom of the pit is excavated. 3.3
Model validation
Since monitoring point D73 is in the middle of the pit, which is more sensitive to the excavation of the pit, the simulated value of the horizontal displacement of monitoring point D73 is selected for analysis with the measured value in this paper, as shown in Figure 4. It can be found that the final displacement value of the measured horizontal displacement data is -1.43 mm, and the numerical simulation result is -1.31 mm, which is a small difference between them. Therefore, it can be proved that the model can reflect the real conditions.
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Figure 4. Comparison between numerical simulation value and actual value of point D73.
Figure 5. Horizontal displacement curve of the enclosure with varying support materials and quantities.
4 ANALYSIS OF RESULTS 4.1
Effect of different support materials and numbers on the horizontal displacement of the enclosure structure
As shown in Figure 5, the displacement of the two steel supports is 6.23 mm, and the maximum displacement is above the bottom of the pit; the displacement of one concrete support and one steel support is 6.08 mm, and the maximum displacement is at the bottom; the displacement of the two concrete supports is 5.62 mm. The maximum displacement is 9.93 mm, and the maximum displacement value appears above the bottom of the pit. In the case of three supports of concrete-steel-steel, the minimum displacement of the enclosure structure is 5.61 mm, and the maximum displacement value appears at the bottom of the enclosure structure. And comparing the overall deformation of the enclosure structure, it can be found that the maximum horizontal displacement of the enclosure structure gradually decreases with the increase of the stiffness of the support structure and the location of the maximum displacement changes from the bottom of the foundation pit to the bottom of the enclosure structure. However, with the increase of the number of supports, the construction process will be hindered, making the construction time longer, so the support scheme suitable for the project should be selected for the actual situation. 4.2
Influence of envelope parameters on the horizontal displacement of the maintenance structure
To study the influence of the change of parameters related to the enclosure structure on the horizontal displacement of the maintenance structure, the parameters related to the thickness and embedment depth of the enclosure structure are changed, other parameters are kept unchanged, and the parameter design scheme is shown in Table 2. Method 1: as shown in Figure 6, when the thickness of the enclosure structure is 0.2 m and 0.4 m, the maximum value is above the bottom of the pit, and the maximum value of deformation in other cases is at the bottom of the enclosure structure. By comparing the overall deformation of the enclosure structure, it can be found that increasing the thickness of the enclosure structure can suppress the displacement of the enclosure structure, which is 127
Table 2.
Design parameter change program.
Method number
Enclosure thickness/m
Embedding depth/m
Method 1 Method 2
0.2, 0.4, 0.6, 0.8, 1.0, 1.2, 1.4 0.6
7.2 0, 3, 5, 7.2, 9, 11
because increasing the thickness of the enclosure structure can increase the lateral stiffness of the maintenance structure and the ability to resist deformation will be gradually enhanced. However, when the thickness of the envelope increases to a certain degree, the ability to inhibit the deformation of the envelope becomes limited. In addition, blindly increasing the thickness of the envelope will also increase the construction cost of pit excavation.
Figure 6. Enclosure horizontal displacement curve at various thicknesses.
Figure 7. Enclosure horizontal displacement curve at various insertion depths.
Method 2: the horizontal displacement of the enclosure structure under the conditions of different insertion depths of maintenance structure is shown in Figure 7 and the case of the constant thickness of the enclosure structure, with the increase of the insertion depth of the enclosure structure, the deformation shape and the maximum displacement value of the enclosure structure are the same, but its maximum displacement position tends to move downward because the active earth pressure will increase, with the increase of the insertion depth.
5 CONCLUSIONS In this paper, based on the model being verified, the effect of changing the relevant parameters of the support and enclosure on the horizontal displacement of the enclosure is analyzed by numerical simulation, and the main conclusions are as follows: 1. With the increase of support quantity and stiffness, the horizontal displacement of the enclosure structure gradually decreases, but with the increase of support quantity, it will
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reduce the construction space of the foundation pit and hinder the construction process, so the support quantity and stiffness should be coordinated with each other during the construction of the foundation pit project so that the displacement of the enclosure structure can be minimized. 2. As the thickness of the pit enclosure increases, it helps to reduce the horizontal displacement of the maintenance structure, but the numerical simulation shows that when the thickness of the pit enclosure increases to a certain degree, the degree of inhibition of the enclosure deformation by increasing the enclosure thickness again becomes relatively weak; with the increase of the enclosure insertion ratio, the deformation shape of the maintenance structure is the same, but, with the increasing insertion depth, its maximum displacement position has a tendency to move downward.
REFERENCES Cheng Xuesong, Li Xinhao, Pan Jun, Zheng Gang. Basal-Heave Stability Analysis of Excavations Considering the Strength Reduction of Retaining Structures [J]. Journal of Tianjin University (Science and Technology), 2019, 52 (S1): 56–62. doi:10.11784/tdxbz201902035. Han JY, Zhao W, Guan YP, Jia PJ. Deformation Characteristics and Key Parameters of Deep Excavation Adjacent to Buildings with Shallow Foundations [J]. Journal of Northeastern University (Natural Science Edition), 2018, 39 (10): 1463–1468. Hu Minyun, Shou Shude, Yuan Jing, Zhang Yong, Zhang Shengzhong. Study on Influence of Excavation Process of Adjacent Foundation Pits [J]. Journal of Zhejiang University of Technology, 2022, 50 (01): 111–118. Liu JG, Zeng YAWU. Application of FLAC3D to the Simulation of Foundation Excavation and Support [J]. Geotechnical Mechanics, 2006 (03): 505–508. doi:10.16285/j.rsm.2006.03.037. Ren Dongxing, Huang Hai, Shao Kang, et al. Displacement and Stress Analysis of a Shanghai Foundation Pit at Excavation Stages [J]. Building Structure, 2022, 52 (16): 116–124. doi:10.19701/j.jzjg.ls210165. Shuaihua Ye, Zhuangfu Zhao & Denqun Wang. (2020). Deformation Analysis and Safety Assessment of Existing Metro Tunnels Affected by Excavation of a Foundation Pit. Underground Space (4). doi:10.1016/j. undsp.2020.06.002. Suhui Yu, Yongchang Geng. Influence Analysis of Underground Excavation on the Adjacent Buildings and Surrounding Soil Based on Scale Model Test [J]. Advances in Civil Engineering. 2019. doi: 10.1155/2019/ 6527175. Wang Chuidong, Wang Shijie, Liu Mingzhu, Liang Liang. Influence of Foundation Pit Excavation on Additional Settlement of Adjacent Existing Building [J]. Journal of Hebei Agricultural University, 2017, 40 (06): 109–113. doi:10.13320/j.cnki.jauh.2017.0135. Wei Hongming. Influence of Foundation Pit Excavation and Precipitation on Settlement of Surrounding Buildings. Advances in Civil Engineering 2021. (2021). doi:10.1155/2021/6638868. Yang T, Liu S, Wang X, Zhao H, Liu Y and Li Y. (2022). Analysis of the Deformation Law of Deep and Large Foundation Pits in Soft Soil Areas. Frontiers in Earth Science. doi:10.3389/FEART.2022.828354. Yong Tan, Bin Wei, Yanping Diao & Xin Zhou. (2013). Spatial Corner Effects of Long and Narrow Multipropped Deep Excavations in Shanghai Soft Clay. Journal of Performance of Constructed Facilities (4). doi:10.1061/(ASCE)CF.1943-5509.0000475. Zhu Chun, Ren Yakun, Tan Xin, Xu Peijun, Zhou Kang. Influence of Adjacent Asymmetric Pit Excavation on the Existing Building and the Optimized Analysis [J]. Geotechnical Engineering, 2021, 35 (06): 400–405 +410. doi:10.3969/j.issn.1007-2993.2021.06.010. Zhu Yanbing, Zhou Xiaohua, Wei Shifeng, Tan Yong. Investigation on Deformation Behaviors of Foundation Pit Adjacent to Existing Metro Stations [J]. Geotechnical Mechanics, 2013, 34 (10): 2997–3002. doi:10.16285/j.rsm.2013.10.023.
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Key technology research on the establishment of the prefabricated component library of assembled structure based on BIM Xue Wang, Lu Zhao, Shangang Wang*, Miao Zhang, Guojiao Wen, Xiaohong Gao & Wen Liu CCCC Wuhan Zhi Xing International Engineering Consulting Company Limited, Wuhan Hubei, China
ABSTRACT: Currently, the design of assembled prefabricated structures in China faces various challenges, such as multiple component sizes and inconsistent standards. These issues hinder the industrialized design and production of prefabricated components. In response to this, this article focuses on the management of prefabricated components. It proposes the establishment of a prefabricated component library for assembled structures utilizing Building Information Modeling (BIM) technology. This approach aims to achieve information integration and sharing throughout the whole life cycle of buildings. Developing a prefabricated component library requires producing, setting up, and creating information for prefabricated components. These components are then organized and managed through various functions, including component information display, storage, inquiry, modification, and deletion. The preliminary design of BIM models can be efficiently completed by accessing prefabricated components from the library. Furthermore, the library can be updated and iterated continuously during the optimization process of the model, ensuring that the library is kept up to date and is effectively utilized. 1 INTRODUCTION Traditional design methods for prefabricated structures often involve multiple component sizes and types, which may hinder the standardization, industrialization, and automation of design and production. In contrast, the BIM-based design method for assembled structures focuses on prefabricated components. A prefabricated component library can be established by creating a complete set of standardized prefabricated component units that meet specific modulus standards and managing these components through database technology. Accessing the appropriate components from the prefabricated component library simplifies the overall design process of the monolithic cast-in-place structure and significantly improves the efficiency of the design stage. The data information contained within BIM models is integral to the entire process, coordinating the joint work of different professional and staff units. It also realizes the linkage updating of detailed design and modification while simultaneously promoting the integration of design, production, construction, and operation and maintenance (Mao et al. 2020; Ren 2022). In contrast to traditional design methods for prefabricated structures, the BIM-based design method for assembled structures enjoys significant advantages. By utilizing prefabricated components from a component library, the creation of which is a crucial factor in implementing this design method (Pei et al. 2022). The overall design process is simplified. Once created, the component library should possess sound organizational and management capabilities (Ye & Deng 2021) and apply to create BIM models. During the BIM-based design process for assembled structures, it is essential to establish clear standards *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-18
for prefabricated components within the BIM model (Wang & Han 2015), including the classification and selection of prefabricated components, the creation of codes and information, and the storage and management of prefabricated components in the component library (Wu et al. 2019). Such standards ensure the traceability of BIM model project records, the uniqueness of prefabricated component coding (Liu et al. 2020), and the completeness and scalability of data information at different stages and enable the transmission and application of component data information at different stages. Traditional design methods are predominantly adopted in domestic prefabricated building construction, and the application of BIM-based prefabricated structural design is still under development. Based on the BIM model and relevant data information during a building project’s design phase, the building’s parametric modeling can be achieved. This can be applied in various facets of the construction process, such as design, manufacturing, construction, and management. This study focuses on creating and applying a BIM-based prefabricated component library, with the results being valuable references for similar projects. The feasibility of establishing and managing a prefabricated component library based on BIM technology is demonstrated by creating prefabricated shear wall structures using the Revit Platform. 2 PREFABRICATED STRUCTURAL DESIGN BASED ON BIM 2.1
Prefabricated structural design process based on BIM
The prefabricated structural design process based on BIM involves various steps, including creating and refining a prefabricated component library, BIM modeling, BIM model analysis and optimization, and BIM model construction and application. The process is depicted in detail in Figure 1. The foundation of BIM-based design for prefabricated structures involves the establishment of a component library which is subsequently used to carry out structural designs using component units contained. Therefore, design units, manufacturers, and construction units must integrate the knowledge of prefabricated building structural systems and construction examples. Furthermore, statistical analysis of prefabricated components should be conducted based on the primary control factors, allowing for the classification of prefabricated components into groups of similar types, and prefabricated components, which are standardized, unified, and versatile. It will be achieved through these procedures. 2.2
Creation procedure of prefabricated components information
The effective utilization of prefabricated components in engineering projects can be achieved by creating a BIM model. BIM allows for integrating prefabricated components into the project, and facilitating information sharing throughout the design, manufacturing, and construction stages. A detailed information depth level table has been developed for prefabricated components in BIM, outlining the information that should be included in each project stage, from conceptual design to operation and maintenance management. This information depth level table is presented in Table 1 (Zhang 2016). The five information depth levels of prefabricated components correspond to the five stages of the application of assembled BIM technology. 1. Information depth L1 corresponds to the schematic design stage. Prefabricated components should include basic information, such as the building’s basic shape, overall size, height, and area, without displaying detailed features or internal information. 2. Information depth L2 corresponds to the preliminary design stage. Prefabricated components should include the building’s main planning features, critical dimensions, and specifications without displaying detailed features or internal information. 131
Figure 1.
Design method of assembled prefabricated structure based on BIM.
3. Information depth L3 corresponds to the construction drawing design stage. Prefabricated components should include the building’s detailed geometric features and precise dimensions without displaying detailed features or internal information but should meet the requirements for guiding construction. 4. Information depth L4 corresponds to the detailed requirements of the construction stage. Prefabricated components should include all design information, especially
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Table 1.
Table of the BIM information depth level of prefabricated components. Prefabricated Component Information Depth Level
Type
BIM Data Information
Geometric Information
Geometric and positioning information of the main prefabricated components Geometric and positioning information of the basic prefabricated components Geometric and positioning information of the secondary prefabricated components Geometrical dimensions and positioning information of complex fabricated joints Detailed design information on prefabricated components Basic information (structural system, service life, seismic fortification intensity, etc.) Physical-mechanical properties (different grades of rebar and concrete, elastic modulus, Poisson’s ratio, materials, etc.) Structural design information of prefabricated components Fire-resistant information of prefabricated components Structural design information of prefabricated components Structural design information of prefabricated components Structural design information of prefabricated components Fire-resistant information of prefabricated components
Non-geometric Information
1
2
3
4
5
non-geometric information. Prefabricated components at this depth level should be able to accommodate engineering changes. 5. Information depth L5 corresponds to the detailed requirements of the operation and maintenance stage. In addition to displaying all design information, prefabricated components should include construction data, technical requirements, performance indicators, and other information. Prefabricated components at information depth L5 contain detailed information that can be used in all stages of the building’s life cycle. Using the conventional metric model.rft and metric volume.rft as templates, prefabricated component families are created in sequence based on the component size names. Nongeometric models are created using shared parameters, and corresponding data information of different depth levels is added according to the various stages of the BIM application. The creation of prefabricated component information includes creating codes, adding geometric information, storing non-geometric information, reviewing information completeness, recoding when calling components in the project, collaborative modeling, and improving shared parameter information for prefabricated components. The primary process is shown in Figure 2.
3 RESEARCH ON THE CREATION OF THE PREFABRICATED COMPONENT LIBRARY OF ASSEMBLED STRUCTURE Prefabricated components are integral parts of the BIM model, with unique characteristics such as reusability, scalability, and independence. Developing a library of prefabricated 133
Figure 2.
Flowchart for creating prefabrication component information.
components involves two essential facets: creating the components and implementing a library management function. This process involves multiple steps, including classification and selection of suitable components, coding, and information creation, auditing and storing the components, and ultimately, the management of the library. 3.1
Classification and coding of the stored prefabricated components
As the example of the creation of a prefabricated component library for the assembled shear wall structural system, it is essential to establish a storage structure for the component library, forming a component system based on components such as exterior wall panels, interior wall panels, composite slabs, stairs, balcony panels, air conditioning boards, and parapets. Components are the fundamental unit of a BIM model. Therefore, it is necessary to systematically classify and code the components to establish a BIM model that efficiently stores and provides access to component information and facilitates the transmission and sharing of BIM information among users. The coding for prefabricated component types is presented in Table 2 (Lu 2022).
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Table 2.
The coding for prefabricated component types.
Type
Feature
Exterior walls Internal
External wall without openings External wall with one opening (high) External wall with one opening (low) External wall with two openings External wall with a door opening External Standard external wall External wall with a balcony panel Interior walls Interior wall without openings Interior wall with a fixed door jamb Interior wall with a door opening in the middle The interior wall of a knife-shaped room Composite slabs Two-way composite slab
Stairs
Code
Illustration
WQ––
Exterior wall without opening-Standard widthFloor height Exterior wall with one window opening-Standard width & floor height-Window width & height
WQC1– –
WQCA– Exterior wall with one window opening-Standard – width & floor height-Window width & height WQC2– – – WQM– –
Exterior wall with two window openings- Standard width & floor height-Left window width & heightRight window width & height One door opening-Standard width & floor height -Door width & height
wy1
/
wy2
/
NQ––
Interior wall without openings-Standard widthFloor height Interior wall with one opening (fixed door jamb)Standard width & floor height-Door width & height
NQM1– – NQM2– – NQM3– – DBSX– – –
One-way compo- DBD– –– site slab Double stairs ST–– Scissors staircase JT––
Balcony panels
Conditioning boards Parapets
Balcony with prefabricated laminated slabs Balcony with fully prefabricated slabs Prefabricated conditioning board Sandwichhermalinsulting panel Sandwich thermal-insulting panel for wall corners Non-thermalinsulting panel Non-thermalinsulting panel for wall corners
Interior wall with one opening (fixed door jamb in the middle)-Standard width & floor height-Door width & height Interior wall with one opening (knife-shaped interior wall)-Standard width & floor height-Door width & height Two-way composite slab-Thickness & Cast-inplace layer thickness-Span & width-Steel bar code-Adjustment width One-way composite slab-Thickness & Cast-in-place layer thickness-Span & width-Steel bar code Double stairs-Floor height-Clear width of the stairway Scissors staircase-Floor height- Clear width of the stairway Balcony panel-Prefabricated laminated-Overhang length-Width-Sealing height
YTB– D– – YTB–B– Balcony panel-Fully prefabricated-Overhang – length-Width-Sealing height KTB– –
Prefabricated conditioning board-Conditioning board length-Conditioning board width
NEQ– J1– NEQ– J2–
Prefabricated parapet-Thermal-insulting panelParapet length-Parapet height Prefabricated parapet-Thermal-insulting panel for wall corners-Parapet length-Parapet height
NEQ– Q1– NEQ– Q2–
Prefabricated parapet-Non-thermal-insulting panelParapet length-Parapet height Prefabricated parapet-Non-thermal-insulting panel for wall corners-Parapet length-Parapet height
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3.2
Creation of prefabricated components information
Creating prefabricated components is a process of information integration. The method of prefabricated coding components based on the Revit Platform mainly includes prefabricated component production, setting, and information creation. (1) When creating prefabricated components, each can be a family file, and its data information in different application stages can be gradually improved. The specific creation process is shown in Figure 3. (2) When setting up, wall families can choose.rft,.rft, and.rft and complete the setting work of prefabricated components by setting parameters such as reference plane, family category, and family parameters.
Figure 3.
Creation process of prefabricated component family.
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(3) The creation of prefabricated component library information includes the input of geometric and non-geometric information (Liu et al. 2022). Creating geometric information mainly involves prefabricated components’ length, width, height, and boundary details. Non-geometric parameters must be made through shared parameters, and some predefined parameters should be reserved for data dynamic expansion. We take the prefabricated exterior wall panels of the assembled shear wall structure as an example. We add geometric and non-geometric information, assign a unique code to the component, and save it to the library, as shown in Figure 4.
Figure 4. Diagrams of storing the prefabricated exterior wall, reinforcing WQ bars, and creating information for the exterior wall (WQ-2728).
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Figure 4.
3.3
(Continued)
Preliminary establishment of prefabricated component library of assembled structure based on BIM
After creating the prefabricated component family file, checking and maintaining the component data information is necessary. The management of the component library should have functions such as displaying, storing, inquiring, modifying, deleting, and displaying views of component information (Ma 2019). When designing the user interface of the management system, the interface can be divided into three parts: component information display area, data information input area, and view browsing area. In the storage parameter input area, the component code, location, type, size, and other information can be entered, and the original parameters can also be modified and adjusted. 138
Figure 5.
Storage interface of prefabricated exterior walls in the prefabricated component library.
In the conditional inquiry input area, files in the library can be searched and called. In the coding input area, prefabricated components that have been incorrectly stored can be deleted from the database. Figure 5 shows the storage interface of prefabricated exterior walls in the assembled component library.
4 RESEARCH ON ESTABLISHING THE BIM MODEL FOR PREFABRICATED SHEAR WALLS Taking the assembled concrete shear wall structure as an example, the corresponding BIM model can be quickly established by calling the prefabricated components. The prefabrication rate of the assembled shear wall residential building to be constructed is 58.77%, and the exploded view of the standard floor is shown in Figure 6. First, based on a specific module, corresponding components can be queried from the component library to complete the preliminary design of the BIM model. By continuously modifying the model to ensure that the design scheme meets the structural design requirements, the prefabricated components in the component library can also be supplemented and improved to meet the design needs. After the preliminary design, the prefabricated
Figure 6.
Modelling diagrams of prefabricated shear walls.
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Figure 6.
(Continued)
components in the BIM model can be recorded, and the BIM model of the structural design scheme can be further designed based on Dynamo. The BIM information of the components should be supplemented and improved according to the information depth level standard at different stages until it can be delivered to the prefabrication factory for production, on-site construction, and subsequent project operation and maintenance. The BIM creation process based on the assembled prefabricated component library is shown in Figure 7.
Figure 7.
Creation process of BIM based on the assembled prefabricated component library.
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5 CONCLUSIONS This article mainly studies creating and applying a prefabricated component library for assembled structures based on BIM. (1) The creation of the prefabricated component library includes creating prefabricated components and implementing management functions for the prefabricated component library. The creation and warehousing of prefabricated components can be realized by classifying, selecting, coding, and creating information for prefabricated components. When designing the management system, functions such as warehousing review, inquiry, modification, and deletion of prefabricated components can be implemented. (2) The prefabricated component library can be effectively utilized by designing assembled structure BIM models and products based on the prefabricated component library. Based on the Revit software platform, the creation of prefabricated components for assembled shear wall structures can be completed. The management functions of the prefabricated component library can be implemented based on SQL Server 2008 R2 and Visual Studio 2012. This indicates that creating and managing prefabricated component libraries based on BIM have specific feasibility.
REFERENCES Liu, H., Hong, J. R., Zhang, M. X., et al. (2020). Research on the Management for the Information of Prefabricated Building Combining BIM and QR Code Technology. Construction Technology (2), 110–114, 118. Lujie. (2022). Establishment and Data Development of Prefabricated Component Library for Prefabricated High-rise Residential Buildings Based on BIM Technology. Railway Construction Technology (01), 34–39. Liu, Z. Q., Wu, Y., Tang, C. C., et al. (2022). Analysis of the Integrated Application of BIM Entity Models in Engineering Structures. Building Technology Development (03), 115–121. Ma, H. L. (2019). Design Research and Application of Assembled Frame Structure Family Library Based on BIM Technology (master’s Thesis, Yangzhou University) https://elib.cqlib.cn:8081/interlibSSO/goto/10/ +jmr9bmjh9mds/KCMS/detail/detail.aspx?dbname=CMFD202001&filename=1019189167.nh Mao, X. Q., Li, W., Mao, H. C. (2020). System Dynamics Analysis of Multi-agent Synergistic Promotion Path of Construction Engineering in BIM Context. Journal of Civil Engineering and Management (06), 80–85. DOI: 10.13579/j.cnki.2095-0985.2020.06.013. Pei, Y. F., Su, R., Su, Q. (2022) Development of BIM-based Management Platform for Medium and Lowspeed Maglev Civil Engineering Components. Railway Standard Design (1), 37–40, 48. DOI: 10.13238/j. issn.1004-2954.202010280006. Ren, X. C. (2022). Application of Prefabricated BIM Technology in the Whole Life Cycle of Building. Journal of Railway Engineering Society (6), 90–94. DOI: 10.3969/j.issn.1006-2106.2022.06.016. Wang, R., Han, T. T. (2015). Standardized Management Research on Ancient Building Components Information Classification and Encoding Based on BIM. Construction Technology (24), 105–109. DOI: 10.7672/sgjs2015240105. Wu, S. G., Wang, L. L., Han, X. D. (2019). Application of BIM Technology in Design and Construction of Multi-Storey Fabricated Steel Structure Residential Buildings. Building Construction (4), 691–693. DOI: 10.14144/j.cnki.jzsg.2019.04.045. Ye, R. R., Deng, X. Y. (2021). A Management Method for Building Component Information Storage Based on IFC Standard. Architecture Technology (4), 469–473. DOI: 10.3969/j.issn.1000-4726.2021.04.023. Zhang, C. (2016). Research on Key Technologies for Design and Construction of Fabricated Structures Based on BIM (master’s Thesis, Southeast University) https://elib.cqlib.cn:8081/interlibSSO/goto/10/+jmr9bmj h9mds/KCMS/detail/detail.aspx?dbname=CMFD201701&filename=1016326504.nh
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Sensor fault classification for bridge SHM using LSTM-based with 1D-CNN feature extraction Yufei Guo* School of Civil Engineering, Chongqing University, China
ABSTRACT: The Bridge Structural Health Monitoring System is critical to ensure the safe operation of bridges. However, false alarms triggered by sensor faults in healthy bridges may cause the evaluation system to fail. Therefore, pre-processing is required to identify and classify sensor faults before identifying structural damage. This paper proposes a deep learning-based method called 1D-CNN-LSTM for classifying sensor faults. The 1D-CNN is used to extract multi-domain features from the original signal, and a Long Short-Term Memory (LSTM) network model is constructed to differentiate sensor faults by using fullyconnected layers and a softmax function with the selected feature vector. Data from six fault conditions are obtained through the three-span continuous beam model. The result shows that the proposed method is successful for sensor fault classification.
1 INTRODUCTION Bridge structural health monitoring (SHM) systems have become the focus of common attention in bridge engineering circles (Shan et al. 2021). SHM systems identify structural damage via abnormal changes in the sensor signals. However, the data anomalies caused by sensor fault and structural damage are coupled. When false alarms are triggered by sensor faults, the monitoring system would be misinterpreted as some kind of structural damage. In addition, sensor fault signals such as deviation and gain types still contain the real information of the structure, the service life of the sensor can be improved based on the use of fault sensors, and the economical structure of the structural health monitoring system can be improved if this type of abnormal signal can be repaired. Therefore, detecting and classifying sensor faults is crucial for the normal operation and longevity of SHM systems. Data-driven is an emerging fault diagnosis method. Heydarzadeh et al. (2016) proposed a two-stage fault detection method to extend the model-based method to a data-driven method for nonlinear problems. Using algorithms such as principal component analysis and wavelet transform to extract features of fault signals with inputting them into neural networks can effectively improve the classification accuracy of machine learning models. For example, Li et al. (2019) obtained statistical correlation coefficients and trained an accompanying vector machine to classify sensor fault types. In recent years, deep learning methods with feature extraction have been used for sensor fault classification, especially the Long Short-Term Memory neural network (long short-term memory, LSTM). LSTM can process time series to capture long-term dependencies and has become an effective model for fault classification of time series signals (Gong et al. 2021; Jalayer et al. 2021). However, the LSTM network can only extract features in the time domain and be vulnerable to the noise of the scene and lacks robustness. On the other hand, one-dimensional convolutional neural network *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-19
(1D-Convolutional Neural Networks, 1D-CNN) has unique advantages in processing time series data (Ozcanli & Baysal 2022; Wang et al. 2021) relying on local feature extraction, parameter sharing, and translation invariance. In this paper, a combination of 1D-CNN and LSTM networks is used to solve the above problems. This new approach that offers a new sensor fault classification method for the bridge structure monitoring system is proposed.
2 SENSOR FAULT CLASSIFICATION RELATED THEORY 2.1
Feature extraction of one-dimensional convolutional neural network method
The one-dimensional convolutional neural network (1D-CNN) is designed to process onedimensional time series data by performing convolution operations on the data as ‘images’. The convolution layer operation is illustrated in Figure 1. However, the convolution operation can greatly increase the dimensionality of the data, which can cause the model capacity to increase sharply and lead to exploding parameters. To prevent this, a pooling layer is added to perform secondary calculations on the feature information output by the convolutional layer. The pooling layer, as shown in Figure 2, is designed to reduce the parameter amount of the feature sequence, improve calculation speed, increase the model’s robustness, and enhance its ability to learn from experience (Li et al. 2020)
Figure 1. Diagram of convolution layer operation .
2.2
Figure 2.
Diagram of pooling layer operation.
Fault classification by LSTM neural network method
Compared with the original signal, the sequence processed by the 1D-CNN is no longer a pure time sequence, but a feature vector containing more information. The feature vector is then fed into the deep learning network model with LSTM in the next step. The cell structure of LSTM can be considered as a complex network unit with a long-term memory function. The cell structure of the LSTM network is shown in Figure 3. The output of the forget gate ft is determined by the output value of the hidden layer at the previous time step ht-1 and the input value of this time step xt-1. The input gate is mainly composed of two parts, one part contains the information of the current time step, which generates a candidate temporary cell state ec t , and the other part controls the proportion of the intermediate variable ct-1 flowing into the input gate. The output gate will determine the 143
Figure 3.
Schematic diagram of LSTM forward propagation.
information ht flowing into the hidden layer at the next moment. By calculating the state of the current cell unit and the output matrix ot of the output gate before updating ht, the mathematical formula of the above forward propagation process is shown in Formula (1). 8 ðxf Þ ðhf Þ ðf Þ > f ¼ s W x þ W h þ b > t t t1 > > > > ðxcÞ ðhcÞ ðcÞ > e > ¼ tanh W x þ W h þ b c t t t1 > > < it ¼ s WðxiÞ xt þ WðhiÞ ht1 þ WðciÞ ct1 þ bðiÞ (1) > > > ct ¼ f t ct1 þ it ec t > > > > ðxoÞ ðhoÞ ðcoÞ ðoÞ > o ¼ s W x þ W h þ W c þ b t t t1 t > > : ht ¼ ot tanh ðct Þ In the formula,Wðxf Þ ,Wðhf Þ ,WðxcÞ ,WðhcÞ ,WðxiÞ ,WðhiÞ ,WðciÞ ,WðxoÞ ,WðhoÞ ,WðcoÞ represent the weight matrix in each forward propagation, bðf Þ ,bðcÞ ,bðiÞ ,bðoÞ represent the bias vector in the forward propagation, tanh and s are activation functions respectively. More details can be found in the reference (Yu et al. 2019) 2.3
Establishment of sensor fault classification model
The sensor data is fed into 1D-CNN for feature extraction to generate a sequence of feature vectors. Compared with the original data sequence, the feature vector has a shorter length and more feature information. Therefore, the feature vector obtained through CNN feature extraction can effectively improve the classification accuracy of the LSTM network model. The steps of the sensor fault classification method based on the 1D-CNN-LSTM model are summarized as follows: 1 Abnormal acceleration data is obtained from sensors under different working conditions. 2 The sensor data is divided into m segments. The length of each segment of data is tl/m, where tl is the total length of the data. 3 The data is divided into a training set and a test set in an 8:2 ratio. The training set data is used to extract features via the 1D-CNN layer. 4 The extracted features are integrated into appropriate dimensions and input to the LSTM layer. 144
5 The output data trained by the LSTM layer is input into the Dropout layer and fully connected layer to enhance its ability to learn nonlinear relationships. 6 The final sequence is fed into the Softmax activation function to calculate values for each sequence. The label is obtained corresponding to each fault from the network. 7 The 1D-CNN-LSTM model is trained for multiple Epochs by using the grid searchcross-validation method to converge the model. Finally, the model performance is tested by substituting the test set data into the trained model.
3 NUMERICAL EXAMPLE VERIFICATION 3.1
Data collection
According to Kullaa’s related research on typical sensor faults (Kullaa 2013), this paper divides sensor faults into five types, constant plus noise (stuck) faults, constant gain faults, constant deviation faults, gain linear drift faults, and electromagnetic interference (noise enhancement) faults. The mathematical formulas of the five fault simulations are shown in sequence in Formula (2): 8 xout ðtÞ ¼ c þ sðtÞ > > > > < xout ðtÞ ¼ b1 xðtÞ þ sðtÞ (2) xout ðtÞ ¼ xðtÞ þ a1 þ sðtÞ > > > xout ðtÞ ¼ b2 xðtÞ þ a2 þ b t þ sðtÞ > : xout ðtÞ ¼ xðtÞ þ sn ðtÞ In the formula, xout(t) represents the actual output signal of the sensor, x(t) is the real signal output of the data, and s(t) represents the actual environmental noise, which is simulated by Gaussian white noise. c, a1, a2 are constant constants, b1, b2 are the gain coefficients of the signal, b is the drift coefficient, and sn(t) represents the noise signal received by the sensor when electromagnetic interference occurs. To verify the applicability of the proposed method in this paper, a three-span continuous girder bridge model was established by using ANSYS, and acceleration response data was extracted. The span of the bridge is 26 m + 40 m + 34 m. Relevant parameters of the bridge were set as follows: elastic modulus is equal to 3.0 1010 N/m2, Poisson’s ratio is equal to 0.3, density is equal to 2500 kg/m3, and section size is equal to 0.6 m 0.25 m. The bridge model can be seen in Figure 4. This experiment assumes that the S4 sensor is faulty, and the structural excitation is white noise ground pulsation, ignoring the damping effect, and applying 10250 white noise ground pulsation excitations with a mean value of 0 and a variance of 1 to the bridge model at the support structure, each incentive contains 100 data points. The resulting acceleration response adds noise with a signal-to-noise ratio of SNR = 20. A total of 1, 025, 000 data points are divided into 5 conditions
Figure 4.
Three-span continuous girder bridge model diagram.
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on average, and each data point contains 205, 000 data points and is divided into 205 segments on average, and each segment of data is an acceleration signal sequence with a length of 1, 000. 5 condition of acceleration data is set as a constant gain fault (F1), a constant deviation fault (F2), a stuck fault (F3), a gain linear drift fault (F4), and an electromagnetic interference fault (F5). To explore whether the proposed method can distinguish between sensor fault and bridge damage, avoiding false alarms triggered. The elastic modulus of 10 units at the middle of the third span is reduced by 20%, and the structural damage condition is added to the sample (D1). 3.2
Model performance verification
After conducting a grid search, K-fold cross-validation and manual adjustment, the acceleration sensor signal prepared in Section 3.1 was used as input for the 1D-CNN-LSTM model. After 70 epochs, the model completed training, and the trained model was saved. The test set was then substituted into the trained 1D-CNN-LSTM model to obtain the classification of five sensor faults and one bridge damage condition. The classification confusion matrix heatmap is shown in Figure 5. This figure shows that the average classification accuracy rate of all working conditions reaches 99.18%. Compared with other working conditions, the accuracy of the constant gain working condition is lower. The main reason is that there are some tiny amplitude data in this condition. F5 and D1 have a strong similarity to this small amplitude, which is easy to be mistakenly classified.
Figure 5. Classification confusion matrix heatmap.
Figure 6. Classification accuracy of three network models.
To verify the necessity of 1D-CNN feature extraction, compared with the different network models without 1D-CNN, in addition to 1D-CNN-LSTM, the LSTM model and Deep LSTM model are built, the same original data are substituted and divided into the same training set and test set to ensure that the performance of the model can be truly compared. Figure 6 shows the classification accuracy and average accuracy of the three neural network models for each fault condition. As shown in Figure 6, the 1D-CNN-LSTM model achieves an average accuracy rate of 99.07%, which is significantly higher than the 75.81% accuracy of the LSTM model and 81.83% accuracy of the Deep LSTM model. The Deep LSTM model uses multiple LSTM layers to learn deep features, but its accuracy rate is still lower than expected. This may be because LSTM can only extract features from time-domain signals, whereas 1D-CNN processing can provide multifield information that helps the network capture more effective features. This is demonstrated by the F5 condition with strong noise (accuracy rate of 28.57% and 40.82%), where the difference in
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the frequency domain is much more obvious than in the time domain, resulting in lower classification accuracy. However, the 1D-CNN-LSTM model outperforms the others in this case. 4 CONCLUSIONS This paper presents an approach for multi-target sensor fault classification, which combines 1D-CNN and LSTM networks. The proposed method utilizes 1D-CNN for feature extraction of time series data, which includes high-dimensional features such as time and frequency domains. The extracted feature sequence is fed into an LSTM network to capture deep features in the time domain. Finally, the fully connected layer and Softmax activation function are used for abnormal data classification. The effectiveness of the proposed method is demonstrated through experiments on a finite element model of a three-span continuous beam bridge, where sensor fault data obtained from a mathematical model is used to evaluate the proposed method. The results show that the proposed method achieves an average classification accuracy of 99.07% for sensor faults and bridge damage. However, the proposed method may be affected by various factors, such as the sensitivity of abnormal data and different types of sensors. Further studies are required to investigate these factors in detail. REFERENCES Gong Wen-feng & Cchen Hui & Wang Dan-wei. (2021). Fast Fault Diagnosis Method of Marine Rotating Machinery with Multi-sensor Monitoring Based on Improved LSTM-SVM, Journal of Ship Mechanics, 25 (09): 1239–1250. https://kns.cnki.net/kcms2/article/abstract?v=ECJfaSgxqGc7IDJiZHDvZ9rUpI0zx3BBX TL5gfRzY_OKJnpcP1UNDtTN8v7eaSglp2tMz10Sces43pYZ6U0tlMIvsz49Vifz4fuatiwWQD3vAFHN QczNGDLonfrDcGVP&uniplatform=NZKPT&language=CHS. Heydarzadeh M & Nourani M. (2016). A Two-stage Fault Detection and Isolation Platform for Industrial Systems Using Residual Evaluation. IEEE Transactions on Instrumentation and Measurement, 65 (10): 2424 – 2432. DOI:10.1109/TIM.2016.2575179 Jalayer M & Orsenigo C & Vercellis C. (2021). Fault Detection and Diagnosis for Rotating Machinery: A Model Based on Convolutional LSTM, Fast Fourier and Continuous Wavelet Transforms. Computers in Industry, 125: 103378. https://doi.org/10.1016/j.compind.2020.103378. Kullaa Jyrki. (2013). Detection, Identification, and Quantification of Sensor Fault in a Sensor Network. Mechanical Systems & Signal Processing, 40 (1): 208–221. https://doi.org/10.1016/j.ymssp.2013.05.007. Li L & Liu G & Zhang L, et al. (2019). Sensor Fault Detection with Generalized Likelihood Ratio and Correlation Coefficient for Bridge SHM. Journal of Sound and Vibration, 442: 445 – 458. https://doi.org/ 10.1016/j.jsv.2018.10.062. Li Shu-jin & Zhao Yuan & Kong Fan, et al. (2020). Application of Convolutional Neural Network in Structural Damage Identification. Damage Identification, 37 (6): 29–37. 10.19815/j.jace.2020.02014. Ozcanli A K & Baysal M. (2022). Islanding Detection in Microgrid Using Deep Learning Based on 1D CNN and CNN-LSTM networks. Sustainable Energy, Grids and Networks, 32: 100839. https://doi.org/10.1016/j. segan.2022.100839 Shan Deshan & Luo Lingfeng & Li Qiao. (2020). State-of-the-art Review of the Bridge Health Monitoring in. Journal of Civil and Environmental Engineering, 2021, 43 (S1): 129 – 134. https://kns.cnki.net/kcms2/article/ abstract?v=ECJfaSgxqGc08B9SuhvXJqppQ0g8qogyps7RKlUBUB0eZUSTD53pUTmU8aFmwbdcrgFg EnbsR4GniyCISh4P17QAmgbMV0BTpMpqPOQEMVdZyNRLsVLf2yyteXS1TGeh&uniplatform=NZKPT& language=CHS Wang X & Mao D & Li X. (2021). Bearing Fault Diagnosis Based on Vibroacoustic Data Fusion and 1DCNN Network. Measurement, 173: 108518. https://doi.org/10.1016/j.measurement.2020.108518 Yu Y & Si X & Hu C, et al. (2019). A Review of Recurrent Neural Networks: LSTM Cells and Network Architectures. Neural Computation, 31 (7): 1235 – 1270. https://doi.org/10.1162/neco_a_01199
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Demolition technology of long-span concrete box girder in the upper span closed frame channel Yanyang Li Beijing Municipal Road & Bridge Co., Ltd., China
Yanhui Cao* Beijing Municipal Road & Bridge Co., Ltd. 2. Beijing University of Technology, China
Jinhua Ye Beijing Municipal Road & Bridge Co., Ltd., China
ABSTRACT: To ensure the safe demolition construction of a 63 m span concrete box girder in an upper closed span frame channel, steel tube columns, and bowl-buckle supports are installed under the box girder to be demolished as temporary load-bearing structures to realize the transformation of box girder supporting structure from existing capping beams to temporary load-bearing structures. Based on the numerical calculation and analysis of different cutting steps, it is determined that when the rigid support is transformed into the flexible support, the way of cutting the middle first is safer for the lower bowl-buckle support to bear the force; the steel tube column structure is erected above the load-bearing wall of the lower closed frame passage, and the gantry crane is erected on the steel tube column structure. Considering the load-bearing capacity of the gantry crane and the steel tube column structure, the concrete mass box girder is divided by a rope saw, and blocks can lift the longspan concrete box girder without a crane. Therefore, reasonable design of temporary support, cutting sequence, and lifting and removing equipment can realize the safe demolition of long-span concrete box girders without affecting the normal passage of closed frame channel, which has reference significance for the demolition construction of similar reinforced concrete structures.
1 GENERAL INSTRUCTIONS With the development of the times, a large number of early-built bridges are subjected to aging and exposed to problems of renewal and reconstruction (Cheng et al. 2022; Chen 2022; Jiang & Lan 2007; Nie et al. 2010; Zong et al. 2009), and the main girder demolition construction is inevitable in the process of bridge reconstruction. In the demolition of long-span concrete box girders in the upper span closed frame channel, the traditional methods of box girder demolition, such as the crane demolition method (Li 2007; Tan 2011), blasting demolition method (Wu & Hua 2011), and bridge erecting machine equipment demolition method (Tang 2004), are not suitable on account of the large structural quality and limited working space. Therefore, this paper discusses adopting the temporary support structure, optimizing the cutting sequence, and the overall design of lifting and removing equipment to realize the smooth demolition of a long-span concrete box girder to accumulate experience for similar projects. *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-20
2 PROJECT OVERVIEW A 63 m span concrete box girder demolition project in Kunming City, Yunnan Province, was conducted for an upper span closed frame channel with a concrete box girder mass of 1437.48 t. The upper span was a 413 m closed frame channel with existing roadway below the bridge deck and above the closed frame channel. The transverse section of the bridge adopts a single box and single chamber section, with a bridge width of 12.75 m, a bottom plate width of 6.75 m, and a wing plate width of 3 m. Figure 1 shows the 63 m span concrete box girder in an upper span of a closed frame channel.
Figure 1. Schematic diagram of 63 m span concrete box girder in an upper Spann closed frame channel (m).
The bridge needs to be replaced by a beam, and the existing concrete box girder should be dismantled and replaced with a steel box girder to retain the substructure. Because of the huge weight of the beam, the current 63 m-span concrete box girder is connected with the supported 20 m-long hollow slab girder at both ends of the longitudinal bridge, resulting in a serious limitation of the working space. Therefore, it bears great significance to explore what kind of equipment and beam dismantling technology to realize the safe dismantling of the box girder.
3 TEMPORARY SUPPORT DESIGN 3.1
Beam segmentation
The beam to be dismantled is divided by rope saw, in which the maximum mass of the wing plate is 20.89 t, the maximum mass of the box chamber beam is 50.43 t, and the division
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Figure 2.
Schematic diagram of beam segmentation mode.
lengths are 1.5, 3.0, 4.5, and 6.0 m, the division width of wing plate is 3 m, and the division width of box chamber beam is 3.375 m, as shown in Figure 2. 3.2
Temporary support structure
The temporary support comprises the full support and steel pipe column structure, in which the full support adopts ordinary bowl-buckle support. The support material is Q235A steel, the diameter is 48 mm, the wall thickness is 3.5 mm, and the arrangement spacing is (60 4 + 90 7 + 60 4) cm. The upright pole of the support under the box girder box room is arranged according to 90 cm (transverse) 60 cm (longitudinal); the upright pole of the support under the wing plate of the box girder is arranged according to 60 cm (transverse) 60 cm (longitudinal); there are j 609 mm steel pipe column structures erected at the end of the box chamber of the box girder, each group of which consists of 8 steel pipe columns with a diameter of 609 mm and a wall thickness of 16 mm; the size of steel tube column foundation is 4.6 m 3.6 m 0. 5 m, and C30 concrete is used for pouring. Figure 3 displays the erection of a temporary support structure. Taking the maximum cut beam block mass of 47.10 t (461.6 kN) supported by a full bracket as an example for calculation, the longitudinal force can be evenly distributed, and the transverse force can be simplified according to the principle of area distribution. Figure 4 depicts the stress calculation model of full support. The bearing capacity of a single upright pole is taken as 18.20 kN for checking calculation and strength analysis: s¼
F 18:20 103 ¼ ¼ 37:2MPa ½s ¼ 210MPa S 4:893 104
(1)
According to the “technical safety code for bowl-buckle steel pipe scaffold in building construction” (Jiangsu Xingxia Construction Engineering Group Co., Ltd. 2017), the stiffness analysis is: ½s0 ¼ 210 0:44 ¼ 92:4MPa
(2)
s ¼ 37:2MPa ½s0 ¼ 92:4MPa:
(3)
After calculation, the strength and stiffness of the full support meet the requirements.
4 ANALYSIS OF THE CUTTING STEP SEQUENCE OF THE BEAM Since the wing plate is in a cantilever state, it can be cut into a comb shape in the transverse bridge direction and then cut in the longitudinal bridge direction to make it fall on the
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Figure 3.
Schematic diagram of the erection of temporary support structure (m).
Figure 4.
Calculation model diagram of full support stress (kN).
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support (Figure 5). The weight of the wing plate beam is small, and safety is easily guaranteed when falling off the frame. 4.1
Design of the cutting step sequence of the beam
First, the end of the beam is cut (step S1) so that the box beam falls on the steel pipe column structure. Because of the large stiffness and small deformation of the steel pipe column, the beam is safer. To ensure the stability of full support in the transformation process of the box girder box room support system, the second cutting of the beam body (step S2) is considered, i.e., we cut the end part (Working Condition 1) or the middle part (Working Condition 2) first, as shown in Figure 6. 4.2
Numerical calculation results
Midas finite element software analyzes the buckling of beam-cutting step sequences. The buckling load comparison under different cutting step sequences is shown in Table 1, and the buckling mode comparison of full support is shown in Figure 7. It can be seen from Table 1 that the buckling load of Working Condition 2 is significantly greater than that of Working Condition 1 when cutting the box girder in S2 and S3.
Figure 5.
Schematic diagram of wing plate cutting.
Figure 6.
Sequence diagram of beam cutting.
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Table 1.
Comparison of buckling loads under different cutting steps. Buckling load/(kN/m2)
Step sequence
Condition 1 (P1)
Condition 2 (P2)
(P2–P1)/P2 100%
S1 S2 S3 S4
194 146 109 111
194 196 115 111
0 25.51 5.22 0
Figure 7.
Comparison of buckling modes of S2 full support in different working conditions.
Especially when cutting in S2, the buckling load of Working Condition 2 is 25% greater than that of Working Condition 1. Therefore, when the beam body is supported by a steel pipe column to full support, the buckling load of cutting the middle first is greater than that of cutting the full support at the end first, and the stability is stronger. As can be seen from Figure 7, when cutting the end first, the box beam body is inclined, the vertical deformation gradually increases from left to right, and the vertical deformation value is larger. However, when cutting the middle first, the vertical deformation value of the beam has little difference, and the stability is better. Therefore, Working Condition 2 is selected as the cutting step sequence of the box chamber beam.
5 SUPPORT STRUCTURE DESIGN OF THE LIFTING EQUIPMENT 5.1
Equipment support structure
The foundation size of the steel tube column structure is 4.6 m 2.5 m 0.5 m. C30 concrete is used for cast-in-place, and flange connection steel bars are used at the bottom of embedded columns during cast-in-place. A single steel pipe column structure consists of four j 325 mm steel pipe columns with the wall thickness being 10 mm. No. 14 groove steel is used as a horizontal connection between the columns, and two I28b I-beams are used as sleeper beams at the top of the columns (as shown in Figure 8). Three groups of single-layer and double-row reinforced Bailey beams are used in the main girder of the track. I 20b @ 30cm I-steel is used as the track distribution beam above the main girder of the track, and No.30 track steel is laid above the track distribution beam as the gantry crane track. Except for the Bailey beam, which is made of steel with a yield strength of 345 Mpa, other steel structures are made of steel with a yield strength of 235 Mpa. Two gantry cranes with a maximum load of 40 t each are selected for beam lifting operation. Figure 9 shows the lifting equipment. 153
Figure 9.
5.2
Schematic diagram of lifting equipment (m).
Verification and calculation of stability
Midas finite element software analyzes the gantry crane track beam and column stress process. The maximum cutting beam mass of 50.43 t for the 63 m span concrete box girder is an example for calculation and analysis. The calculation results of each component under load are shown in Table 2. Table 2.
Analysis of calculation results of each component.
Component
Maximum internal force/MPa
Yield stress/MPa
Deformation state
Bailey beam Distribution beam Upright post Horizontal unit
195.1 108.9 27.3 12.4
345 235 235 235
Elastic Elastic Elastic Elastic
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deformation deformation deformation deformation
It can be seen from Table 2 that the stress of the rail beam and column of the gantry crane is in an elastic state during construction, and the structure is safe and reliable.
6 CONCLUSIONS The main conclusions are drawn as follows: 1) Considering the load-carrying capacity of lifting and removing equipment, the block size of the box girder is determined. The full support combined with the steel pipe column structure is used as a temporary support system, which can effectively support the beam mass in the demolition process. 2) Based on the buckling analysis results of the support, when the temporary bracing system is transformed from being rigid to flexible, the deformation of the middle part of the concrete box girder is smaller than that of the end part, the buckling load of the temporary support is larger, and the demolition process is safer. 3) The method of removing the beam with the gantry is adopted, and the supporting structure of the gantry is designed. By analyzing the calculation results, the steel pipe column structure erected above the load-bearing wall of the closed frame channel and the gantry crane erected on it can realize the safe demolition of the beam block when the space on the side of the bridge is limited.
REFERENCES Chen D.F. (2022) Research on Bridge Reconstruction Scheme and Traffic Organization of Underpass Expressway [J]. China Water Transport, (06): 135–137. DOI:10.13646/j.cnki.42-1395/u.2022.06.045 Cheng X.C., Xiong H., Zhang T. (2022) Reconstruction and Reuse Technology of Old Bridges [J]. Construction Technology (Chinese and English), 51 (15): 25–28. https://kns.cnki.net/kcms2/article/abstract? v=3uoqIhG8C44YLTlOAiTRKibYlV5Vjs7iJTKGjg9uTdeTsOI_ra5_XSOb-oa76NOgEuPwqUBn4SZV flyM4qdR7d-Vr4XZGHFz&uniplatform=NZKPT. Jiangsu Xingxia Construction Engineering Group Co., Ltd., et al. (2017) Safety Technical Specification for Bowl-buckle Steel Pipe Scaffolding in Building Construction: JGJ 166–2016 [S]. Beijing: China Building Industry Press. https://wenku.so.com/d/c9adb4736d2bb2006a9f269fb79eba78 Jiang Y. F., Lan W. Y. (2007) Research on Key Technology of Bridge Integral Jacking [J]. Architectural Structure, 37 (S1): 547–549. DOI: 10.19701/j.jzjg.2007.s1.164 Li Y. J. (2007) Beam Replacement Construction Design of 2 # Bridge of Beijing-Qinhuangdao Railway Under the Main Road of Beijing East Third Ring Road [J]. Railway Construction, (4): 35–37. https://kns.cnki.net/ kcms2/article/abstract?v=3uoqIhG8C44YLTlOAiTRKgchrJ08w1e7aLpFYbsPrqEh_mcCMsSo9aR-OQc uPOzFpJ5mg-g19Pm4atFibC-OgqvcWVKW7MqD&uniplatform=NZKPT Nie J. G., Tao M. X, Fan J. S., et al. (2010) Study on the Application of Steel-concrete Composite Structure in Bridge Reinforcement and Reconstruction [J]. Journal of Disaster Prevention and Mitigation Engineering, 30 (S1): 335–344. DOI: 10.13409/j.cnki.jdpme.2010.s1.032 Tang J. S. (2004) Demolition of a Super-large Bridge [J]. Bridge Construction, (3): 54–57. https://kns.cnki.net/ kcms2/article/abstract?v=3uoqIhG8C44YLTlOAiTRKgchrJ08w1e7eeyE9jLkqq8FPJvAPincTFzxT9ZPx zg1ARskSUH5UYuOtlaaEjfFGhIp2Bu4cjNJ&uniplatform=NZKPT Tan W. (2011) Construction Technology of Dismantling T-beam of the Existing Line by Crane and the Military Beam [J]. Shanxi Architecture, 37 (14): 159–160. DOI: 10.13719/j.cnki.cn14-1279/tu.2011.14.067 Wu G., Hua L. H. (2011) Demolition Scheme Design of a 3-span Continuous Box Girder Bridge [J]. World Bridge, (2): 77–80. https://kns.cnki.net/kcms2/article/abstract?v=3uoqIhG8C44YLTlOAiTRKgchrJ08w 1e7tvjWANqNvp9yiZ-tbsVRKGWNJ5Y0TPhsbrgit_8mrOHbKVa4w4HvjhZZH8wv2rrl&uniplatform= NZKPT Zong Z. H., Xia Z. H., Chen Y. Y., et al. (2009) Research Status and Case Analysis of Longitudinal Joints in Widening and Reconstruction of Existing Bridges [J]. Journal of Fuzhou University (Natural Science Edition), 37 (02): 248–260. https://kns.cnki.net/kcms2/article/abstract?v=3uoqIhG8C44YLTlOAiTRKgchrJ08w1e75TZJapv oLK3xcAZ5O8oshRQD48-KcjHKYH4v3SWPbKVyPggyS0MpECwbteA2L10X&uniplatform=NZKPT
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Design and application of the three-dimensional model for seepage deformation of diversion tunnel Zhijing Xu*, Peng Zhao, An Dong & Yuhuan Gao Power China Guiyang Engineering Corporation Limited, Guiyang, China
ABSTRACT: Aiming at the problems of the concealed construction environment, closed construction space and crisscrossing tunnel layout, this paper puts forward the solution of developing a three-dimensional monitoring system for tunnel engineering. In this paper, based on WebGIS technology, according to the engineering hydrogeological conditions, combined with the regional pumping and water pressure test data, a three-dimensional monitoring system for tunnel engineering is developed by using the vue framework, which realizes the basic functions of three-dimensional visualization of tunnel model, sensor positioning, sensor data analysis and so on. This can provide timely, accurate and intuitive information for the safety management and operation scheduling of tunnels.
1 INTRODUCTION The construction environment of the diversion tunnel is hidden, the construction space is closed, and the tunnel layout is crisscrossed. These characteristics bring difficulties to the reasonable formulation of the construction progress plan and the observation of the overall construction appearance of the project. Therefore, it is very necessary to simulate the whole construction process of the diversion tunnel with advanced construction progress visualization methods to make the formulation of the construction organization design more scientific and reasonable (Luo 2021). The research on visualization of construction progress abroad mainly focuses on industrial and civil construction and road and bridge engineering. In terms of visualization research of underground cavern construction progress in the domestic hydropower field, the main achievements at this stage are the 3D dynamic demonstration system of the underground cavern group construction process based on 0penGL and GIS, and the dynamic visualization analysis method of underground cavern group construction process based on 3D geological model (Gong 2021). Aiming at the problems of low informatization and single software function in the safety monitoring of diversion tunnels in China, this paper designs and develops a three-dimensional monitoring system for diversion tunnels, which realizes the functions of three-dimensional display, sensor positioning and data analysis of diversion tunnels. The system can provide effective support for the monitoring and management of diversion tunnels (He et al. 2022). 2 KEY TECHNOLOGIES 2.1
BIM + GIS
BIM (Building Information Modeling) supports the fine expression of three-dimensional models of buildings and contains rich semantic information, which is the product of applying information technology to the construction industry. BIM is a digital expression of the *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-21
physical and functional characteristics of the project facilities, which can express the detailed elements such as the structural components and products of the diversion tunnel, that is, the three-dimensional embodiment of the framework and internal detailed composition of the diversion tunnel in the local environment and rectangular coordinate system (Sun et al. 2021). GIS can study large-scale spatial objects and abstract architectural objects, and can globally display, evaluate and analyze the overall outlines of shapes, facades and external spaces closely related to geographical elements through a unified geographical coordinate system. BIM focuses on the three-dimensional expression of the structural framework and internal detailed composition of the diversion tunnel; GIS uses a unified geographical coordinate system to present the geological environment and geographical environment of the diversion tunnel (Jin & Envelope 2022). By using their advantages, the local structure and the whole structure of the diversion tunnel project can be combined, and the interaction and influence between the engineering entities including personnel and machinery and the surrounding macro-geography and geological environment are comprehensively considered, which provides a visual three-dimensional environment for the structural safety monitoring of the diversion tunnel and provides technical support and basis for users, managers and decision-makers (Lv et al. 2021). Real-time deformation monitoring information, such as deformation data, ground subsidence data and sensor monitoring data, is integrated with the BIM tunnel model and GIS data of the surrounding environment, and the safety risk of the diversion tunnel is comprehensively analyzed, which is displayed in the form of three-dimensional visualization on the Web monitoring platform to facilitate dynamic management of the diversion tunnel (Sun & Shi 2022). 2.2
Key technologies for establishing dimensional parametric models
The construction of the headrace tunnel is a cyclic process consisting of drilling, blasting, ventilation, smoke dissipation, safety inspection, initial support, slag removal, and other operations. Therefore, a simulation model of its construction progress is established based on the idea of hierarchical and modular modeling. At the same time, Visual C ++ 6.0 program development software is used to program the tunnel construction schedule simulation model, and the construction schedule simulation calculation is performed according to the construction organizational design. The construction schedule obtained through simulation calculation can be converted into construction information that can be called by CATIA and saved to an Excel file. The data processing function of Excel can be used to query or modify the schedule. The commonly used tunnel cross-section types include circular, city gates and horseshoe shapes. The main control parameters of the tunnel shape are extracted as userdefined variables, and the correlation between the control parameters is established (Li 2021). A three-dimensional parametric model is created by translating, scanning, and rotating a two-dimensional sketch. Parametric methods can not only create a threedimensional model that is easy to modify, but also automatically update the model when parameters are modified, thereby improving the efficiency of model modification. Designers only need to modify or redefine control parameters, which can quickly complete the construction or modification of 3D models (Zhang & Ding 2022). The concept of parametric design is expressed as follows: F ðD; X Þ ¼ 0 where F (f1, f2, . . . , fn) is a series of parametric equations, and n is the number of equations; D (d1, d2,.., dn) is a variable of the F function that represents the constraint relationship between structural dimensions; X (x1, x2, . . . , xn) is a variable of the F function that represents the position, size of the structure. The advantages of parametric modeling methods mainly include (1) the variability of the model. Traditional models are built based on
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known design data and may require large-scale modifications to the model whenever the scheme changes, resulting in repetitive and cumbersome work. Applying the idea of parametric design to the modeling process can automatically update the model by modifying the corresponding parameters when the scheme changes, which greatly improves the modeling efficiency of designers; (2) the reusability of the model. Due to the structural similarity of many hydraulic structures, new 3D models can be constructed by modifying parameters. (3) supports concurrent design. Because some models need to follow strict precedence relationships, most of the subsequent work needs to rely on data from previous designs, while parametric design can effectively support concurrent design (Li & Chen 2022).
3 SYSTEM OVERALL DESIGN 3.1
Architecture design
The overall architecture of the system is divided into four layers, which are the data layer, service layer, business layer and presentation layer from bottom to top. The bottom layer is the data layer, which provides data support for the system, stores and manages twodimensional data through PostgreSQL, and manages three-dimensional model data through the file management system. The upper layer is the service layer, which obtains the data stored in the PostgreSQL database through nodejs and publishes the three-dimensional model data through IIS. Further up is the business layer, which mainly implements various business operations (Zhang et al. 2021). The top layer is the presentation layer, which generally uses a browser to provide a user interface, and is the layer where the system interacts with users, as shown in Figure 1.
Figure 1.
3.2
Overall system architecture.
System data organization
Data organization is the foundation of the system. The data of the three-dimensional monitoring system of the diversion tunnel consists of three parts: basic geographic data, three-dimensional model data and tunnel thematic business data. Among them, the basic geographic data is located at the bottom of the data structure, reflecting the most basic landform information; As the attachment of tunnel thematic business data, threedimensional model data is a fine and intuitive display of diversion tunnel, which is located in the middle layer; Thematic business data is attached to three-dimensional model data, which is dynamically loaded according to different needs of users, and is located at the top 158
level. Basic data includes two types: (1) Image data, which comes from the “sky map” and is used to reflect the real landform. (2) DEM data, describing the three-dimensional morphology of landforms, is used for terrain modeling in three-dimensional scenes. Threedimensional model data is mainly the three-dimensional model of the diversion tunnel, which is displayed intuitively, stored through the file system and published on IIS (Cui et al. 2022). Tunnel thematic business data includes attribute information, operation data, data information recorded by various monitoring equipment and safety standard data related to the safety of diversion tunnels.
4 SYSTEM FUNCTION REALIZATION The working area of a diversion tunnel project is located in the southeast of a plain, which is the transition zone between the plain and the mountainous area, and the middle part is the northeast distribution of the Yilan-Yitong basin. Generally speaking, the terrain in the area is high on both sides, low in the middle, high in the northeast and low in the southwest (Zhou & Hu 2021). Geomorphological units mainly include valley accumulation landforms (floodplain terrace), denudation accumulation landforms (wave platform) and structural denudation landforms (middle and low hills). The altitude in the area is generally 200 m to 400 m, and the highest peak is 934.2 m above sea level. The elevation of the valley is generally about 200 m, and in mountainous and hilly areas, the main feature of the ridge is round or wide and gentle without great ups and downs. The natural vegetation in the area is undeveloped, mostly secondary forests and artificial forests. The landform is alternating hills and valleys, and the length of the hole in this section is 15 km. The terrain in the area is generally high in the south and low in the north, and the limestone section is high in the northeast and low in the southwest. This section of the line runs southwest. Geomorphological units belong to low hills and valley floodplains. Ups and downs, the bedrock is exposed at the top of the mountain, and cultivated land is in the valley. The highest peak is 484 m above sea level, and the valley is 200 m above sea level, with a relative height difference of 270 mm. Generally, the peak is 200 m to 410 m above sea level. Vegetation is relatively undeveloped. In mountainous and hilly areas, the main feature of the ridge is round or wide and gentle without great ups and downs. The natural vegetation in the area is undeveloped, mostly secondary forests and artificial forests. The amount of water passing through the valley varies greatly with the season (Huang 2021). The buried depth of the tunnel is 44 m to 186 m. The strata crossed by this section are exposed completely, including Upper Paleozoic, Mesozoic and Cenozoic. The lithology is mainly Carboniferous and Devonian limestone, Carboniferous tuff, Triassic tuff and Jurassic tuff. The types of groundwater in the project area are loose rock pore water, carbonate karst fissure water and bedrock fissure water. Groundwater in the project area is mainly supplied by atmospheric precipitation, surface runoff is supplied by groundwater in the dry season, and river water supplies groundwater for a short time in the wet season, and groundwater is mainly shallow circulation. Affected by topographic cutting, groundwater is often mainly short-distance runoff and often discharged from the surface in the form of springs to replenish river water (Dai et al. 2021). The pumping test results are shown in Table 1, and the rock mass permeability results are shown in Table 2. According to the analysis of water pressure test results, the shallow rock mass of general rock mass (non-structural fracture zone, joint dense zone and dissolution zone) has strong water permeability, which is weak to medium water permeability. It is considered that this is mainly due to the influence of weathering unloading joint cracks on the surface rock mass. The permeability of deep rock mass is mainly weak, which shows that the permeability of rock mass tends to weaken with the increase of depth. The water permeability of rock mass in the cavern is mainly weak. Water pressure in the structural fracture zone and joint dense
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Table 1.
Table of the pumping experiment results.
Number Pumped test section
Water inflow Q (m3/ d)
Down The thickdeep S ness of the (m) aquifer (m)
Pumping hole radius T (m)
Filters live long L (m)
C1 C2 C3 Average
87.12 145.61 172.08
1.34 2.88 3,75
0.074 0.074 0.074
5.5 6.5 5.5
Table 2.
55.26 55.26 55.26
Pumping af- The permeability fects the ra- coefficient of the dius R (m) aquifer layer K (m/d) 36.19 64.54 85.24 61.99
7.34 5.02 5.16 5.84
Water permeability of the rock mass.
Formation code name
Rate of deLithology cay
C1-2m C1-2m
Limestone Weak Limestone Minute
C1-2m
Limestone Fresh
Permeability rate (lu)
Permeability of the cavity site (lu)
9.5 – 9.8 9.8 – 10.2 10.9 – 17.7
10.9 – 16.0
Water permeability Weak Weak – medium Medium / medium
The top of the hole is buried Statistical deep (m) depth (m) 31.7
10.0 – 19.0 19.0 – 25.0 30.0 – 50.0
zone shows that its water permeability is mainly moderate, which is also in line with the actual law (Zhu 2021). The main interface of the system consists of four parts, the top is the system name and navigation bar, the left is the various parts of the diversion tunnel and sensors, and the middle is the main window of the view. The main content is to load the diversion tunnel model on the three-dimensional earth. To highlight the model, the image is temporarily closed. The upper right corner is the statistics chart of the number of sensors, and the bottom is the basic operation toolbar. The toolbar includes model transparency, zoom in, zoom out, full map, true north view, two-dimensional linkage, model showing and hiding, image showing and hiding, roaming and so on. The sensor positioning function can locate the sensor in the model by clicking the sensor in the tree structure on the left, and highlighting the sensor. In the pop-up dialogue window, the measured curve of the corresponding sensor in the recent period is displayed, and the maximum, minimum and average values of the sensor in a period are given on the right. The state of the sensor is displayed quickly and intuitively, which is convenient for users to analyze and evaluate the security state and its development trend. The measuring point information input function can import data in batches or enter data individually. By uploading the parameter file before uploading the data, you can calculate whether there are abnormal values in the sensor values according to the parameters. The data analysis function is the core function of the system. When data analysis is turned on, all the data in the database are read by default. By selecting the sensor on the left and selecting the period, you can query the data of the corresponding period of the sensor and draw it as a line chart. The single-value curve and multi-value curve can be selected, and the multi-value curve is beneficial to the comparison of measured data of the same sensor. Among them, line charts are drawn by ECharts, whose abbreviation comes from Enterprise Charts, a pure Javascript chart library, and the bottom layer relies on the lightweight Canvas class library ZRender to provide intuitive, vivid, interactive and highly personalized data visualization charts. The patrol inspection function is used to
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configure patrol inspection configuration information, input patrol inspection information, upload inspection situations and scene pictures (Chen 2021).
5 CONCLUSION This system can be regarded as a preliminary exploration of BIM + GIS in water conservancy. WebGIS technology is introduced, and a three-dimensional monitoring system of diversion tunnel is established with maps, three-dimensional models and thematic data as the core, which realizes the basic functions of three-dimensional visualization of tunnel model, sensor positioning and sensor data analysis, and provides timely, accurate and intuitive information for tunnel safety management and operation scheduling. Through this system, the workload of operation and maintenance can be reduced to some extent, and effective support can be provided for the monitoring and management of water conservancy facilities.
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Seismic response analysis of large-span structure considering rotational ground motion under uneven settlement sites Haolin Han & Hao Zhang* Shenyang Jianzhu University, Shenyang, China
Hongnan Li Dalian University of Technology, Dalian, China
ABSTRACT: In order to study the influence of rotational ground motion on structural seismic response under the uneven settlement, A large-span industrial plant structure model is established to analyze the seismic response of the translational ground motion independently and translational and rotational components of the ground motion coupled. The results show that, under uneven settlement sites, the responses in all directions will be increased, and with the compressibility of subsoil increasing, the influence increases more. Therefore, when analyzing the seismic response of long-span structures under the uneven settlement, the adverse effects of the ground motion rotation component should be considered.
1 INTRODUCTION With the development of China’s economy, the shortage of land resources is becoming more and more obvious, and the development and utilization of underground space are inevitable. However, the development and utilization of underground space will cause the surrounding surface to sink, thus causing uneven settlement of surrounding buildings. For large-span space structures, uneven settlement of the site will have a greater impact on the structure, which is more easily to occur large deformation and affect its seismic performance (Wang 2013). Li (2013) studied the hazard of site uneven settlement on frame structures and proposed specific reinforcement measures. Liu and Zheng (2004) used an elastic support model to analyze the effect of the uneven settlement on steel frame structures, and the results showed that the greater the base stiffness, the greater the effect of the uneven settlement on the structure. Wang (2016) investigated the effect of site uneven settlement on the seismic performance of steel frame structures. The results showed that the settlement difference between adjacent columns is the main factor causing the change in structural internal forces, structural nodal acceleration and rod stress response. Yan and Wang (2017) analyzed the seismic response of a large-span structure under an uneven settlement, and the results showed that considering uneven settlement would increase the structural nodal acceleration and rod stress response. Xu (2022) studied the seismic performance of a steel frame structure under uneven settlement conditions and found that the ground vibration intensity, settlement area and settlement amount all have a great influence on the seismic response of the structure.
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-22
When an earthquake occurs, the ground motion has a rotational component in addition to the translational component. However, since it is difficult to measure the rotation component, most scholars use the finite difference method and elastic fluctuation theory method to derive the rotation component (Li 2006). Newmark (1969) proposed the traveling wave method, which was the first time that the rotational component was calculated from the translational component. Li and Sun (2001) considered the dispersion effect of surface waves in the calculation of the rotation component, which made the analysis results more consistent with the actual. Zhang et al. (2015) analyzed the structural response of a large-span asymmetric space structure under the multi-dimensional ground motion, and the results showed that the rotational component has a large influence on the displacement and bottom internal force of the structure. Han et al. (2022) analyzed the seismic response of reinforced concrete frame structures considering the rotational component and the results showed that considering the rotational component would increase the displacement of the top floor of the structure and the torsional effect of the structure. At present, when analyzing the seismic response of large-span structures under the uneven settlement, only the influence of the translational component is considered. The effect of the rotational component is not considered. In this paper, A large-span industrial plant structure model is established to analyze the seismic response of the translational ground motion independently and the translational and rotational components of the ground motion coupled, and analyze the influence of the rotational ground motion on the dynamic response of the large-span space structure under uneven settlement site.
2 BUILD A STRUCTURAL MODEL This paper selects a representative engineering example as the research object, which is a typical single-span double-slope portal frame structure. The roof slope is 5%, the span is 60 m; the longitudinal length is 150 m, the spacing between two spans is 6 m, the steel column in the main structure is Q345B H-beam, of which the lower column is H650 400 16 20 and the upper column is H400 380 10 18, the steel truss is Q345B I-beam, the roof frame is Q235B H-beam. The support between columns uses Q235B steel, the cross-sectional specifications of upper and lower chords are HW300 300, the cross-sectional specifications of diagonal webs are HW150 150, and the cross-sectional specifications of vertical webs are HW150 150 and HW100 100. The dimensions and plan layout of the structure are shown in Figure 1.
Figure 1.
The structure size and the layout.
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The paper uses SAP2000 finite element analysis software for modeling analysis. Only the self-weight and seismic effects of the structure were considered, and the supports were fixed. Figure 2 shows the finite element model of the structure.
Figure 2.
Finite element model of the structure.
3 MODAL ANALYSIS The modal analysis of the structure can determine the vibration characteristics of the structure initially, so as to better obtain the structural force performance. In this paper, we adopt the Ritz vector method in Sap2000 software to conduct a modal analysis of the structure. Table 1 gives the first four natural vibration characteristics of the structure, and the results show that the structure has the largest mass participation coefficient in the X direction, followed by the Y direction, and t in the Z direction is the smallest. From this, it can be obtained that the structure is most likely to excite the vibration in the X direction, followed by the Y direction, and the Z direction is the smallest. Table 1.
Natural vibration characteristics of structure.
Vibration
Period (s)
Frequency (hz)
Vibration forms
1 2 3 4
0.537019093 0.477638868 0.381804344 0.346592553
1.862131185 2.093631963 2.619142545 2.885232215
X-direction translation XY plane torsion X-direction translation X-direction translation
4 SELECTION OF GROUND VIBRATIONS AND ACQUISITION OF ROTATIONAL COMPONENTS 4.1
Selection of ground vibrations
The seismic design intensity of this structure located is at 7 degrees (0.15 g), and the Classification of Construction Site is Class II. According to the requirements of China’s structural seismic design code, the El-Centro wave (Figure 3), Kobe wave (Figure 4) and artificial wave (Figure 5) were selected.
Figure 3.
El-Centro wave.
Figure 4.
Kobe wave.
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Figure 5.
Artificial wave.
4.2
Acquisition of rotational components
This paper is based on the theory of elastic half-space. And it assumed that the propagation medium of seismic waves is uniform and elastic. and the frequency domain method is used to obtain the rotational ground motion. the specific calculation steps are as follows (Zhao & Luo 2014): € ð tÞ (1) We decompose the translational component of the actual ground motion into €u ðtÞ, w and €v ðtÞ. € ðwÞ and €v ðwÞ from €u ðtÞ, w € ðtÞ and €v ðtÞ by Fourier transform. (2) We get €u ðwÞ, w Uðf Þ (3) We obtain the apparent velocity of the seismic wave according to C ¼ sin a. C is the apparent velocity, and a is the incidence angle of the seismic wave at the surface. (4) The Fourier spectrum of rotational acceleration is obtained by the following formula: € ðwÞ w sin a C
(1)
€ ðwÞ w cos a C
(2)
€ x ðwÞ ¼ iw j € y ðwÞ ¼ iw j € z ðwÞ ¼ j
iw ½€v ðwÞcos a €u ðwÞsin a 2C
(3)
€v ðwÞ a ¼ arctan €v ðwÞ
(4)
€ y ðwÞ € x ðwÞ; j where i is the imaginary part, and w is the circular frequency of seismic wave. j € z ðwÞ is the Fourier spectrum of rotational acceleration. and j 4 SETTING OF UNEVEN SETTLEMENT CONDITIONS The foundation soil of the site where this structure is located is highly compressible soil, and according to the Code for the design of the building foundation. the settlement difference between adjacent column bases is 0.002 L (L is the distance between the centers of adjacent column bases). This paper takes 18 mm. The settlement mode is an inclined settlement. The number of each column is shown in Figure 6 and the Sedimentation of each column under different operating modes is shown in Table 2. In addition, this paper sets two working conditions (S1: Only the translational component of ground motion is considered; S2: Both the translational and rotational components of the ground motion are considered) and selects 3 control nodes on the control section of the structural grid, and extracts the dynamic response analysis results of the nodes. Figure 7 shows the node location distribution.
Figure 6. Table 2.
Number of each column. Sedimentation of each column under different operating mode (mm).
Operating mode
1
2
3
4
...
12
13
14
15
...
22
23
24
25
26
No settlement occurred Uneven settlement
0 450
0 432
0 414
0 396
...
0 252
0 264
0 216
0 198
...
0 72
0 54
0 36
0 18
0 0
165
Figure 7.
Selected node location distribution.
5 STRUCTURAL SEISMIC RESPONSE ANALYSIS 5.1
Structural displacement response
Table 3.
The peak displacement of the control nodes under no settlement occurred. S1
S2
Seismic wave
1
2
3
1
2
3
El-Centro wave Kobe wave Artificial wave
45.86 23.79 33.88
46.03 23.18 33.67
47.03 23.81 33.92
48.09 24.19 34.55
47.18 23.87 34.37
48.29 24.77 35.26
Table 4.
The peak displacement of the control nodes under uneven settlement. S1
S2
Seismic wave
1
2
3
1
2
3
El-Centro wave Kobe wave Artificial wave
54.26 26.66 37.90
54.81 26.39 38.30
54.98 27.26 39.75
63.65 30.79 43.45
64.15 30.73 44.03
63.96 31.71 44.54
When settlement does not occur after the action of rotational ground motion is considered, the displacement increase of the El-Centro wave is 4.48%, the Kobe wave is 4.03%, and the Artificial wave is 3.97%. When settlement occurs, after the action of rotational ground motion is considered, the displacement increase of the El-Centro wave is 17.31%, the Kobe wave is 16.44%, and the Artificial wave is 15.00%. Comparing the displacement response of the structure under the selected three groups of ground motions, it can be found that the rotation component of the ground motion has a great impact on the displacement response of the structure under the inclined settlement. 5.2
Structural torsional response
Table 5.
The torsion angle of the control nodes under no settlement occurred. S1
S2
Seismic wave
1
2
3
1
2
3
El-Centro wave Kobe wave Artificial wave
0.000136 0.000045 0.000099
0.000289 0.000079 0.000255
0.000106 0.000043 0.000077
0.000145 0.000048 0.000107
0.000314 0.000086 0.000276
0.000115 0.000047 0.000077
166
Table 6.
The torsion angle of the control nodes under uneven settlement. S1
S2
Seismic wave
1
2
3
1
2
3
El-Centro wave Kobe wave Artificial wave
0.000155 0.000051 0.000117
0.000336 0.000092 0.000303
0.000121 0.000051 0.000092
0.000173 0.000056 0.000134
0.000382 0.000010 0.000346
0.000135 0.000056 0.000105
When settlement does not occur after the action of rotational ground motion is considered, the torsion angle increase of the El-Centro wave is 8.74%, the Kobe wave is 8.14%, Artificial wave is 8.22%. When settlement occurs, after the action of rotational ground motion is considered, the displacement increase of the El-Centro wave is 13.72%, the Kobe wave is 10.24%, and the Artificial wave is 14.11%. Comparing the displacement response of the structure under the selected three groups of ground motions, it can be found that the rotation component of the ground motion has a great impact on the displacement response of the structure under the inclined settlement.
6 CONCLUSIONS (1) When settlement does not occur, the rotational component will increase the displacement response and torsional response of the structure, so it is not sufficient to consider only the translational component in the seismic design of large-span structures. (2) After considering the ground vibration rotation component, the displacement response and torsional response of the structure in all directions are further increased, so when the tilted settlement occurs in the structure, especially in the large-span structure, it is more necessary to calculate the effect of the rotation component to get more accurate results.
ACKNOWLEDGMENT The authors would like to thank the financial support from the National Key R&D Program of China (2018YFD1100402), the State Key Program of the National Natural Science Foundation of China (51738007) for carrying out this research, the Program of the Educational Department of Liaoning Province (LJKMZ20220949) and the Program of Shenyang Bureau of Science and Technology (RC220171)
REFERENCES Han M, Liu Y B, Du H K, et al. (2022) Analysis of the Influence of Rotational Ground Motion on the Dynamic Response of the Structure [J/OL]. Journal of Vibration Engineering.:1–8. http://kns.cnki.net/kcms/ detail/32.1349.TB.20220120.1142.002.html, 2022-01-20. Li Y. Discuss the Hazards and Countermeasures of Ground Foundation Differential Settlement in the Frame Building [J]. Urbanism and Architecture, 2013. 118(14): 215. DOI: 10.19892/j.cnki.csjz.2013.14.187. Li H N. Multidimensional Seismic Theory of Structures [M]. Beijing: Science Press, 2006. Li H N, Sun L Y. (2001) Rotational Components of Earthquake Ground Motions Derived from Surface Waves [J]. Earthquake Engineering and Engineering Vibration. 2001(01): 15–23. DOI: 10.13197/j. eeev.2001.01.003
167
Liu C, Zheng G. Analysis of the Influence of Ground Unequal Settlement on the Structure with Elastic Support Model Method [J]. Journal of Building Structures, 2004(04): 124–128. DOI: 10.3321/j.issn:10006869.2004.04.020. Newmark N M. Torsional in Symmetrical Building [C]. Proceedings of the 4th World Conference on Earthquake Engineering, Santiago: Chile, 1969. Wang B. Exploration and Practice of the Development and Utilization of Urban Underground Space [D]. Beijing: China University of Geosciences (Beijing), 2013. Wang C. The Influence of Uneven Settlement on Internal Force and Seismic Behavior of Steel Frame Structure [D]. Hefei: Hefei University of Technology, 2016. Xu F Z. Research on the Seismic Performance of Frame Structures with Uneven Settlement of Foundation [D]. Yinchuan: Ningxia University, 2022. Yan X Y, Wang X Z. Seismic Response Analysis of Long-span Grid Structure Considering Uneven Settlement [J]. Journal of Fuzhou University (Natural Science Edition), 2017, 45(04): 486–492. DOI: 10.7631/issn.10002243.2017.04.0486. Zhang J, Li H N, Wang L C. The Research of Multi-dimensional Seismic Response of Large Span Asymmetric Spatial Structure [J]. Earthquake Engineering and Engineering Dynamics, 2015, 35(03): 8–16. DOI: 10.13197/j.eeev.2015.03.8.zhangj.002. Zhao S W, Luo Q F. Statistical Analysis of Rotational Components in Wenchuan Earthquake Near-field Region [J]. Journal of Tongji University (Natural Science), 2014, 42(1): 9–12. DOI: 10.3969/j.issn.10003835.2013.14.032.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Numerical simulation of mechanical properties of fiber-reinforced recycled concrete beam-to-column joints under monotonic loading Tianbei Kang & Ye Yuan Department of Civil Engineering, Shenyang Jianzhu University, Shenyang, Liaoning Province, China
Jinghai Zhou* Department of Green and Livable Rural Construction Institute, Shenyang Jianzhu University, Shenyang, Liaoning Province, China
Yibo Liang, Jingtong Qu & Liwei Jin Department of Civil Engineering, Shenyang Jianzhu University, Shenyang, Liaoning Province, China
ABSTRACT: The mechanical properties, damage forms, and deformation characteristics of waste fiber recycled concrete beam-to-column joints under monotonic loadings are studied by numerical simulation. The design variables are the replacement rate of recycled coarse aggregates and the volume fraction of waste fibers. The results indicate that when the replacement rate of recycled aggregates is 50%, incorporating a certain amount of waste fibers can effectively improve the tensile strength of recycled concrete. The ultimate bearing capacity of the beam-to-column is best when the volume fraction of waste fibers is 0.12%. 1 INTRODUCTION The production of recycled coarse aggregate (RCA) concrete has decelerated the consumption of natural resources while continuing the consumption of vast amounts of construction and demolition waste. Adding fibers can inhibit the development of cracks in the cement matrix and improve the strength of recycled concrete (Carneiro et al. 2014; Yin et al. 2015). Beam-column joints are divided into corner joints, side joints, and middle joints in the frame. Najafghilipour et al. (2019) established the ultimate shear strength formula corresponding to the calculation of the core tensile strength of the joint. Candido et al. (2018) used fiber-reinforced concrete materials to reinforce the frame structure in key areas. Su et al. (2016) found that adding high-toughness fiber can effectively control the number of cracks in the core area of the joint. The same conclusion was reached by Du et al. (2016). This paper used waste fiber (WF) as reinforcing fibers for recycled concrete. The mechanical properties, damage forms, and deformation characteristics of the beam-to-column joint of waste fiber recycled concrete under monotonic loads were investigated by numerical simulation. The simulations correspond to experiments from Zhou et al. (2019), and Zhou et al. (2022). 2 ESTABLISHMENT OF NUMERICAL MODEL 2.1
Numerical model
The numerical model was established by ABAQUS, and the specimen design by Zhou et al. (2022) was considered. The specimens’ fitting ratio and compressive strength are shown in Table 1, and the dimensions and reinforcement of the models are shown in Figure 1. *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-23
169
Table 1.
Mix proportion and compressive strength of specimens.
Serial
Natural Cube Cement/ Sand/ aggregate/ RCA/ Water/ Strengthfcu,k/ kgm3 kgm3 kgm3 kgm3 kgm3 Nmm2
Axial compressive strengthfck/ Nmm2
NCJ RCJ-19-0.12 RCJ-50-19-0.12 RCJ-50-19-0.08 RCJ-50-19-0.16 RCJ-100-19-0.12
390 390 390 390 390 390
29.24 29.23 29.22 29.22 29.22 29.19
Figure 1.
709 709 709 709 709 709
1156 1156 578 578 578 0
Size and reinforcement of specimen.
0 0 578 578 578 1156
Figure 2. mode.
195 195 205 205 205 215
43.77 43.75 43.70 43.69 43.70 43.64
Boundary conditions and loading
A force-controlled loading system was used to apply a constant axial force N at the top of the column and a monotonic loading at the beam end. The axial compression ratio of each specimen was 0.4. The specimen was deemed to yield when the load-displacement curve at the beam end had an obvious inflection point. The finite element model of the beam-tocolumn joint was mainly composed of concrete and a reinforcement cage. The C3D8R element was used for concrete, a three-dimensional 8-node hexahedron linear reduced integral part, and the T3D2 component was used for reinforcement. The boundary conditions and loading methods of the finite element model of beam-to-column joints were the same as those of the experiment, as shown in Figure 2. 2.2
Constitutive relationship
The stress-strain of waste fiber recycled concrete under uniaxial compression (Zhou & Liu 2013) is as follows: 8 2 e e > > K > > > s e e0 e0 > > > Ri sin g section : ¼ e ; 0 e0 < 1 > f > c < 1 þ ðK 2Þ e0 (1) e > > > s e > e0 > > Descending section : ¼ ; 1 2 > > f e > c e e 0 > : a 1 þ e0 e0
170
where fc is the compressive strength; e0 is strain corresponding to the peak stress of concrete, e0 = 0.00159; a, K, characteristic parameters of RCA replacement ratio r, and WF content lf are calculated by Formula (2): 8 5 4 3 > < K ¼ 0:288r2 0:3r þ 0:965ð100lf Þ 7:913ð100lf Þ þ 23:19ð100lf Þ 2 (2) 28:53ð100lf Þ þ 1221lf þ 1:815 > : a ¼ 0:37r 45:34ð100lf Þ4 þ 30:52ð100lf Þ3 65:74ð100lf Þ2 þ 46:32ð100lf Þ þ 1:386 The stress-strain constitutive relationship of recycled WF-reinforced concrete under uniaxial tension (Guo 1997): ( y ¼ 1:2x 0:2x6 ; x 1 x (3) ; x>1 y¼ a1 ðx 1Þ þ x where x = e/et; a1 = 0.312ft2; a1 is the parameter of the descending section of the concrete uniaxial tensile stress-strain curve; ft is the peak tensile stress of concrete; et is the corresponding peak tensile strain of concrete. The longitudinal reinforcement in this study is HRB400, and the stirrup is HPB300. The stress-strain curve of the reinforcement is a two-line model with an elastic-plastic hardening section.
3 RESULT ANALYSIS 3.1
Load-displacement curves
The influence of RCA on the load-displacement curve at the beam end of specimens is analyzed, as shown in Figure 3. The ultimate bearing capacity obtained by simulation is 57 N, 58 kN, and 52 kN, with errors of experiment date 4.6%, 9.0%, and 9.2%, respectively. In the elastic stage, the load-displacement curves of the push and pull sides coincide, and the RCA has little influence on the elastic stage. With the increase of load, when the RCA content is 100%, the yield state is reached first. The ultimate bearing capacity of the RCJ-50-19-0.12 specimen reaches 53 kN, which is 9.6% and 11.22% higher than that of RCJ-19-0.12 and RCJ-100-19-0.12.
Figure 3.
The influence of the RCA replacement ratio.
The influence of WFs on the load-displacement curve at the beam end of specimens RCJ50-519-0.08, RCJ-50-19-0.12, and RCJ-50-19-0.16 is analyzed, as shown in Figure 4. The simulated value of the curve is higher than the test value. The simulated ultimate bearing capacity is 48 N, 58 kN, and 52 kN, with errors of 9.5%, 9.0%, and 9.1%, respectively. When the content of WFs is 0.12%, the yield load of the beam end in the joint area is 14.1% and 2.82% higher than that of the yield load with the content of WFs of 0.08% and 0.16%. 171
Figure 4.
3.2
The influence of different WF content.
Characteristic load
The comparison of beam end cracking load Pcr, yield load Py, ultimate load Pu, and failure load Pm of specimens with different variables obtained from simulation is shown in Figure 5. The standard part NCJ is added as a reference.
Figure 5.
Characteristic load.
Figure 5 (a) shows the characteristic load of beam-to-column joints under different replacement rates of RCAs. When the RCA content is 50%, the cracking load, yield load, ultimate load, and failure load of specimen RCJ-50-19-0.12 are increased compared with RCJ-19-0.12 and RCJ-100-19-0.12, which are close to the characteristic load value of the specimen NCJ. The addition of RCA has a negative impact on the ultimate bearing capacity of the beam end in the core area of the waste fiber-reinforced recycled concrete beam-column joint. With the increase of the replacement rate of RCAs, the fluidity of fresh concrete becomes worse, but when the RCA content is about 50%, it benefits strength and bearing capacity. The WFs can improve the ultimate bearing capacity of the beam end. The results show that when the volume fraction of WFs is 0.12%, the ultimate bearing capacity and strength will be improved compared with other volume fractions, as shown in Figure 5(b). 3.3
Shear deformation of the core area
There is no shear failure in the beam-to-column specimen’s core area, and each joint specimen’s failure mode is beam-end bending failure. The shear angle of the core area of the joint is calculated by measuring the change of the diagonal length of the core area of the beam-to172
column joint g to calculate the shear deformation of the core area of the joint specimen. The value of shear angle g can be calculated according to the following formula: pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi a2 þ b2 jd1 þ d2 j þ jd3 þ d4 j (4) g¼ ab 2 where a and b represent the length and width of the core area; d1, d2, d3, and d4 represent the change of diagonal length of the core area after the deformation of the specimen. The shear angle of each node is calculated through the g. Then, the shear deformationload curve of the node core area of the specimens is compared and analyzed under different RCA replacement ratios and volume fractions of WFs. The standard illustration NCJ is added for comparison and analysis. The shear deformation-load curves of the core area of each node are shown in Figure 6.
Figure 6.
Shear deformation-load curves. (a) Replacement rate of RCAs (b) Volume fraction of WFs.
It can be seen from the comparison of specimens at each node that when the shear angle is less than 3 10-3 rad, the curve slope of each specimen is high, and the curve tends to be steep. As shown in Figure 6 (a), the growth rate of the curve slows down, with the load reaching about 30 kN. After the load reaches about 40 kN, the curve in Figure 6 (b) tends to be relatively flat. It can be seen that the volume of WFs has a significant impact on the shear deformation-load curve in the core area of the node. According to the numerical simulation results, incorporating WFs can improve the shear failure deformation of frame beamcolumn joints and enhance the ductility of concrete.
4 CONCLUSIONS The following conclusions can be drawn from the analysis of numerical simulation results: (1) All the beam-to-column specimens eventually undergo plastic hinge damage at the end of the beam under monotonic loading. (2) The addition of RCAs could be more conducive to the ultimate bearing capacity. However, when the RCA content is 50%, adding a certain amount of WFs can effectively improve the tensile strength of concrete. Therefore, the addition of WFs has a beneficial effect on the ultimate bearing capacity of the beam end. (3) As the reinforcing fiber of recycled concrete, WFs can delay the cracks’ generation, control the cracks’ development, and improve the bearing capacity of beam ends in the connection area. The results show that the ultimate bearing capacity of the beam end is the best when the volume fraction of WFs is 0.12%. 173
REFERENCES Candido L., Micelli F. (2018) Seismic Behavior of Regular Reinforced Concrete Plane Frames with Fiber Reinforced Concrete in Joints. Bulletin of Earthquake Engineering, 16 (9): 4107 – 4132. https://doi.org/ 10.1007/s10518-018-0325-9. Carneiro JA, Lima PRL, Leite MB. (2014) Compressive Stress-strain Behavior of Steel Fiber Reinforcedrecycled Aggregate Concrete. Cement and Concrete Composites, 46 (4): 65 – 72. https://doi.org/10.1016/j. cemconcomp.2013.11.006. Du YF, Wang SL. (2016) Experimental Research on the Seismic Bearing Performance of Fiber Composite Reinforced Recycled Concrete Frame Joints. Journal of Building Structures, 37 (4): 40 – 46. DOI: 10.14006/ j.jzjgxb.2016.04.006. Guo ZH. (1997) Strength and Deformation of Concrete – Experimental Basis and Constitutive Relation. Tsinghua University Press, Beijing. ISBN: 7-302-02609-2. Najafgholipour M.A., Arabi A.R. (2019) A Nonlinear Model to Apply Beam-Column Joint Shear Failure in the Analysis of RC Moment Resisting Frames. Structures, 22: 13 – 27. DOI: 10.1016/j.istruc.2019.07.011. Su J, Liu JP, Li W, Chen M. (2016) Experimental Research of Shear Behavior of Frame Joints by Fiber Reinforced Concrete Under Reversed Cyclic Loading. Journal of Earthquake Engineering and Engineering Vibration, 36 (2): 36 – 41. DOI: 10.13197/j.eeev.2016.02.36.suj.006. Yin S, Tuladhar R, Shanks RA, et al. (2015) Fiber Preparation and Mechanical Properties of Recycled Polypropylene for Reinforcing Concrete. Journal of Applied Polymer Science, 41866: 1 – 10. https://doi.org/ 10.1002/app.41866. Zhou JH, Liu D. (2013) Constitutive Relationship of Waste Fiber Recycled Concrete. Concrete, (2): 54–58. DOI: 10.3969/j.issn.1002-3550.2013.02.015. Zhou JH, Kang TB, Wang FC. (2019) Pore Structure and Strength of Waste Fiber Recycled Concrete. Journal of Engineered Fibers and Fabrics, 14: 1 – 10. https://doi.org/10.1177/1558925019874701. Zhou JH, Jin LW, Qu JT, Sun H, Kang TB, Yuan Y, and Liu Y. (2022) Experimental Research on Waste Fiber Recycled Concrete Beam-to-Column Joints Under Monotonic Loading. Advances in Materials Science and Engineering, 2022: 2240624. https://doi.org/10.1155/2022/2240624.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Prediction and experimental research on fatigue life of corroded prestressed concrete beams Lizhao Dai* & Jingjin Liu School of Civil Engineering, Changsha University of Science and Technology, Changsha, China
ABSTRACT: The combined effects of repeated load and corrosion can cause fatigue damage to prestressed concrete beams, which would reduce their service life. The fatigue cracks growth size of the strand and cumulative residual strain of concrete were taken as damage parameters in the present study, a fatigue life prediction method of corroded prestressed concrete beams was proposed, and the method comprehensively considered the influence of fatigue crack growth of steel strand, interface corrosion fatigue bond degradation and fatigue damage of concrete. Then the rationality of the fatigue life prediction method was verified by the experimental data. Results show that the experimental values of fatigue life are in good agreement with the predicted values, which proves that the proposed prediction method could effectively predict the fatigue life of corroded prestressed concrete beams.
1 INTRODUCTION Prestressed concrete structures have long been considered good bonding properties and service capabilities due to their lightweight, high strength, and good overall compactness. The existing bridge structure has been in an environment of high chloride ion content for a long time, and at the same time, it is subjected to the repeated action of vehicle load. These combined effects accelerate the corrosion of prestressed tendons in the structure, cause the concrete protective layer to crack, and lead to structural damage and fatigue accumulation. Corrosion of prestressed reinforcement and concrete cracking would affect the safety and service life of the structure. Some scholars have studied the fatigue test of prestressed concrete beams and analyzed the failure mode of cross-section under fatigue load, crack width, deflection variation, and fatigue life. Through the axial tensile fatigue test, Yu et al. (2014) found that corrosion seriously affected the fatigue life of steel strands, and the attenuation of fatigue life under the same load changed exponentially with the increase in the corrosion rate. Du et al. (2020) found that the fatigue failure mode of PC beams is the fatigue fracture of the bottom tensile steel bar on the pure bending section by carrying out the fatigue test of prestressed concrete beams. Under the action of constant amplitude fatigue load, the main crack width, stiffness degradation, and maximum displacement show three stages of development. The above discussions mainly studied the fatigue life and failure mode of corroded PC beams under fatigue load through experiments. Currently, there are also some theoretical models for calculating fatigue life both domestically and internationally, but most of these models only consider a single influencing factor. For example, based on the Paris formula, the fatigue life at a given frequency is calculated by considering the fatigue crack initiation *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-24
175
and fatigue crack propagation caused by corrosion (Guo et al. 2019). According to the test data, the S-N curve of the corroded steel strand is fitted to predict fatigue life (Miarka et al. 2022). However, considering a single influencing factor does not accurately reflect the stress state of the concrete structure in the actual situation, the fatigue performance degradation of corroded PC beams is a multi-factor process. The combined effect of corrosion and fatigue would cause bond strength degradation between steel strands and concrete, the fatigue damage of concrete in the compression zone of PC beams, and the stress concentration at the root of rust pits. These will lead to stress redistribution inside the PC beam, increase the actual stress of concrete and force bars, and accelerate the fatigue crack growth rate. The PC beam may have fatigue damage before reaching the expected service life. Therefore, it is necessary to consider multiple factors simultaneously to improve the accuracy of fatigue life prediction of corroded PC beams. In this paper, a fatigue life prediction method is proposed, which can comprehensively consider the influence of fatigue crack propagation of steel strands, corrosion fatigue bond degradation of the interface and fatigue damage of concrete. Then, the rationality of the method is verified by the fatigue test data of corroded PC beams.
2 FATIGUE LIFE PREDICTION METHOD FOR CORRODED PC BEAMS 2.1
Fatigue crack propagation model of corroded steel strand
The existing research on fatigue crack propagation is mostly focused on steel bars and steel wires, and there is little research on corrosion crack propagation of prestressed steel strands. Because the steel strand is made of 7 steel wires, the fatigue crack propagation model of steel wire is used for analysis in this paper. The formula is as follows (Liao et al. 2022): da=dN ¼ C ðK Kth Þm
(1)
where da/dN is the crack growth rate; C and m are the coefficient and index of crack propagation, respectively. DKth is the threshold value of crack propagation, and DK is the amplitude of the stress intensity factor; a is the crack depth; N is the number of fatigue cycles. The equation of crack propagation life can be obtained by modifying Equation (1): ð Ns ð ac 1 Ns ¼ dN ¼ (2) m da 0 ai cðK Kth Þ where ai is the equivalent initial crack size; ac is the critical crack size. The crack propagation threshold value DKth, crack propagation parameters C and m can be expressed as (Zheng et al. 2017) kth ¼ 1:6 103 sy þ 6:42 1:87R
(3)
C ¼ 6:18 1013 sy 3:31 1010
(4)
m ¼ 6:79 104 sy þ 2:38
(5)
The calculation formula of stress intensity factor amplitude DK proposed by Liu et al. (2009) is suitable for life prediction and crack propagation analysis of long cracks. rffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi n h a ioffi K ¼ Y s p a þ d 1 exp ðKt2 1Þ (6) d
176
where Ds is the stress amplitude of the steel strand; a is the crack depth; d is pit depth; Kt is the stress concentration factor. The expression of the shape correction coefficient of the stress intensity factor of the steel wire is expressed as (Qiao et al. 2017): Y
a
a a 2 a 3 a 4 ¼ 0:587 þ 1:133 13:45 þ 64:44 53:63 D D D D D
(7)
Under cyclic loading, the root of the rust pit is prone to stress concentration, which promotes the initiation and propagation of fatigue cracks in the root of the rust pit, resulting in a decrease in the service life of the PC bridge structure. The expression of the relationship between the pit depth and the stress concentration factor is as follows: Kt ¼ 2:663 ðd þ 0:0624Þ0:353 :
2.2
(8)
Assessment of corrosion fatigue bond degradation
2.2.1 Fatigue bond stress Both corrosion and repeated loading can cause degradation of bond strength. Therefore, the effective bond stress of corroded steel strands after experiencing cyclic loading must be reevaluated. In this paper, the equivalent bilinear model proposed by Wang et al. (2017) is used to calculate the average bond stress tave of uncorroded steel strands. Wang et al. (2017) also proposed a degradation prediction model of the bond stress with the change in the corrosion rate: 1:0h 6% Rc ðhÞ ¼ : (9) 2:03e0:118h h > 6% Fatigue would lead to the gradual deterioration of bonding performance and even lead to fatigue bonding failure. This paper uses a formula reflecting the relationship between fatigue load and bond strength of steel bars to calculate. The formula is as follows: 1:0 1gn 4 Rf ðnÞ ¼ (10) 0:0685log10 n þ 1:274 lgn > 4 Based on the above discussion, the average bond strength taven of corroded steel strands after cycles n is expressed as follows: taven ¼ Rc ðhÞRf ðnÞtave
(11)
Therefore, the effective bond force Feb between corroded steel strand and concrete after fatigue load cycles n is expressed as: Feb ¼ leb Lp taven
(12)
where leb is the effective bond length; Lp is the circumference length of the steel strand section. 2.2.2 Calculation method considering residual strain of reinforcement under cyclic loading Under cyclic loading, the degradation of bond strength will lead to the bond-slip between the steel strand and concrete, and the bond-slip would lead to the residual strain of the steel strand and steel bars. With the increase of fatigue cycles, the cumulative residual strain in the steel strand and steel bars increases significantly, which increases the actual stress of steel 177
strands and steel bars and leads to fatigue failure of PC beams without reaching the expected service life. Therefore, the influence of residual strain on the fatigue life of PC beams must be considered. When the cycle is assumed to be n, the strains of the steel strand and steel bar in the bondslip zone based on the plane section assumption are ep(n) and es(n) respectively. The strains of the steel strand and steel bar considering residual strain are expressed as: ( 0 ep ðnÞ ¼ ep ðnÞn ¼ 1 (13) 0 ep ðnÞ ¼ ep ðnÞ þ epr ðn 1Þn > 1
0
es ðnÞ ¼ es ðnÞn ¼ 1 es ðnÞ ¼ es ðnÞ þ esr ðn 1Þn > 1
(14)
0
where epr(n1) and esr(n1) are the residual strain of steel strand and steel bar when the cycles reach n-1, respectively. By assuming that the deformation value of steel strands and steel bars is evenly distributed in the length range of the cracking section and the tensile limit state section, the residual strains of the steel strands and steel bars are caused by the strain difference between the reinforcement and the concrete, the residual strain formulas of the steel strand and the steel bar after n cycles are expressed as: sffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi ffi 2
Ap 2Afp ðnÞ
epr ðnÞ ¼
þ
Ap 2Afp ðnÞ
2pdp Ebf ðnÞWpr;n
þ
npf fct ðnÞAfp ðnÞ
fct ðnÞ (15)
2Ebf ðnÞ s ffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi 2
esr ðnÞ ¼
2.3
AS 2AfS ðnÞ
þ
AS 2AfS ðnÞ
þ
2pds Ebf ðnÞWsr;n
2Ebf ðnÞ
nsf fct ðnÞAfS ðnÞ
fct ðnÞ :
(16)
Fatigue damage of concrete
As an uneven composite material, concrete has original defects inside. Cyclic loading would gradually develop these defects, resulting in residual strain inside the concrete, further causing fatigue damage accumulation. Some scholars regard the cumulative residual strain of concrete ecr reaching a certain value as the basis for the failure of concrete under cyclic loading (Wei et al. 2016). ecr 0:4
fc Eb
(17)
where Eb = b Ec, Ec and Eb are the initial compressive elastic modulus and initial flexural compressive deformation modulus of concrete respectively. b is the initial flexuralcompressive elastic modulus ratio of concrete. The cumulative residual strain of concrete ecr in the compression zone of PC beams under cyclic loading is expressed as: fc ecr ¼ Ec
(
)0:5382 m X
1 ni lg 4:0935ar;i 8:5576 : i¼1
178
(18)
When ecr is determined, the flexural compressive fatigue elastic modulus of concrete Ebf is expressed as: Ebf ¼
sfc max sfc max Eb
2.4
þ ecr
:
(19)
Calculation process of fatigue life prediction for corroded PC beams
The fatigue bonding degradation and the fatigue damage of concrete can cause stress redistribution in PC beams. Although the PC beam is subjected to constant amplitude cyclic loading, it would produce variable amplitude fatigue stress in the beam. Therefore, it is appropriate to analyze the fatigue crack growth of steel strands in stages. Assuming that the characteristics of the section and the material are stable in each stage, the stress calculation of the fatigue bending control section of the beam is carried out according to the linear elastic method. When each stage of calculation is completed, the relevant parameters are updated. The expression of fatigue crack propagation stage analysis of corroded steel strands is expressed as: ð au 1 Ns ¼ (20) m da al cðK Kth Þ where Ns is the number of fatigue cycles in a calculation stage; al and au are the lower limit and upper limit of integral respectively. To clearly illustrate the proposed prediction method, Figure 1 shows the calculation flow chart of the fatigue life prediction method for corroded PC beams.
Figure 1.
Calculation flow chart of fatigue life prediction.
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3 EXPERIMENTAL PROGRAM AND MODEL VALIDATION 3.1
Specimen details
The experimental program consists of testing six prestressed concrete beams (S1-S6). All beams are designed with a rectangular cross section of 200 350 mm, and a total length of 3800 mm. A single 15.2 mm Grade 1860 prestressing strand with a concrete cover of 75 mm on the bottom side was used in all beams. Each beam was also reinforced with two 16 mm deformed reinforcing bars in the tension zone, two 10 mm bars in the compression zone, and 8 mm closed stirrups at 100 mm spacing. In addition, the stirrup spacing at the end of the beams was decreased to 70 mm. The thickness of the protective layer of deformed reinforcing bars was 30 mm. To measure the slip of the mid-span steel strand, a hollow area of 500 60 100 mm was set at the bottom of the beam span. Figure 2 shows the beam details.
Figure 2.
Reinforcement and details of beams.
The beam S1 was uncorroded and used as a control specimen. The electrochemical method was used for accelerating the corrosion of the steel prestressing strand in beams S2 to S6. The beams were cured for 28 days before being subjected to the electrochemical corrosion process. A local corrosion groove was installed at a distance of 650 mm from the beam end, and the groove length was 400 mm. Different local corrosion rates were obtained by controlling the power-on time and current size. After the test, the corrosion rates of corroded beams S2-S6 were 4.6%, 10.4%, 14.6%, 19.7% and 36.5% respectively. 3.2
Loading procedure
The fatigue tests were conducted on the MTS servo-hydraulic system. The upper and lower limits of cyclic loading are 0.5Pu (70 kN) and 0.1Pu (15 kN), respectively, and Pu is the ultimate load of non-corroded specimens under static loading. The stress ratio is 0.2. The cyclic loading was interrupted periodically at a predetermined number of cycles and then, a static cycle was applied between zero and Pmax to verify the condition of the beams. Each time the fatigue tests were interrupted for static loading, the orientation of the cracks was observed, and crack width was also measured during the static loading. When any of the main bars in the test beam has fatigue fracture, that is, the change of the test load exceeds the limit, the actuator will automatically unload to prevent the test beam from continuing to bear the load, and the test will be terminated immediately. If the beam has not been damaged after 2 million cycles, the beam continues to be loaded until static failure occurs. 3.3
Concrete crack development
The concrete crack development under the fatigue test is shown in Figure 3. The numbers in Figure 3 represent the number of cycles when cracks are generated. Whether it is a corroded beam or an uncorroded beam, multiple cracks appear in the bottom tension zone during the
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first static loading process, and the crack spacing is about 100 mm to 200 mm. During the cyclic loading process, the crack width and depth increase with the number of cycles. Because the test beams are subjected to local rather than complete beam corrosion, the development of cracks on both sides is asymmetric. After the fatigue tests, compared with the uncorroded beam S1, the corroded beams S2-S6 produce 3 to 4 vertical cracks in the corroded area. The reason for the above phenomenon is that the expansion of corrosion products leads to concrete cracking, reducing the tensile properties of concrete in this area. Therefore, vertical cracks appear in the corroded area under cyclic load. Throughout the fatigue stage, the crack evolution of the test beam conforms to the threestage development law of rapid increase, stable development, and rapid change. The rapid increase and rapid change stages of cracks account for the first 10% and the last 10% of fatigue life, respectively. It can be seen from Figure 3 that the rapid increase stage of beams S1-S3 basically ends within 200, 000 cycles, and the rapid increase stage of test beams S4-S5 ends within 50, 000 cycles.
Figure 3.
Crack development of test beams.
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3.4
Failure modes and fatigue life
Figure 4 shows the failure modes of each test beam. Both the uncorroded beam S1 and the slightly corroded beam S2 do not occur fatigue failure after 2 million cycles, and then the static load failure test is carried out. The characteristics of the failure are that the steel strand is broken, the concrete near the loading point is crushed, and an unrecoverable main crack is formed in the mid-span, which is ductile damage. The results show that the cyclic loading will not substantially impact the static loading failure mode of beams.
Figure 4.
Failure modes of test beams. (1) S1 (2) S2 (3) S3 (4) S4 (5) S5 (6) S6.
The failure of the test beams S3-S5 occurred in the fatigue stage. The failure was a sudden fracture of the steel strand wire in the mid-span, accompanied by a brittle sound, the deflection suddenly became larger, and a wider main crack was formed in the mid-span of the beams, which was a brittle fatigue failure. The test beam S6 has severe corrosion of steel bars due to improper anti-rust treatment of steel bars before electrochemical corrosion. When the cyclic loading was applied to 20, 000 cycles, the steel bar was suddenly broken, an unrecoverable main crack was formed in the corroded area, and the crack extended to about 93% of the height of the beam section. The fatigue life and failure mode of each test beam are shown in Table 1. It can be seen from Table 1 that the corrosion rate has a significant effect on the fatigue life of beams. With the increase of corrosion rate, the fatigue life of the PC beam decreases rapidly. The reason for the above phenomenon is that the rust pit will deepen with the increase of the corrosion rate, resulting in a more severe stress concentration, thereby accelerating the fatigue fracture of the steel strand. Table 1.
Corrosion rate, fatigue life and failure modes of test beams.
No.
S1
S2
S3
S4
S5
S6
Corrosion rate (%) Fatigue life (104) Decline ratio (%) Failure modes
0 200 1 Steel strand ruptured
4.6 200 1 Concrete crushed
10.4 119.3 40.4 Steel strand ruptured
14.6 86.5 56.8 Steel strand ruptured
19.7 40.9 79.6 Steel strand ruptured
36.5 2.1 99 Steel bar ruptured
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3.5
Model validation
To verify the rationality of the fatigue life prediction method in this paper, the above fatigue data were used to verify the prediction method. Figure 5 shows the comparison of fatigue life between the test results and the predicted results. Test beam S6 was not included in the fatigue life prediction analysis because of the test result error. The average error was 5%, indicating that the experimental values of fatigue life are in good agreement with the predicted values. When the corrosion rate is 19.7%, the error is the largest, which is 8%. These test errors may be caused by the inhomogeneity of material properties, improper operation of the cyclic loading test, and the uncertainty of the analytical calculation model. The prediction results show that the fatigue failure of beams S1 and S2 does not occur after 2 million cyclic loadings, and the fatigue failure modes of beams S2-S5 are all steel strand fractures, which are consistent with the experimental phenomena.
Figure 5.
Comparison between test value and predicted value of test beams.
In addition, to further verify the rationality of the fatigue life prediction method in this paper, the fatigue test results of corroded PC beams carried out by Zhang et al. (2016) and Su et al. (2022) are used to verify the prediction method. The results are shown in Table 2. The results show that the average error is 4%, which is in good agreement with the experimental value. It can be seen from the above discussion that the theoretical model established in this paper can reasonably predict the fatigue life of corroded PC beams. Table 2.
Literature test parameters and prediction results.
Corrosion rate (%)
Test fatigue life Fmax (kN) Fmin (kN) N (104)
Predicted fatigue life Npre (104)
Npre =N Reference
0 1.3 4 3.7 2.5 5.6 4.1 6.8 8.2 10.8
85.7 85.7 81.0 101.0 85.7 85.7 94 94 94 94
190.2 106.5 25.4 83.1 39.6 23.1 163.9 98 56.4 42.8
0.966 0.973 0.981 1.004 1.008 1.065 1.006 0.902 0.940 1.065
14.3 14.3 13.5 16.9 14.3 14.3 18.8 18.8 18.8 18.8
196.9 109.4 25.9 82.8 39.3 21.7 162.9 108.7 60.0 40.2
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Zhang et al. (2016)
Su et al. (2022)
4 CONCLUSIONS 1. A fatigue life prediction method of corroded PC beams was proposed in this paper, which comprehensively considers the influence of fatigue crack growth of steel strands, interface corrosion fatigue bond degradation, and fatigue damage of concrete. 2. The test results show that the corrosion rate has a significant effect on the fatigue life of beams, the fatigue life of PC beams decreases rapidly with the increase of corrosion rate. Compared with the uncorroded beam, the fatigue life of the beam with a corrosion rate of 19.7% is reduced by about 80%. 3. The rationality of the method was verified by the experimental data. The results show that the experimental values were in good agreement with the predicted values. The theoretical model established in this paper can reasonably predict the fatigue life of corroded PC beams.
REFERENCES Du, Y. X., Wei, J., Yuan, J., Lai, Y. F. & S, D. H. (2020) Experimental Research on Fatigue Behavior of Prestressed Concrete Beams Under Constant-amplitude and Variable-amplitude Fatigue Loading. Construction and Building Materials, 259: 119852. Guo, Z. Z., Ma, Y. F., Wang, L. & Zhang, J. R. (2019) Modelling Guidelines for Corrosion-fatigue Life Prediction of Concrete Bridges: Considering Corrosion Pit as a Notch or Crack. Engineering Failure Analysis, 105: 883–895. Liao, X. X. & Li, Y. D. & Qiang, B. & Wu, J. & Yao, C. R. & Wei, X. (2022) An Improved Crack Growth Model of Corrosion Fatigue for Steel in Artificial Seawater. International Journal of Fatigue, 160: 106882. Liu, Y. M. & Mahadevan, S. (2009) Fatigue Limit Prediction of Notched Components Using Short Crack Growth Theory and an Asymptotic Interpolation Method. Engineering Fracture Mechanics, 76 (15): 2317– 2331. Miarka, P. & Seitl, S. & Bilek, V. & Cifuentes, H. (2022) Assessment of Fatigue Resistance of Concrete: S-N Curves to the Paris’ Law Curves. Construction and Building Materials, 341: 127811. Qiao, Y., Miao, C. Q. & Sun, C. Z. (2017) Evaluation of Corrosion Fatigue Life for Corroded Wire for the Cable-supported Bridge. Journal of Civil and Environmental Engineering, 39 (04): 115–121. Su, X. C., Ma, Y. F., Wang, L., Guo, Z. Z. & Zhang, J. R. (2022) Fatigue Life Prediction for Prestressed Concrete Beams Under Corrosion Deterioration Process. Structures, 43: 1704–1715. Wang, L., Zhang, X. H., Zhang, J. R. & Yi, J. (2017) Simplified Model for Corrosion-induced Bond Degradation Between Steel Strand and Concrete. Journal of Materials in Civil Engineering, 29 (4): 04016257. Wei, J., Li, S. L.,, Dong, R. Z., Liu, X. C. & Wu, Z. Q. (2016) Fatigue Damage Constitutive Model of Concrete Considering the Effect of Residual Deformation. Journal of Hunan University (Natural Sciences), 43 (7): 57–61. Yu, F., Jia, J. Q., Yao, D. L., Wu, F. (2014) Experimental Analysis of Fatigue Properties of the Corroded Prestressing Strand. Journal of Harbin Engineering University, 35 (12): 1487–1491 + 1502. Zhang, W. P. & Liu, X. G. & Gu, X. L. (2016) Fatigue Behavior of Corroded Prestressed Concrete Beams. Construction and Building Materials, 106: 198–208. Zheng, X. L. & Xie, X. & Li, X. Z. & Qian, L. Q. & Shen, Y.G. (2017) Estimation Model for Steel Wire Crack Propagation and its Application in Calculation of Pre-corrosion Fatigue Life. China Civil Engineering Journal, 50 (03): 101–107.
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Size effect of fly ash geopolymer concrete at different temperatures Shuangxing Wang* The Xi’an University of Architecture and Technology, Xi’an, China
ABSTRACT: Fly ash geopolymer concrete, as a new green building material, has good high-temperature resistance. The size effect phenomenon widely exists in various materials, and it is of great engineering significance to explore the size effect behavior of fly ash geopolymer concrete, a high-temperature resistant material, at high temperatures. This paper studies the size effect rule of fly ash geopolymer concrete cubes with different proportions at different temperatures. The research results show that the compressive strength of fly ash geopolymer concrete has an obvious size effect, and under high temperatures, its size effect will weaken with the increase in temperature. And the modified Bažant size effect theoretical formula can reflect the size effect behavior of compressive strength at high temperatures.
1 INTRODUCTION With the continuous advancement of the national urbanization process and the rapid development of the construction industry, concrete is widely used in the construction of structures due to its good durability, frost resistance, impermeability, weathering resistance, corrosion resistance, and other advantages. In China, the number of concrete projects per year is more than the total of the world. However, it is accompanied by the rapid consumption of cement materials, which will produce a large amount of CO2 during the preparation of concrete, accounting for one-fifth of the national industrial carbon emissions, and has a huge impact on the environment and climate (Duxson et al. 2007). A large number of mineral admixtures is used to partially replace cement or alkali-activated geopolymer concrete is used to completely replace cement to reduce the carbon emissions of the cement industry (Ding et al. 2016; Papadakis et al. 1991). Using fly ash to replace part of cement in concrete has become an important direction in civil engineering research. Compared with ordinary concrete, fly ash geopolymer concrete has better high-temperature resistance and can be regarded as a good building material under the background of frequent fire (Hai et al. 2020; Kong & Sanjayan 2010; Shaikh & Vimonsatit 2014). Some achievements have been made on the high-temperature performance of concrete (Ahmad et al. 2018; Kowalski 2010; Omer 2009), but the high-temperature performance of geopolymer concrete still needs further study. Therefore, this paper selects fly ash geopolymer concrete and studies its mechanical properties at high temperatures. The size effect phenomenon widely exists in concrete materials. In practical engineering, the size of the structure is usually large, while under laboratory conditions, small size specimens are generally used. There are still problems in predicting the performance of largesize structures from small-size specimens. Therefore, it is of great engineering significance to explore the size effect behavior of fly ash geopolymer concrete, a high-temperature resistant material, at high temperatures. *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-25
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2 TEST 2.1
Specimen model
The side length of the fly ash geopolymer concrete cube is 70.7 mm, 100 mm and 150 mm respectively. Each size of the test piece contains four temperature classes of 20 C, 200 C, 400 C, and 600 C. Three groups of test pieces of each specification are made, so the total number of fly ash geopolymer concrete cube test pieces is 36 groups. 2.2
Raw materials and mix ratio
The components of fly ash, slag, and silica fume are shown in Table 1. Water glass is made of liquid water glass, the original modulus is 3.3, the coarse aggregate is made of 5 to 12 continuous graded gravel, the fine aggregate is made of natural river sand screened according to the standard, and the test water is made of ordinary tap water. The combination of fly ash geopolymer concrete is shown in Table 2. Table 1.
Chemical composition table determined by XRF %.
Composition
SiO2
Al2O3
Fe2O3
CaO
Na2O
SO3
Fly ash Slag Silica fume
52.2 28.46 94.33
29.9 13.42 0.272
5.71 0.369 0.133
4.26 45.09 1.73
0.862 0.524 0.361
1.41 1.8 0.124
Table 2.
Ratio design of fly-ash-based geopolymer concrete kg•m3.
Fly ash
Slag
Water glass
NaOH
Water
Fine aggregate
Coarse aggregate
200
200
60
24
150
670
1030
2.3
Preparation and curing of test pieces
Before making concrete, sodium hydroxide solution is prepared at a certain concentration and mixed with a water glass to form an alkali solution of a different modulus. It is mixed evenly and sealed for 24 hours. At the beginning of preparation, an electronic scale is used to calculate and weigh all kinds of raw materials with corresponding mass at one time according to the mix proportion in Table 2. First, the slag and fly ash of corresponding quality is weighed and put into the mixer for mixing for 3 min pouring the weighed fine aggregate into the mixer for mixing for 1min, pouring the mixed solution of sodium hydroxide solution and water glass prepared in advance the day before into the mixer for mixing for 1 min, making it fully mixed, and finally pouring the coarse aggregate, continuously mixing until it is evenly mixed, finally putting it into the mold and place it on the vibration table for vibration and compaction. The test piece shall be demoulded 24 hours after molding and put into the standard curing room ((20 1) C, relative humidity 95%) for curing. The curing period is 28 days. 2.4
Test device and test method
The high-temperature test and compressive strength test of fly ash geopolymer concrete were carried out on an XL-21 high-temperature resistance furnace and TYA-2000 pressure testing
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machine respectively. The test device is shown in the figure. The compression test is carried out according to the standard test method. To avoid the impact of the loading rate on the compressive strength of the test piece, the test piece is loaded at 0.5 MPa/s.
Figure 1.
XL-21 high temperature resistance furnace.
Figure 2.
TYA-2000 pressure testing machine.
3 TEST RESULT The compressive strength of the fly ash geopolymer concrete cube in this test is shown in the following table. The number of test pieces in the table is in the form of Cx-y, where C represents cube test pieces, x represents different temperature grades, and y represents the size of test pieces. For example, C20-70.7 represents the fly ash geopolymer concrete test piece with a side length of 70.7 at 20 C. The test results are shown in Table 3. Table 3.
Average compressive strength of fly ash geopolymer concrete (MPa).
Test piece number
C20-70.7
C200-70.7
C400-70.7
C600-70.7
C20-100
C200-100
Compressive strength Test piece number Compressive strength
33.96 C400-100 35.11
36.02 C600-100 15.77
33.61 C20-150 26.2
13.8 C200-150 45.97
30.34 C400-150 39.19
41.33 C600-150 16.69
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4 SIZE EFFECT OF FLY ASH GEOPOLYMER CONCRETE The change in compressive strength of fly ash geopolymer concrete with different sizes at the same temperature is shown in Figure 3. It can be seen from the figure that the compressive strength of fly ash geopolymer concrete has an obvious size effect, and the size effect under normal temperature and high temperature shows different rules. At normal temperature, its compressive strength decreases with the increase in size. At high temperatures, its compressive strength increases with the increase in size. The degree of this increase can be expressed by the slope k, and the size of k decreases with the increase in temperature. In other words, the size effect becomes insignificant with the increase in temperature.
Figure 3.
Linear regression analysis of cube compressive strength of fly ash geopolymer concrete.
According to the size effect theory based on the energy release criterion (Bažant & Planas 1997), the size effect of quasi-brittle materials is caused by the dissipation of strain energy during macro-crack propagation in specimens under load. Based on the energy balance and deformation coordination conditions, the relationship between the compressive strength of concrete specimens and the size of concrete specimens under normal temperature is as follows: 0
Bft sN ¼ pffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi 1 þ D=D0
(1)
where sN = compressive strength; D = the size of specimens; B and D0 = geometric constants 0 dependent on concrete material; ft = tensile strength. By converting the formula, a logarithmic curve describing the size effect can be obtained: lgðsN =Bfc Þ ¼ lgð1 þ D=D0 Þ1=2 : 0
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(2)
However, in the case of high temperatures, the formula needs to be modified, assuming that the double logarithmic curve under high temperatures is: 0
lgðsN =Bfc Þ ¼ lgð1 þ D=D0 Þm
(3)
where m = the reaching coefficient. Figure 4 shows the comparison between the theoretical values and the actual values corresponding to different m values. It can be found that when m is between 0.3 and 0.7, the data points of pressure resistance degree at 200 C are in good agreement with the curve. When m ranges from 0.2 to 0.5, the data points of resistance to pressure at 400 C are in good agreement with the curve. When m ranges from 0.2 to 0.5, the data points of resistance to pressure at 600 C are in good agreement with the curve. It is concluded that the modified Bažant size effect theoretical formula can reflect the size effect behavior of compressive strength at high temperatures.
Figure 4.
Comparison of test data points and theoretical values.
5 CONCLUSIONS In this paper, the compressive strength of fly ash geopolymer concrete cubes at different temperatures and sizes is tested. Through the analysis of the test results, it is found that the compressive strength of fly ash geopolymer concrete has an obvious size effect phenomenon, and has different size effect rules under normal temperature and high-temperature conditions. Under high-temperature conditions, the apparent degree of size effect decreases with the increase of temperature. And the modified Bažant size effect theoretical formula can reflect the size effect behavior of compressive strength at high temperatures. This provides an important reference for designers to carry out the high-temperature resistance design of relevant structures in the future.
REFERENCES Ahmad, Shamsad, Rasul, Mehboob, Adekunle, Saheed Kolawole, et al. Mechanical Properties of Steel Fiberreinforced UHPC Mixtures Exposed to Elevated Temperature: Effects of Exposure Duration and Fiber Content [J]. Composites, Part B. Engineering, 2019, 168 (Jul. 1): 291 – 301. DOI: 10.1016/j.compositesb.2018.12.083. Bažant ZP, Planas J. Fracture and Size Effect in Concrete and Other Quasibrittle Materials [J]. Epfl, 1997. DOI: http://dx.doi.org/,
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Duxson, P., et al. “The Role of Inorganic Polymer Technology in the Development of ‘Green Concrete’” Cement & Concrete Research 37. 12 (2007): 1590 – 1597. DOI: 10.1016/j.cemconres.2007.08.018, Ding Y, Dai J G, Shi C J. Mechanical Properties of Alkali-activated Concrete: A State-of-the-art Review [J]. Construction & Building Materials, 2016, 127 (Nov. 30): 68 – 79. DOI: 10.1016/j.conbuildmat.2016.09.121, Hai Y, Ghq C, Vk B, et al. Spalling Behavior of Metakaolin-fly Ash Based Geopolymer Concrete Under Elevated Temperature Exposure [J]. Cement and Concrete Composites, 106. DOI: 10.1016/j. cemconcomp.2019.103483, Kong DLY, Sanjayan JG. Effect of Elevated Temperatures on Geopolymer Paste, Mortar and Concrete. Cement and Concrete Research 2010; 40 (2): 334 – 339. DOI: 10.1016/j.cemconres.2009.10.017, Kowalski R. Mechanical Properties of Concrete Subjected to High Temperature, Architecture Civil Engineering Environment [J]. Silesian University of Technology, 2010. http://www.mendeley.com, Omer, Arioz. Retained Properties of Concrete Exposed to High Temperatures: Size Effect [J]. Fire and Materials, 2009. DOI: 10.1002/fam.996, Papadakis V G, Vayenas C G, Fardis M N. Fundamental Modeling and Experimental Investigation of Concrete Carbonation [J]. ACl Materials Journal, 1991, 88 (4): 363 – 373. DOI: 10.14359/1863, Shaikh F. U. A. and Vimonsatit V. Compressive Strength of Fly-ash-based Geopolymer Concrete at Elevated Temperatures. Fire and Materials, 2014. DOI: 10.1002/fam.2240,
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
The key problems of arch supported roof structural system design of a sports building in Xiaogan Yifeng Wu*, Han Ji, Jun Dong, Hongsheng Li, Wei Wang & Zhifang Li Central-South Architectural Design Institute Co., Ltd, Wuhan, China
ABSTRACT: A sports building is composed of a gymnasium and a natatorium and the roof shape is a free-form curved surface. The overall architectural scheme takes an “arch supported roof” as the shape, emphasizing the structural logic of taking a longitudinal arch as the main part and a transverse truss as the auxiliary part. The structural design cooperates with the architect to adopt the structure system of the Vierendeel truss arch with a plane truss perpendicular to the arch. To ensure the indoor effect of the building, the plane truss is not provided with a vertical secondary truss, and the top chord continuous purlin and the bottom chord cross-arranged continuous steel tie rod are used as the out-of-plane support of the plane truss. In the design, the key issues such as the arch foot horizontal thrust, the 245 m super long plane temperature effect, the stability of the Vierendeel truss arch and the roof plane truss, and the feasibility of the steel tie rod as the lower chord support of the truss were analyzed and studied. Based on the analysis results, engineering judgments were made and the corresponding structural measures were taken. The results show that the structural design is reasonable, feasible, safe and reliable.
1 INTRODUCTION This project is located in Xiaogan City, Hubei Province. The building is composed of an 8052-seat gymnasium and a 1503-seat natatorium, and the total construction area is about 52875 m2. The outer contour of the gymnasium is an arc shape with a diameter of about 130 m, and the natatorium is a trapezoid shape with a width of 45 m to 88 m and a length of 160 m. The combination of the two venues forms a plane shape similar to “9”. The main area of the building is 2 floors above the ground with 3–4 floors locally, and the swimming pool area is partially equipped with a basement. The shape of the roof is a free-form curved surface, the maximum projected size of the roof is about 125 m 293 m, and the vertex elevation is 34.40 m. The architectural rendering effect is shown in Figure 1. The design service life of the structure is 50 years with a structural importance factor of 1.1 (GB50068 2018), the seismic precautionary intensity is 6 degrees (GB50011 2016), the design basic acceleration of ground motion is 0.05 g, the site category is Class III, and the seismic fortification category is the key fortification category. The project has completed the wind tunnel test, and by comparing the wind tunnel test results with the wind load calculation results of the code (GB50009 2012), data such as base shear and roof wind suction in the wind tunnel test are less than the code results. The design wind load is taken as per the envelope of the code load and wind tunnel test results. The basic wind pressure is taken as 0.4 kPa based on 100 years, and the ground roughness is Class B.
*Corresponding Author: [email protected] DOI: 10.1201/9781003425823-26
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Figure 1.
Architectural rendering effect.
2 STRUCTURAL SYSTEM The main functions of the building are the spectators’ stand and subsidiary functional rooms. The main structure adopts a reinforced concrete frame structure, the floor adopts a reinforced concrete beam and slab system, and the seismic rating of the frame is Grade 2. To cooperate with the architectural effect, no seismic joint is set between the two venues, and the longest unit of the concrete structure is about 245 m. In the design, a U-shaped groove is added, an expansion post-cast strip is set, and temperature reinforcement is added considering the temperature load condition to resist the adverse effects of temperature. The overall architectural scheme takes “arch supported roof” as the shape, emphasizing the structural logic of taking a longitudinal arch as the main part and a transverse truss as the auxiliary part. The following two requirements are put forward for the structural design (1) The main arch is arranged along the longitudinal direction, which needs to be set in accordance with the roof’s architectural surface. The form of a truss arch can be adopted, but the diagonal web member is not allowed; (2) The auxiliary truss is arranged horizontally. To ensure the simplicity of the structure, it is not expected to use the space truss and the longitudinal secondary truss is not allowed in the indoor area. The interior rendering effect of the building is shown in Figure 2.
Figure 2.
Interior rendering effect of the building.
192
Figure 3.
Roof structure system.
The roof structure system is shown in Figure 3. The structure adopts the form of a unidirectional plane truss. The truss span above the stadium is about 22 m to 49 m with a truss height of 2.5 m, and it is mainly supported on 26 concrete frame columns, a Vierendeel truss arch and several leaning columns arranged in combination with the curtain wall structure. The truss span above the swimming pool is about 33 m to 60 m with a truss height of 2 m to 3 m, and it is mainly supported on the Vierendeel truss arch, several leaning columns, some frame columns in outdoor areas, herringbone and V-shaped steel pipe columns. The supporting system of the roof is shown in Figure 4.
Figure 4.
Supporting system of the roof.
To solve the problems caused by the two requirements of the architect, the following solutions are proposed in the structural design: (1) The rise span ratio of the arch in the natatorial area is relatively small (about 1/10), the arch effect is weak, and the force form is inclined to the Vierendeel truss beam. To reduce the span of the structure, combined with the building function, several V-shaped steel pipe columns are set as the support points of the Vierendeel truss arch to improve the structural efficiency; (2) The secondary structure adopts plane truss and cannot set vertical support truss, which makes the out-of-plane stability of the truss become a key problem. Considering that the upper chord of the truss uses
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continuous purlins perpendicular to the truss to form a constraint on the structure, and the purlins are designed as compression-bending members; cross-continuously arranged steel tie rods are set in the lower chord as the out-of-plane support of the lower chord of the truss. Out-of-plane calculated length of the bottom chord member is taken as the unsupported length between the steel tie rods (Tang et al. 2003). Purlins and steel tie rods are connected to support or rigid areas to ensure that the structure has reliable anchorage points to improve the overall stability of the structure. The anchorage areas of purlin and steel tie rods are shown in Figure 5.
Figure 5.
Anchorage areas of purlin and steel tie rod.
The total length of continuous Vierendeel truss arch 1 (Figure 6) is 293.9 m, the span is 120.8 m + 82.6 m + 58.8 m + 31.7 m (cantilever part), while Vierendeel truss arch 2 (Figure 7) has a total length of 176.2 m, the span is 16.6 m (cantilever part) + 32.2 m + 50.4 m + 58.8 m + 18.2 m (cantilever part). The upper and lower chords are made of a 500 100 0mm box-type section and the vertical web members are made of a 500 500 mm box-type section, the vertical web members and chord members are connected by using the rigid connection. When the Vierendeel truss arch passes through the floor, a structural hole is set on the floor to ensure the arch is not connected with the floor. The roof plane truss chord adopts D245 8 – D600 30 mm round steel pipe, while the web member adopts D95 4 – D299 14 mm round steel pipe. The joints are all welded with penetration, and the support joints are welded by hemisphere. V-shaped and herringbone-shaped steel pipe columns are made of D400 20 – D600 30 mm round steel pipe, and both column top and column base joints adopt pin shaft connection. Leaning columns are made of 300 150 – 400 250 mm box-type section and the steel tie rod adopts 550 class steel with a diameter of 45mm.
Figure 6.
Vierendeel truss arch 1 elevation.
Figure 7.
Vierendeel truss arch 2 elevation.
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3 KEY PROBLEMS IN STRUCTURAL DESIGN 3.1
Arch foot horizontal thrust
As the main supporting structure of the roof truss, the Vierendeel truss arch will generate a certain horizontal thrust at the arch foot. The schematic diagram of the horizontal thrust at the arch foot and corresponding solutions are shown in Figure 8.
Figure 8.
Horizontal thrust at the arch foot and corresponding solutions.
Part of the arch foot of the gymnasium is directly located on the spread foundation, and the maximum horizontal thrust is about 3500 kN. In the design, a continuous strip foundation is set between the two spread foundations of the arch foot and linear pre-stressed reinforcement is built in to balance the horizontal force of the arch foot (Ji et al. 2018). The applied value of pre-stress is the arch foot thrust under the condition of 1.0 dead load plus 0.5 live loads. Pre-stress is tensioned in batches. The order of the batch tensioning is as follows (1) After the concrete reaches the design strength and the backfill on both sides of the strip foundation is completed, the pretension force reaches 50%; (2) After the construction of the main structure (excluding the metal roof system) is completed and the falsework of roof structure is removed, the pretension force reaches 100%. Due to the existence of pre-stress, before the thrust of the arch foot is completely transmitted to the strip foundation, the force form of the strip foundation is a compression-bending component. To prevent instability, the design requires that the backfill process on both sides should be completed before the prestress tension, and several anchor rods should be set every 20 to 30 m to ensure the lateral stability of the strip foundation. The arch foot at the end of the natatorium is located on the top of the basement. Due to the 1m height difference between the basement top elevation and the spread foundation top elevation, it is difficult to set the pre-stressed balance thrust, and the total span of the arch is 141.4 m which makes the adverse effect of temperature large. A sliding bearing is set at the arch foot at the end of the natatorium to release the horizontal thrust under gravity load and temperature effect. The displacement limit of the sliding bearing is 250 mm, which is taken as the envelope displacement of each working condition under elastoplastic time history analysis of rare earthquakes. 3.2
Super long plane temperature effect
The maximum plane size of the structure on the second floor is 245 m. The conventional treatment method is to set structural joints between the gymnasium and the natatorium to divide the structure into two units with moderate plane size (Luo et al. 2020). However, since the junction area between the two venues is designed as the main entrance, setting a fulllength structural joint here will have a great impact on the architectural effect, and the architect hopes that the structure can maintain as a whole. Due to the particularity of the building function, the natural large floor openings are formed in the competition field, training field and swimming pool (Figure 9), which will release the temperature effect to a certain extent, so the structure is not treated with a joint.
195
Figure 9.
Second floor plane.
The temperature effect of the second floor slab is analyzed during the design (He et al. 2021), and the stress results of the slabs under heating and cooling conditions are shown in Figure 10. The results show that the stress of most slabs is between 0.7 Mpa and 1.2 Mpa, there is a local stress concentration effect at the edge and corner of a few floor openings, and the tensile stress is large. In the design, this part of the slab is reinforced, and the double layer with two-way reinforcement is adopted to improve the resistance of the temperature effect of the floor. The reinforcement ratio of each layer is above 0.25%. In addition, radial reinforcement is added at the corner to resist the stress concentration effect.
Figure 10.
3.3
Stress results of the second floor slab/Mpa. (a) Heating condition (b) Cooling condition.
Out-of-plane stability of Vierendeel truss arch
The out-of-plane stability of the Vierendeel truss arch in the area above the stadium is provided by the roof plane trusses and supporting columns. In the conventional large-span roof, sliding supports are often set along the span direction to release the adverse effects of earthquakes and temperatures (Wu et al. 2020). Considering the need to use the connection 196
between the plane truss and the supporting column as the out-of-plane support of the Vierendeel truss arch, the column top supports are set as fixed hinge bearings, and the axial stiffness of the plane truss is used to provide lateral restraint for Vierendeel truss arch. The out-of-plane stability of the Vierendeel truss arch in the area above the natatorium is provided by the roof plane trusses, supporting columns and herringbone-shaped columns. The simplified structure schematic diagram is shown in Figure 11. The two Vierendeel truss arches form a stable mega-frame through rigid connection with plane trusses, and out-ofplane stability is further improved through the herringbone-shaped columns and supporting columns.
Figure 11.
Simplified structure schematic diagram.
In the design, the stability analysis of the Vierendeel truss arch is carried out, and the overall model is adopted for the analysis, the area load in the model is deleted, and the load is converted to the top of the Vierendeel truss arch in the form of point load according to the load area. The influence of initial geometric defects is considered in the design (Yu et al. 2021). The first buckling mode of each Vierendeel truss arch is taken as the initial defect shape, and the maximum value of the defect is taken as 1/300 of the arch span (Song & Liu 2022). The stability analysis considering only geometric nonlinearity and both geometric and material nonlinearity is carried out respectively. The structural failure modes are shown in Figures 12 and 13, and the analysis results are shown in Table 1.
Figure 12.
The structural failure mode of Vierendeel truss arch 1.
Figure 13.
The structural failure mode of Vierendeel truss arch 2.
Table 1.
Stability analysis result of Vierendeel truss arch.
Condition
Checking term
Geometric nonlinearity
Safety factor Failure displacement/mm Failure deflection span ratio Geometric and material non- Safety factor linearity Failure displacement/mm Failure deflection span ratio
197
Vierendeel truss arch 1
Vierendeel truss arch 2
11.4 910 1/130 4.1 300 1/393
17.3 1080 1/55 6.9 500 1/120
The results show that the minimum safety factor of arch structure is 11.4, which is greater than the limit value of 4.2 under the condition of only considering geometric nonlinearity; In consideration of both geometric and material nonlinearity, the minimum safety factor of arch structure is 4.1, which is greater than the limit value of 2.0 and meets the requirements of Chinese Code (JGJ7 2010). 3.4
Out-of-plane stability of roof plane truss
As mentioned above, the top chord of the plane truss uses the continuous purlin perpendicular to the truss as the out-of-plane support, and the bottom chord uses the crosscontinuously arranged steel tie rods as the out-of-plane support. When the truss has an outof-plane instability trend, the steel tie rod on one side (as shown in green in Figure 14) tends to become shorter and quit working, while the steel tie rod on the other side (as shown in blue in Figure 14) tends to become longer and provides tension force for the unstable truss to return to the original position to ensure the out-of-plane stability of the truss.
Figure 14.
Steel tie rod support mode schematic diagram.
The stability analysis of the roof plane truss is also carried out in the design. The analysis adopts the overall model and considers the initial geometric defects of the structure. The initial geometric defects distribution adopts the first three order buckling modes of the structure respectively, and the maximum value of defects takes 1/300 of the span of the truss. The analysis results are shown in Table 2. The results show that the stability of the roof plane truss meets the code requirements and has a certain degree of redundancy. Table 2.
Stability analysis result of roof plane truss.
Condition
Initial geometric defects distribution
Safety factor
Limit
Geometric nonlinearity
Buckling Buckling Buckling Buckling Buckling Buckling
5.97 6.08 6.08 3.27 3.30 3.29
4.2
Geometric and material nonlinearity
mode mode mode mode mode mode
1 2 3 1 2 3
2.0
To verify the reliability of cross-continuously arranged steel tie rods as the out-of-plane support of the bottom chord, a typical truss of the roof structure was selected, and three models were compared: (1) the original steel tie rods as the out-of-plane support; (2) the outof-plane displacement of steel tie rod connection points (completely out-of-plane support) is directly constrained; (3) the steel tie rods deletion on both sides of the truss (no out-of-plane support). The comparative analysis is for the whole elastoplastic process considering geometry and material nonlinearity and the initial geometric defects (the first order buckling mode of the truss is selected for distribution). The structural area load of the roof is deleted in the model, and the load is loaded into the upper chord node of the selected plane truss in the form of a point load. Analysis results under the three calculation assumptions are shown in Table 3. 198
Table 3.
Stability analysis results under the three calculation assumptions.
Loading step
Lateral displaceModel ment/mm
Vertical displacement/mm
Safety Factor
1
1 2 3 1 2 3 1 2 3 1 2 3 1 2 3
123 124 125 222 232 233 324 324 327 371 372 377 421 421 421
1.20 1.20 1.20 2.40 2.40 2.40 3.48 3.35 3.28 4.34 4.26 4.18 4.94 4.85 4.67
2
3
4
5
22 21 23 41 41 42 63 59 82 76 69 110 86 79 141
It can be seen from the table that: (1) when loading in the first two steps, the structural members are in the elastic state, and the calculation results are basically consistent under the three calculation assumptions. With the increase of the load, some structural members enter into plastic state, and the calculation results vary significantly according to different calculation assumptions; (2) when the structure fails, the vertical displacement under the three calculation assumptions is completely consistent (about 1/138 of the span of the truss), and the lateral displacement is different according to the lateral constraint conditions, indicating that the structural failure is mainly caused by vertical deformation; (3) although the structural failure vertical displacement is the same under the three calculation assumptions, the safety factor of the model 3 is slightly lower than those of the other models, indicating that the lateral displacement of the structure has adverse effects on the structural safety, and it is necessary to constrain the lateral displacement of the truss in the design; (4) in the case of structural failure, the lateral displacement difference between model 1 and model 2 is small (only 8%), but the difference between model 3 and model 1 is large (about 64% to 74%), indicating that the cross-continuously arranged steel tie rods can be effectively used as the out-of-plane support of the truss; (5) the safety factors of the structure under the three calculation assumptions are basically the same, and all meet the requirements of the code. In model 1, because the steel tie rod is connected to the trusses on both sides, the safety factor is slightly improved due to partial vertical constraints on the failure trusses.
4 CONCLUSIONS 1) The structural design of the project fully fits the architectural design concept using the form of “arch supported roof”, a roof structure using a unidirectional plane truss system without a vertical secondary truss to meet the architectural interior effect. 2) For the treatment of horizontal thrust at the arch foot, combined with the rise span ratio and force form of the arch, the pre-stressed reinforcement is used to balance the horizontal thrust at the arch foot for the part with obvious arch effect, and the form of sliding bearing is used to release the horizontal thrust for the part with weak arch effect.
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3) Several large openings are set on the second floor of the building, which weakens the adverse effects of the temperature effect. Therefore, the main structure of the project is 245 m without structural joints. According to the analysis results, a variety of structural measures are adopted to resist and release the adverse effects of the temperature effect. 4) In the design, the out-of-plane stability of the Vierendeel truss arch is ensured by strengthening the restraint of the support, setting the herringbone-shaped column, and adopting a rigid connection between the roof truss and the arch. The calculation results show that the stability of the Vierendeel truss arch is good. 5) The roof structure adopts the structure system of the plane truss without setting the vertical secondary truss, the design adopting the upper chord continuous purlin plus the bottom chord cross-continuously arranged steel tie rods form as the roof truss out-ofplane support and the steel tie rod as the truss out-of-plane support is analyzed. The analysis results show that the roof structure has good stability and the cross-continuously arranged steel tie rods can be effectively used as the out-of-plane support of the truss.
REFERENCES GB50009. (2012). Load Code for Design of Building Structures. Peking, China. (in Chinese) GB50011. (2016). Code for Seismic Design of Buildings. Peking, China. (in Chinese) GB50068. (2018). Unified Standard for Reliability Design of Building Structures. Peking, China. (in Chinese) He, X.H., Wen, S.Q., Wang, X., et al. (2021) Analysis and Strategy of the Effect of Ambient Temperature on the Qin-tai Museum, 51 (S2): 171–175. (in Chinese) JGJ7. (2010). Technical Specification for Space Frame Structures. Peking, China. (in Chinese) Ji, H., Li, T., Sun, Z.M., et al. (2018) Design and Analysis of the Composite Arch Structure of Hohhot East Passenger Station, 48 (17): 1–7,12. DOI:10.19701/j.jzjg.2018.17.001 (in Chinese) Luo, G.F., Li, H.S., Li, T., et al. (2020) Structure Design on the Stadium of Huangshi Olympic Sports Center, 50(08): 113–119. DOI:10.19701/j.jzjg.2020.08.020 (in Chinese) Song, H.J., Liu, Q. (2022) Stability Analysis and Anti-continuous Collapse Analysis on Complex Supporting Roof Structure of Jiangmen Railway Station, 52 (01): 17–23. DOI:10.19701/j.jzjg.2020.03.050 (in Chinese) Tang, H., Zhou, D., He, Z.C. (2003) Structural Design on Guangzhou Gymnasium Roof. Building Structure, 33 (1): 51–54. DOI:10.19701/j.jzjg.2003.01.014 (in Chinese) Wu, Y.F., Fan, H.B., Li, H.S., et al. (2020) Structural Design of Large-span Roof in Area A of China (Huai’an) International Food Expo Center, 50 (08): 127–131. DOI:10.19701/j.jzjg.2020.08.020 (in Chinese) Yu, H.R., Li, W.B., Luo, S.S. (2021) Stability Analysis on a Bi-directional Cylindrical Thin-shell Structure Supported by Arches, 27 (02): 34–40, 69. DOI:10.13849/j.issn.1006-6578.2021.02.034 (in Chinese)
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
An explainable two-stage data-driven approach for risk modelling in tunnel construction Fenghua Liu, Wenli Liu, Yangyang Chen* & Yafei Qi School of Civil and Hydraulic Engineering, Huazhong University of Science and Technology, Wuhan, Hubei, China
ABSTRACT: Ensuring the safety of employees and adjacent buildings during tunnel construction is vital, and considerable effort is undertaken to reduce the risk represented by the ground settlement. In the past decades, there have been many solutions for modeling the ground settlement risk using machine learning or deep learning algorithms. Despite their satisfactory performance, the existing research needs to pay more attention to the sequence characteristics of tunnel construction and being uninterpretable. To this end, we aim to address the following question: How can we accurately model the ground settlement risk using complex sequence features and improve its explainability? We proposed a hybrid two-stage data-driven approach to improve the accuracy and explainability of the ground settlement’s prediction. Our proposed approach includes (1) linear trend prediction using seasonal autoregressive integrated moving average model (SARIMA), (2) nonlinear residuals prediction using the deep neural network (DNN), and (3) the posthoc explain technique using Shapley additive explanation (SHAP). Our proposed approach (ARIMA-DNN) is validated in a real-life tunnel construction in Wuhan, China. The prediction results with R2 of 0.895, RMSE of 1.085, MAE of 0.841, and MAPE of 9.07% show the superiority and applicability of the ARIMA-DNN method.
1 INTRODUCTION In tunnel construction using a Tunnel Boring Machine (TBM), there is soil disturbance during the excavation of the TBM, which causes an unavoidable risk of ground settlement (Qian et al. 2016). Despite the considerable effort that site managers have undertaken to reduce the risk, unacceptable financial loss and deaths still occur, making it a challenge to ensure the safety of the workers and surrounding structures in tunnel construction (Zhu et al. 2022). With the booming of machine learning (ML) and deep learning (DL), it has been widely used in the risk analysis of tunnel construction. For example, Ling et al. (2022) suggested the random forest to predict the ground settlement using geological messages and operational data of TBM, achieving better prediction accuracy than multiple regression. Zhang et al. (2021) proposed a novel expanding deep learning model and took constructed and constructed data into prediction. The results show that the expanded long short-term memory neural networks outperformed expanded artificial neural networks, etc. Despite the wide adoption of ML and DL, there are several drawbacks in the existing research. The ground settlement has sequence characteristics because soil properties are usually continuous variables, and the excavation of surrounding points will influence the ground settlement at a point. Most research randomly divides the dataset to get training and testing sets and ignores the sequence nature of the ground settlement. Adding to the mix, various ML and DL algorithms are considered ‘black boxes’, whose outputs are difficult to understand how and why they were generated. Their lack of interpretability undermines site managers’ trust in using the ML model to assist or replace human safety decisions. *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-27
201
Against these contextual backdrops, we aim to address the following research gap: How can we accurately model the ground settlement risk with complex sequence features and improve its explainability? With this in mind, we develop a two-stage data-driven approach to predict the ground settlement and improve its accuracy and explainability. Our proposed approach includes (1) linear trend prediction using seasonal autoregressive integrated moving average model (SARIMA), (2) nonlinear residuals prediction using the deep neural network (DNN), and (3) the posthoc explain technique using Shapley additive explanation (SHAP). Its structure and application will be introduced in the following sections in detail. 2 DEVELOPMENT OF A HYBRID ARIMA-DNN MODEL The autoregressive integrated moving average (ARIMA) model has been widely utilized for sequence forecasting. ARIMA is a composite model consisting of the autoregressive (AR) model, the integrated (I) model, and the moving average (MA) model. The resulting differential autoregressive moving average model, known as ARIMA (p, d, q), involves the inclusion of three parameters: p, which represents the autoregressive term; d, which signifies the order of differencing applied to obtain a smoother sequence; and q, which denotes the number of lags for the moving average. A deep neural network (DNN) is a classical deep learning model, which comprises the input layer being the initial layer, multiple intermediate hidden layers, and the output layer being the concluding layer. The output from one layer is utilized as the input to the next hidden layer, and this iterative process is continued until the final output of the network is obtained. Several scholars have demonstrated proficiency in utilizing ARIMA and DNN models to investigate safety and security issues in tunnel construction (Li et al. 2021; Liu et al. 2023). In this study, we propose a two-stage hybrid model, which combines the principles of an ARIMA model and a DNN model, to predict sequence data. In the initial phase of the study, the original dataset undergoes partitioning into training and testing sets in a predetermined ratio. Following this, the ARIMA model is applied to the training data to estimate and capture the underlying linear characteristics of the dataset. After the first stage, residual values are obtained by computing the discrepancies between the ARIMA forecasts and the observed values from the test dataset. These residuals are subsequently utilized for training the DNN model, which endeavors to ascertain the nonlinear association between the residual values and the actual test set data. Finally, the ARIMA prediction results are combined with the residual mapping obtained by DNN learning to obtain the final sequence prediction results. The details of the procedure above are illustrated in Figure 1.
Figure 1.
The technical route of the ARIMA-DNN method in this study.
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3 CASE STUDY 3.1
Project overview
This project is the section from Garden Expo Park North Station to Garden Expo Park Station of Wuhan City Rail Transit Line 7 Phase I Project. The underlying bedrock is composed of Silurian sandstone and shale. The tunnel section was constructed using the shield tunneling method. The total length of the tunnel is 1691 meters, the inner diameter of the tunnel is 5.6 meters, and the outer diameter is 6.2 meters. According to the survey results, the covering soil layer along the construction site consists primarily of the modern artificial fill layer, a Quaternary River alluvial deposit layer, and a sand layer. The shield machine traverses the strata, mainly including silty clay, silty sand, and silty clay with silt (sand), as shown in Figure 2.
Figure 2.
Schematic diagram of the longitudinal section of the tunnel.
As stated by Suwansawat et al. (2006) and Liu et al. (2020), the ground settlement resulting from tunnel excavation is primarily influenced by three types of characteristics: tunnel geometry parameters, shield machine operation parameters, and geological condition parameters. For this study, thirteen elements, consisting of twelve characteristics and one goal, were gathered. The tunnel geometry parameter pertains to the depth of the cover (H). The shield machine operating parameters can be derived from the shield machine database and mainly comprise advance rate (AR), thrust force (TF), chamber pressure (CP), grout filling (GF), cutterhead rotation speed (CRS), cutterhead torque (CT), screw conveyor speed (SCS), tail grease (TG) and foaming agents (FA) (Liu et al. 2020). The weighted average soil compressible modulus (ACM-ES) and the weighted average soil Poisson’s ratio (APR-m) obtained from exploration data represent the geological condition parameters. Table 1 illustrates the top ten samples from the complete dataset for training the ARIMA-DNN model. The target output under consideration is a ground settlement (Smax). Table 1.
Data samples.
H Sample (m)
ACMEs (MPa)
AR APR- (mm/ m (-) min)
CP GF CRS TF (kN) (bar) (m3) (rpm)
CT SCS TG (kN•m) (rpm) (kg)
FA (L)
Smax (mm)
1 2 3 4 5 6 7 8 9 10
18.46 18.20 17.55 21.80 17.29 16.11 16.76 16.27 17.32 19.57
0.33 0.32 0.31 0.30 0.31 0.32 0.30 0.30 0.30 0.31
5380.00 7315.00 7937.00 8449.00 9530.00 9559.00 10236.00 11713.00 10857.00 11782.00
1776.00 2423.00 1814.00 2120.00 2280.00 2344.00 2439.00 2375.00 2717.00 4551.00
25.89 46.41 19.28 51.99 50.88 40.00 57.32 65.00 30.00 54.44
3.36 6.72 9.50 9.77 7.20 8.80 10.87 10.67 11.45 19.88
13.94 13.94 13.93 13.93 13.92 13.92 13.91 13.91 13.90 13.90
10.50 19.00 19.00 21.00 27.00 18.50 16.50 17.50 26.50 29.50
0.20 0.24 0.36 0.67 0.85 1.25 1.08 1.05 1.21 1.24
203
4.94 4.35 4.50 2.00 4.06 4.00 4.00 4.00 3.00 2.50
1.00 1.10 1.00 1.50 1.00 1.50 1.00 1.10 1.50 1.20
4.89 5.41 6.25 6.34 7.22 7.35 7.59 7.68 7.01 7.01
35.40 48.39 31.22 55.09 53.28 40.00 64.79 87.00 35.00 60.77
3.2
Construction of the hybrid ARIMA-DNN model
3.2.1 Anomalies detecting monitoring data Monitoring data collected by the sensors may have anomalies due to the uncertainty of tunnel construction. Hence, detecting and removing the anomalies affecting ML’s trend learning is necessary when using the sequence feature to predict tunnel-induced ground settlement. We suggest the Extend Isolated Forest (EIF) implements anomaly detection of monitoring data, which can gain better-detecting accuracy than IF and other algorithms (e.g., random forests) (Hariri et al. 2019). The tree number of EIF is set to 2000, which makes the detection results stable. Figure 3 shows the results of the anomaly detection. The average anomaly score is near 0.5, indicating that the sensors work properly, and the monitoring data are not significant abnormalities (Liu et al. 2008). Meanwhile, several data points (e.g., No.19, 68) achieve an anomaly score of more than 0.6, which can be considered potential anomalies. Therefore, we drop out these potential anomalies in prediction to achieve higher precision.
Figure 3.
The anomalies detection of the monitoring data.
3.2.2 Hyperparameters setting Due to the prediction that the tunnel-induced settlement is a “multiple input-single output” regression, the DNN is designed by funnel type with decreasing ratio of 0.5, avoiding the accuracy loss caused by the large difference of neurons between the last hidden layer and the output layer (Liu et al. 2023). Furthermore, the SARIMA is utilized better to catch the linear trend among the monitoring data. To contrast, the DNN and ARIMA are selected as baselines to show the superiority of the hybrid ARIMA-DNN. We use Grid Search to find the best combination of hyperparameters. For DNN, the hyperparameters are hidden layers N, neurons of the first hidden layer n, epoch e, and batch size Nbs . For SARIMA, the hyperparameters are the autoregressive, difference, moving average order of the trend item ðp; d; qÞ and the seasonal item ðP; D; QÞ. We use the R2 and the Bayesian Information Criteria (BIC), which can be computing by (1), as the key performance indicator (KPI) of the DNN and SARIMA, respectively (Chicco et al. 2021; Choi et al. 2021). The best hyperparameters settings are shown in Table 2. 8 Pn ðybi yi Þ2 < 2 R ¼ 1 Pi¼1 n 2 (1) i¼1 ðy yi Þ : BIC ¼ klnðnÞ 2lnðLÞ Table 2.
The best hyperparameters of the proposed approach and baselines.
Algorithm
Hyperparameters
ARIMA-DNN
(p, d, q) *(P, D, Q) = (0, 1, 1) *(0, 0, 0) ðN; n; e; Nbs Þ ¼ (3, 64, 170, 64) ðN; n; e; Nbs Þ ¼ (3, 128, 200, 32) (p, d, q) *(P, D, Q) = (0, 1, 1) *(0, 0, 0)
DNN SARIMA
Highest KPI Value
204
0.895 0.616 1413.655
3.3
Prediction results
3.3.1 Prediction accuracy The sequence dataset is divided into training and testing sets with a ratio of 0.7. With the best hyperparameters shown in Table 2, the ARIMA-DNN makes the prediction, as illustrated in Figure 4. The ARIMA well catches the trend feature of the training set and makes the prediction. The prediction is constant due to the single use of the moving average model determined by the data feature. Then, the DNN accurately fits the characteristics of the nonlinear residuals, which has an R2 of 0.896 in residual prediction. Consequently, the ARIMA-DNN accurately splits and fits the sequence monitoring data’s linear and non-linear features, leading to a high-precision prediction. The comparison of ARIMA-DNN and baselines is shown in Figure 5. The ARIMADNN outperforms the DNN and ARIMA with the R2 of 0.895, RMSE of 1.085, MAE of 0.841, and MAPE of 9.07%. It is calculated by Formula (2). The lower KPIs of the DNN and ARIMA indicate that a single linear regression approach or a nonlinear regression approach is hard to use well enough to learn the features of the sequence data. Our proposed method lets the linear and nonlinear algorithms focus on what they are skilled in and achieve better prediction performance. 8 Pn > bi yi Þ2 > i¼1 ðy 2 > R ¼ 1 > P > n 2 > > i¼1 ðy yi Þ ffi > sffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi > > n > 1X > > > ðybi yi Þ2 < RMSE ¼ n i¼1 (2) n > 1X > > jybi yi j MAE ¼ > > > n i¼1 > > > n > > ybi yi 1X > > > y 100% : MAPE ¼ n i¼1
i
3.3.2 Prediction explainability The explainability of the proposed method is divided into two parts. In linear trend learning, the SARIMA is a transparent-designed model because formulas can represent its computing process; hence, it is inherently interpretable (Love et al. 2022). In nonlinear residual learning, the DNN, known as a ‘black box’ model, is an opaque model in which its computing formulas cannot analyze the output. Therefore, post hoc explainable techniques are required to generate explanations of the predictions. This research introduces the SHAP as a key post-hoc technique to explain the prediction of the DNN (Lundberg et al. 2017). The SHAP computes the contribution of the inputs toward output based on the Shapley value defined by game theory. The Shapley value can be computed by Formula (3). 8 1X > > ji ¼ jS j!ðn 1 jS jÞ!½f ðS⋃figÞ f ðS Þ > SNnfig > n! < Þ x ¼ hx ðx M > X > > > Þ ¼ j0 þ f ðxÞ ¼ gðx ji xi :
(3)
i¼1
where ji is the Shapley value for feature i. f is the prediction method. S and N are a subset are the and a full set of features, respectively. n is the number of elements in N. x and x
205
Figure 4.
The prediction of the ARIMA-DNN in the training and testing set.
interpreted and simplified input. M is the number of simplified inputs. j0 is the base Shapley value when all inputs are missing. Figure 6 shows the contribution of the inputs toward output in the prediction of the DNN. Figure 6 (a) shows the initial results of the SHAP, the importance of the features is represented by the horizontal axis, and Figure 6 (b) is the average of the SHAP value. The generation of settlement is more dependent on the shield machine operation parameters, and it can be seen from Figure 6 that for this project, AR, GF, and CT have a greater influence 206
Figure 5.
The prediction comparison between ARIMA-DNN, DNN, and ARIMA.
Figure 6.
The post-hoc explainability of the DNN.
on ground settlement. This indicates that reasonable control of shield operating parameters can effectively control the magnitude of ground settlement. SHAP provides an effective solution for shield operators to drive the shield better and ensure project quality.
4 CONCLUSIONS This paper addresses the following question: How can we accurately model the ground settlement risk using complex sequence features and improve its explainability? To solve this issue, we proposed a two-stage data-driven approach to predict the ground settlement and improve its accuracy and explainability. Our proposed approach consists of (1) the SARIMA used to learn linear trends, (2) the DNN used to learn nonlinear residuals, and (3) the SHAP used to post-hoc explain the outputs. We use a real engineering case in Wuhan, China, to validate the effectiveness and feasibility of our approach. Our results comprise: (1) the ARIMA-DNN outperforms the single linear or nonlinear method (i.e., ARIMA or DNN) and gains prediction precision with R2 of 0.895, RMSE of 1.085, MAE of 0.841, MAPE of 9.07%. (2) The shield machine operating parameters significantly affect the
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ground settlement. For this study case, the parameters with the highest SHAP values are advance rate, grout filling, and cutterhead torque. The results show that our proposed approach can accurately predict ground settlement and explain how the model generates its prediction. This helps the site manager understand the important influencing factors of ground settlement and targeted safety decisions.
ACKNOWLEDGMENTS The authors gratefully acknowledge the support provided by the National Natural Science Foundation of China (Grant Nos. U21A20151, 52192664, and 72171094) and the Fundamental Research Funds for the Central Universities No. 2021XXJS076.
REFERENCES Chicco, D., Warrens, M. J., & Jurman, G. (2021). The Coefficient of Determination R-squared is More Informative Than SMAPE, MAE, MAPE, MSE, and RMSE in Regression Analysis Evaluation. Peerj Comput Sci. 7, e623. https://doi.org/10.7717/peerj-cs.623. Choi, C. Y., Ryu, K. R., & Shahandashti, M. (2021). Predicting City-level Construction Cost Index Using Linear Forecasting Models. J Constr. Eng. Manag. 147(2), 04020158. https://doi.org/10.1061/(ASCE) CO.1943-7862.0001973. Hariri, S., Kind, M. C., & Brunner, R. J. (2019). Extended Isolation Forest. IEEE T. Knowl. Data Eng. 33(4), 1479–1489. https://doi.org/10.1109/TKDE.2019.2947676. Ling, X., Kong, X., Tang, L., Zhao, Y., Tang, W., & Zhang, Y. (2022). Predicting Earth Pressure Balance (EPB) Shield Tunneling-induced Ground Settlement in Compound Strata Using Random Forest. Transp. Geotech. 35, 100771. https://doi.org/10.1016/j.trgeo.2022.100771. Li, B., Wang, E., Shang, Z., Liu, X., Li, Z., Li, B., Niu, Y., & Song, Y. (2021). Optimize the Early Warning Time of Coal and Gas Outburst by Multi-source Information Fusion Method During the Tunneling Process. Process Saf. Environ. Prot. 149, 839–849. https://doi.org/10.1016/j.psep.2021.03.029. Liu, W., Li, A., Fang, W., Love, P. E., Hartmann, T., & Luo, H. (2023). A Hybrid Data-driven Model for Geotechnical Reliability Analysis. Reliab. Eng. Syst. Saf. 231, 108985. https://doi.org/10.1016/j. ress.2022.108985. Liu, W., & Ding, L. (2020). Global Sensitivity Analysis of Influential Parameters for Excavation Stability of Metro Tunnel. Autom. Constr. 113, 103080. https://doi.org/10.1016/j.autcon.2020.103080. Liu, F. T., Ting, K. M., & Zhou, Z. H. (2008). Isolation Forest. IEEE. 2008 8th Int. Conf. on Data Mining pp. 413–422. https://doi.org/10.1109/ICDM.2008.17. Love, P. E., Fang, W., Matthews, J., Porter, S., Luo, H., & Ding, L. (2022). Explainable Artificial Intelligence: Precepts, Methods, and Opportunities for Research in Construction. Preprint arXiv:2211.06579. Lundberg, S. M., & Lee, S. I. (2017). A Unified Approach to Interpreting Model Predictions. Proc. 31st Int. Conf. Neural Inform. Process. Syst. pp. 4768e4777. https://doi.org/10.48550/arXiv.1705.07874. Qian, Q., & Lin, P. (2016). Safety Risk Management of Underground Engineering in China: Progress, Challenges, and Strategies. J. Rock Mech. Geotech. 8, pp 423–42. https://doi.org/10.1016/j. jrmge.2016.04.001. Suwansawat, S., & Einstein, H. H. (2006). Artificial Neural Networks for Predicting the Maximum Surface Settlement Caused by EPB Shield Tunneling. Tunn. Undergr. Sp. Tech. 21(2), pp 133–150. https://doi.org/ 10.1016/j.tust.2005.06.007 Zhang, N., Zhou, A., Pan, Y., & Shen, S. L. (2021). Measurement and Prediction of Tunnelling-induced Ground Settlement in Karst Region by Using Expanding Deep Learning Method. Meas. 183, 109700. https://doi.org/10.1016/j.measurement.2021.109700. Zhu, Y., Zhou, J., Zhang, B., Wang, H., & Huang, M. (2022). Statistical Analysis of Major Tunnel Construction Accidents in China from 2010 to 2020. Tunn. Undergr. Sp. Tech. 124, 104460. https://doi.org/ 10.1016/j.tust.2022.104460.
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Experimental study on the collapse resistance of precast monolithic reinforced concrete beam-column subassembly with additional connecting steel bars Yihua Zeng* Key Laboratory of Concrete and Prestressed Concrete Structures of the Ministry of Education, Southeast University, Nanjing, China
Yanpeng Shen School of Civil Engineering, Southeast University, Nanjing, China
ABSTRACT: Recently the topic of the progressive collapse of concrete structures has received extensive attention. With the worldwide adoption of precast concrete structures, the collapse resistance performance of precast concrete structures is of high importance. In this paper, two 1:2 scaled precast concrete beam-column subassemblies were designed, fabricated and experimentally studied. The main factor considered herein is the role of the additional connecting steel bars. By result analysis, it is found that the existence of the additional steel bars has little influence on the resistance mechanisms i.e. the compressive arch mechanism and the tensile catenary mechanism. Moreover, it is found that the additional steel bars significantly improve the collapse resistance of the specimens. In this study, an increase of 28% in the compressive arch action stage and an increase of 76% in the tensile catenary action stage were observed.
1 INTRODUCTION In the traditional structural design of reinforced concrete structures, the main focus of designers is on the permanent and live loads, as well as the conventional factors such as actions due to winds and earthquakes. However, in recent years, there have been numerous incidents of building damage or failure leading to collapse due to accidental loads such as explosions, impacts, and fires. When these accidental events are not taken into account in the traditional design and if a concrete structure cannot withstand the specific initial damage, the failure of the initial damage will spread through the structure i.e. progressive collapse and pose a huge threat to human life safety. Since the collapse of the Ronan Point apartment in 1968, many countries have learned from the lesson and made corresponding modifications to some building regulations, such as incorporating design factors to increase the structural integrity and robustness to resist progressive collapse into building codes or design guidelines (e.g. ACI 2019; GSA 2016). Many scholars have also conducted experimental (e.g. Qian & Li 2014; Qian et al. 2019; Su et al. 2009) or theoretical (e.g. Hayes & Woodson 2005; Harry & Lu 2019; Yu & Tan 2014a) research on traditional cast-in-place reinforced concrete frame structures. Precast concrete structures, referring to those concrete structures where prefabricated components are manufactured in a factory and then transported to the site for assembly and *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-28
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casting, have many advantages due to industrialized production such as construction convenience, short construction period and environmental friendliness. Recently several researchers have conducted studies on the progressive collapse resistance of wet connections (e.g. Kang & Tan 2015) or dry connections (e.g. Qian et al. 2019) of precast concrete structures by using similar research methods applied in the research work of traditional castin-place concrete structures. These studies have shown that ordinary precast concrete structures and cast-in-place structures have similar resistance mechanisms. However, there are significant differences in the progressive collapse resistance performance between precast and cast-in-place reinforced concrete structures. On the one hand, traditional cast-in-place concrete frame joints are synchronized with other components by casting whereas the joints in precast ones are formed by reinforcement connection (such as grouted sleeve connection) and cast-in-place concrete. A large number of studies have shown that the progressive collapse resistance performance of precast concrete frames is inferior compared with cast-inplace concrete frame structures, leading to a greater risk of collapse when encountering accidental events (Nimse et al. 2014; Lin et al. 2019; Yu & Tan 2014b). On the other hand, precast reinforced concrete structures often have numerous reinforcing bars in the joint area. The connection forms of these reinforcing bars that may have significant effects on the progressive collapse resistance performance of the corresponding structures deserve more research attention (Adam et al. 2018). This paper focuses on the progressive collapse resistance performance of a precast reinforced concrete frame structure with additional connecting bars. The failure mode, loaddisplacement curve, and the effect of the additional connecting bars on collapse resistance were analyzed based on the experimental test results of two beam-column subassemblies.
2 EXPERIMENTAL PROGRAM 2.1
Specimen design
The prototype structure of the experiment adopts a five-story reinforced concrete frame structure designed according to the current standards in China. The height of the first floor is 4 meters, while the height of the other floors is 3.3 meters. The frame has six spans in the horizontal direction and five spans in the vertical direction. The length of each span for both directions is 6 meters. C40 concrete and HRB 400-type steel bars were used in the prototype structure. As shown in Figure 1, the two-span beam-column subassembly was chosen as the test zone in the consideration of the central column loss scenario.
Figure 1.
Prototype structure.
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In the study, two 1/2 scale precast beam-column subassemblies (i.e. SP1 and SP2) were fabricated, with their configuration details respectively shown in Figures 2 and 3. In specimen SP1, the upper longitudinal rebars are continuous and the lower longitudinal bars are bent and extended into the columns (Figure 4 (a)). Compared with SP1, U-shaped hollows were formed at the precast beam ends of specimen SP2 and additional connecting steel bars were anchored within the hollows (Figure 4 (b)). The U-shaped hollows were reserved for placing connecting steel bars within a length range of 700 mm on both sides of each beam. Within the anchorage zone, each steel bar extended into the inner side of the longitudinal bars of the columns. The space of the stirrups within the beams for SP1 is 200 mm and a densification for stirrups is applied for SP2 within its hollow ranges with a space of 100 mm. During the production of the two specimens, the beam and column sections were prefabricated and cured. Thereafter the beams and columns were assembled, followed by the pouring of concrete into top layers, beam-column joints as well as the hollows. The height of the precast beams is 120 mm and the top layer is 80 mm high. To improve the bonding properties, the surface between the prefabricated component and the cast-in-place concrete was treated by chiseling before assembling. The thickness of the concrete cover for both specimens is 15 mm.
Figure 2.
Configuration details of specimen SP1.
Figure 3.
Configuration details of specimen SP2.
Figure 4.
Configuration of longitudinal rebar and additional steel bar.
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2.2
Materials
The concrete grades for prefabricated components and cast-in-place parts are C40 and C50, with the average compressive strength of the standard cubic specimen at 28 days at 40.78 MPa and 50.06 MPa respectively. The steel bars used in the test are HRB400 grade steel bars, with a diameter of 12 mm for longitudinal reinforcement and 6 mm for stirrups. The tested average yield and ultimate strength of the longitudinal reinforcements are 552 MPa and 658 MPa, respectively. 2.3
Test set-up
A 100 T hydraulic servo actuator with a unidirectional stroke of 1000 mm and an actuation system for displacement control were adopted for loading. The edge columns on both sides of the specimen were placed on two pinned supports sitting on rollers, which provided the flexibility of horizontal sliding and rotating of the supports. To simulate the boundary conditions of fixed ends on both sides, each cap of the side column was bolted to the braced reaction frame through two tie connections within which a load cell was installed to record the horizontal forces. Considering that the actual floor of the building would provide significant lateral restraint during the structural collapse. A cross-shaped steel frame, as a restraint device, was also bolted to the lower cap of the middle column stub to avoid unnecessary out-of-plane rotation of the middle column under large deformation. A preloading was conducted to ensure the system works properly, followed by the monotonic vertical loading of the specimen up to failure. The hydraulic servo actuation system was used in the test to record the bearing capacity and the middle-joint displacement of the tested specimen. The experimental setup is shown in Figure 5.
Figure 5.
Test set-up.
3 TEST RESULTS 3.1
Compressive arch action stage
The loading process can be divided into two stages i.e. the compressive arch action (CAA) stage when the vertical displacement is relatively small and the tensile catenary action (TCA) stage when a large displacement is applied. For ease of description, the cross-sections of the beam near the central column and the side columns are defined as sections A and B respectively. The end section of the U-shaped hollows in specimen SP2 is defined as
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section C, and the cross-section positions of the left and right spans are distinguished by the letters L and R, as shown in Figure 5. In the CMA stage, symmetrical features were observed for each specimen. To avoid repetition, the description for the left and right beams was not specified herein. For specimen SP1, bending cracks appeared around sections A and B at a displacement of 4 mm and steel yielding were observed under a displacement of 15 mm. When the displacement reached 50 mm, the bending cracks near section A extended horizontally along the beam axis to the prefabricated and cast-in-place interface for approximately 4 cm. At a displacement of 80 mm, the bottom concrete near section B and the top concrete near section A tended to crush with cracking sounds when the peak load in the arch action had been reached. It is noticeable that the compressed concrete near sections A and B crushed and started to spill when the displacement arrived at 200 mm. When more displacement was applied, more and more cracks appeared and the crushing of the concrete became more severe. The maximum height of cracks, however, remained almost constant. The loading process of specimen SP2 is similar to that of SP1 except that the peak load during arch action is higher due to the strengthening of the additional steel bars. 3.2
Catenary action stage
In the TCA stage, there were differences between specimens SP1 and SP2. For specimen SP1, two bottom longitudinal steel bars at section LA broke at the same time at a displacement of 316 mm, causing a sudden drop in the specimen’s resistance by nearly 50%. Afterward, the specimen resistance steadily increased with displacement. When the displacement reached 340 mm, vertical cracks extending from the top surface to the beam axis were observed in the middle part of both the left and right beams, indicating that the tensile area of the specimen gradually expanded. Later, three top steel bars near section LB broke simultaneously at a displacement of 515 mm, and the three top steel bars near section RB broke afterward. The bearing capacity still showed an increasing trend even when the test was terminated at 720 mm. Regarding specimen SP2, there was no significant difference in crack development compared to SP1. When the displacement was loaded to 358 mm, three top steel bars RB section ruptured simultaneously, resulting in a sharp decline of the load. Subsequently, the three top steel bars near the LB section ruptured successively during the displacement at around 360 mm-400 mm. When the displacement reached 420mm, it was found that the crack width at the LC and RC sections became significantly wider and developed upward obliquely. A small in-plane rotation occurred between the AC and CB segments, indicating the occurrence of a local plastic hinge at section C. This was due to the change in the steel bar area at the section. At a displacement of 440 mm, the two bottom steel bars on the LC section ruptured and the two bottom steel bars in the RC section ruptured at a displacement of 608 mm, followed by the test termination. The failure modes of SP1 and SP2 are shown in Figure 6.
Figure 6.
Failure modes of specimens SP1 and SP2.
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3.3
Load-displacement curves
Figure 7 shows the load-displacement curves of the specimens. It can be seen that the development trends for SP1 and SP2 share similarities. As the loading displacement increases, the load-bearing capacity increases gradually up to the ultimate strength at the CAA stage and decreases to a relatively low value which corresponds to the turning point of the TCA stage. The load-bearing capacity then maintains the increasing trend up to its ultimate strength. The ultimate load capacity for SP1 in CAA and TCA stages is 80.19 kN and 59.40 kN, respectively. Such values for SP2 are 102.63 kN and 104.40 kN. Compared with SP1, the load capacity of SP2 is 28% higher in the CAA stage and 76% higher in the TCA stage. This is mainly attributed to the additional connecting steel bars. Though an additional weak point i.e. the plastic hinge at section C existed in SP2, its load-bearing capacity increased at a higher rate than that of specimen SP1 in the TCA stage due to a better connection and higher stress contribution. The curve of the horizontal load versus the displacement helps to understand the role of additional connecting steel bars in enhancing the load-bearing capacity. As shown in Figure 7(b), especially in the TCA stage, the presence of additional steel bars increased the horizontal axial force of specimen SP2 when compared to specimen SP1 and then increased the load-bearing capacity.
Figure 7.
Load-displacement curves.
4 CONCLUSIONS In conclusion, the two precast concrete beam-column subassemblies with or without additional connecting steel bars used in this study exhibited similar resistance mechanisms i.e. CAA and TCA stages, and similar development trends in these two stages. It is found the existence of the additional connecting steel bars resulted in additional plastic hinges and thereafter lead to a different failure mode. It is observed that the additional connecting steel bars help the specimen to sustain higher stress and thus increase the ultimate load-bearing capacities in both CAA and TCA stages. Findings in this paper may contribute to improving the design of precast concrete structures to resist progressive collapse in case the central column loss scenario is considered. ACKNOWLEDEMENT The authors would like to acknowledge the financial support from the National Natural Science Foundation of China (Grant No. 52008089). 214
REFERENCES ACI Committee (2019). Building Code Requirements for Structural Concrete (ACI 318 – 19) and Commentary. https://www.concrete.org/ Adam J, Parisi F, Sagaseta J, et al. (2018). Research and Practice on Progressive Collapse and Robustness of Building Structures in the 21st Century. Eng. Struct. 173: 122–149. https://doi.org/10.1016/j. engstruct.2018.06.082 General Services Administration (GSA) (2016). Alternate Path Analysis & Design Guidelines for Progressive Collapse Resistance. https://www.gsa.gov/ Harry O & Lu Y (2019). Simplified Theoretical Model for Prediction of Catenary Action Incorporating Strength Degradation in Axially Restrained Beams. Eng. Struct. 191 (219–28. https://doi.org/10.1016/j. engstruct.2019.04.043 Hayes JR, Woodson SC, Pekelnicky RG, et al. (2005). Can Strengthening for Earthquake Improve Blast and Progressive Collapse Resistance. ASCE J. Struct. Eng. 131 (8): 1157–1177. https:/doi.org/10.1061/(ASCE) 0733-9445(2005)131:8(1157) Kang S & Tan KH (2015). Behaviour of Precast Concrete Beam-column Sub-assemblages Subject to Column Removal. Eng. Struct. 93: 85–96. https://doi.org/10.1016/j.engstruct.2015.03.027 Lin K, Lu X, Li Y, et al. (2019). A novel Structural Detailing for the Improvement of Seismic and Progressive Collapse Performances of RC Frames. Earthq. Eng. Struct. Dyn. 48 (13): 1451–1470. https://doi.org/ 10.1002/eqe.3208 Nimse R, Joshi D & Patel P (2014). Behavior of Wet Precast Beam Column Connections Under Progressive Collapse Scenario: An Experimental Study. Int. J. Adv. Struct. 6 (4): 149–159. https://doi.org/10.1007/ s40091-014-0072-3 Qian K, Li B, Ma J (2014). Load-Carrying Mechanism to Resist Progressive Collapse of RC Buildings. ASCE J. Struct. Eng. 141 (2): 4014107–4014101. https://doi.org/10.1061/(ASCE)ST.1943-541X.0001046 Qian K, Liang S L, Fu F, et al (2019). Progressive Collapse Resistance of Precast Concrete Beam-column Subassemblages with High-performance Dry Connections. Eng. Struct. 198: 109552. https://doi.org/10.1016/j. engstruct.2019.109552 Su Y, Tian Y, Song X (2009). Progressive Collapse Resistance of Axially-restrained Frame Beams. ACI Structural J. 106 (5): 600–607. https://doi.org/10.14359/51663100 Tan K & Yu J (2013). Structural Behavior of RC Beam-Column Sub-assemblages Under a Middle Column Removal Scenario. ASCE J. Struct. Eng. 139 (2): 233–250. https://doi.org/10.1061/(ASCE)ST.1943541X.0000658 Yu J & Tan K (2014a). Analytical Model for the Capacity of Compressive Arch Action of Reinforced Concrete Sub-assemblages. Mag. Concrete Res. 66 (3): 109–126. https://doi.org/10.1680/macr.13.00217 Yu J & Tan K (2014b). Special Detailing Techniques to Improve Structural Resistance Against Progressive Collapse. ASCE J. Struct. Eng. 140 (3): 04013077. https://doi.org/10.1061/(ASCE)ST.1943-541X.0000886
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Experiment on the fatigue behavior of steel bridge deck pavement structures paved by high-content hybrid steel fiber reinforced self-compacting concrete Hua Zou China Railway 24th Bureau Group Southwest Construction Co., Ltd. Sichuan, China
Yu Pang* Master student, School of Civil Engineering, Chongqing Jiaotong University, Chongqing, China
Long Feng, Jiyun Zhang, Jiansong Liu & Chengyang Wang China Railway 24th Bureau Group Southwest Construction Co., Ltd. Sichuan, China
Qingguo Yang Professor, School of Civil Engineering, Chongqing Jiaotong University, Chongqing, China
Ying Li & Xuefeng He School of Civil Engineering, Chongqing Jiaotong University, Chongqing, China
ABSTRACT: Fatigue tests were carried out on steel bridge deck pavement structures paved by high-content hybrid steel fiber reinforced self-compacting concrete to study the durability of this kind of pavement structure. The results show that the fatigue performance of the pavement with the addition of hybrid steel fibers is significantly better than that of the plain concrete structure and that the hybrid steel fiber reinforced concrete and the steel structure can work better together. Besides, the pavement structure paved by selfcompacting concrete with a steel fiber admixture of 2% P and 6% DX is superior to that with a steel fiber admixture of 1% P and 6% DX in terms of working ability with cracks under the design load.
1 INTRODUCTION Ordinary concrete bridge decks are often subjected to scores of cracks before reaching their expected service life due to the long-term fatigue loading of vehicles. Sometimes, they may produce larger cracks that compromise traffic safety. To overcome the difficulties caused by fatigue, scientific researchers and engineering practitioners worldwide are actively exploring solutions. The addition of fibers to concrete is one of the most important methods. Mulheron et al. (2015), Matthew (2016), Charalambidi et al. (2020), and Wang et al. (2020) investigated the fatigue durability of basalt fiber, carbon fiber, and polypropylene fiber reinforced concrete. Their results demonstrated that fiber reinforcement could enhance concrete fatigue performance and toughness. Mohammadiet et al. (2005) conducted bending fatigue and static bending tests on plain concrete and flat crimped steel fiber specimens at the fiber volume fractions of 0, 1.0, 1.5, and 2.0%. They found that the probability distribution of concrete’s fatigue life depended on the fatigue stress level applied, fiber volume fraction, and fiber lengthto-diameter ratio. Fataar et al. (2021) conducted fatigue tests on pre–slipped, hooked–end steel *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-29
fiber reinforced concrete (SFRC) at a single fiber level and found that the fiber hook provided the greatest resistance to fatigue loading. Ahsan Parvez (2014) compared the test results of SFRC beams with those of plain reinforced concrete beams. As the fiber content and the fiber length-to-diameter ratio increased, the steel and concrete stresses decreased significantly, increasing the fatigue strength. Al-Azzawi1 and Karihaloo (2017) found that the uniform distribution of steel fibers greatly affected the fatigue strength of ultrahigh-performance fiberreinforced concrete (UHPFRC). Yeswanth Paluri (2021) investigated the flexural fatigue behavior of RAP-based concrete with and without steel fibers and demonstrated that adding fibers increased the fatigue strength of RAP-based concrete by 50-65%. Single fiber-reinforced concrete is not well adapted to complex bridge decks and pavement loading situations, and an increase in the single fiber content is not necessarily beneficial. For example, Banjara et al. (2018) conducted fracture and flexural fatigue experiments on plain concrete and SFRC specimens with different volume fractions (0.5%, 1%, and 2%) of steel fibers. They concluded that the load-carrying capacity and tensile strength decreased when steel fibers were added to the concrete at a volume fraction greater than 1%. Based on the above results, many studies have been carried out on hybrid steel fibers (a mixture of different types of steel fibers) reinforced concrete. Mohod et al. (2019) studied the fatigue damage of concrete pavements. The experimental results suggested using hybrid fibers, i.e., a mix of the waste tire and industrial steel fibers mixed in a definite proportion to improve the mechanical properties of concrete. By studying the mechanical behavior of ultra-highperformance concrete reinforced with different hybrid shapes of steel fiber, Ye (2012) found that steel fibers can improve the compressive strength and ultimate flexural strength of ultrahigh-performance concrete and enhance ductility. Meanwhile, self-compacting concrete has attracted a lot of attention due to its advantages, which include no need for vibration, a large decrease in the time necessary for concrete placement, and an enhanced working environment and safety. Wang (2021) compared the flexural strength and fatigue resistance of rubber self-compacting concrete, plain concrete, and steel fiber-reinforced rubber selfcompacting concrete. The results demonstrated that adding steel fibers to rubber selfcompacting concrete boosted its flexural strength. Compared to plain self-compacting concrete, the steel fibers greatly enhanced the resistance to crack initiation. Goel (2014) explored the flexural fatigue performance of plain self-compacting concrete and steel fiber-reinforced self-compacting concrete containing three fiber volume fractions using S-N relationships. The flexural fatigue performance of SCC, as well as SFRSCC, was found to be better than conventionally vibrated concrete (CVC) and fiber-reinforced concrete (CVFRC). However, it is one-sided to study and judge the fatigue properties of pavement structure based solely on the bridge deck pavement material. The fatigue resistance of the pavement structure under actual loading can only be determined when studied in conjunction with steel bridge deck plates. To improve the fatigue behavior of asphalt pavement on an orthotropic bridge deck, Ma (2019) investigated the factors influencing the fatigue limit of asphalt pavement on an orthotropic bridge deck, and the results showed that the stress and strain of the pavement decreased significantly with the increase of the elastic modulus of the overlay layer. To analyze the effects of the thickness of a super-toughness concrete (STC) layer on the fatigue performance of a lightweight composite bridge deck. Zhan (2019) built a serial local finite element model and found that the lightweight composite bridge deck structure could strongly enhance the local stiffness of the steel bridge deck. Xin (2022) investigated the fatigue damage mechanism and life prediction on a carbon-fiber-reinforced polymer (CFRP) reinforced bridge on an asphalt pavement under high temperatures. The girder crack resistance, stiffness, and fatigue damage were cumulative and analyzed. The results showed that the CFRP plate effectively constrains concrete cracking and the crack’s development in an asphalt layer construction under high-temperature conditions. The fatigue life of the test beam reinforced with CFRP plates is somewhat prolonged. Xu (2021) proposes a new orthotropic steel-concrete composite bridge-type deck system. Two reference decks (reinforced concrete deck and orthotropic steel deck) were also used for 217
comparative research. The results proved that the composite deck could effectively control the crack initiation and propagation in the concrete, postpone the yielding of the steel bars and steel plates, and improve the fatigue performance of the welded joint of the deck. Shao et al. (2018) developed a composite deck of lightweight steel–ultrahigh performance concrete (UHPC). It was found that for the orthotropic deck slab (OSD) details in the composite deck, the values of the vehicle-induced stress ranges were significantly reduced when a thin UHPC layer was added, indicating that the fatigue-cracking risks in the details could be eliminated. Composite decks were found to depict better the genuine strength and fatigue resistance of pavements and decks in the research. Moreover, much study has been conducted on steel fiber-reinforced concrete decks. By employing the finite element method (FEM), Jibobai (2013) assessed fatigue durability of the root-deck fatigue in the trough-deck welded joints over the diaphragms. The results revealed that the pavement–deck interaction should be considered in the stress range analysis, especially in thin decks. The durability of the root–deck fatigue would be improved if SFRC pavement is applied. Chen (2021) conducted fatigue performance tests on SFRC composite girder under high cycle negative bending action and found that the interlayer cyclic slip was developed slowly in the SFRC composite girder during cyclic loading and the subsequent ultimate static loading phase, indicating that steel fibers in concrete facilitated interlayer interaction and inhibited the development and maximum width of cracks. Su (2018) investigated the effects of SFRC, the reinforcement ratio, and the shear connector type and found that the use of SFRC had an obvious restraining effect on crack-width development. Ye et al. (2021) developed a steel-SFRC composite bridge deck model with U-ribs and SFRC composite structure with a volume fraction of 1.5% by ANASYS. It was concluded that the thickness and flexural strength of the SFRC overlay determined the fatigue performance of the composite bridge deck under wheel loading. As steel fibers play a significant part in enhancing the fatigue performance of concrete, single fibers do not deliver the desired results, and the composite deck is more reflective of the true fatigue state. Besides, the steel fibers in the composite deck are still essential to enhancing fatigue and flexural strength and strengthening the slip resistance between the deck and the concrete pavement. Steel fiber reinforced self-compacting concrete offers many construction advantages and is superior to conventional vibrated concrete in flexural fatigue. YANG Q, RU N, HE X, et al. (2022) proposed a new type of self-compacting concrete with a mixture of ultra-short and ultra-fine steel fibers with ordinary steel fibers and discussed its advanced nature. This paper further explores this new type of self-compacting concrete with a mixture of ultra-short and ultra-fine steel fibers with ordinary steel fibers. Fatigue experiments are carried out on its composite structure with steel bridge panels.
2 CONFIRMATION OF THE MOST UNFAVORABLE LOADING MODE AND POSITION 2.1
The most critical loading position
A numerical model of concrete steel bridge deck pavement structure paved by steel fiber reinforced self-compacting concrete was developed in Abaqus to obtain the most critical section of the steel bridge deck pavement, thereby deriving a deck pavement specimen that is more in line with the actual working stresses. A real project bridge was modeled with the specific dimensions shown in Table 1. 2.1.1 Model geometry parameters and model establishment The model section geometry of the steel bridge panel is 6 m x 12 m. The model includes 10 U-shaped stiffening ribs (10 600 mm) and 4 diaphragms (3 400 mm). The diaphragm is 500 mm in height and 12 mm in thickness. The upper portion of the steel bridge panel is a 218
Table 1.
Dimensions of the pavement model.
Components Steel bridge panel thickness (mm) The thickness of U-shaped stiffening ribs (mm) Width of the upper opening of U-shaped stiffening rib in (mm) Width of the lower opening of U-shaped stiffening ribs (mm) Height of U-shaped stiffening ribs (mm) Distance between side ribs and the edge of steel bridge (mm)
Size 14 8 300 170 280
Components The thickness of the diaphragm (mm) Spacing of U-shaped stiffening ribs in transverse arrangement (mm) Spacing between horizontal diaphragms (mm) Height of horizontal diaphragms (mm) The cross-sectional span of the steel bridge (mm)
Size 12 600 4000 500 6000
300
concrete steel bridge deck pavement structure paved by steel fiber-reinforced self-compacting concrete, as detailed in Table 1. 2.1.2 Model building and meshing The model dimensions and the model meshing are displayed in Figure 1.
Figure 1.
A schematic diagram of model dimensions and model meshing.
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2.1.3 Model calculation results The most critical loading positions can be identified by loading the orthotropic steel bridge panels at different loading positions in the transverse and longitudinal directions, and comparing and analyzing the key control indicators of the pavement and the structure. The loading response of the model reveals that when the longitudinal action is in the middle of the span. The center of the transverse load action is in the center of the fifth stiffening rib. The maximum transverse tensile stress occurs at the paving layer above the welded side ribs of the U-shaped stiffening ribs, as shown in Figure 2. It can be concluded that the transverse tensile strain in this area is often very large, and longitudinal cracks can easily form in this location.
Figure 2.
2.2
The most dangerous cross-section of a bending fatigue specimen.
Experimental loading mode and method
There are currently two types of load control methods for indoor fatigue testing that are widely used by the academic community: fatigue testing with controlled stresses and fatigue testing with controlled strains. When the steel bridge deck pavement is under load, the paving layer and the steel plate work together. The service life of the steel plate will often be longer than that of the paving layer because the steel plate enjoys higher strength and is not directly under the load. Considering such circumstances, it can be roughly concluded that the bending and tensile modulus of the steel bridge deck remains constant over the years of regular use. The stress-loading mode is practical and reasonable. 2.2.1 The method for loading the specimen Various fatigue tests for composite beams have emerged at home and abroad in recent years. However, there are currently several problems, including unavailable test conditions, inadequate controllability, and overly complex fatigue damage criteria. Among these are the more well-known German and Danish composite beam test models. The test form used by the German Otto-Graf Institute differs somewhat from that used in Denmark, the most significant difference being the loading location: the German Otto-Graf Institute loads on a steel plate. In contrast, the Danish scholar loads on the pavement’s surface using a rectangular steel plate and restrains the roller bearings at the ends of the composite beam. However, the maximum tensile stresses in both test methods were found on the surface of the paving layer above the middle roller and were parallel to the specimen’s long axis. The maximum tensile strains and tensile stresses in the paving layer when loaded on the steel plate and paving were calculated based on the above two different forms of fatigue loading, resulting in Figure 3. According to Figures 1–3, the load response in terms of maximum tensile stresses and tensile strains is the same whether loaded on steel plates or pavers. The maximum tensile 220
Figure 3.
A comparison of fatigue tests on steel plates and pavement layers.
stresses and tensile strains resulting from loading on steel bridge plates will also be greater than the maximum tensile stresses when loaded on pavers, meaning that the maximum tensile stresses and tensile strains calculated by the model of the University of Stuttgart in Germany will be greater than those by the Danish model. As the response on the pavement is greater while loading on the steel bridge deck, the final calculations of this loading method are relatively more dangerous, and the conclusions obtained are safer; therefore, a simpler and more controllable test method could be considered. 2.2.2 Selection of the specimen Based on the results of the previous analysis, it is clear that under the wheel load, when the vehicle load is located longitudinally in the middle of the span and transversely at the top of the side ribs of the U-shaped stiffening ribs (as shown in Figure 2 at A), as the ribs of the Ushaped stiffening ribs have a stable support function for the bridge deck steel plates, the surface of the paving layer exhibits the maximum transverse tensile stress at A where the load position is selected. The fatigue testing of composite beams is conducted by repeatedly applying forces to the specimen, gradually accumulating surface and internal damage, with the final damage occurring at a certain level. The procedure is a streamlined version of an accelerated fatigue test performed on a short section of pavement on a steel bridge deck, simulating the effects of removing a steel plate and its pavement from the most vulnerable spot on the steel deck to conduct a fatigue test. 2.2.3 The size of the specimen Drawing on the fatigue test models of composite beams at home and abroad, a simpler and more practical fatigue test model for composite beams was designed based on the actual operation of the MTS testing system in China, combined with the local stress damage characteristics of orthotropic steel deck pavements. As demonstrated in the previous section, when the load is applied to the steel bridge deck, the maximum tension value of the paving layer will occur above the U-shaped stiffening ribs. Therefore, when selecting the fatigue specimen for the composite beam, a 300 mm span section of the bridge deck was intercepted with the top of the rib on one side of the U-rib as the center, and the width of the beam was taken to be 100 mm. Due to the small size of the intercepted part and the smaller area where the U-rib attaches to the deck plate, the removed piece can be assumed perpendicular to the plate to simplify the model. As the maximum tensile stress is greater when acting on the steel deck, the girder is loaded by inverting the top and bottom, with the load being transmitted through the section of rib in the middle of the 221
specimen in practice. To facilitate the support of the specimen, the long axis of the composite beam specimen is extended by 80 mm, half of each side is removed, and two steel plates of the same thickness and width as the deck plate are welded at both ends, as can be seen in Figure 4 (d). The fatigue specimen for the concrete steel bridge deck pavement structure paved by steel fiber reinforced self-compacting concrete is 380 mm in length and 100 mm in breadth.
Figure 4. Dimensions of the fatigue specimen model of the concrete steel bridge deck pavement structure paved by steel fiber reinforced self-compacting concrete studied in this paper.
*The fatigue specimen size and loads of the concrete steel bridge pavement with selfcompacting hybrid steel fiber investigated in this paper are small, making it ideal for use in the school’s MTS test system. 2.3
Materials and proportioning
Materials of the concrete steel bridge deck pavement structure paved are by steel fiber reinforced self-compacting concrete. Steel plates: Q345 steel. Bonding layer: SKO bonding layer. Paving layer: The two types of concrete steel bridge deck pavement structures paved by steel fiber reinforced self-compacting concrete listed in the study by Yang et al. [23] (1%OS + 6% MS, 2% OS + 6% MS; OS is shear type ordinary steel fiber and MS is melt-drawn type ultra-short ultra-fine steel fiber) are chosen for this study. Four specimens of the same mix ratio are compared with two volume fractions.
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2.4
Experimental loading
2.4.1 Loading force and frequency Regarding the estimated traffic volume of the Huangpu Bridge in Guangzhou, the equivalent conversion of all vehicles passing through one wheel track zone of the deck pavement to an axle weight of 180 kN over a design life of 15 years would yield the axle weights of 2.34 107 times, which would make it impractical to carry out the fatigue test in terms of equipment and time. In studies by Zhang and Shi (2013), and Ge and Huang (2002), the test loads were determined to be 6 kN, 7 kN, and 8 kN. The loading force in the test is set at 8 kN to shorten the test duration and possibly compare the fatigue test results with similar tests for composite beams on steel bridge decks in China. Before applying the cyclic load, a force of 100-200 N is repeatedly applied to the specimen to remove the test piece due to uneven errors caused by poor contact. The waveform used in the fatigue tests is the non-intermittent sinusoidal waveform. The fatigue test results of the Runyang Yangtze River Bridge composite beam demonstrate [26] that, in the frequency range of 5-l0 HZ, any loading frequency has little effect on the test results. The test frequency for this study is reduced to 4 HZ due to equipment limitations. Referring to the bending fatigue test of the bridge deck pavement material, the group of Southeast University proposes that the load waveform is a sinusoidal waveform under intermittent sinusoidal fatigue loading. The vertical deformation curve of SFC is sinusoidal, the response frequency is equal to the frequency of the load, and the loading time is related to the test load and vertical deformation, as shown in Figure 5.
Figure 5.
The load-deflection curve for the bending fatigue test.
2.4.2 Loading times and damage guidelines The fatigue damage criterion is defined in the following three ways: a) The surface of the paving layer begins to crack. b) The paving material begins to delaminate, or there is a slippage or more visible separation on the joint surface. c) The overall deflection or deflection difference diverges significantly from the original value. When the loading times exceed 1.2 107, the throttle should be closed immediately, and the test should be halted. The fatigue life of materials and structures is the number of fatigue loading cycles a material or structure undergoes when it reaches a critical fatigue state. In this paper, the total number of load cycles or loading time elapsed between the start of loading. The occurrence of ultimate damage to the specimen is referred to as the fatigue life of the structure or material. The letter n usually denotes it.
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2.4.3 Experimental equipment The loading test of the concrete steel bridge deck pavement structure specimens paved by steel fiber reinforced self-compacting concrete is conducted at room temperature on the 4# MTS system dynamic fatigue testing machine at Chongqing Jiaotong University. The main technical parameters are as follows: Force measuring transducer: 250 kN. Frequency range: 2 Hz to 4 Hz. Travel of the loading head: 50 mm. Pressure sensors: 0 to 25 kN. The MTS records the number of loads actions and displays the load waveform in realtime. The test is halted after certain loading times to observe the change in appearance and crack development. The fatigue loading times recorded by the MTS system and the external morphology of the fatigue specimen are utilized to determine whether the specimen meets the fatigue damage criteria. The test load is therefore applied at a frequency of 4 Hz and a loading force of 8 kN. The fatigue loading diagram is shown in Figure 6.
Figure 6. Fatigue loading diagram for the specimen of the concrete steel bridge deck pavement structure paved by steel fiber reinforced self-compacting concrete.
3 RESULTS AND ANALYSIS Based on the number of fatigue loads recorded by the fatigue system, the cracking behavior of six specimens is compared and analyzed, and the crack development mechanism is discussed. These specimens include the ordinary self-compacting concrete specimen, the self-compacting concrete specimen mixed with 1% ordinary steel fibers, the self-compacting concrete specimen mixed with 2% ordinary steel fibers, the self-compacting concrete specimen with 6% ultra-short and ultra-fine steel fibers, the self-compacting concrete mixed with 1% ordinary steel fibers (P) and 6% ultra-short and ultra-fine (DX), and the self-compacting concrete specimen mixed with 2% ordinary fibers (P) and 6% ultra-short and ultra-fine (DX). 3.1
Fatigue damage patterns
3.1.1 Self-compacting concrete specimens without the addition of steel fibers During the fatigue loading of the self-compacting concrete specimen without adding steel fibers, a crisp sound is heard before a crack develops at the bottom of the concrete specimen when the number of load cycles is 65,400. As the loading continues, the crack develops, and at 105,000 loads, the longitudinal length of the crack was approximately 25 mm, almost half the height of the paving layer; at 140,400 loads, the crack is found to be on the verge of penetrating the whole specimen, so an additional 1,000 loads are added, and the specimen is carefully observed. After another 400 loads, the crack develops to the joint surface and penetrates the
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concrete specimen. The throttle is closed immediately, and the load is terminated. The total number of loads on the plain concrete specimens is 140,630. 3.1.2 Self-compacting concrete specimens with single steel fibers During fatigue loading of the self-compacting steel fiber concrete specimens with the addition of a single steel fiber, no cracks are observed after applying preload to the specimens, indicating that the prepared concrete has no visible cracks and that the specimens are placed in a flat position. As the load and the fatigue life increase, the damage accumulates, and cracks emerge in a relatively weak area of the specimen, generating a narrow but clear crack in the span of the specimen. The self-compacting concrete specimens with 1% and 2% ordinary steel fibers in the paving layer are loaded for 82, 500 and 99, 000 cycles, respectively, when a loud noise is heard, and a crack is seen at the bottom of the concrete. The crack continues to develop as the loading continues, but since the steel fibers have high tensile strength, the specimens can continue to withstand a certain amount of cyclic loading. At 210, 500 loads, the crack in the 1% steel fiber specimen is about to penetrate the whole specimen, and the crack can be seen to become more visible at the moment when the load is being applied downwards; at 235, 600 loads, the joint surface of the 2% steel fiber specimen becomes detached, and the throttle is closed to stop the loading. A comparison reveals that the self-compacting concrete specimens with 1% and 2% ordinary steel fibers produce narrower cracks than the self-compacting concrete specimens without adding steel fibers. The initial cracks are narrower due to the crack-arresting effect produced by the chaotic distribution of steel fibers within the concrete. The self-compacting concrete specimens with an admixture of 6% ultra-short and ultrafine steel fibers are subjected to fatigue loading. When the number of load cycles reaches 145, 000, it is observed that cracks start to appear at the bottom of the concrete on the left side of the specimen near the spanning plate. The cracks gradually grow as the loading continues, but the cracks develop more slowly compared to the self-compacting concrete specimens mixed with 1% and 2% ordinary steel fiber. The steel fiber volumes in these two specimens are lower in admixture, and the fibers are relatively easier to pull out during the fatigue cyclic loading. At 398, 000 cycles, the main crack is about to penetrate, and a smaller crack is found next to the main crack when the loading is stopped. A comparison reveals that the crack path of the self-compacting concrete specimen with an admixture of 6% ultra-short and ultra-fine steel fibers becomes somewhat curved because the chaotic distribution of the high admixture of steel fibers within the concrete causes the crack to deviate from its original path and become curved as it progresses to the steel fibers. This is shown in Figure 7.
Figure 7. Fatigue crack diagram for self-compacting concrete specimens mixed with 6% ultra-short ultra-fine steel fiber in mid-span.
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3.1.3 Hybrid steel fiber reinforced self-compacting concrete specimens The self-compacting concrete specimen with a steel fiber admixture of 1% P + 6% DX and the self-compacting concrete specimen with a steel fiber admixture of 2% P + 6% DX beginning to crack at 217, 000 and 292, 000 load cycles, respectively, with the former showing a crack at the bottom span center; the latter does not crack first at the span center, but rather a flat crack appears first slightly near the right-hand side. As the loading continues, the cracks grow slowly and significantly slower than those in plain concrete and self-compacting concrete specimens with a single fiber. At 525, 000 and 775, 000 cycles, respectively, the cracks in the paving layer are almost close to the joint surface, and some small cracks also appear in other weak areas next to the main crack. A comparative observation reveals that both the self-compacting concrete specimen with a steel fiber admixture of 6% DX and the specimens with a steel fiber admixture of 1% P + 6% DX and with a steel fiber admixture of 2% P + 6% DX produce one or more cracks in the weak zone beyond the span. Still, the cracks produced by the hybrid steel fiber reinforced self-compacting concrete specimens are much narrower than those by the single steel fiber reinforced self-compacting concrete specimen with a high volume of ultra-short ultra-fine steel fiber, mainly because the two large numbers of uniformly and irregularly distributed steel fibers coordinate with each other. They play a bridging role at different times, preventing the crack from expanding together. In addition, the bond between the matrix and the steel fibers slows the pull-out of the steel fibers so that the best crack-arresting effect can be achieved in paving structures with hybrid steel fibers. The fatigue life of the specimens increases significantly when the amount of ordinary steel fibers in the hybrid steel fibers is increased by 1%.
4 DISCUSSION 4.1
A comparison of fatigue damage cracks
As can be seen from the location of the initial cracks and the form and number of cracks for the six types of specimens in Table 2, the first cracks in the specimens with 6% DX and the hybrid steel fiber specimens with 2% P + 6% DX are produced first in the weak zone outside the span, which does not exclude the possibility that the fibers are not
Table 2.
Fatigue damage cracks in steel fiber reinforced self-compacting concrete specimens.
Specimen number
Program
Location of initial cracks
Crack pattern
Number of cracks
1 2 3 4
S–0 P1 P2 DX6
Wide, straight Wide, diagonal Thin, diagonal Thin, twisting
1 1 1 2
5 6
P1+DX6 P2+DX6
Mid-span Mid-span Mid-span Left side near the mid-span Mid-span Middle right
Subtle, twisting Subtle, twisting
2 3
Note: S is plain concrete; P is shear type ordinary steel fibers; DX is melt-drawn type ultra-short ultra-fine steel fibers; 1%, 2%, and 6% steel fiber contents are represented by 1, 2, and 6 respectively, e.g., P1 + DX6 stands for 1% ordinary steel fiber plus 6% melt-drawn type ultra-short ultra-fine steel fiber mixed concrete specimens.
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completely evenly distributed in the concrete specimens due to the high volumes of steel fibers. Thus, in future experiments, attention should be paid to the issue of how to achieve a uniform distribution of the steel fibers in the concrete. It can be concluded from the crack pattern that the higher the volume of steel fibers, the narrower the crack width and the more twisting the crack path, which is because the steel fibers prevent the crack from developing within the concrete and allow the crack direction to change when the concrete cracks. Regarding the specimens mixed with a higher volume of steel fibers (6%DX, 1%P+6%DX), two cracks are produced, and a small crack is also found next to both main cracks. When the steel fiber is mixed with 2%P+6%DX, even two narrow cracks appear, which can almost close after the load is removed. 4.2
A comparison of fatigue life
The final fatigue damage results for the six paving programs above are shown in Table 2.
Table 3. Specimen number
Fatigue test results for steel fiber reinforced self-compacting concrete specimens.
Program
Frequency (Hz)
Load (kN)
Damage Temperature pattern
1
S–0
4
8
20 C
2
P1
4
8
20 C
3
P2
4
8
20 C
4
DX6
4
8
20 C
5
P1+DX6
4
8
20 C
6
P2+DX6
4
8
20 C
Number of Number cracks of loads (million) (million)
Cracking in the 6.54 mid-span Cracking in the 8.25 mid-span Joint surface 9.90 separation Cracking in the 14.50 mid-span Cracking in the 21.70 mid-span Cracking in the 29.20 span
14.06 21.05 23.56 39.80 52.50 77.50
An analysis of Table 3 and Figure 8 shows that from S-0 to P2, the fatigue loading times at initial cracking are not particularly increased, but the final fatigue loading times increase by 67.6%. From P2 to DX6 to P1+DX6 to P2+DX6, the increase in the fatigue loading times at initial cracking and the final fatigue loading times become gradually greater, with the change from P1+DX6 to P2+DX6 being the most significant. The bending fatigue life of the P2+DX6 specimen is significantly better than the other types of specimen, as its loading times at cracking delay by 0.35 times compared to the P1 +DX6 specimen and by 3.5 times compared to the plain concrete specimen. The final fatigue loading times of P2+DX6 specimen increase by 47.3% compared to the P1 +DX6 specimen and by 450.0% compared to the plain concrete specimen. Before the specimens are damaged, the pavement layer is well-bonded to the steel bridge plate. None of the joints exhibited cracks or relative sliding, except for the P2 specimen, which exhibits joint surface separation; none exhibited cracks or relative cracks sliding. P2 +DX6 specimen also performs best with cracks, and the concrete can work with the steel plate, indicating that the P2+DX6 program is preferable. (The specimens mixed with steel fibers are less able to enhance the number of initial cracks, but fatigue damage is improved
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Figure 8.
Fatigue loading times of steel fiber reinforced self-compacting concrete specimens.
more, and ultra-short and ultra-fine steel fibers can enhance both the number of initial cracks and the ultimate fatigue loading times.)
5 CONCLUSION The results of the fatigue tests on the hybrid steel fiber reinforced self-compacting concrete specimens indicate that the fatigue performance of the pavement with the addition of hybrid steel fibers is significantly better than that of the plain concrete pavement and that the hybrid steel fiber reinforced concrete and the steel structure can work better together. The hybrid steel fiber reinforced self-compacting concrete pavement with a steel fiber admixture of 2% P and 6% DX is even better than the pavement with a steel fiber admixture of 2% P and 6% DX. It can work better with cracks under the design load. Its paving layer can meet the durability requirements, thus confirming that the hybrid steel fiber reinforced selfcompacting concrete pavement with 2% ordinary steel fibers + 6% ultra-short and ultrafine steel fibers can meet the requirements. A further increase in the volume can significantly improve fatigue performance, and SFRC is more resistant to fatigue than ordinary concrete pavement.
CONFLICTS OF INTEREST The authors declare that they have no conflicts of interest regarding the publication of this paper. 228
ACKNOWLEDGMENTS The author sincerely thanks China Railway 24th Bureau Group Xinyu Engineering Co., Ltd. for financial support.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Structural performance design analysis of a high-rise office building beyond the limit Xiaofang Cao, Ling Huan* & Erhong Hu Nanchang Hangkong University, Jiangxi Province, China
ABSTRACT: A high-rise office building with a height of 100.5 meters beyond the limit has three irregularities: floor discontinuity, sudden change of bearing capacity, and local irregularity. The project carries out seismic design according to the performance-based design concept to explore whether the performance of the overrun structure meets the expected standards. PKPM and MIDAS are used to calculate the overall and local structure and analyze the internal force and deformation of the structure under elastic and elastic-plastic conditions. The result analysis shows that the structure meets the code design and reaches the preset level C seismic performance target, which has good seismic performance. The component layout is safe and reasonable. Finally, according to the results of seismic performance analysis, corresponding seismic strengthening measures are taken for the weak parts of the structure.
1 INTRODUCTION The traditional aseismic design can only ensure the fortification goal of “strong earthquake does not collapse” of the structure, but cannot predict and reduce the degree of damage and damage of the structure, as well as the economic loss and repairability after the earthquake. The performance-based seismic design concept can not only predict the extent of the structure entering the elastic-plastic stage during the earthquake, but also effectively control the damage of the structure, which have the expected seismic function, and minimize the damage caused by the earthquake. It is a new seismic design based on the existing seismic theory and economic conditions (Wang 2014). Taking an overrun project as an example, this paper focuses on the performance-based design of its structure overrun. The performance-based structural design method is applied to the actual seismic analysis and design, which provides a certain reference for the over-limit high-rise seismic design method.
2 PROJECT OVERVIEW 2.1
Basic information
The project is located in Honggutan District, Nanchang City. The building covers an area of 26450 m2 and is used for a commercial office. The office building belongs to a complex highrise structure beyond the limit. The total height of the building is 93.8 meters. There are 23 floors above the ground and 1 floor below the ground. The height of the first floor above the ground is 3.6 meters. The large openings on the floor and the second floor connect to form a commercial hall. The second floor is 2.2 meters high. The design of the first and second floors above the ground is an inclusive staggered structure. The other standard floors are *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-30
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4.2 meters high. The design service life of the building structure is 50 years, the structural safety class is Class II, the project is a 6-degree seismic fortification, the design basic seismic acceleration is 0.1g (China Construction Industry Press 2008), the building seismic fortification class is Class C, the building site class is Class II, the design earthquake group is Group 1, and the characteristic period is 0.35 s. 2.2
Structural design
The tower is designed with a cast-in-place concrete frame-core tube structure system, and the main floor plan size of the structure is 41.1 m 31.2 m, of which the plane size of the interlayer is 38.1 m 18.35 m, the height-width ratio is 3.05, within the specification limit, the wall thickness of the core tube 1-2 layers is 400 mm, and the wall thickness of the 3-23 layers is 350 mm and 300 mm. The frame columns of the main building run through from the bottom to the top, and the section size gradually decreases with the height of the floor. The staggered columns on the second floor are short columns, as shown in Figure 1. The first floor has partially through columns, as shown in Figure 2. The concrete strength grade of each component is C50 - C30, and the concrete wall column grade of the bottom reinforcement layer is C50, which gradually decreases with the increase of the floor height. According to the “Technical Regulations for Concrete Structures of High-Rise Buildings” (China Building Industry Press 2011), it is determined that the seismic rating of the frame structure is Grade III, the seismic rating of the shear wall of the core tube is Grade II, and the embedded end of the superstructure is the basement roof.
Figure 1.
1 - 8 layer structure model diagram.
Figure 2.
Floor plan of the first floor.
3 STRUCTURE OVERRUN AND SEISMIC PERFORMANCE TARGET 3.1
Identification of overrun
To judge the out-of-gauge situation of the building structure, we can start from the horizontal and vertical irregularity of the structure, and preliminarily evaluate and calculate the structural system according to the engineering characteristics. According to the provisions of “Technical Key Points for Special Review of Seismic Fortification of Over-gauge High-rise Building Engineering” (Ministry of Housing and Urban-Rural Development of the People’s Republic of China 2006), we can know that the structure has the following irregular options 1. The floor slab is discontinuous, and the height difference of partial floor staggering is greater than the beam height, which belongs to the horizontal irregularity item.
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2. The bearing capacity mutation occurs, and the interlayer of 2.2 meters high appears on the second floor, and the floor height mutation results in the sudden change of the bearing capacity of this floor, which belongs to the vertical irregularity item. 3. The structure has a total of 6 through-story columns, and the torsion displacement ratio of a certain floor is greater than 1.2, which belongs to other irregularity items in the overrun. It is comprehensively judged that the project belongs to the over-limit high-rise building with serious irregular structure. 3.2
Determination of performance objectives
According to 4.12 of “Key Points of Technical Requirements for Special Review of Seismic Fortification of Over-gauge High-rise Building Projects”, the seismic performance grade of the structure needs to comprehensively consider the geographical factors and scale and use of the building as well as the irregular out-of-gauge situation of the structure, and discuss with the owner, after ensuring the economy and reliability of the design, the seismic performance target grade of the tower is determined as Grade C. “The Technical Specification for Concrete Structures of High-rise Buildings” indicates that each component of the structure plays different roles under different earthquake actions, so it is necessary to assign performance levels to them respectively, as shown in Table 1 for seismic performance target C. Table 1.
Seismic performance target C. Earthquake level
Key components
Ordinary vertical member Energyconsuming component
Performance level Overall structural performance target Bottom shear wall, outer frame beam column Piercing column at the bottom of 1-4 floors Shear walls, frame columns
Frequent Fortified earthearthquake quakes 1 Intact and undamage Elasticity
Elasticity
Elasticity
Shear wall coupling beam, concrete frame beam, and other beams
Elasticity
Frame beam
Elasticity
Rare earthquakes
3 Mild damage
4 Moderate damage
Shear elasticity, bending non-yielding Flexural elasticity Bending failure Shear elastic bending non-yielding Shear non-yielding, bending allowable yielding, plastic deformation control Shear elasticity and bending resistance
Shear and bending resistance without yielding
Some components allow yielding and control plastic deformation Most of them are allowed to yield and control plastic deformation
4 STRUCTURAL CALCULATION AND ANALYSIS 4.1
Elastic analysis under frequent earthquake
4.1.1 Mode decomposition response spectrum method The code requires that at least two mechanical models should be used for the calculation and analysis of complex out-of-limit structures. In this paper, two software, Satwe and Midas Building, are used for small earthquake elastic analysis of the structure, and the results are shown in Table 2. 233
Table 2.
Small earthquake calculation results. Calculation software
Calculate the number of modes Natural period 1st translation period T1 Second translation period T2 1st torsion cycle Tt
PKPM X
Y 2.66 (Y) 2.36 (X) 2.28 (T)
MIDAS X
Y 2.64 (Y) 2.11 (X) 2.25 (T)
1st torsion cycle Tt /1st translation period T1 0.857 0.852 effective mass coefficient 96.76% 97.37% 91.02% 93.42% First layer shear force under earthquake (KN) 72261.98 59500.07 82420.46 62136.30 Minimum shear weight ratio 0.80 0.81 0.81 0.81 Maximum interstory displacement angle under 1/4492 1/3217 1/4918 1/3248 seismic load Maximum displacement ratio under seismic force 1.19 1.24 1.15 1.22 Maximum interlayer displacement ratio under 1.15 1.21 1.17 1.22 seismic force Ratio of minimum floor shear capacity 0.65 0.65 0.73 0.66 Minimum stiffness ratio 1.0 1.0 1.0 1.0
Summary: from the data, it can be seen that the structural cycle, minimum shear weight ratio, inter-story displacement angle, stiffness ratio, and other indicators meet the requirements of the specification, and the difference between the calculation results of Satwe and Midas Building is within 20%. The results of the two are consistent, and the established model is reliable. In addition, the ratio of torsional displacement to bearing capacity exceeds the specification limits, but the overall structure and various components are in an elastic state, which meets the performance target requirements under small earthquakes and can be used for calculation and analysis under subsequent moderate and large earthquakes. 4.1.2 Elastic time-history analysis method In this paper, TH099TG035 (natural wave), TH095TG035 (natural wave) and RH2TG035 (artificial wave) were selected according to the Code for Seismic Design of Buildings (China Construction Industry Press 2016) for supplementary structural elasticity time history analysis. The purpose is to take the envelope value under the results of elastostatic analysis and time history analysis to balance the safety and economy of the structure. The peak acceleration of seismic wave is 18 mm/m2, and the two-way input is conducted in the main direction of X and Y respectively (the ratio of seismic wave peak in the main and secondary directions is (1:0.85). According to the results of elastic time history analysis, it can be seen that the specification requirements are met: According to Table 3, the difference between multiple groups of time history curves and the seismic influence coefficient curve adopted by the mode decomposition response spectrum method is less than 20%, which is consistent in a statistical sense; The average bottom Table 3. Period
Comparison of seismic wave acceleration spectrum and gauge spectrum data. TH099TG035/ gauge spectrum
T1 = 2.85 – 9.07% T2 = 2.45 – 4.87% T3 = 2.35 – 2.74%
TH095TG035/gauge RH2TG03/gauge spectrum spectrum
Seismic wave average/ gauge spectrum
– 10.39% 12.77% 0.06%
– 11.10% – 0.28% – 4.50%
– 13.83% – 8.73% – 10.82%
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Table 4.
Results of elastic time history analysis. Maximum inter-story displacement angle
/
Earthquake wave X TH099TG035 1/3416 TH095TG035 1/3338 RH2TG03 1/3533 Average value of the seismic wave 1/3456 CQC 1/3960
Y 1/2649 1/3055 1/2477 1/2899 1/3112
Maximum total base shear (KN) X 2538.85 2686.49 3910.17 3088.50 2922.72
Y 3409.17 2455.31 3272.64 3045.72 2963.27
Maximum total base shear/CQC X 87% 92% 133% 105% /
Y 115% 83% 110% 102% /
shear force of three groups of time-history waves is not less than 80% of the mode decomposition response spectrum method that is not more than 120%, and the output result of a single seismic wave is not less than 65% and not more than 135%, as shown in Table 4. The average base shear of seismic waves is compared with the maximum shear value of the base under the CQC method, and it can be seen from the results that it is necessary to adjust the seismic amplification coefficient to amplify the floor shear under the CQC method, considering the amplification factor of 1.25 times. 4.2
Floor stress analysis under seismic fortification
In the architectural design, the large opening of the first floor weakens the overall rigidity and stability of the floor and cannot ensure the effective transmission of the horizontal force in the plane and good shear resistance. Therefore, PKPM software is used to check the unyielding of the weak floor under the action of a moderate earthquake. Without considering the influence of wind load, and assuming that the floor is elastic modulus, Figure 3 shows the stress analysis results of the second floor under the action of a moderate earthquake. The concrete strength grade of the floor is C30. From Figure 3, it can be seen that the maximum tensile stress in the X direction of the floor is 2.15 Mpa, and the maximum tensile stress in the Y direction is 2.35 MPa, both of which are greater than the design value of concrete tensile strength 2.01 MPa, and the maximum tensile stress of the floor is far less than the yield stress of the reinforcement (Fu et al. 2008). Therefore, the floor slab meets the performance target of shear resistance and non-yielding under medium earthquakes.
Figure 3.
4.3
Stress diagram of 2-story floor slab.
Dynamic elastoplasticity under large earthquake
The seismic design goal of a large earthquake is the bottom line of seismic fortification of the building structure, and this paper uses the Midas building to analyze the dynamic 235
elastoplastic analysis of the structure under the action of rare earthquakes to test the component performance and structural deformation under the large earthquake. Three sets of time-history waves were selected with a peak acceleration of 125 cm/s2 and a characteristic period of 0.40 s. The coefficient ratio of seismic waves in the X and Y directions was 1:0.85 by taking the artificial wave RH4TG040 as an example to carry out the elastic-plastic analysis under the rare earthquake. Comparing the maximum floor shear force of the structure in the elastic state and the elastoplastic state, the base shear force in the X direction under the large earthquake was 13894 KN, which was 4.75 times that under the small earthquake, and the base shear force in the Y direction was 11214 KN, which was 3.78 times that under the mode decomposition reaction spectrum, indicating that the structure played the role of dissipating seismic energy under the large earthquake, and had good energy dissipation and energy absorption capacity (Liu et al. 2022; Yang et al. 2022), and also confirmed that the tower had a good safety reserve level. RH4TG040 artificial wave X-to-maximum interlaminar displacement angle of 1/656 and Y direction maximum interlaminar displacement angle of 1/504 do not exceed the elastoplastic specification limit (1/100), so the deformation under the large earthquake meets the requirements of the specification and meets the seismic fortification target of the large earthquake. 4.3.1 Plastic hinge state of frame beam and column By checking the development of the plastic hinge at each moment of frame beam and column under the action of seismic wave, the development of plastic hinge at the moment of maximum inter-story displacement angle (X direction: 19 s; Y direction: 19.12 s) is shown in Figure 4. The whole frame structure has almost no plastic deformation in the X-axis direction, and only 1.1% of the columns and 0.5% of the beams enter the second yield state, which mainly occurs on the upper floors. The plastic deformation of the frame beam in the Y direction is larger than that of the frame column. 78.1% of the beam hinges are in the state of yielding to yield, and a few of the beam hinges (accounting for 4.9%) are in the limit state. The bottom frame beam is severely damaged, but there are only a few plastic hinges on the upper floor and lower part of the frame column. Under rare earthquakes, most of the concrete frame columns are undamaged, the frame beams are moderately damaged, and a few are in the stage of serious damage. Therefore, the structural design meets the seismic design concept of a strong wall limb and weak coupling beam (China Construction Industry Press 2015). 4.3.2 Plastic hinge state of shear wall Under the loading of seismic waves, the damaged state of the core tube shear wall is shown in Figure 5. The damage degree of the Y-axis of the building shear wall is more serious than that of the X-axis, and the damage is mainly concentrated in the middle and lower floors. Due to the determination of the seismic performance target, the frame-core wall shear wall under a large earthquake can have a few serious failures without affecting the stability of the structure. According to the statistics, the proportion of serious failures of components is 10.3%. The plastic hinge of the shear wall is mainly concentrated on the shear wall connecting beam, which is the first component in the structure to produce shear yield deformation, and the damage is serious, and the damage gradually weakens with the increase of height. In architectural design, designers give priority to the role of connecting beams, and it intends to fully consider the project funds on the premise of ensuring structural safety. According to the analysis of the damage degree of frame columns and shear walls, each component can meet the performance requirements of performance level 4 under rare earthquakes.
236
Figure 4
Figure 5. diagram.
Yield state of frame structure.
Shear wall damage
5 MAIN SEISMIC STRENGTHENING MEASURES According to the performance analysis and calculation results of the structure, anti-seismic strengthening measures should be taken for the out-of-limit situation in the design. (1) The slenderness ratio of the cross-story column shall be strictly controlled. The higher the slenderness ratio is, the more flexible the column is than the ordinary frame column, which is prone to instability and damage. Therefore, the strength of the cross-story column shall be guaranteed, and the seismic performance target shall be set for key components. (2) Split-level columns are generally short columns, which are prone to brittle shear failure, and it is necessary to control the axial compression ratio of split-level columns that are not easy to be too small, and at the same time, it needs to strengthen the column stiffness (Wang 2014), and improve the seismic level of split-level columns during design. (3) If the floor is opened in a large hole, it is necessary to thicken the opening area as a whole, double reinforcement in both directions and specifically increase the reinforcement of the weakly connected plate according to the stress analysis of the floor slab to meet the effective transmission of horizontal force under large earthquakes and coordinate the deformation of components. (4) For the sudden change of the bearing capacity of the first floor caused by the height difference of adjacent floors, the thickness of the shear wall of the first floor can be thickened or the concrete grade of the 12 wall columns can be increased to improve the shear bearing capacity of the first floor, and the bearing capacity needs to be increased by 1.1 times during the calculation.
6 CONCLUSION In this paper, aiming at the overrun situation caused by local split-layer, SATWE and MIDAS building are used to calculate and analyze the structure as a whole and part, and verify that the structure has good safety performance. The damage degree of ordinary vertical components, energy-dissipating components and key structures of the structure under the action of multiple, fortified and rare earthquakes can reach the seismic grade of the preset seismic performance target C, and the internal force and deformation of the overall structure are also within the specified range. According to the structural layout and seismic analysis, the weak points are found, and corresponding structural measures are taken to strengthen them. Using the advantages of performance-based design to distinguish the traditional seismic design, according to each performance level, the damage degree of each 237
component is judged and prevented, and the degree of damage to the structure under the earthquake is controlled, achieving different fortification objectives with multiple risk hazard levels. It provides a reference for the design method of over-limit high-rise seismic resistance.
REFERENCES China Construction Industry Press, (2008) Classification Standard for Seismic Fortification of Building Engineering. Beijing. GB50023-2008 [S]. China Building Industry Press, (2011) Technical Specification for Concrete Structures of High-rise Buildings. Beijing. JGJ3-2010 [S]. China Construction Industry Press, (2016) Code for Seismic Design of Buildings. Beijing. GB 50011-2010 [S]. China Construction Industry Press, (2015) Code for Design of Concrete Structures. Beijing. GB 50010-2010 [S]. Fu, C., Liu, C., Li, Y. (2008) Stress Analysis of Weak Connection Concrete Floor of High-rise Buildings [J]. Building Structure, 38 (3): 106–1 10. 10.19701/j.jzjg.2008.03.024. Liu, Z., Hu, X., Liu, T., Li, R., Chen, T. (2022) Design of a High-rise Structure with Large-span Connection [J]. Building Structure, 52 (S2): 46–50. 10.19701/j.jzjg.22S2616. Ministry of Housing and Urban-Rural Development of the People’s Republic of China, (2006) Technical Points for a Special Review of Seismic Fortification of High-rise Buildings Beyond the Limit. Beijing JZ [2006] No. 220 [A]. Wang, X. (2021) Seismic Analysis of a Super High-rise Building Based on Performance Requirements [D]. Master’s thesis, Shandong University of Architecture, 10.27273/d.cnki.gsajc.2021.000369. Wang, S. (2014) Seismic Performance Analysis of Frame Structures with Short Columns Formed at Staggered Floors [D]. Master’s thesis, Taiyuan University of Technology. https://kns.cnki.net/kcms/detail/detail. aspx?FileName=1014418345.nh&DbName=CMFD2015 Yang, C., Cao, L., Xia, Lu. (2022) Performance-based Design and Damage Analysis of Super-high Shear Wall Structure of a Project [J]. Shanxi Architecture, 48 (21): 66–70. 10.13719/j.cnki.1009-6825.2022.21.017.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Study on stress intensity factor of steel wire with double surface cracks Hongsheng Xu & Ying Yang* Changsha University of Science and Technology, Changsha, China
ABSTRACT: The wire crack stress strength factor is an important parameter for the fatigue fracture life evaluation, fatigue crack growth analysis and wire fracture strength evaluation of steel wire. In this paper, the crack propagation analysis software FRANC3D is introduced first, and then the stress intensity factor of steel wire with surface double cracks under different tensile loads is simulated based on the submodel method. The research subjects of double cracks are parallel double cracks and collinear double cracks. Taking the stress intensity factor of a surface single crack as the control parameter, the influence of the size, position and tensile load of the auxiliary cracks on the interaction between cracks is discussed when the size of the main crack is constant. The results show that when the depth of the main and auxiliary cracks of the parallel double cracks is equal and the distance between the parallel double cracks is closer, the interaction between the main and auxiliary cracks is only shielding. With the increase of the depth ratio or parallel spacing of the main and auxiliary cracks, the shielding effect between the parallel double cracks will gradually weaken; When the main and auxiliary crack depths of collinear double cracks are equal and the angle between collinear double cracks is smaller, the interaction between main and auxiliary double cracks is only manifested as enhancement, and the enhancement effect between main and auxiliary double cracks will gradually weaken with the increase of the ratio of main and auxiliary crack depth or circumferential angle. It is more effective to evaluate the fatigue life and fracture strength of double-crack steel wire by using the variation law of stress intensity factor of double-crack steel wire.
1 INTRODUCTION The cable is an important force component of the cable-stayed bridge, and its safety and durability will directly affect the life of the whole bridge structure. Fatigue failure is possible on all components which are subjected to cyclic loads and have high stress levels. Due to the failure of the cable protection system, the wet environment will corrode the steel wire, and the corrosion will lead to the fatigue crack initiation of the steel wire (Wang & Zheng 2019). The crack stress intensity factor is an important parameter for fatigue fracture analysis, fatigue crack propagation life evaluation and steel wire fracture strength evaluation. At present, many scholars have done a lot of research on the stress intensity factor of a single crack of steel wire and achieved a lot of research results. Mahmoud obtained the polynomial of the stress strength factor for the propagation of the single crack of steel wire under tensile load and evaluated the fracture strength of steel wire by using the average value of the measured fracture toughness according to the measured critical crack depth (Mahmoud 2007). Based on the propagation characteristics of the cable steel wire crack ‘first round and then flat’ of the actual bridge, Zeng et al. proposed a practical geometric correction coefficient *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-31
239
of stress intensity factor fitting. The formula of the circular front crack and the straight front crack was fitted to evaluate the fracture strength of the cracked steel wire of the main cable of the suspension bridge (Zeng et al. 2009). Qiao et al. simulated and analyzed the stress intensity factor of steel wire with surface crack based on the submodel method. According to the crack propagation process, the stress intensity factor shape correction coefficient expression of steel wire with surface crack is obtained by fitting (Qiao et al. 2017). The fatigue cracks generated by the steel wire often appear together with multiple cracks, so there is an interaction of multiple cracks during crack initiation and propagation. Lam et al. studied the influence of multi-crack elastic interaction on the stress intensity factor, and the results showed that the interaction between microcracks can have an enhanced or shielding effect on the stress intensity factor according to the position and direction of microcracks (Lam et al. 1991). Jiang et al. studied double unequal parallel cracks in finitewidth plates subjected to remote tensile loads. They found that due to the crack interaction, the stress intensity factor at the tip of the double crack decreases at the same time. When the length difference between the double cracks is large, the short crack is dormant, and the effect is negligible (Jiang et al. 1990; Jiang et al. 1992). It can be found that most of the previous studies on the interaction between multiple cracks are finite wide plates and infinite thin plates, and there is not much research on slender cylinders. In this study, the finite element method was used to study the variation of the crack stress intensity factor under the influence of the size, distance and circumferential angle of the parallel double cracks with the common bisector and the collinear double cracks in the same plane under different tensile loads. The enhancement or shielding effect between the double cracks on the surface of the steel wire is revealed, and the focus is on determining whether and how the stress intensity factor of the double cracks of the steel wire is affected.
2 CRACK PROPAGATION ANALYSIS SOFTWARE FRANC3D At present, the calculation methods of stress intensity factor are mainly divided into the analytical method, numerical method and direct measurement method. However, due to the complex boundary conditions under actual working conditions, the analytical method is difficult to apply. The method of determining the stress intensity factor by directly measuring the stress and displacement near the crack tip has higher precision requirements for the detection equipment and increases the cost of detection. There is no systematic method for the analytical method of the stress intensity factor of multiple cracks, so the stress intensity factor of multiple cracks is solved by the finite element method. Franc3D adopts adaptive mesh generation technology, which can always maintain the high quality of the crack tip mesh, and can carry out multi-crack, multi-condition and load step analysis. It is also the only software that can simultaneously calculate the stress intensity factors of three crack types in anisotropic materials. The calculation flow chart for crack growth analysis to calculate stress intensity factor and fatigue life is shown in Figure 1. In recent years, many scholars have combined FRANC3D software based on the submodel method to calculate the stress intensity factor of components. FRANC3D can be combined with finite element analysis software such as ABAQUS and ANSYS to calculate the fracture parameters such as stress intensity factor, J integral, M integral and crack propagation life at the crack tip, which is quick and easy to calculate (Jia et al. 2004). This section mainly introduces its application in the crack stress intensity factor. The energy expression for calculating the stress intensity factor in FRANC3D software is (Chen et al. 2006): ! ð ð ð2Þ ð1Þ @q @q ð1;2Þ ð1Þ @ui ð2Þ @ui M ¼ G sij þ sij ds G W 1;2 d1j ds (1) @x1 @x1 @xj @xj
240
Figure 1.
FRANC3D calculation and analysis process.
In Formula (1), G is the integral loop around the crack tip where: ð1Þ ð2Þ
ð2Þ ð1Þ
W ð1;2Þ ¼ sij eij ¼ sij eij :
(2)
The relationship between M-integral and the stress intensity factors of three different basic crack forms is: ð1;2Þ ¼ 1 n K ð1Þ K ð2Þ þ 1 n K ð1Þ K ð2Þ þ 1 þ n K ð1Þ K ð2Þ : M I I II II III III E E E 2
2
(3)
For composite cracks with multiple crack characteristics, the calculation formula is as follows: qffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi ð2Þ 2 (4) Kequ ¼ KI þ bII KII2 þ bIII KIII DKequ ¼ Kequ max Kequ min :
(5)
In Formula (5), DKequ is the equivalent stress intensity factor, bII and bIII are the weighting factors.
3 CALCULATION MODEL OF DOUBLE CRACK STEEL WIRE Corroded steel wires are studied by simplifying the rule that uses semi-oval cracks instead of actual surface pits (Jiang et al. 2009, 2018; Mahmoud 2007; Nakamura et al. 2013; Sun 2018). According to the corrosion degree of the steel wire, the measured macroscopic defect depth range is about 0.1 mm to 0.6 mm (Jiang et al. 2018; Li et al. 2014; Nakamura et al. 2013). In this paper, the steel wire calculation model was studied by semi-oval cracks according to the simplified rule, and the crack depths were 0.2 mm, 0.4 mm and 0.6 mm. The sample parameters are based on the performance of the cable steel wire, as shown in Table 1, and the size and position of the double cracks on the surface of the steel wire are shown in Table 2. The tensile load in the steel wire calculation model is taken according to the steel wire fatigue test specification (GB/T17101-2019.2019). The stress ratio is 0.1, and the stress range is 420 MPa, 390 MPa, 360 MPa and 330 MPa. The corresponding maximum values of 467 MPa, 433 MPa, 400 MPa and 367 MPa are obtained. In ABAQUS, the relevant load and boundary conditions are defined. The lower section constrains the translation and rotation in the XYZ direction, the upper section constrains the rotation in the XYZ direction and the translation in the XY direction, and the translation in 241
the Z direction is not fixed. The analysis step applies a uniform tensile load in the Z direction to the upper section. When meshing the steel wire model, the steel wire matrix is divided into three regions: the crack-free part at both ends and the crack-containing part in the middle. The mesh size of the crack-free part along the axial direction is 0.5 mm, the circular section layout dimensions are 0.3 mm and 0.2 mm respectively, and the mesh attribute is C3D8R. After the stress analysis of crack-free steel wire in ABAQUS, the ABAQUS port is connected in Franc3D to intercept the middle area of the steel wire model. Then, double cracks are inserted in the middle of the model, and the mesh is automatically divided. The size of the main crack in the middle of the double crack model is unchanged, and the auxiliary crack varies according to the depth of 0.6 mm, 0.4 mm and 0.2 mm respectively. The axial distance between the parallel double cracks is distributed according to 0.5 mm, 1.0 mm and 1.5 mm, and the angle between the collinear double cracks is distributed according to the vertical and opposite sides, as shown in Figure 2. Table 1.
Parameters of high strength steel wire.
Material
E/Gpa
fu /Mpa
fy /Mpa
n
Diameter/mm
Length/mm
High strength galvanized steel wire
210
1670
1860
0.3
7
260
Table 2.
Size and location of double cracks on the steel wire surface.
Groups
Numbering
Crack size/mm
Types of crack
A1
1
H = 0.6 b = 2.84
Single crack
A2
2
H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0:5 q ¼ 0 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 1:0 q ¼ 0 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 1:5 q ¼ 0 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0:5 q ¼ 0 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 1:0 q ¼ 0 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 1:5 q ¼ 0 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0:5 q ¼ 0 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 1:0 q ¼ 0
3 4 5 6 7 8 9 10
A3
11 12 13 14 15 16
H2 ¼ 0:6
b2 ¼ 2:84
H2 ¼ 0:6
b2 ¼ 2:84
H2 ¼ 0:6
b2 ¼ 2:84
H2 ¼ 0:4
b2 ¼ 2:34
H2 ¼ 0:4
b2 ¼ 2:34
H2 ¼ 0:4
b2 ¼ 2:34
H2 ¼ 0:2
b2 ¼ 1:66
H2 ¼ 0:2
b2 ¼ 1:66
H1 ¼ 0:6 b1 ¼ 2:84 H2 ¼ 0:2 d ¼ 1:5 q ¼ 0
b2 ¼ 1:66
H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0 q ¼ 90 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0 q ¼ 180 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0 q ¼ 90 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0 q ¼ 180 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0 q ¼ 90 H1 ¼ 0:6 b1 ¼ 2:84 d ¼ 0 q ¼ 180
H2 ¼ 0:6
b2 ¼ 2:84
H2 ¼ 0:6
b2 ¼ 2:84
H2 ¼ 0:4
b2 ¼ 2:34
H2 ¼ 0:4
b2 ¼ 2:34
H2 ¼ 0:2
b2 ¼ 1:66
H2 ¼ 0:2
b2 ¼ 1:66
Parallel double cracks
Double collinear cracks
242
Figure 2.
Double crack finite element model.
4 STRESS INTENSITY FACTOR ANALYSIS OF STEEL WIRE DOUBLE CRACK 4.1
Concept of interaction between double cracks
Based on the stress field model of elastic crack tip given by Westergaad and Sneddon, Irwin defined the concept of stress intensity factor (SIF) in 1956. In the process of crack initiation and propagation, it may be affected and acted by adjacent cracks, that is, there is an interaction between multiple cracks under fatigue load (Dundar et al. 2015). However, for multiple cracks, driving forces, such as stress intensity factors (SIFs), should be affected by the interaction between multiple cracks, and ignoring crack interactions will result in an inaccurate estimation of fatigue levels of engineering components (Kim et al. 2015; Mahadevan et al. 2001; Tan et al. 2015). Therefore, it is of great significance to study the interaction between multiple cracks. The early work of Lam and colleagues showed that for the configuration of two identical and parallel microcracks, the space around the microcracks can be divided into the so-called “enhanced” and “shielded” regions (Lam et al. 1991). The term “enhancement” refers to the increase of stress intensity factor. It can be expressed numerically by KI =KI 0 >1, where KI 0 is the mode I stress intensity factor of a single crack. However, the term “shielding” refers to the reduction of stress intensity factor or KI =KI 0 in mathematical interpretation. 4.2
Calculation results of single crack stress intensity factor
It can be seen from the finite element model that the cracked steel wire model bears the uniform tensile load, so the stress intensity factor at the crack tip is mainly type I and KI is considered to be a parameter to evaluate the interaction between double cracks. Figure 3 shows the stress strength factors of each point of crack under four different tensile loads of a single crack with a depth of 0.6 mm, of which point A is the leftmost point of the surface crack, point B is the middle position of the crack, and point C is the rightmost end of the crack. It can be seen from Figure 4 that the changing trend of the stress intensity factor at the crack tip is consistent under different stress amplitudes, and the stress intensity factor at each point of the crack front end first increases and then decreases from point A to point C, and reaches a maximum value at point B. The shape of the equivalent crack front is semielliptical, and the changing trend of the stress intensity factor is the same as that of the linear crack (Qiao et al. 2017). Under tensile load, the stress intensity factor at the deepest point of the crack front edge is always greater than that of the outer surface, and the crack starts to expand from the center of the front edge curve. When the crack extends to a certain length, the crack has the same propagation rate in the depth direction and the outer surface direction, and the curvature of the crack front curve reaches the maximum. After that, the propagation velocity of the crack in the outer surface direction exceeds that in the depth direction, and the curvature of the crack front curve decreases gradually until the fracture occurs (Yang 2005). 243
Figure 3.
4.3
The stress intensity factor value of single crack tip under different tensile loads.
Stress intensity factor calculation results of parallel double cracks
The parallel double cracks studied in this paper consider the relationship between the different sizes of double cracks and the interaction between cracks. The interaction between the main crack and the auxiliary crack is mainly discussed. The size of the auxiliary crack is less than or equal to the main crack, which means that the main crack is more “dangerous” than the auxiliary crack. Therefore, in the following analysis, we focus on the influence of the auxiliary crack on the stress intensity factor of the main crack to determine the shielding and enhancement effects between the double cracks. Figure 4 shows that the stress intensity factor ratio of the crack tip A, C points and the deepest point B at the crack front of the parallel equal-length double cracks has the same change trend under different tensile loads, and its value increases with the increase of the axial distance, approaching 1. There is a small deviation between the calculated value of the stress intensity factor at point A and the stress intensity factor at point C. This is mainly because the mesh of the semielliptical crack is not symmetrical when the model has meshed although the geometric model is completely symmetrical. Therefore, in the follow-up study, one of the two points A and C can be selected for analysis. When the axial distance is the minimum value of 0.5 mm (about the depth of the main crack), the minimum value KI =KI 0 is about 0.67 at the maximum tensile load, which means that when the parallel double cracks are closer, the interaction between the main and auxiliary cracks is shielded. When the axial distance between parallel double cracks increases, KI =KI 0 gradually increases to about 1, which means that with the increasing distance between parallel double cracks, the shielding effect between cracks is also weakened. Until KI =KI 0 tends to be 1, there is no interaction between the two parallel cracks. These results indicate that the axial distance between two parallel cracks plays a key role in crack interaction. At the same time, it can be observed from Figure 4 that a larger tensile load will produce a greater shielding effect, but under a smaller tensile load, the KI =KI 0 value and the changing trend tend to coincide, so the interaction between different tensile loads on cracks remains to be evaluated. It can be seen from Figure 5 that the KI =KI 0 value of the parallel double cracks is less than 1, indicating that the stress intensity factor at the front end of the main crack is smaller than 244
Figure 4. Comparison of stress intensity factors of equal length parallel double cracks and single cracks under different loads.
Figure 5. Comparison of stress intensity factors between unequal parallel double cracks and single cracks under different loads.
245
that of the single crack when there is an auxiliary crack. Under the same tensile load, when the size of the main crack and the axial distance between the parallel double cracks are kept constant, the value of KI =KI 0 increases with the decrease of the depth and length of the auxiliary crack, indicating that the shielding effect between the double cracks is gradually weakened. When the depth ratio of the main and auxiliary cracks is in the range of 1–1.5, the value of KI =KI 0 changes most obviously, indicating that when the two cracks are close in size, the shielding effect between the cracks is strong; The interaction between the two cracks is weak when the depth of the auxiliary crack is reduced to half of the depth of the main crack. In summary, when the depth of the auxiliary crack in the parallel double cracks is less than one-third of the main crack, or the axial distance between the main and auxiliary cracks is greater than three times the depth of the main crack, it can be considered that the two parallel cracks do not affect each other. In other cases, the auxiliary crack has a weakening effect on the stress intensity factor at the tip of the main crack, that is, there is a shielding effect between the two cracks. 4.4
Stress intensity factor calculation results of collinear double cracks
It can be seen from Figure 6 that the KI =KI 0 value of collinear double cracks is greater than 1, indicating that when the two cracks are collinear, the stress intensity factor at the front end
Figure 6. Comparison of stress intensity factors between unequal collinear double cracks and single cracks under different loads.
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of the main crack is larger than that of the single crack. Under the same tensile load, when the size of the main and auxiliary cracks is kept unchanged, as the circumferential angle between the collinear cracks continues to increase, that is, the collinear crack tip continues to stay away, the value of KI =KI 0 continues to decrease, indicating that the enhancement effect between the collinear cracks gradually decreases with the increase of the circumferential angle. When the circumferential angle between the collinear double cracks is 180 (paired side distribution), the value of KI =KI 0 tends to be stable near 1, and the change of the depth ratio of the main and auxiliary cracks has little effect on the interaction between the cracks. When the angle between the auxiliary crack and the main crack is 90 and the depth ratio of the main crack to the auxiliary crack is in the range of 1–1.5, the change of KI =KI 0 is prominent, indicating that the enhancement effect between the equal-length double cracks is obvious. In general, when the main and auxiliary cracks are less than one-third of the main crack depth or the main and auxiliary cracks are distributed on the opposite side, it can be considered that there is no interaction between the two collinear cracks.
5 CONCLUSION In this paper, the stress intensity factor of steel wire with surface double cracks is simulated and analyzed based on the submodel method. By taking the stress intensity factor of the surface single crack as the control group, the influence of the size, position and tensile load of the auxiliary crack on the interaction between the cracks is discussed when the size of the main crack is unchanged. The interaction between different positions and sizes of double cracks in strengthening or shielding is determined. The main conclusions are as follows: 1) When the depth of the main and auxiliary cracks of the parallel double cracks is equal and the distance between the parallel double cracks is closer, the stress intensity factor of the main crack is less than that of the single crack, that is, the interaction between the main and auxiliary cracks is only shielding. With the increase of the depth ratio of the main and auxiliary cracks or the parallel spacing, KI =KI 0 gradually increases to 1, which means that the shielding effect between parallel double cracks will gradually weaken. When the depth of the auxiliary crack in the parallel double cracks is less than one-third of the main crack, or the axial distance between the main and auxiliary cracks is greater than three times the depth of the main crack, it can be considered that the two parallel cracks do not affect each other. 2) When the depth of the main and auxiliary cracks of the collinear double cracks is equal and the angle between the collinear double cracks is smaller, the stress intensity factor of the main crack is greater than that of the single crack, so the interaction between the main and auxiliary cracks is only enhanced. With the increase of the depth ratio or circumferential angle of the main and auxiliary cracks, that is, the collinear crack tip keeps away, and the value KI =KI 0 decreases to 1, which means that the enhancement effect between the main and auxiliary cracks will gradually weaken. When the auxiliary crack depth of collinear double cracks is less than one-third of the main crack depth or the main and auxiliary cracks are distributed on the opposite side, it can be considered that there is no interaction between the two collinear cracks. 3) If two parallel cracks are close and share the same vertical double sectors, only the shielding effect exists. In this case, it would be too conservative or even unreasonable to simply merge them into larger cracks through combination rules. If the angle between two collinear cracks in the same section is small, only the enhancement effect exists. At this time, the component safety of collinear multi-crack steel wire should be considered more. Therefore, it is of great significance to determine and study the enhancement, shielding or non-interaction effects between double cracks and apply them to actual structures with similar multi-crack configurations. 247
4) In parallel double cracks and collinear double cracks, the larger tensile load will make the value KI =KI 0 of the main crack change obviously, but under the smaller tensile load, the changing trend tends to coincide. Therefore, the effect of different tensile loads on the interaction between double cracks remains to be evaluated.
REFERENCES Chen, J. Zhao, S. S. (2006). Fracture Mechanics. M. Science Press, Beijing. Dundar, H. Ayhan, AO. (2015). Three-dimensional Fracture and Fatigue Crack Propagation Analysis in Structures with Multiple Cracks. J. Computers & Structures.158, 259–273. GB/T17101-2019. (2019). Hot-dip Zinc or Zinc-aluminium Coated Steel Wires for Bridge Cables. S. Standards Press of China, Beijing. Jiang, Z. D. Zeghloul, A. Bezine, G. Petit, J. (1990). Stress Intensity Factors of Parallel Cracks in a Finite Width Sheet. J. Engineering Fracture Mechanics. 35 (6), 1073–1079. Jiang, Z. D. Petite, J. Bezine, G. (1992). An Investigation of Stress Intensity Factors for Two Unequal Parallel Cracks in a Finite Width Plate. J. Engineering Fracture Mechanics. 42, 129–138. Jia, X. M, Wang, Q. Z. (2004). Three Dimensional Fracture Analysis Software FRANC3D J. Chinese Journal of Computational Mechanics. 21 (6), 764–768. Jiang, J. H. Ma, A. B. Weng, W. F. Fu, G. H. Zhang, Y. F. Liu, G. G. Lu, F. M. (2009). Corrosion Fatigue Performance of Pre-split Steel Wires for High Strength Bridge Cables J. Fatigue & Fracture of Engineering Materials & Structures. 32 (9), 769–779. Jiang, C. Wu, C. Jiang, X. (2018). Experimental Study on Fatigue Performance of Corroded High-strength Steel Wires Used in Bridges. J. Construction & Building Materials 187 (3), 681–690. Kim, E. K. Choi, H. Park, K. Kang. W, H. (2015). Deterministic and Probabilistic Investigation on Multiple Crack Interactions in a Semi-Infinite Domain. J. Mathematical Problems in Engineering. 2015 (1), 1–9. Lam, K. Y. Phua, S. P. (1991). Multiple Crack Interaction and its Effect on Stress Intensity Factor. J. Engineering Fracture Mechanics. 40, 585–592. Li, S. L. Xu, Y. Li, H. Guan. X, C. (2014). Uniform and Pitting Corrosion Modeling for High-Strength Bridge Wires. J. Journal of Bridge Engineering. 19 (7), 1. Mahadevan, S. Shi, P. (2001). Corrosion Fatigue Reliability of Aging Aircraft Structures. J. Progress in Structural Engineering and Materials.3 (2), 188–197. Mahmoud, K. M. (2007). Fracture Strength for a High Strength Steel Bridge Cable Wire with a Surface Crack. J. Theoretical and Applied Fracture Mechanics. 48 (2), 152–160. Nakamura, S. I. Suzumura, K. (2013). Study on Fatigue Strength of Corroded Bridge Wires J. Journal of Bridge Engineering. 18 (3), 200–209. Qiao, Y. Miao, C. Q. Sun, C. Z. (2017). Crack Stress Intensity Factor of Wire with Surface Crack based on Sub-model Method. J. Chinese Journal of Computational Mechanics 34 (2), 238–243. Sun, B. (2018). A Continuum Model for Damage Evolution Simulation of the High Strength Bridge Wires Due to Corrosion Fatigue. J. Journal of Constructional Steel Research. 146, 76–83. Tan, J. T. Chen, B. K. (2015). Prediction of Fatigue Life in Aluminium Alloy (AA7050-T7451) Structures in the Presence of Multiple Artificial Short Cracks. J. Theoretical and Applied Fracture Mechanics.78, 1–7. Wang, Y. Zheng, Y. Y. (2019). Research on Corrosion Fatigue Performance and Multiple Fatigue Sources Fracture Process of Corroded Steel Wires. J. Advances in Civil Engineering. 2019, 1–25. Yang, F. P. (2005). Experimental Research on Fatigue Growth of Surface Cracks in a Round Bar. D. Shanghai Jiao Tong University. Zeng, Y. Chen, A. R. Ma, R. J. (2009). Fracture Strength of Wires with Cracks in Suspension Bridge. J. Journal of Tongji University (natural science). 37 (8), 1010–1013.
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Intelligent building and equipment installation technology
Taylor & Francis Taylor & Francis Group http://taylorandfrancis.com
Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Utilization of recycled construction waste filler in urban greenway Peichen Cai, Xuesong Mao*, Xiaoyong Lai & Qian Wu College of Highway, Chang’an University, Xi’an, P.R. China
Xiang Li Jiangsu Suhuaiyan Expressway Management Co., Ltd, Huai’an, P.R. China
Xiaojun Shi Xuzhou Transportation Planning, Design and Research Institute Xuzhou, P.R. China
ABSTRACT: Sponge City is an innovative approach to promote the construction of green buildings, develop low-carbon cities, and create smart cities. This paper proposes an improvement to the green road pavement structure using recycled construction waste materials, based on the concept of a sponge city. Taking into account the secondary crushing of bricks in recycled construction waste materials, the paper proposes a design method for the gradation composition of the recycled material for the previous concrete base course. The best mix ratio is determined through trial mix tests and strength tests, considering the road performance of the recycled material. The research results show that the design index of the recycled material pervious concrete base course mixture should have a design porosity of not less than 25%, a permeability coefficient of not less than 0.35 cm/s, and a 7-day unconfined compressive strength of not less than 3.5 MPa. Furthermore, the best mix proportion is 0.65 water-cement ratios and 167.4 kg/m3 cement consumption when the design porosity is 25%, as determined through trial mixing tests and strength tests.
1 GENERAL INSTRUCTIONS With China’s rapid economic development and urbanization, the amount of construction waste generated by urban infrastructure and highway projects is increasing (Li 2022; Li & Mao 2021; Lv et al. 2021; Wei et al. 2023). In recent years, the annual output has ranged from 1.55 billion to 2.4 billion tons, with a current stock of over 20 billion tons. Urban waste accounts for approximately 40% of this total (Chang et al. 2021). The three main sources of construction waste are the demolition of old buildings (58%), construction of new buildings (36%), and decoration of new buildings (6%). Due to the lack of effective management and disposal methods, most construction waste is transported to the suburbs for open stacking or simple landfill without treatment. This practice harms the urban and ecological environment and causes social problems (Xue et al. 2022; Xie et al. 2022). In light of the urgent need to dispose of construction waste and the push for sponge city construction, developing construction waste recycling materials for permeable greenway paving is significant. This approach addresses the recycling of construction waste and deepens the sponge city construction concept. Therefore, this paper aims to utilize construction waste recycling materials in the permeable concrete base of urban greenways, based on the concept of the sponge city. It focuses on *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-32
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the technical challenges associated with the construction of urban greenways and provides solutions for the application of construction waste recycling materials in such areas.
2 PROPOSING THE CONCEPT OF SPONGE CITY SELECTING As its name suggests, a sponge city is a city that can absorb, store, and purify water like a sponge, and release water in a scientific manner to make it functional. Sponge cities represent a new generation of urban rainwater and flood management concepts, which means that they possess good adaptability to environmental changes and can cope with natural disasters caused by rainwater. This adaptability can also be referred to as “water-elasticity”. During the construction of urban greenways, the “sponge city” concept is incorporated. This involves utilizing the greenway pavement structure to facilitate rainfall infiltration, retention, storage, and discharge, as well as leveraging the natural purification effects of vegetation, soil, and wetlands to improve water quality. By employing a combination of natural and artificial sponge designs, multiple sources are created to control runoff and pollution caused by rainwater, while simultaneously enhancing the city’s aesthetic appeal.
3 IMPROVEMENT OF GREENWAY PAVEMENT STRUCTURE The concept behind sponge city roads is to incorporate road motorways, sidewalks, green belts, and other low impact development (LID) facilities, such as permeable pavement, sunken green spaces, and rain wetlands, to achieve the “sponge” function of urban roads. This involves infiltration, purification, regulation, and storage, as well as ecological drainage, while ensuring that traffic function and safety are met. 3.1
Permeable slow lane
Permeable bricks are suitable for paving permeable slow-moving roads. The surface course consists of 6 cm permeable brick, the screed course is made of 3 cm medium coarse sand, and the base course is made of 25 cm graded crushed stone. Drainage hoses are embedded in the base course to connect with the roadside green belt, as illustrated in Figure 1.
Figure 1.
3.2
Optimal design of the previous slow track.
Slow traffic road with construction waste recycling materials as permeable base
Most of the current research and projects focus on preparing previous concrete base courses using natural aggregates. However, in recent years, construction waste recycling materials have been widely used as a substitute for natural aggregates in the subgrade, base course, and even surface course materials. This paper aims to study the application of construction waste
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recycling materials in the greenway pavement structure of the previous concrete base course, based on the concept of a sponge city. Through a series of tests, this study proposes a reasonable grading composition and technical indicators for road performance to address the key technical challenges of permeable pavement using construction waste recycling materials. This will expand the ecological permeable pavement’s “green” meaning and deepen the connotation of “sponge city.” 4 APPLICATION TECHNOLOGY OF RECYCLED MATERIAL PERMEABLE BASE 4.1
Grading selection
Grading 1 is typically chosen by referencing various documents, including the Code for Design of Highway Drainage (JTG/T D33-2012), Technical Specification for Pervious Asphalt Pavement (CJJT 190-2012), Code for Design of Highway Cement Concrete Pavement (JTG D40-2011), and others. Grading 2, on the other hand, is selected by using the median grading of No.56 cement stabilized macadam drainage base recommended by AASHTO. Table 1 displays both Grading 1 and Grading 2. Table 1.
Test grading.
Sieve size (mm)
31.5
26.5
19
16
13.2
9.5
4.75
2.36
0.075
Grading 1 Grading 2
100 –
91 100
70 95
52 75.9
43.5 56.7
27 37.5
5 5
3 2.5
– –
4.2
Grading correction considering secondary crushing of construction waste
Due to the high brick content in recycled construction waste materials, which is susceptible to the secondary crushing, the selected typical grading may change, resulting in reduced porosity and compromised drainage performance. Therefore, the selected typical grading is adjusted in accordance with the correction principle outlined by Lai et al. (2022). This principle stipulates that the content of particles less than 2.36 mm should not exceed 5%, while the content of particles less than 4.75 mm should not exceed 10%. To achieve this, each barrier in the grading is increased or decreased accordingly. The compaction tests must adhere to the aforementioned principles, and both grading 1 and grading 2 need to be rectified. The grading curve is illustrated in Figures 2(a) and (b). These figures indicate that while correcting grading 1 and grading 2, grading 1 undergoes significant alterations, whereas grading 2 remains relatively stable, with only a substantial change observed at 16 mm. Consequently, grading 2 will serve as the testing grading in the subsequent investigation. 4.3
Trial mixing test of recycled permeable base material
The recommended water-cement ratio range for preparing pervious concrete using natural aggregates is 0.25–0.35. However, due to the high-water absorption rate of building waste recycling materials, a water-cement ratio range of 0.40–0.80 with a variation step of 0.10 is selected for this test. The test will include water-cement ratios of 0.40, 0.50, 0.60, 0.70, 0.80, and 0.90. The range will be further reduced through a trial mixing test, with a variable step size of 0.05 to determine the best water-cement ratio accurately. Table 2 shows the calculated cement and water consumption during the trial mixing test. The materials used in the test have a compact bulk density of 1508.5 kg/m3 and an apparent density of 2518.9 kg/m3.
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Figure 2.
Table 2.
Grading curve. (a) Grading 1; (b) Grading 2.
Cement dosage and water consumption for trial mix test. MG = 1433.08 kg/m3, Rvoid = 25%
Category Water-cement ratio R Cement content wc (kg/m3) Water content ww (kg/m3)
0.40 225.4 90.2
0.50 198.0 99.0
0.60 176.5 105.9
0.70 159.2 111.4
0.80 145.0 116.0
0.90 133.2 148.0
Figure 3. State of trial mixes with different water-cement ratios. (a) – (f) The water-cement ratio is 0.40, 0.50, 0.60, 0.70, 0.80, and 0.90 respectively.
To conduct the trial mixing test, this study prepares the mixture based on the cement and water consumption values provided in Table 2. The state of the mixture can be observed in Figure 3. Upon examination of Figure 3, it is evident that a water-cement ratio of 0.40 results in a dry cement paste, leaving some construction waste recycling materials uncovered. The trial mixture is dry and scattered as a result. Conversely, a water-cement ratio of 0.90 produces a thin cement paste, causing some of the paste to sink into the gaps of the construction waste recycling material, resulting in a significant collapse. Through the trial mixing test, the suitable water-cement ratio range (approximately 0.50–0.80) is narrowed down, with a variation step of 0.05. The molded test pieces are then subjected to strength testing at water-cement ratio values of 0.55, 0.60, 0.65, 0.70, and 0.75, respectively, to determine the optimal mix ratio.
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4.4
Strength test of recycled permeable base material
Strength tests were conducted on molded test pieces with varying water-cement ratios of 0.55, 0.60, 0.65, 0.70, and 0.75. The results are presented in Table 3 and Figure 4. It is evident from the data that the 7-day unconfined compressive strength initially increases and then decreases with an increase in the water-cement ratio (i.e., a decrease in cement consumption). This is primarily due to the fact that although a higher initial cement consumption leads to a lower water-cement ratio, it affects the hydration reaction of cement, thereby impacting its bonding effect. As the water-cement ratio and cement consumption reach their optimal range, the strength also reaches its maximum. However, when the water-cement ratio is relatively high and cement consumption is low, the strength decreases significantly. Therefore, the optimal water-cement ratio lies between 0.60 and 0.65. Table 3.
Strength test results.
Water-cement ratios R Cement content wc (kg/m3) Water content ww (kg/m3) 7-day unconfined compressive strength (MPa)
Figure 4.
0.55 186.6 102.63 3.6
0.60 176.5 105.9 5.1
0.65 167.4 108.8 4.8
0.70 159.2 111.4 4.4
0.75 151.8 113.9 3.5
Variation relationship between water-cement ratio and strength value.
Upon comparing the strength values corresponding to water-cement ratios of 0.60 and 0.65, it is observed that although the strength value for the latter is slightly lower, it meets the requirements of the Technical Specification for Pervious Asphalt Pavement (CJJT 190-2012) that the strength value of skeleton void cement stabilized macadam should not be lower than 3.5 MPa. Considering project cost and construction workability, the best mix ratio is 0.65 water-cement ratio and 167.4 kg/m3 cement consumption. 4.5
Drainage performance test of recycled pervious concrete base material
In this test, the constant head method was utilized. This involved maintaining a consistent water pressure while measuring the amount of water passing through the test piece within a specific time frame. The permeability coefficient was then calculated based on these measurements. A self-made, simple permeability coefficient measuring device was employed, as depicted in Figure 5. The permeability coefficient of the test piece, as determined by the permeability test, was only 0.10 cm/s, even with the best mix ratio. According to the Code for Design of Highway Drainage (JTG/T D33-2012), the permeability coefficient of the drainage base should not be less than 300m/d, which is equivalent to 0.35 cm/s. Additionally, the permeability coefficient of the drainage base in the Code for Design of Highway Cement Concrete Pavement (JTG D40-2011) should be 1.0 cm/s. Unfortunately, the permeability coefficient obtained in this test did not meet the minimum requirements of these relevant specifications. After testing the test piece, a mallet was used to gently break it, as shown in Figure 6. The figure reveals that the internal bricks of the test piece were severely damaged. The broken 255
Figure 5.
Permeability test.
Figure 6.
Internal conditions of the test piece.
fine bricks were then filled with a cement slurry to fill the gaps in the test piece, resulting in a dense structure. This, in turn, led to a small permeability coefficient, which was insufficient for achieving effective drainage.
5 DISCUSSION 5.1
Technical performance of recycled permeable concrete base
At present, there is no clear specification for the recycled material permeable concrete base course. This section outlines the technical requirements for the drainage base course, as specified in the Technical Specification for Permeable Asphalt Pavement (CJJT 190-2012), the Code for Design of Highway Drainage (JTG/T D33-2012), and the Code for Design of Highway Cement Concrete Pavement (JTG D40-2011). Additionally, this paper draws on a series of experimental studies on recycled material permeable concrete base course to establish the following technical performance requirements: 1) The grading selection should take into account the secondary crushing of recycled bricks and be corrected accordingly. 2) The mix design should be carried out using the volume method and follow the guidelines outlined in Section 4.2 to determine the optimal water-cement ratio and cement dosage. 3) The 7-day unconfined compressive strength of the recycled permeable concrete base should not be less than 3.5 MPa. 4) The design void ratio of the recycled material permeable concrete base should not be less than 25%, and the permeability coefficient should not be less than 0.35 cm/s. 5.2
Suggestions for improving the drainage capacity of recycled permeable concrete base
In conclusion, replacing the permeable concrete base material is essential, and there are two options available. 1) During grading correction, it was discovered that the 16 mm grade (25% by mass) is severely damaged, while the 13.2 mm grade (19% by mass) and 9.5 mm grade (19% by mass) contain high levels of bricks. Therefore, natural aggregate will replace these three grades of materials. 2) The permeable concrete base material will consist of recycled construction waste. Considering that the main factor affecting its drainage performance is the content of bricks, the second scheme will remove all bricks from the construction waste recycled material and use them for the permeable concrete base.
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6 CONCLUSIONS In this paper, the road structure of the existing greenway is optimized, and the construction waste recycling material is used as the slow-moving road of the permeable base. Considering the secondary crushing of bricks, the grading correction is carried out, and the mixed mix proportion design method of the permeable base is obtained. The strength test and drainage performance test of the permeable base material are carried out, and the following conclusions are obtained: 1) The design indexes of recycled material pervious concrete base course mixture are put forward. The design porosity should be no less than 25%, and the permeability coefficient should be no less than 0.35 cm/s. The 7-day unconfined compressive strength should not be less than 3.5 MPa. 2) Through trial mixing test and strength test, it is determined that the best mix proportion is 0.65 water-cement ratios and 167.4 kg/m3 cement consumption when the design porosity is 25%. 3) The secondary crushing of bricks in the construction waste recycling materials will affect their drainage performance, so bricks need to be treated. One approach is to substitute severe crushing gravels and high brick content with natural crushed stone aggregate. Another method is to eliminate all bricks from the construction waste recycling materials and only retain high-strength materials like concrete blocks. This study draws on previous concrete technical standards and existing research findings to propose technical requirements for the raw material indicators, material gradation, and mixture design indicators of construction waste recycling material for the previous concrete base. However, the construction technology and quality control of recycled materials used for urban greenways require further exploration, making it a future research direction.
ACKNOWLEDGMENT The authors express their gratitude for the support received from the Research and Development Project of the Ministry of Housing and Urban-Rural Development (2020-K-078) and the Zhejiang Provincial Department of Housing and Urban-Rural Development 2020 Urban and Rural Planning Project Service Procurement Project (CTZB-2020050374 (3))
REFERENCES Chang, J. W., Du, G. J., Li, H. K., & Du, J. L. (2021). Wide Market Space for Green and Low Carbon Resource Utilization of Construction Waste. Environmental Economy 7, 38–41. Lai, X. Y., Mao, X. S., Wu, Q., Zheng, H. H., Ye, J. H., Tang, X. L., & Shi, X. J. (2022). Application of Building Waste Recycled Filler in Greenway Subgrade. Subgrade Engineering 5, 76–80. Li, X. (2022). Application of Construction Waste Recycled Aggregate in Greenway Base. A Thesis Submitted for the Degree of Master, Chang’an University, Xi’an. Li, X., & Mao, X. S. (2021). Overview of the Application of Construction Waste in Road Engineering. In: World Transportation Engineering Technology Forum (WTC2021), Xi’an, 342–345. Lv, Z. B., Zhang, K., Du, L. B., & Hua, Y. T. (2021). Current Situation and Prospect of Construction Waste Resource Disposal in China. Cement Technology 3, 94–98. Xie, L., Zhu, J. G., & Benson, D. (2022). Partnership Building Government-led NGO Participation in China’s Grassroots Waste Governance. Geoforum 137, 32–41. Xue, Y. X., Arulrajah, A., Narsilio, G. A., Horpibulsuk, S., & Chu, J. (2022). Washed Recycled Sand Derived From Construction and Demolition Wastes as Engineering Fill Materials. Construction and Building Materials 358, 129433. Wei, Y. H., Zhang, L., & Sang, P. D. (2023). Exploring the Restrictive Factors for the Development of the Construction Waste Recycling Industry in a Second-tier Chinese City: A Case Study From Jinan. Environmental Science and Pollution Research International 30, 46394–46413.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Surveying and mapping ancient buildings with 3D laser scanning technology Dongyang Huang* Guangzhou Maritime Hydrographic Department, NGCS, MOT, Guangzhou, Guangdong, China
ABSTRACT: The traditional method of surveying and mapping ancient buildings requires a significant amount of manpower and material resources, with complex procedures, low efficiency, and a high risk of damage to the buildings. However, 3D laser scanning technology provides a non-contact and accurate way to obtain the spatial geometric information of ancient buildings. This technology can quickly reconstruct 3D models and 2D thematic maps through dense point cloud data. This paper explores the application of 3D laser scanning technology in ancient architecture surveying and mapping, using the Nanjing Jiming Temple Herbalist Buddha Tower as an example. The results demonstrate that this technology can efficiently obtain the status information of ancient architecture, which is crucial for achieving 3D visualization of ancient architecture and establishing digital archives.
1 INTRODUCTION Surveying and mapping of ancient buildings involve measuring, collecting, processing, and updating relevant geometric, physical, cultural, and other information. The surveying and mapping results play a vital role in the maintenance, repair, reconstruction, publicity, exhibition, and education of ancient buildings (Ran 2009). However, the traditional method of surveying and mapping is not suitable for modern and ancient architectural surveying and mapping due to the large and irregular shape of ancient buildings. Additionally, frequent contact with buildings during surveying and mapping increases the risk of damage to architectural relics. 3D laser scanning technology is a technological revolution and innovation after GPS.
2 OVERVIEW OF 3D LASER SCANNING TECHNOLOGY 2.1
Principle of three-dimensional laser scanning technology
The point cloud data collected by the 3D laser scanner is based on the scanning coordinate system (Liao 2017). The scanning coordinate system is defined as follows: the origin of the coordinate system is the place where the laser beam is emitted, the X-axis is the zero direction of the horizontal rotation axis of the scanner, the Z-axis is the zenith direction (theoretical vertical axis) when the scanner is horizontal, and the Y-axis has a right-handed coordinate relationship with the X-axis and Z-axis. The measurement principle of 3D laser scanning technology is shown in Figure 1. *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-33
Figure 1.
Measuring principle diagram of 3D laser scanning technology.
In Figure 1, a represents the horizontal angle measured by the scanner, q represents the vertical right angle measured by the scanner, and S represents the distance from the coordinate origin to the monitoring point. Equation (1) can be used to represent the coordinates of the monitoring point in the scanning coordinate system. 8 < x ¼ S cos q cos a y ¼ S cos q sin a (1) : z ¼ S sin q
2.2
Technical process
The process of ancient building surveying and mapping using 3D laser scanning technology involves two main steps: field data acquisition and office data processing, as illustrated in Figure 2.
Figure 2.
Operation flow chart of 3D laser scanning technology.
3 CASE ANALYSIS 3.1
Project overview
The Herbalist Buddha Pagoda is a magnificent structure located in the Jiming Temple in Nanjing. It has seven levels and eight sides, double eaves, and copper temple tiles. The bottom of the pagoda is 7 m high, the tower is 44.8 m high, and the pagoda temple is 11 m high. The objective of this task is to use 3D laser scanning technology to collect data from the Herbalist Buddha Tower, draw relevant drawings of the tower body with professional
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software, and establish 3D digital archives to provide technical support for the later maintenance of the tower. 3.2
Field data collection
Based on the scanning purpose and accuracy requirements, and taking into account the surrounding environment of the Herbalist Buddha Tower, eight measuring stations have been set up 20 meters away from each of the tower’s eight corners. However, due to the tower’s height, the scanner cannot capture the upper parts of the tower when it is too close to the body (Huang 2011; Li 2014; Yuan 2009). To address this issue, four additional stations have been established 40 meters away from the tower body, and 12 control points have been marked with steel nails to ensure visibility between two pairs of stations. This arrangement guarantees complete scanning coverage of all surfaces of the Herbalist Buddha Tower. After laying out the control points, the CORS network RTK is used to measure the known control points in the vicinity, and observations are only carried out after verification. Each control point is observed twice, and the average of the two observations is taken as the final value. Finally, the plane coordinates and elevations of each control point are obtained in the CGCS2000 coordinate system, facilitating the later conversion of the scanning point cloud data to the geodetic coordinate system.
Figure 3.
Layout diagram of measuring station.
The Leica ScanStation C10 scanner is used for point cloud data collection. Target balls are uniformly laid between adjacent stations, and the scanner is erected, leveled, and started. During scanning, the instrument resolution is set to medium resolution and the field of view is set to panorama at 10 m from the tower body. At 40 m away from the tower, the instrument resolution is set as high resolution, and the field of view is set as user-defined to ensure that the point cloud spacing on the tower surface is 2 cm. After the scanning is completed, the scanning object is photographed with a CCD camera for later texture mapping. 3.3
Point cloud data processing
The collected point cloud data is imported into the Cyclone software matched with the instrument for processing, which includes point cloud splicing, point cloud denoising, and point cloud unification (Wang 2014). To ensure the accuracy of point cloud registration on the surface of Herbalist Buddha Pagoda, the method of sequence splicing is used for the data of eight stations on the inner side (Tang 2011; Wang 2011; Zhang 2014). The point stage is mainly used to fine process
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massive point cloud data, while the polygon stage mainly includes the steps of smoothing the polygon surface, repairing the holes and gaps of the polygon, simplifying the polygon, etc. 3.4
3D modeling
The point stage is primarily utilized for refining large-scale point cloud data. The key steps involve filtering out noise points, rectifying point cloud coordinates, importing and coloring point cloud data, resampling the point cloud, and encapsulating the point cloud. This stage serves as a crucial foundation for subsequent modeling. The polygon stage encompasses various steps, such as smoothing the polygon surface, repairing holes and gaps in the polygon, and simplifying the polygon. This stage significantly impacts the quality of the model surface generation. Therefore, the processing work must be meticulous and cautious to ultimately achieve a complete and seamless polygon model. 3.5
Texture mapping
Texture data acquisition involves taking pictures on the spot with a CCD camera, and texture processing should also be carried out, including selecting textures, processing perspective relations, processing texture details, removing occlusion and noise, etc. Finally, in 3Dmax, the processed maps are mapped face by face using the method of manual interaction (Zhang 2011). The true color 3D model of the Herbalist Buddha Tower is shown in Figure 4.
Figure 4.
3.6
Three-dimensional model of Herbalist Buddha Tower.
Accuracy verification
The total station opposite observation method was used to spot check eight control side lengths, and the results were compared with the side lengths obtained by coordinate back calculation using CORS network RTK measurement. To verify the quality of the 3D laser scanning results, ten characteristic points were selected from the Herbalist Buddha Tower, and their coordinates were measured using a prism-free total station. The results were compared with the coordinate values measured in the 3D model, and the statistical results are presented in Table 1.
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Table 1.
Statistics of 3D model accuracy.
Point number
DX
DY
plane
DZ
S01 S02 S03 S04 S05 S06 S07 S08 S09 S10
0.114 0.052 0.122 0.034 0.127 0.102 0.131 0.13 0.162 0.108
0.127 0.062 0.106 0.048 0.14 0.126 0.096 0.101 0.013 0.145
0.174 0.084 0.165 0.062 0.177 0.152 0.135 0.155 0.141 0.167
0.078 0.169 0.159 0.074 0.078 0.101 0.156 0.132 0.062 0.134
Statistical data in Table 1 reveals that the maximum error of the plane position of the 10 characteristic points is 0.189 m, and the maximum error of the elevation is 0.186 m. This meets the Technical Specifications for Urban 3D Modeling, which require the precision of the plane size and elevation of the fine model to be no less than 0.2 m. 3.7
Thematic map drawing
AutoCAD software is used to extract the two-dimensional plan of the tower body. These plans, which contain entity information, are crucial in the protection of cultural relics. Typically, in the protection of cultural relics, buildings’ front elevation, side elevation, top plan, section, cross section, and profile are drawn. The section, cross section, and profile serve to display the internal structure of the building and relevant information. Due to environmental limitations on the site, the internal structure of the Herbalist Buddha Tower was not scanned in this research. Therefore, AutoCAD software was utilized to draw the top plan and north elevation of the Herbalist Buddha Tower, as shown in Figure 5.
Figure 5.
Top view and north elevation of Herbalist Buddha Tower.
4 CONCLUSION With its unique advantages, 3D laser scanning technology has been successfully applied to survey and map ancient buildings. Through our application, we have summarized the following benefits:
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1. The non-contact measurement of three-dimensional laser scanning technology accurately obtains spatial geometric information of ancient buildings. This method quickly reconstructs three-dimensional models and two-dimensional thematic maps through dense point cloud data. It is fast, accurate, and does not cause damage to ancient buildings. Its application in complex and irregular ancient buildings has incomparable advantages. 2. In this study, we used 3Dmax software to process texture images and then mapped them to the 3D model of the ancient building using Geomagic Studio software. However, we found that the final mapping effect was not realistic enough. Therefore, further research is needed to improve the mapping algorithm. The emergence of 3D laser scanning technology has brought technological innovation to ancient building surveying and mapping. We believe that with the continuous reduction of equipment prices and the continuous improvement of post-processing software functions, it will have broad application prospects.
REFERENCES Huang Chengliang, Xiang Juan. Research on Measurement Method of 3D Laser Scanning Technology Applied to Building Modeling [J]. Urban Survey, 2011 (1): 87–90. Li Min. Application of 3D Laser Scanning Technology in Surveying and Mapping of Ancient Buildings [J]. Beijing Surveying and Mapping, 2014 (1): 111–114. Liao Zhongping, Yu Zebin, Liu Ke, et al. Application of 3D Laser Scanning Technology in Building Modeling [J]. Beijing Surveying and Mapping, 2017 (2): 84–87. Ran Qibing. Research on Surveying and Mapping Methods for Ancient Building Restoration [J]. Journal of Fujian Institute of Engineering, 2009, 7 (1): 19–21. Tang Yuyang, Du Boyi, Ding Yanhui. Discussion on the Application of Three-dimensional Laser Scanning Data in the Protection of Cultural Relics [J]. Journal of Beijing Institute of Architecture and Engineering, 2011 (4): 1–6. Wang Tianming, Wang Yanmin, Huang Ming. Modeling of Ancient Buildings Based on 3D Laser Scanning Technology [J]. Surveying and Mapping Bulletin, 2014 (S1): 146–150. Wang Jingwei, Yang Fenglei, Guo Qiuying. Mapping Method of Ancient Architectural Relics Based on Prism Free Total Station [J]. Journal of Shandong Jianzhu University, 2011, 26 (2): 178–181. Yuan Ming, Wang Jianhui, Zhou Zui, et al. A 3D Modeling Method for Ancient Architecture – Taking Suzhou Huqiu Tower as an Example [J]. Modern Surveying and Mapping, 2009, 22 (1): 43–47. Zhang Jian, Dong Yuhua. Fine 3D Reconstruction of Large Ancient Buildings Based on 3D Laser Scanning Technology [J]. Journal of Henan University of Urban Construction, 2014 (3): 58–62. Zhang Qifu, Sun Xianshen. 3D Laser Scanning Measurement Method and Prospect [J]. Beijing Surveying and Mapping, 2011 (1): 39–42.
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Building Integrated Photovoltaic (BIPV): Applications and development Shilong Jia* & Yinchu Ma Department of Civil Engineering, Shenyang Jianzhu University, Shenyang, China
Jinghui Gao Beijing JH Eco-Energy Technology Co., LTD, Beijing, China
ABSTRACT: The technology of photovoltaic curtain walls has become increasingly mature, representing the future development trend of curtain walls. Photoelectric curtain walls, which prioritize environmental protection and energy conservation, have broad development prospects. In the design of ultra-low energy consumption buildings, curtain wall systems are often used for exterior decoration to enhance the facade’s effect. However, the connecting structure between the curtain wall system and the main body, typically composed of metal materials with high thermal conductivity, poses a risk of generating thermal bridges. This study focuses on the connecting structure of the curtain wall dry hanging system, investigates its impact on the heat transfer of exterior walls in ultra-low energy consumption buildings, and proposes reasonable optimization suggestions. The study provides a reference for designing heat cutoff bridges for future ultra-low energy consumption buildings.
1 INTRODUCTION Practitioners in the construction industry are familiar with photovoltaic buildings, as they have likely encountered some form of photovoltaic and building combination (Ruan 2020). Architecture and energy have played crucial roles in human history and social development. Buildings have been the primary source of human energy consumption since the Industrial Revolution, and their share continues to increase. In fact, building-related activities such as air conditioning equipment maintenance, building demolition, and construction waste disposal account for over 40% of annual human energy output. Traditional energy production methods, such as fossil fuels, oil, coal, and hydropower, have devastating environmental impacts from extraction to transportation, conversion, and final energy use (Liang 2020). The combination of solar energy and buildings is one of the few investment options that can achieve both “steady growth” and “carbon reduction,” and it is strongly supported by the government, leading to rapid development. To expand the market space for carbon neutrality, the key is to realize “two alternatives” (Van der Borg 2003). The first alternative involves replacing traditional fossil fuels with electric energy for energy consumption, while the second alternative involves replacing traditional fossil fuels with renewable energy for energy production (Jiao 2010). We believe that the demand for photovoltaic power generation will grow rapidly in the future, particularly in terms of power generation.
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-34
2 THE DESIGN PRINCIPLE OF BIPV The connection structure comprises Hafen groove embedded parts, switching parts, and bolt groups. The combination relationship is illustrated in Figure 1. The Hafen groove embedded parts, which are 300 mm in length, are fixed in the structural beam or column. The adapter parts (360 100 8 mm) are fixed in the groove embedded parts through the bolt group, and the adapter supports the entire curtain wall dry hanging system through the primary and secondary keels.
Figure 1.
2.1
Connection structure diagram.
Mathematical model
To facilitate simulation and analysis, a model unit is established, and a typical wall mounted with a connecting structure is selected for the model unit. The wall specification is 1750 1450 (length width, mm), and the geometry of the model unit is shown in Figures 2 to 3.
Figure 2.
2.2
Plane Diagram of model element.
Simulation results
The influence of connection structures on heat transfer of external walls is mainly calculated through the simulation of two working conditions of model units with and without connection structures. The difference in heat transfer can represent the impact of connection structures on the heat transfer of walls.
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Figure 3.
Model unit 3D model diagram.
Based on the simulated temperature nephogram (Figures 4 and 5), it is evident that the thermal insulation continuity of the area with connecting structures on the exterior wall is damaged, and the thermal bridge is apparent. The heat flow cloud diagram of the exposed part of the adapter shows that the local heat flow of the exposed part of the adapter reaches 78 W/m2, resulting in fast heat dissipation. Table 1 shows that the presence of a connecting structure increases the total heat dissipation of the enclosure structure, and the heat transfer increment caused by a single connecting structure is 1.726 W. The heat dissipation of the exposed portion of the adapter accounts for 94.5% of the total heat dissipation increment.
Figure 4. Cloud chart of temperature distribution along the cross and longitudinal section of the connecting structure (Unit: K).
Figure 5.
Cloud chart of heat flow distribution in the exposed part of the adapter (Unit: W/m2).
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Table 1. Comparison table of heat dissipation between external walls with and without connecting structures. Without connecting structures Wall heat dissipation/qw (W) Heat dissipation from exposed parts of connectors/ql (W) Total heat dissipation/qt (W)
With connecting structures
Increment
8.478 0
8.572 1.632
0.094 1.632
8.478
10.204
1.726
The influence range of heat transfer of the connecting structure on the external wall is analyzed using the simulation calculation method of the model unit. The external wall is partitioned into different subregions, and the heat dissipation is compared between the two external wall forms with or without the connecting structure. The difference in heat dissipation represents the impact of the connecting structure on the external wall. Figure 6 shows the exterior wall partition diagram of the model unit, and Table 2 presents the comparison of heat dissipation in each subregion. The connection structure is only present in the Wall1 area (750 450 mm), indicating that the heat bridge generated by the connection structure only affects the range of 750 mm in length and 450 mm in height centered on the adapter.
Figure 6.
External surface zoning diagram of model unit.
Table 2.
Comparison of heat dissipation in different areas.
Wall
Heat dissipation without the connection
Heat dissipation with the connection
1.166 0.959 1.217 1.465
1.291 0.949 1.208 1.459
Wall Wall Wall Wall
1 2 3 4
2.3
Optimization measures
Increment 0.125 0.01 0.009 0.006
Simulation results reveal that the exposed part of the adapter in the connection structure experiences high temperature, leading to a large heat flow. The local heat flow reaches 78 W/m2, with the exposed portion of the connector accounting for 94.5% of the total heat dissipation 267
increment. Insulating the exposed portion of the connector can effectively reduce the impact of the thermal bridge on the external envelope, as shown in Figure 7. The embedded parts and adapters are made of metal, which has fast heat transfer. To prevent heat transfer, high strength polyurethane is used for heat cutoff treatment.
Figure 7.
Schematic diagram of proposed thermal insulation measures.
To optimize the connection form and analyze the feasibility of the measures, the following steps were taken based on the original simulation model: 1. A high-strength polyurethane thermal insulation gasket, 10 mm thick, was added between the embedded part and the connecting piece. 2. Rock wool insulation was added to the exposed part of the adapter, with the thickness of the insulation layer for the exposed part determined through simulation calculation and analysis. Due to the limited installation gap between the embedded parts and the connecting parts, the thickness of the thermal insulation gasket remained unchanged at 10 mm. The simulation calculation analysis focused on the impact of adding 10 mm, 20 mm, and 30 mm thick rock wool insulation to the exposed part. Table 3 shows a comparison of the simulation results. Table 3.
Comparison of effects of different thermal insulation measures.
Thermal insulation measures
Total heat dissipation
No measure 10.204 Insulation thick- 9.565 ness: 10 mm Insulation thick- 9.411 ness: 20 mm Insulation thick- 9.311 ness: 30 mm
Maximum external surface heat flux
Heat dissipation increment
Reduction ratio of heat dissipation increment
78.5 32.3
1.726 1.087
— 37.0
29.0
0.933
45.9
25.5
0.833
51.7
3 BIPV PROJECT EXAMPLES The China Pavilion at the Beijing International Horticultural Exposition is situated in the core landscape area of the China Beijing International Horticultural Exposition Park, located in the southwest of Yanqing District, Beijing. This green building is brimming with vitality. The roof is equipped with photovoltaic modules that utilize cadmium telluride thin film cells to generate power. This technology ensures a light transmittance of 40% while considering the irregular shape and landscape color design. The design and construction details fully demonstrate the concept of low-carbon ecological green development.
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The photovoltaic system has an installed capacity of 80 kW and generates approximately 83, 000 KWH of electricity annually. This is equivalent to reducing 26.91 tons of standard coal and 70.5 tons of carbon dioxide emissions each year. The green electricity generated powers the internal lighting, power, and other energy needs of the entire building, implementing the green building concept from passive energy saving to active power generation. Additionally, the building roof and curtain wall components are delicately connected to replace traditional curtain wall glass and achieve the integration of photovoltaic architecture. This has significant promotion value (Yang 2022).
4 CONCLUSIONS By conducting simulation calculations and analyzing the connecting structure of the curtain wall system, as well as simulating optimization measures, the following conclusions have been drawn: (1) The connecting structures of curtain wall systems in ultra-low energy consumption buildings have significant thermal bridges on the exterior walls, with each connecting structure having a heat transfer increment of 1.726 W. Therefore, heat cutoff measures should be added to this area during the design of ultra-low energy consumption buildings. (2) The heat transfer enhancement caused by connecting structures is mainly due to some metal components exposed in the air. Thermal insulation wrapping can effectively reduce heat transfer increment. When the exposed part of the connecting structure is wrapped with 20 mm thick thermal insulation material, its heat dissipation is reduced by 45.9% compared to when no measures are taken, and the heat dissipation increment generated by each connecting structure is reduced to 0.933 W. (3) The heat cutoff measures for each connecting structure can be adjusted based on the total energy consumption control objectives of the project and the number of connecting structures used in the design of ultra-low energy consumption buildings. With the new development trend of urban energy conservation and emission reduction, the demand for green environmental protection has increased. Solar photovoltaic building integration has become the new trend in solar energy application power generation. The photoelectric integration technology of BIPV can not only save energy but also reduce emissions, making it a necessary technology and product to achieve the “double carbon” construction goal. In the future, we can expect to see more photovoltaic curtain walls, glazed roofs, and photovoltaic sheds around us (Wang 2022). This will empower buildings, make cities more beautiful, and improve our quality of life.
REFERENCES Jiao Y, Song Q, Liu W H. Practical Simulation Model of Photovoltaic Cell and Simulation of Photovoltaic Power Generation System [J]. Power Grid Technology, 2010(11):204–208. Liang S G, Fu X C, Liang S L. Current Situation and Prospect of Green Energy-saving Curtain Wall Design [J]. China Building Decoration,2020(01):72–74. Ruan X l. Photovoltaic Architecture—A Trip Back to the Past [J]. China Building Metal Structure,2020 (05):40–44. Van der Borg N J C M, Jansen M J. Energy Loss Due to Shading in BIPV Application [R]. Osaka, Japan, Third World Conference on Photovoltaic Energy Conversion, 2003. Wang Q, Wang G G. Discussion on the Application of Building Photovoltaic Integration Engineering [J]. Intelligent Building Electrical Technology, 2022 (02): 163168. DOI: 10.13857/j.carolcarroll nki cn11-5589/tu 2022.02.026. Yang P, Zhang Y, Xu S F, Ling M L, Yang Z X. Photovoltaic Curtain Wall Engineering Practice of Preliminary Exploration with Zero Carbon Building [J]. Building Science and Technology, 2022 (5): 5356. DOI: 10.16116/j.carolcarroll nki JSKJ. 2022.05.013.
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Design method of municipal water supply and drainage pipeline based on BIM technology Yuhong Gan Taiyuan Water Supply Design & Amp; Research Institute Co., Ltd, Taiyuan, Shanxi, China
Haibin Yu* Taiyuan Municipal Engineering Design and Research Institute, Taiyuan, Shanxi, China
ABSTRACT: Urban municipal pipelines ensure the city’s essential functions, such as material transportation, energy transfer, information transmission, etc. They are the infrastructure for the normal operation of the city. With the accelerated pace of urban construction and the rapid expansion of cities, the number of elements of water supply and drainage network systems is increasing. The accidents are caused by unreasonable network planning, extensive management, and slow response. BIM technology focuses on microstructure and information expression, which can reduce the loss of model information in the transmission process of municipal stages by establishing an information flow model. This paper proposes a research method of multi-scale modeling integrated with BIM, which ensures the integrity of semantic information of BIM models at different levels and converts BIM model data into Cesium-supported tile data to improve the visualization efficiency of the model. Finally, different water supply and drainage pipe network models are applied in the pipe network management system. It is to meet the management and application requirements of the multi-dimensional municipal water supply and drainage pipe network data model in different stages of the pipe network management business. Thus, the multiscale semantic data model of municipal water supply and drainage pipe network is constructed. The problems of optimizing business analysis and improving visualization efficiency are solved.
1 INTRODUCTION With the growth of the urban population, new pipes are added to the municipal water supply and drainage network, which leads to a heterogeneous system with components of different ages, materials, and sizes, thus making engineering design a difficult task. The water supply and drainage pipe network has many pipelines and a complex layout (Devi Priya Munagala 2020). Thus, it is inevitable that the pipelines will overlap, collide, and the space is too narrow to lay the pipelines only by drawing the plan to guide the construction. BIM technology can guide the design, and its visibility, collaboration, and other characteristics can be adopted to optimize the pipeline (Eleonora Troiani 2020). Using BIM technology to participate in the design process of the water supply and drainage network of construction projects, with the help of BIM technology, a three-dimensional model of the overall network can
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-35
be constructed. The layout of the network can be displayed intuitively. The intersection points of the water supply and drainage network with HVAC, electrical, and other professional pipelines and the conflict points with buildings and structural wall beams can be presented intuitively. Information sharing can be strengthened (Bryan Franz 2019). The automatic layout of components can reduce designers’ workload, adjust the design plan in the shortest time, and fundamentally improve design efficiency (Zheng 2019). The building BIM model is linked to the water supply and drainage BIM model to realize the collaborative design. BIM technology is used to solve the problem of cross collision between pipelines. It counts the quantities and provides the possibility for collaborative design, collision check and quantities statistics based on BIM technology (Alan V. Hore 2018). The Revit software is used to design the building water supply and drainage for the residential project of Block A8 in the central area of Wuda District, Wuhai City. It synthesizes the pipelines of various professional models to generate the pipeline system diagram, which breaks through the traditional two-dimensional design (Zhang 2020). The BIM technology is applied to carry out water supply and professional drainage design. It realizes data sharing among various specialties, reduces collision problems, avoids rework, and improves design quality by sharing data (Gang 2019). BIM technology is applied to design large public buildings’ electromechanical (water supply and drainage, heating, and ventilation) systems. It checks out and solves the collision problems between electromechanical models and between electromechanical and building models. It improves construction efficiency and applies BIM technology to optimize the design to maximize the utilization of building space (Babar Ali 2020). Although the above literature optimizes the design process, problems still need to be solved, such as cumbersome operation in the model construction process and increased difficulty in construction and laying (Zhu 2019). In this paper, the program algorithm can greatly simplify the process of related modeling operations. Based on the secondary development platform of BIM technology, it automates the heavy workload and repetitive work in the modeling process, makes up for the shortcomings of some commercial plug-in functions, enriches the way of automatic optimization of pipelines, and facilitates the design of municipal water supply systems. To improve the municipal water supply, drainage, and other buried pipeline system layout, reasonable design, and upgrade the pipe network, this paper provides a strong guarantee for the safe operation of the pipe network.
2 DESIGN OF DYNAMIC TRANSLATION MODEL BASED ON BIM The stereoscopic 3D effect design based on a two-dimensional display can only vaguely observe the project’s pipelines and can only know that the project has pipelines. When the user wants to view the pipe points of the project and the overall spatial position relationship or specific characteristics of the pipelines, the transparency factor can be set to 0.1 or other values (Xu 2020). The problem of model deviation is solved by the model matrix’s dynamic loading method to create a planar matrix. First, it should get the offset as follows: 2 3 2 3 0 1 fa a h ¼ 4 b 5 ¼ 4 1 0 fb 5 (1) 1 1 fg g In Formula (1), we set the matrix h and dynamically adjust the size to obtain the translation matrix model matrix (Al Hattab Malak 2021). For point and line elements in other spaces, the kernel density estimation method is used to estimate the density of the distribution of elements in the whole study area, as shown in 271
Formula (2). lðtÞ ¼ m
n X
rs ðft Þ2
(2)
i¼1
In Formula (2), l(t) is the kernel density function, the sample, the smoothing parameter, and the parameter (Hui 2021). The effect after the center position of the model is adjusted by the translation matrix is shown in Figure 1.
Figure 1.
Translation matrix adjustment model.
We load the tile model data, use JavaScript, HTML, and CSS programming languages, and set the transparent value attribute according to the requirements. The range is between [1,0]. The value 0 indicates that the building elements are fully transparent and the engineering components can be seen, while 1 indicates no transparency (Cui 2019). The visual effect is displayed in Figure 2.
Figure 2.
Visualization effect of loading tile model.
If you want to know the general situation of the pipeline in the whole project, you can set the transparency property to close to 0 to see the desired effect (Pastor 2020). The pipeline system model is translated to the corresponding position of municipal engineering. That is, the pipeline system model of the three-dimensional tile is consistent with the coordinates of the municipal engineering model, thus solving the problem of model offset.
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3 SIMULATION TEST ANALYSIS 3.1
Simulation environment
The three-dimensional view of the municipal engineering pipeline collision is displayed in Figure 3.
Figure 3.
Collision area of the municipal engineering pipeline.
The main problems in the collision area in Figure 3 are as follows: 1. Collision between fire hydrant pipe A and sewage pipe B. 2. Collision between rainwater pipeline C, domestic water supply pipeline D, and wastewater pipeline E (Rasifaghihi N 2020). If the collision of these pipelines is not found in time in the design, it is easy to cause a construction rework and delay the project’s construction progress. 3.2
Comparison of optimization effects
Because the purely manual method is verified by repeated accounting, its modeling result is completely correct, so the modeling condition of this method is used as the criterion to judge the modelling accuracy. The calculation formula of relative error y is shown in Formula (3): y¼
n X ai b
i
i¼1
(3)
gi
In Formula (3), when i is taken as 2 and 3, respectively, it represents the number of component instances using the method in this paper and the number of component instances using modeling, and n represents the number of component instances using the purely manual method (Liu 2020). A comparison of calculation results is shown in Table 1. Table 1.
Comparison of modeling effects.
Parameter
Not optimized
This article optimizes
Modeling quantity Time consumption/min Relative error% Correction time/s
4530 12.2 6.3 76
5149 11.1 2.5 37
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3.3
Pipe collision detection
In the design stage, the designers of various disciplines design separately. Therefore, they need help finding the design problems in a short time. Many problems are not found until the installation and construction, which is likely to directly lead to the rework of the project, resulting in a great waste of manpower and material resources. Using the legend with a high visualization effect can improve the detection efficiency more clearly, as shown in Figure 4. The most important purpose of applying Revit software in building water supply and drainage engineering is to check the intersection or collision of pipelines and solve the design problems from the source. BIM technology can use Revit or Navisworks software to check the collision of pipelines.
Figure 4.
3.4
High visual detection effect.
Pipeline optimization
The effect of layout adjustment and engineering optimization after pipeline collision is displayed in Figure 5.
Figure 5.
Optimized pipeline design.
Considering the operability, the small pipe diameter avoids the large pipe diameter. The branch pipe avoids the mainline pipe. The metal material pipe avoids the non-metal material pipe. The low pressure gives way to the high pressure, and the fewer accessories give way to the more accessories. The upper wind is in the middle, and the lower water is in the space. Pressurized flow gives way to a non-pressure flow pipe because the non-pressure flow pipe is a gravity flow pipe, which needs a certain slope to drain smoothly.
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4 CONCLUSIONS This paper studies the secondary development of municipal water supply and drainage pipeline design and analyzes the data format, elements, and families of Revit software. Then, it summarizes Revit’s secondary development workflow, which provides a reference scheme for solving the problem of pipeline collision in municipal pipe network design and explores the method of model optimization in practice. Due to the limited practical project experience and engineering practice experience, there are still many areas for improvement in the research process. In future work, we should enrich the index screening methods, consider the problem from multiple perspectives, and try to cover a wider range of the index system.
REFERENCES Alan V. Hore, Barry Mc Auley, Roger P. West. BIM Macro Adoption Study: Establishing Ireland’s BIM Maturity and Managing Complex Change. International Journal of 3-D Information Modeling (IJ3DIM), 2018, 7(1). Al Hattab Malak. The Dynamic Evolution of Synergies between BIM and Sustainability: A Text Mining and Network Theory Approach. 2021, 37. Babar Ali, Hafiz Zahoor, Abdur Rehman Nasir, et al. BIM-based Claims Management System: A Centralized Information Repository for Extension of Time Claims. 2020, 110. Bryan Franz, John Messner. Evaluating the Impact of Building Information Modeling on Project Performance. Journal of Computing in Civil Engineering, 2019, 3(3): 1–9. Cui K, Jing X. Research on Prediction Model of Geotechnical Parameters based on BP Neural Network. Neural Computing Applications, 2019, 31(12): 8205–15. Devi Priya Munagala, Munagala Devi Priya, Kone Venkatesh. Feasibility Study and Implementation of BIM in Small Scale Projects. IOP Conference Series: Materials Science and Engineering, 2020, 912(6). Eleonora Troiani, Abdul-Majeed Mahamadu, Patrick Manu, Ernest Kissi, Clinton Aigbavboa, Akponanabofa Oti. Macro-maturity Factors and Their Influence on Micro-level BIM Implementation within Design Firms in Italy. Architectural Engineering and Design Management, 2020, 16(3). Gang Xu, Yabo He. Research on Decoration Design of Green Building Based on BIM Technology. IOP Conference Series: Earth and Environmental Science, 2019, 242(6). Hui Xin, Zheng Yudong, Yan Haijun. Water Distributions of Low-pressure Sprinklers as Affected by the Maize Canopy Under a Centre Pivot Irrigation System. 2021, 245(prepublish): 106646. Junxiang Zhu, Xiangyu Wang, Peng Wang, et al. Integration of BIM and GIS: Geometry from IFC to Shape File Using Open-source Technology. 2019, 102: 105–119. Liu H, Zhao Y, Li H, et al. Individual Water-Saving Response Based on Complex Adaptive System Theory: Case Study of Beijing City, China. Water, 2020, 12(5): 1478. Pastor D J, Fullerton T M. Municipal Water Consumption and Urban Economic Growth in El Paso. Water, 2020, 12(10): 2656. Rasifaghihi N, Li S, Haghighat F. Forecast of Urban Water Consumption Under the Impact of Climate Change. Sustainable Cities Society & Natural Resources, 2020, 52(101848). Zhao Xu, Lu Zhang, Heng Li, et al. Combining IFC and 3D Tiles to Create 3D Visualization for Building Information Modeling. 2020, 109. Zhang Ruixue, Tang Yuyan, Wang Liang, Wang Zeyu, Zhang Jiansong. Factors Influencing BIM Adoption for Construction Enterprises in China. Advances in Civil Engineering, 2020, 2020. Zheng X, Lu Y, Li Y, et al. Quantifying and Visualizing Value Exchanges in Building Information Modeling (BIM) Projects. Automation in Construction, 2019, 99: 91–108.
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Lateral stability analysis of single-column pier bridge based on geometric nonlinearity Wenxue Wang* & Jianlei Ma China Communication North Road and Bridge Co., Ltd, Beijing, China
ABSTRACT: It has become a trend to solve the modern complex bridge structure by computer finite element analysis of the bridge. The continuous solution region is discretized into a set of entities consisting of a finite number of elements, which are connected in some way. According to the force and structure of the actual bridge structure, different shapes of elements, materials, and connection forms are adopted to simulate. The finite element analysis method has the advantages of high efficiency, quickness, and accuracy and can solve most structural analysis problems. In this paper, the geometric nonlinear finite element model of the construction process of the single-column pier is established, and the dynamic characteristics are analyzed. The instability forms of the bridge tower in the construction process are compared and analyzed. The change law of the instability forms in the construction process of the single-column pier is summarized. The main girder’s section simulation model is established using the fluid simulation software. The three-component force coefficients of the main girder are simulated and calculated. The action forms of the wind load on the main girder under different wind attack angles are summarized, which provides the calculation basis for the static wind stability, galloping, and flutter stability of the single-column pier during construction. This study reflects the stability of the single-column pier bridge to a large extent, which is of great significance to the seismic and wind resistance of the bridge structure.
1 INTRODUCTION The bridge structure is an overhead building structure with a certain bearing capacity. The previous stone and wood materials are gradually replaced by cement, steel, and other materials with better performance. With the development of prestressed concrete and highstrength steel, the development of material elastic-plastic and limit theory and the research field of soil mechanics has made a great breakthrough. Bridge construction technology has made a great leap (Lee 2020). Due to the continuous growth of the span of modern bridges, they have become more and more slender and soft. The damping ratio of the structure is getting lower and lower. Therefore, long-span bridges are more sensitive to the wind, and the wind load has become a factor that must be addressed in the construction of bridges (Joseph 2020). Masrilayanti et al. (Masrilayanti 2021) studied the stability under the influence of geometric and material nonlinearity and used the vault displacement to judge whether it is unstable. It is found that the geometric nonlinearity alone does not greatly impact the stability coefficient under different working conditions. But when the double nonlinearity is considered, the safety factor is significantly reduced, which indicates that the geometric nonlinearity cannot be ignored. The study also did not consider the impact of the *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-36
construction process. Park et al. (Park 2020) studied the stability of extreme points of a crescent-shaped concrete-filled steel tubular arch bridge. Using ANSYS software, the most unfavorable condition of the structure is obtained through the eigenvalue analysis. Then, the double nonlinear analysis is carried out under the most unfavorable condition. The Riks method is used for the iterative solution. The instability mode and stability safety factor of the bridge is obtained. The instability mode is lateral instability. The safety factor of stability is related to the dead load, and the ultimate bearing capacity of the bridge is reduced by 3% due to the consideration of geometric nonlinearity. Yu et al. (Yu 2019) proposed the Riks method to calculate the stability safety factor by iterative convergence. Compared with the N-R method, the possibility of singularity in the calculation results is reduced. Jigar and Atul (Jigar 2020) studied the problem of support buckling in many aspects. The idea of starting from axial stress and lateral displacement was put forward. The mathematical formula realtime analysis program was combined with the monitoring program, which was the first breakthrough in combining the two. In this paper, focusing on the nonlinear stability of the single-column pier bridge, the method of nonlinear geometric simulation is applied to study the construction stability of the singlecolumn pier bridge with an engineering example. Considering the nonlinear factors, the influence of horizontal wind on the stability of arch ribs is studied. The relationship between the wind tension and the horizontal angle and the stability safety factor of arch ribs is revealed.
2 CALCULATION OF LATERAL STABILITY OF BRIDGE WITH GEOMETRIC NONLINEARITY The reference height k is determined by the height k of the arch rib vault from the water surface or the ground. The design reference wind speed md at the reference height of the component can be calculated according to the following formula. k m (1) md ¼ f f 10 s10 where ff is the wind resistance risk coefficient, which is taken as 1.02; s is the surface roughness, which is taken as 0.16. The design wind speed can be calculated in the construction stage using the following formula. msd ¼ fsf md
(2)
where msd and fsd are the design wind speed and wind resistance risk coefficient in the construction stage. The static wind load mainly manifests as the aerodynamic drag on the arch rib for the single-column pier bridge. The wind load has little effect on the deformation of the main arch and the main girder, so the main instability state that may be caused is the lateral buckling of the main arch. When calculating the static wind load of the bridge, only the transverse wind load caused by the static gust load is considered (Saeed 2019). When the wind load acts on the finite element model, the static wind load is converted into the equivalent static gust wind load, which is loaded on the beam element for analysis and calculation (Nathaniel 2019). Under transverse wind, the equivalent static gust load of the arch rib shall be calculated as per Formula (3). hm2g lg Tan qg ¼ pffiffiffi kK
(3)
where qg is the wind load on the member per unit length (N/m). h is the air density. 1.25kg/m3 is taken. lg is the equivalent static gust wind speed of the member at the reference height (m/s). 277
K is the resistance coefficient of the member, and Ta is the projected area m2/m of the member per unit length in the downwind direction. lg ¼ $t mg
(4)
where $t is the equivalent static gust coefficient, which is related to the horizontal loading length of the structure. The calculation results are displayed in Table 1. Table 1.
Wind speed of each section.
Segments
1
2
Reference height (m) Section length (m) Loading length (m) Design wind speed at the construction stage (m/s) Equivalent static gust wind speed (m/s)
15 17.5 16.2 30.5 29.8
16.1 17.5 17.3 31.1 30.6
...
11
12
132.2 22 201.6 32.5 40.2
135.6 22 203.5 33.4 41.3
The value of the arch rib resistance coefficient CD relates to the section. When the arch rib section is a single-limb arch rib, it is determined by a wind tunnel test or numerical wind tunnel simulation. Jirawat and Virote (Jirawat 2020) studied the variation of the resistance coefficient of the arch rib of a bridge along the span direction through numerical wind tunnel simulation. The load-deformation curve of the structure can be obtained by solving the above equilibrium equations. Most of the bridges belong to nonlinear buckling in real projects. However, the difference between the finite element calculation results obtained by linear buckling analysis and the upper limit of the ultimate load in the nonlinear buckling analysis results is manageable. The characteristic value problem of this method is easy to solve, and its structural mechanic concept is clear. Therefore, the calculation of linear instability still has some practical value.
3 EXPERIMENTAL TEST 3.1
Test platform
For the static wind stability analysis of the bridge construction’s whole process, the wind stability study is carried out from the construction stage to the completion stage of the singlecolumn pier bridge under the wind attack angles of 5 , 0 , and 5 . The loading step length is 20 m/s, and the step length is 10 m/s when the wind speed exceeds 100 m/s. When the threecomponent force coefficient norm is less than the allowable value, the iteration stops, and the static wind instability is determined. Then, we increase the wind speed step to the next wind speed level, obtain the torsion angle, and calculate the three-component force coefficient under this torsion angle (Liu 2017). We take the form of a Euclidean norm. The formula is: n X ln ðRT Þ ln ðRT1 Þ i¼1
ðln ðRT 1 ÞÞ2
2 = espn
(5)
where espn is the error threshold; ln is the drag, lift, and lift moment coefficients. During the construction of the double cantilever, two working conditions of symmetrical loading and asymmetrical loading of wind load shall be considered. Among them, the asymmetrical coefficient is 0.5. The schematic diagram of the wind load loading of the main beam is presented in Figure 1.
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Figure 1.
3.2
Asymmetric loading of the wind load.
Experimental test results
(1) Symmetrical loading of wind load In the double-cantilever construction stage, a symmetrical wind load is applied to the south cantilever. The wind-induced response of the main beam under wind attack angles of 5 , 0 and 5 is shown in Figure 2.
Figure 2.
Wind-induced response under symmetrical wind load.
As displayed in Figure 2, during the construction stage, the main girder box girder is erected from the No.0 girder to the No.2 girder. The number of box girder segments is small. The response of the main girder unit to the static wind load could be clearer, and the lateral displacement can be ignored. Under the wind attack angle of 0 , the wind load produces a downward lift force on the main girder. The main beam produces a downward vertical displacement, but the effect of the lifting moment on the main beam is not obvious. With the increase in wind speed, the main girder has almost no torsion. The influence of the lifting force on the near wind side is greater than that of the lifting moment. The force of the No.1 cable on the near wind side also increases slowly. There is almost no static wind instability in this construction stage, with the progress of the main girder erection, the length and
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structural stiffness change. The effect of wind load on the main beam becomes increasingly obvious. (2) Asymmetric wind loading The asymmetric wind load is applied to the construction stage, and the asymmetry coefficient is 0.5. The wind-induced response of the main girder under the wind attack angle of 0 is displayed in Figure 3.
Figure 3.
Wind-induced response under asymmetric wind load.
In the case of asymmetric wind load, the wind-induced response of the girder is consistent with that of the symmetrical wind load. However, the transverse displacement of the girder under asymmetric wind load is slightly greater than that under symmetrical wind load. The torsion angle of the girder is slightly smaller than that under a symmetrical wind load. Therefore, in the process of double cantilever construction, there is a slight difference in the static wind instability speed between symmetrical and asymmetric loading of the wind load.
4 CONCLUSIONS This paper analyzes the stability of the construction stage based on the instability mode of typical conditions and the stability safety factor under the influence of nonlinearity. The reasons for the change in the stability safety factor under wind load and its influence on construction stability are analyzed. Through the simulation calculation of wind load, the relationship between the horizontal angle and the stability safety factor is revealed, and the relationship between the stability safety factor and the horizontal angle of the gust is established. The results verify that the improvement of the stability safety factor is not obvious, and the stability safety factor increases first and then decreases when the horizontal angle of the gust increases. Currently, the practical application of the beam and arch combination bridge with asymmetric side span is less. This paper is only based on a bridge to study. The dead weight and external load are limited. It is supposed that the asymmetric form is applied to highway bridges or replaced by structural materials with larger dead weights. In that case, it still needs further checking whether it can be realized or promoted. For this kind of asymmetric structure, whether the concept of the limit difference of the side span can be introduced is
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worth considering. That is, whether there is an upper limit for the difference between the spans on both sides. If so, how much the maximum difference reaches will this type of structure no longer be applicable.
REFERENCES Jigar P Variyavwala, Atul K. Desai. Dynamic Performance of High-Speed Rail Cable StayedBridge with Altered Track Assembly. International Journal of Recent Technology and Engineering (IJRTE), 2020, 8(5). Jirawat J, Virote B 2020 Vortex Induced Vibration and Flutter Instability of Two Parallel Cable-stayed Bridges. Wind and Structures, 30(6). Joseph Vianny X, Vimala S, Swathini S 2020 Behavior of Asymmetric Cable Stayed Bridge for Long Span. International Journal of Innovative Technology and Exploring Engineering (IJITEE), 9(7). Lee Y, Jang M, Kim S, Kang Y 2020 A Study on the Long-Term Measurement Data Analysis of Existing Cable Stayed Bridge Using ARX Model. International Journal of Steel Structures, 20 (republish). Liu A, Lu H, Fu J 2017 Analytical and Experimental Studies on Out-of-plane Dynamic Instability of Shallow Circular Arch Based on Parametric Resonance. Nonlinear Dynamics, 87(1): 677–694. Masrilayanti M, Nasution Ade P, Kurniawan R, Tanjung J, Sarmayenti S 2021 Fragility Curve Analysis of Medium Cable Stayed Bridge. Civil and Environmental Engineering, 17(1). Nathaniel J, Johannes K, Tom T B 2019 Wester, Sebastian Wegt, Klaus Schiffmann, S-uad Jakirlic, Michael Hölling, Joachim Peinke, Cameron Tropea. Insights into the Periodic Gust Response of Airfoils. Journal of Fluid Mechanics, 876. Park K, Kim D, Hwang E 2020 Correction to: Investigation of Live Load Deflection Limit for Steel Cable Stayed and Suspension Bridges. International Journal of Steel Structures, 20(4). Saeed Najmadeen M 2019 Simultaneous Force and Deformation Control of Cable Arch Stayed Bridges. Kufa Journal of Engineering, 10(4). Yu X, Chen D, Bai Z 2019 A Stability Study of the Longest Steel Truss Deck Cable-stayed Bridge during Construction. KSCE Journal of Civil Engineering, 23(4), 1717–1724.
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Design and application of cable crane for Zhangjiajie Grand Canyon Glass Bridge Qiang Yi*, Junlong Zhou & Xiaomin Liu China Construction Sixth Engineering Bureau Co., Ltd., Tianjin, China
Chuan Yan Central Research Institute of Building and Construction Co., Ltd. MCC Group, Beijing, China
Wenbin Geng & Penglin Xie China Construction Sixth Engineering Bureau Co., Ltd., Tianjin, China
ABSTRACT: Zhangjiajie Grand Canyon Glass Bridge is a space cable plane ground anchored suspension bridge with a main span of 430 m, which is the highest and longest pedestrian landscape glass bridge in the world. Considering the topographic characteristics of the bridge site, the stiffening beam segments are hoisted and spliced by a cable crane. Both the construction site conditions and the transportation conditions are poor, the terrain elevation difference is large, and there are also other problems at the bridge site. In addition, it is clear that there are difficulties in project implementation and characteristics of the bridge structure. In view of the factors mentioned above, the cable crane of this project is designed as one main cable with a span of 60 + 440 + 60 m, which can meet the lifting requirements of the stiffening beam segments and reduce the occupation of the construction site. The main cable is anchored with ground anchors; the two towers are assembled with large standardized Bailey pieces, saving steel consumption, and shortening the construction period with fast assembly and disassembly; the wind cables are set on the towers to increase their wind resistance of them; two lifting cranes are erected on the main cable, and can be lifted independently. The spacing between the two lifting cranes is 5 m. The total lifting capacity is 60 t, which can meet the requirements of the lifting and also improve the flexibility and stability of the cable crane.
1 INTRODUCTION The bridge deck of Zhangjiajie Grand Canyon Glass Bridge is more than 280 meters from the bottom of the valley. It is the highest and longest glass bridge in the world, for pedestrian sightseeing, bungee jumping, ropeway, stage, science, and education (Wan 2017). The bridge is a space cable plane ground anchored suspension bridge with a main span of 430 m, with a bridge length of 536 m. The bridge deck width gradually changes from 15 m at both ends to 6 m. Tunnel anchorages are used on the west bank and gravity anchorages are used on the east bank (Luo 2017; Wang 2017). The whole bridge is designed with 4 tower columns, with a maximum of 40.65 m. The substructure of the bridge tower is pile group foundations. The stiffening beam consists of 7 types of segments, and the maximum lifting weight of all
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-37
segments is about 45 t (Fang 2017; China Railway Major Bridge Reconnaissance & Design Institute Co., Ltd 2015). The bank slope of Zhangjiajie Grand Canyon is steep, and the canyon section is “U” shaped. The original geomorphic unit at the bridge site is the peak forest landform in the karst landform. Karst caves and joint fissures are mainly developed vertically. The geological conditions are complex. For long-span suspension bridges and arch bridges, cable cranes are often used for hoisting construction (Chang 2016; Fang 2022; Wu 2017; Zhao 2018). Considering the structural characteristics of the bridge and the geographical environment of the project, the cable hoisting scheme is adopted to construct the stiffening beam, and a set of cable cranes is arranged at the bridge axis (Guo 2017; Ma 2017). The cable crane is the key construction facility to complete the hoisting construction. Aiming at the key and difficult points of the project construction, the key design parameters of the cable crane are analyzed and then determined.
2 KEY AND DIFFICULT POINTS IN THE DESIGN AND CONSTRUCTION OF CABLE CRANE The bridge structure of the project is novel, the construction site conditions are poor, the construction is difficult, and the construction period is tight. The erection of the bridge’s main cable, the adjustment of the main cable space plane, and the installation of a stiffening beam are the key points of this project. The cable crane is one of the main construction facilities, which is crucial to the smooth implementation of the project. The key and difficult points in the design and implementation of cable crane are as follows: 1. The project is located on both sides of the steep banks of Zhangjiajie Scenic Area’s canyon, and the traffic is inconvenient. In particular, construction machinery cannot pass through the existing village road on the east bank, so a construction road is needed. Because it is difficult to transport large components and large construction machinery, the components used in the cable crane need to be lightweight and standardized to facilitate transportation and assembly. 2. The elevation of the construction site varies greatly and the terrain is complex. The layout of the cable crane should be combined with the plane layout of the whole construction site to minimize its impact on other processes and the surrounding vegetation environment. 3. The main span of the bridge adopts a longitudinal and transverse stiffened beam structure system, which is prefabricated and installed in sections. The hoisting bearing capacity of the cable crane must be more than 45 t, under the condition of meeting the overall light weight. 4. The main cable of the bridge is a spatial cable plane, and the stiffening beam is more than 280 meters from the ground. The installation of the stay cable requires high stability during the lifting operation of the cable crane. 3 KEY TECHNOLOGY OF CABLE CRANE DESIGN 3.1
Overall design
Considering the site conditions and project construction requirements, the whole cable crane is designed as one group of the main cable, and the cable span is designed as 60 + 440 + 60 m. The overall layout of the cable crane is shown in Figure 1. The design of one single group of the main cable and a small side span reduces the impact on the surrounding environment and other processes. The whole hoisting system is composed of the tower frame and its foundation, tower top cable saddle, lifting crown block, cable system (main cable, lifting cable system, towing cable system, wind cable system), east 283
Figure 1.
General layout of cable crane (Unit: m).
and west bank anchorage, etc. The cable crane is equipped with two Bailey pieces towers. Bailey piece components meet the requirements of lightweight and standardization and are convenient for transportation and assembly in mountainous areas. The system adopts double lifting points for hoisting, with a maximum net lifting weight of 60 t, so as to meet the requirements of the stiffening beam segments for hoisting and splicing for the maximum lifting weight, and hoisting stability. The included angle between the side span rear wind cable and the horizontal plane is about 23 . 3.2
Cable system design
The main component parameters of the cable system are as follows: 1. Main cable The cable crane is designed with one group of the main cable, which is composed of 6 hemp core wire ropes with a diameter of 47.5 mm. The sag is 44 m at the maximum lifting load, and the vertical span ratio is 1/10. The main cable is supported on the tower top by a cable saddle and connected to the main ground anchors. The cable saddle is designed as an adjustable structure, and the main cable can slide on the cable saddle when it is loosened, while the main cable cannot slide when it is clamped. 2. Hoisting crown blocks and lifting points Two hoisting crown blocks are erected on the main cable with the spacing of two hoisting crown blocks of 5 m, and two lifting points can be lifted independently. Each crown block movable pulley block is composed of 5 pulley pieces. By setting two lifting cranes and two lifting points, the requirements for the bearing capacity and the specification of a single crane during lifting construction are reduced, and the stability and flexibility of lifting are improved. The total weight of crown blocks, two lifting points, and slings is about 20 t. 3. Towing cables and lifting cables Wire rope with a diameter of 28 mm is used for the towing cables. One towing cable is equipped for the whole bridge. Connect four short steel wire ropes with a diameter of 24 mm between the two crown blocks to make them move synchronously. A steel wire rope of diameter 19.5 mm is used for lifting cable, and each lifting point is equipped with 10 steel wire ropes. The whole bridge is equipped with 2 hoisting winches. The towing cable and the lifting cables are installed under the main cable of the cable crane, without a supporting device. Their threading and winding principle are shown in Figure 2 and 3, respectively.
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Figure 2.
Threading and winding principle of towing cable.
Figure 3.
Threading and winding principle of hoisting cable.
4. Wind cable The tower wind cables of the cable crane are divided into front wind cable, rear wind cable, and side wind cable. One end of the wind cable is fixed to the tower top, and the other end is fixed to the gravity anchor. Each group of wind cables of the tower is composed of 4 steel wire ropes with a diameter of 24 mm.
3.3
Tower design
The tower is composed of the tower top tripod, the tower body, and the tower bottom tripod. The width of the tower in the transverse direction of the bridge is 4.52 m, and the width in the longitudinal direction is 2.58 m. The height of the two towers on both banks is 47.8 m, and the tower body is designed to be assembled with 11 large Bailey trusses. The pinhole spacing of the bailey piece is 4 m 2.14 m. The chord is made of 12 # channel steel. The Bailey pieces are connected by special riding bolts and right-angle bolts. The connection between the tower and the foundation is hinged to ensure that the tower only bears axial force, but not bending moment. The single-section tower structure is shown in Figure 4. The design of a large Bailey piece assembly tower has the advantages of fast assembly and disassembly, low cost, safety, and reliability.
Figure 4.
Schematic diagram of single-section tower structure.
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3.4
Tower foundation and anchor design
The tower foundation is designed as C30 reinforced concrete strip foundation. The foundation is 8 m long, 5 m wide, and 2 m high. The main cable anchor is 8 m long, 8 m wide, and 3 m high, and is poured with C30 concrete. 4 APPLICATION OF CABLE CRANE 4.1
Tower and foundation construction
Considering the geological conditions of the site, the slag is replaced and filled for the foundation treatment to ensure that the bearing capacity of the foundation is greater than 200 kPa. After the foundation treatment is completed, the reinforced concrete strip foundation of the tower will be constructed. The first three sections of the tower are hoisted and spliced with a 25 t truck crane. Before assembling the tower, temporary support is set for the tower bottom tripod to temporarily fix the connection between the tower and the foundation. After the completion of the tower assembly, the temporary support will be removed. The method of temporary tower support is shown in Figure 5.
Figure 5.
Schematic diagram of temporary consolidation of the tower.
After the assembly of the three tower sections at the bottom is completed, a special rocker boom is erected at the upper end of the tower. Bailey pieces are lifted and installed by a 5 t winch at the bottom of the tower. The rocker boom gradually rises with the increase of the tower. During the assembly of the tower, the wind cable should be tensioned in time to ensure the safety of the tower construction. The verticality of the tower should be checked every 12 meters of assembled height, and any problems found should be adjusted and corrected in time. After the completion of the tower assembly, a comprehensive inspection should be carried out. The tower top cable saddle can be installed after the tower is assembled and inspected as qualified. The surface of the cable saddle and baffle should be smooth without burrs and edges; when threading, a Teflon plate should be installed on the contact surface between the cable saddle and the rope, and apply grease to reduce the friction to protect the main cable and the cable saddle. 4.2
Cable installation
The cables should be protected from soil and other pollution during installation. Before threading and hanging the cable, the pilot cable must be threaded and hung first. The pilot cable crossing the valley is used to pull the 11 mm diameter thin steel wire rope from the west 286
bank to the east bank, and then the 19.5 mm diameter steel wire rope is pulled to the east bank. The east bank winch is used to tighten the 19.5 mm diameter steel wire rope. The traction rope bypasses the traction wheel at the tower top on both sides, and both ends enter the winch. Finally, the 47.5 mm diameter main cable is pulled back and forth by the winch on both banks through the traction rope until the cable installation is completed. Personnel is not allowed to enter and exit the area below when the main cable is put on. The technical requirements for cable installation are as follows: 6 cable wire ropes should be paralleled to each other, and the sag error should be less than 50 mm. Once the cable is put through, a comprehensive inspection should be organized to check whether the winch, drum, motor, and brake are in good condition; whether the cable is damaged or polluted; whether the contact between the cable and the pulley is consistent; and whether the anchor point is reliable. 4.3
Installation of crown blocks and lifting points
Before the installation of the crown blocks, it should be confirmed that the dimensions of each part meet the design requirements: the structure is tight, stable, and reliable; the running wheels on the same line should be straight and consistent, without tilting and twisting; the sheave groove should be consistent with the main cable steel wire rope to ensure smooth and flexible traveling of the crown blocks, free from hanging by the rope and abrasion of the steel wire rope. During the installation of lifting points, the upper and lower pulley blocks should be paralleled and horizontal and the wire ropes should be correctly wound. The ropes should be arranged in order, without cross-twisting, and should be consistent with the wheel groove. The lifting and falling should be flexible, without friction. After the installation of crown blocks and lifting points, the gear train should be filled with calcium base grease, and the wire rope should be coated with surface grease to protect and reduce friction. After the installation of the cable crane crown blocks and lifting points is completed, a comprehensive inspection should be carried out. After the requirements are met, the crane should run with no load, and the acceptance procedures should be handled after it is confirmed to be normal. 4.4
Test lifting inspection of cable crane
Before the formal hoisting and use of the cable crane, the trial hoisting system should be carried out to check the hoisting capacity and the working state of the system. The trial lifting of the cable system includes the determination of lifting weight, the selection of weights, systematic observation, and the collection and sorting of test data. After the trial lifting test is qualified, it shall be put into construction.
5 CONCLUSIONS The cable crane of Zhangjiajie Grand Canyon Glass Bridge has the following advantages: 1. Safe and reliable: The force transmission path of the whole cable crane system is clear, and the force of each component is clear, which is convenient for design and calculation, and can effectively ensure the safety of the whole system; 2. Cost-efficient: The traditional cable crane mostly uses the universal bar assembly tower, and the cable crane in this project uses the large Bailey piece assembly tower, which greatly saves the steel consumption and labor input, and simplifies the requirements for tower assembly construction machinery. Compared with the traditional cable crane, the cost of a cable crane in this project is reduced by more than 50%;
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3. Short construction period: Only one type of Bailey piece is used for the whole tower, which is superior to the use of universal bars; it is easy to assemble and disassemble. After the construction of Zhangjiajie Grand Canyon Glass Bridge is completed, the main materials and equipment of the cable crane can be used for other bridge construction, which improves the turnover times of materials and equipment and reduces the construction cost. The cable crane designed for this project and its installation experience can provide references for similar projects.
ACKNOWLEDGMENTS This work was supported by the Science and Technology Research and Development Project of CSCEC (Project Number: CSCEC-2021-Z-30).
REFERENCES Chang W, Zhu D S, Liu D J, Liang J D, Li Q D, Qiu D Y. (2016) The System Design of Cable Cranes Applied in Erection of Long-span Suspension Bridge in Mountainous Region. Journal of Chongqing Jiaotong University (Natural Science). 35(5):13–16. DOI: 10.3969/j.issn.1674-0696.2016.05.04. China Railway Major Bridge Reconnaissance & Design Institute Co., Ltd. (2015) Construction Drawing Design Documents of Zhangjiajie Grand Canyon Glass Floor Bridge Project. Wuhan, in China. Fang X Z, Wang Z B. (2017) Design of Stiffening Girder of Zhangjiajie Grand Canyon Glass Floor Bridge. Bridge Construction. 47(2):78–82. DOI: 10.3969/j.issn.1003-4722.2017.02.014. Fang N P, Xu X, Wang S. (2022) Design of Heavy Cable Crane for Zigui Changjiang River Highway Bridge. Bridge Construction. 52(1):116–123. DOI: 10.3969/j.issn.1003-4722.2022.01.016. Guo K, Xuan S Y, Ma L, Wang G. (2017) Stiffening Girder Erection Techniques for Zhangjiajie Grand Canyon Glass Bridge. Wold Bridge. 45(3):15–19. DOI: 10.3969/j.issn.1671-7767.2017.03.004. Luo H Y, Wang Z B. (2017) Design of Foundations of Zhangjiajie Grand Canyon Glass Floor Bridge. Bridge Construction. 47(4):96–100. DOI: 10.3969/j.issn.1003-4722.2017.04.017. Ma L, Guo K, Sun S K. (2017) Principal Construction Techniques of Zhangjiajie Grand Canyon Glass Floor Bridge. Bridge Construction. 47(3):99–104. DOI: 10.3969/j.issn.1003-4722.2017.03.018. Wan T B. (2017) Key Techniques of Design of Special Shape Glass Floor Suspension Bridge over Zhangjiajie Grand Canyon. Bridge Construction. 47(1):6–11. DOI: 10.3969/j.issn.1003-4722.2017.01.002. Wang Z B. (2017) Design of Cable Systems of Zhangjiajie Grand Canyon Glass Floor Bridge. Bridge Construction. 47(3):83–87. DOI: 10.3969/j.issn.1003-4722.2017.03.015. Wu Y H, Liang S, Zhao W D. (2017) Design and Application of Bidirectional Moving Cable Crane of Tuoba Bridge. Bridge Construction. 45(3):15–19. DOI: 10.3969/j.issn.1003-4722.2017.02.021. Zhao C F, Shi H Q, Niu Y Z. (2018) Innovative Technology and Application of Cable Suspension System for Stiffening Girder of Long-span Suspension Bridge. High Way. 63(2):92–99. https://kns.cnki.net/kcms/ detail/detail.aspx?dbcode=CJFD&dbname=CJFDLAST2018&filename=GLGL201802019.
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A comparison study on different suction bucket pre-piling templates for offshore wind jacket foundation Zhenya Tian Yangjiang Offshore Wind Power Laboratory, Yangjiang, China
Rongsheng Zhang Ocean College, Zhejiang University, Zhoushan, China
Ronghua Zhu* Ocean College, Zhejiang University, Zhoushan, China Yangjiang Offshore Wind Power Laboratory, Yangjiang, China
Jiezhan Chen China Energy Engineering Group Guangdong Power Engineering Co., Ltd, Guangzhou, China
Hanqiu Liu Ocean College, Zhejiang University, Zhoushan, China
Meiyang Zhang & Xiang Sun Yangjiang Offshore Wind Power Laboratory, Yangjiang, China
ABSTRACT: The pre-piling template has been applied to ensure the accuracy and efficiency of pile-driving of offshore wind pre-piling jacket foundations. In contrast to the conventional pre-piling template with mudmat, two different types of suction bucket prepiling templates (i.e. the pre-piling template with the manual working platform (PPTP) and without the manual working platform (PPT)) were introduced to adopt the unique soft clay geological condition in China. This paper presents a comparison study on the structure strength and the suction bucket foundation bearing capacity of PPTP and PPT based on an offshore wind project in the South China Sea. The results showed that the PPT not only has a lightweight advantage but also can reduce the risk and time of installation of suction buckets. The PPT was successfully applied to Yangjiang Qingzhou offshore wind farm project for the pile installation of the pre-piling jacket foundation. The results of this work can provide a reference for the design practice of the pre-piling templates for the OWT prepiling jacket foundation.
1 INTRODUCTION Offshore wind power has achieved rapid development in China in the last decades. The offshore wind turbine (OWT) pre-piling jacket foundation has been frequently used as the water depth of the offshore wind farm (OWF) exceeds 30 m (Zhang et al. 2022). For the prepiling jacket foundation, three or four steel piles should be hammered into the seabed previously with the help of the pre-piling template before the installation of the jacket structure
*Corresponding Author: [email protected] DOI: 10.1201/9781003425823-38
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(Jalbi et al. 2020; Jiang 2021). According to the design standard (DNVGL 2016), the vertical inclination angle of the embedded pile cannot be greater than 0.5 , and the tolerance of distance between each pile should be less than 100 mm. Therefore, the pre-piling template has a great impact on the efficiency and accuracy of offshore pile installation. Zhang et al. (2021) carried out the construction design and analysis of the offshore wind multi-pile pre-piling template, which is used for piling on the hard seabed. Whilst the prepiling template is vital for the installation process of the OWT jacket foundation, there are few academic reports on the pre-piling template with suction buckets for soft clay seabed, especially the comparison study of two types of suction bucket pre-piling templates (i.e., the pre-piling template with the manual working platform (PPTP) and without the manual working platform (PPT)). Based on the marine environment and geological conditions of a specific offshore wind farm in the South China Sea, this paper investigates the structure strength and foundation bearing capacity concerned by engineering practice of PPTP and PPT for comparison.
2 MARINE AND SOIL DATA The preliminary designs of PPTP and PPT are based on the marine environment and geological condition of a specific offshore wind farm in the South China Sea. The four-pile jacket foundation was adopted as the OWT support structure, which has good fatigue performance and anti-typhoon capability. Two load cases (i.e., the ultimate limit state (ULS) and serviceability limit state (SLS)) are set to cover the extreme and operating conditions of the template platform, and their corresponding marine data are listed in Table 1. Table 1.
Marine data for ULS and SLS load cases.
Load case
Water depth (m)
Maximum wave height (m)
Current speed at surface (m/s)
Wind speed at 10 m above sea level (m/s)
ULS SLS
40 40
9.0 2.5
1.50 0.95
45.5 20.0
As shown in Table 2, the main geological condition of the site is silty clay. The previous geological survey showed that the strength of soil near the mud surface was relatively low. If the conventional pre-piling template with mudmat was used, it was prone to cause settlement Table 2. Soil layer No. 1 2 3 4 5 6 7
Soil data of the offshore wind farm.
Soil l ayers Silty clay Silty clay Silty clay Silty clay Silty clay Medium dense sand Silty clay
The thickness Saturated Undrained shear of the soil unit weight Relative density (%) strength (kPa) layer (m) (kN/m3)
Effective internal friction angle ( )
11.2 10.1 8.8 4.1 10.7 6.6
16.5 17.2 16.8 18.3 18.4 20.8
10 20 30 55 60 0
0 0 0 0 0 37
7.8
18.6
70
0
65
290
and tilt during pile-driving, which would exert an adverse effect on the efficiency and accuracy of installation. Comparatively, the pre-piling template with suction buckets has better bearing capacity and on-site stability performance than the pre-piling template with mudmat in the soft soil ground (Kim et al. 2012). When the suction bucket is installed, the inclination of the pre-piling template can be controlled by changing the suction pump flow (Zdravkovic et al. 2001). In terms of construction convenience, the pre-piling template could be moved simply by pumping water into the buckets after the pile-driving completion (Andersen et al. 2005) In the design stage, considering that the offshore wind farm is located in a typhoon-prone area, it is necessary to take into account the safety of the pre-piling template structure and the bearing capacity of the suction bucket foundation under extreme service conditions. So, this paper will take the analysis and discussion on these two aspects.
3 NUMERICAL MODEL SETUP AND LOAD CASE 3.1
PPTP model
The horizontal spacing distance between the suction bucket centres of the PPTP is 20 m, and the diameter of the suction bucket is 7 m with a wall thickness of 30 mm. According to the design input of the OWT jacket structure, the horizontal spacing distance between the piling sleeves is 30 m, and the diameter of the pile sleeve is 3.6 m. The PPTP is mainly fabricated by steel tubes. The top of the PPTP platform is constructed with H-shaped steel. The diameter of the main support tube of the PPTP is changed from 2.0 m to 1.0 m by cone transition. Given the weak stiffness in the horizontal direction between adjacent suction buckets, there are three layers of horizontal reinforcement steel pipes (0.63 m in diameter) and vertical steel pipes (0.42 m in diameter) between the main columns to strengthen the horizontal stiffness of the whole structure. The four vertical steel tubes (1.0 m in diameter) in the splash zone are reinforced with two layers of “K” type nodes. Four pile guiding ports are arranged around the PPTP platform. The axis of the pile guiding port is aligned with the underwater piling sleeve. The pile guiding port is used for guiding the pile into the underwater piling sleeve. The manual working platform of the PPTP can be used to lay out pile-driving equipment. Since many typhoons pass through the construction site every year, it is hard to evacuate PPTP in a timely manner. Therefore, when the PPTP is in place, the top of the manual working platform is often laid with steel plates. Before the typhoon comes, the steel plates laid on top of the platform are removed to reduce the wave loads on the platform. The SACS software was applied to perform the structural analysis of the pre-piling template. The numerical model of PPTP is shown in Figure 1.
Figure 1.
The PPTP model.
Figure 2.
The PPT model.
291
Figure 3. model.
The suction bucket
3.2
PPT model
The dimensions of the underwater substructure of PPT are the same as the PPTP. The horizontal spacing distance, diameter, and wall thickness of the suction bucket are the same as well. The OWT jacket piles’ horizontal spacing between the piling sleeves is 30 m, and the pile sleeve diameter is 3.6 m, the same as PPTP. The diameter of the main underwater support tube transits from 2 m to 1.5 m. There are three layers of horizontal reinforcement steel pipes (0.5 m in diameter) and vertical steel pipes (0.42 m in diameter) between the main columns. Four diagonal braces connect the support tubes to the vertical pipe. A part of the vertical pipe is located above sea level, mainly used for lifting and recycling PPT. The numerical model of PPT is shown in Figure 2. 3.3
Suction bucket model
The internal loads of each pile head of the PPTP and PPT were extracted to further analyze the foundation bearing capacity of the suction bucket by Plaxis 3D finite element software. The Mohr-Coulomb soil constitutive model was used to model the nonlinear behaviour of soil (Wang et al. 2021). The suction bucket structure dimension of PPTP is 7 m 0.03 m 15 m (7 m in diameter, 0.03 m in wall thickness, 15 m in height). The suction bucket structure dimension of PPT is 7 m 0.03 m 10 m (7 m in diameter, 0.03 m in wall thickness, 10 m in height). The soil domain has a horizontal length of 30 m and a depth of 30 m to minimize the boundary effects. The finite element model of the suction bucket is shown in Figure 3. 3.4
Loading conditions and load cases
Morison’s equation is used to calculate the wave loads on slender members, which can be expressed as (API 2000): F ¼ CM
rpD2 rD u_ þ CD ujuj 4 2
(1)
where CM is the hydrodynamic inertia coefficient, 1.6 for smooth members, 1.2 for fouling members with marine growth; CD is the hydrodynamic drag coefficient, 0.65 for smooth members, 1.05 for fouling members with marine growth; u is the horizontal velocity of water particle; u_ is the horizontal acceleration of water particle; D is the diameter of the structure component submerged in the water; r is the density of seawater, taken as 1028 kg/m3. The sea current loads are calculated according to the following formula (API 2020): 1 FD ¼ CD rAUc2 2
(2)
where CD is the hydrodynamic drag coefficient, which is 1.2; A represents the projected area of the unit length member perpendicular to the direction of the sea current; Uc denotes the current speed. In Consideration of the operation process and potential typhoon events during the service period of the pre-piling template, both SLS and ULS analysis load cases are set as listed in Table 3 (API 2020). The SLS load cases are divided into four subcases, which correspond to the situation that the four steel piles are installed in sequence. ULS load case is the typhoon condition when the pre-piling platform is in place. For SLS, the load factors of components of permanent load, environmental loads, and live load are all taken as 1.0. For ULS, the load factor of component of environmental loads is taken as 1.35, and the rest is identical. Due to the plane symmetry of the pre-piling template, the loading directions are set as 0 and 45 degrees, respectively, as shown in Figure 4. 292
Figure 4.
Table 3.
Sketch of loading direction.
SLS and ULS load cases.
Load case
Description
Load combinations and load factors
SLS
Installation of 1st pile
1.0 permanent load + 1.0 (operating wind + wave + current) + 1.0 live load 1.0 permanent load + 1.0 (operating wind + wave + current) + 1.0 live load 1.0 permanent load + 1.0 (operating wind + wave + current) + 1.0 live load 1.0 permanent load + 1.0 (operating wind + wave + current) + 1.0 live load 1.0 permanent load + 1.35 (extreme wind + wave + current)
Installation of 2nd pile Installation of 3rd pile Installation of 4th pile ULS
Typhoon condition
4 RESULT AND DISCUSSION 4.1
Structure strength analysis
The wind, wave, current and permanent loads are loaded on the two pre-piling template models according to the pre-set load cases. By comparison of the analysis results under various load cases, it can be found that the ULS case is the dominant load case. Therefore, this paper shows typhoon condition (ULS) analysis results only. The structure strength analysis result of the PPTP is shown in Figure 5. The maximum unity check (UC) value of the PPTP is 0.88, which occurs at the position where the pile guiding port is connected with the vertical column. The structure strength analysis result of the PPT is shown in Figure 6. The maximum UC value of the PPT is 0.758, which occurs at the position where the diagonal support tube is connected with the main column. Both of the maximum UC values of the two pre-piling templates are less than 1, meeting the design requirements. The structural deformation diagram of the PPTP is shown in Figure 7. The maximum deformation of the PPTP occurs at the edge of the deck beam, with a value of 27 mm. The structural deformation diagram of the PPT is shown in Figure 8. The maximum deformation of PPT occurs at the diagonal brace, with a value of 120 cm. Both structural deformations of the two pre-piling templates meet the design requirements. The UC value of the PPTP is close to that of the PPT. But the structural deformation of the PPTP is much smaller than that of the PPT. It indicated that the structural stiffness of the PPTP is greater than that of the PPT.
293
Figure 5.
UC value of the PPTP.
Figure 6.
UC value of the PPTP.
Figure 7.
Deformation of the PPTP.
Figure 8.
Deformation of the PPT.
4.2
Suction bucket analysis
The soil displacement patterns around the suction bucket for two pre-piling templates under tension and compression load cases are shown in Figures 9–12. The largest soil displacement occurs under the bucket top plate of the PPTP and PPT. The suction bucket failure model of the PPTP is shown as the rotational displacement field of soil formed in the middle and lower part of the bucket. An eccentric rotation centre appears at a depth of 10 m below the top plate of the suction bucket. The wedge-shaped active soil pressure zone and passive soil pressure zone are generated between 010 m below the mud surface. Different from the PPTP, a rotational displacement failure pattern is displayed at the base of the PPT suction bucket. The rotation centre is also located at the same position. The active soil pressure zone and passive soil pressure zone are formed around the suction bucket and extend to a larger range. Due to the asymmetry of load on the suction bucket, both the active and passive soil pressure zones around the suction bucket are asymmetric too. Given the weak stiffness of the shallow soil layer in this OWF, there is a large displacement at the top of the suction bucket. Thus, it is necessary to strengthen the structural rigidity of the top plate to avoid local buckling failure and reduce the soil displacement below the bucket in the detailed structure design stage (Skau 2019).
294
Figure 9. Displacement result of the PPTP suction bucket under tension.
Figure 10. Displacement result of the PPTP suction bucket under compression.
Figure 11. Displacement result of the PPT suction bucket under tension.
Figure 12. Displacement result of the PPT suction bucket under compression.
5 ENGINEERING APPLICATIONS The cross-sectional area of the PPT near the sea level is smaller than that of the PPTP, leading to smaller marine loads on the structure. As a result, the suction bucket height of the PPT is smaller than that of the PPTP, with a height of 10 m can meet the foundation bearing capacity requirements. Comparatively, the suction bucket height of the PPTP is 15 m, with a dead weight of 1200 tons. The dead weight of the PPT is about 25% lighter than that of the PPTP. The PPT has an obvious lightweight advantage. Finally, the pre-piling template without the manual working platform (PPT) was selected. It was successfully applied to the Yanjing Qinzhou OWF project in Guangdong province, China. It takes 26 hours to install the OWT jacket foundation, including durations of PPT installation and positioning in place, installing four steel piles, and pulling out the suction bucket from the seabed. The installation of steel piles by utilizing PPT in the Yanjing Qinzhou OWF project is presented in Figures 13–14. The field practice shows that although there is no manual working platform and the pile guiding port for PPT, which exerts little impact on the construction efficiency. Since the dead weight of PPT is lighter than that of PPTP, the utilization of PPT will reduce the crane requirements and ship costs. Furthermore, the suction bucket foundation of the PPT is shorter than that of the PPTP, which could also reduce the risks and time of the installation of the suction bucket.
295
Figure 13. The PPT was installed in the Yangjiang Qingzhou OWF project.
Figure 14. Jacket piles installed in Yangjiang Qinghzhou OWF project.
6 CONCLUSIONS This paper presents a comparison study of structure strength analysis and suction bucket analysis of two types of pre-piling templates (PPTP and PPT). The PPT was selected and successfully applied to the Yangjiang Qingzhou OWF project. The main conclusions are summarized below: l
l
l
l
The pre-piling template with suction buckets is very suitable in the South China Sea with a wide distribution of soft clay seabed from the perspective of efficiency and accuracy of installation. Through numerical analysis, it was determined that the PPTP and PPT under ULS (typhoon load case) both meet the design criteria for structural strength and foundation bearing capacity. Since the PPT without the working platform is less affected by wave loads, the dead weight of the PPT is about 25% lighter than that of the PPTP, and the height of the PPT suction bucket is 5 m shorter than that of the PPTP under the same marine and geological conditions. The engineering case presented noticeable advantages in terms of the selection of crane ships and installation of suction buckets by utilizing PPT. The PPT will have more advantages in deeper offshore wind projects than PPTP.
ACKNOWLEDGEMENTS The authors would like to acknowledge the support from the Key-Area Research and Development Program of Guangdong Province (No. 2021B0707030001; No. 2022B0101100001) and the Guangdong Province Science and Technology Special Funds and Science (No. SDZX2021005)
REFERENCES Andersen, K. H., Murff, J. D., Randolph, M. F., Clukey, E. C., Erbrich, C. T., Jostad, H. P., . . . & Supachawarote, C. (2005). Suction Anchors for Deepwater Applications. In Proceedings of the 1st International Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth (pp. 3–30). https://doi.org/ 10.1201/NOE0415390637
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API (American Petroleum Institute). (2000) Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms—Working Stress Design. API Recommended Practice 2A-WSD (RP 2A-WSD). Jalbi, S., & Bhattacharya, S. (2020) Concept Design of Jacket Foundations for Offshore Wind Turbines in 10 Steps. Soil Dynamics and Earthquake Engineering, 139, 106357. https://doi.org/10.1016/j. soildyn.2020.106357 Jiang, Zhiyu. “Installation of Offshore Wind Turbines: A Technical Review.” Renewable and Sustainable Energy Reviews 139 (2021): 110576. https://doi.org/10.1016/j.rser.2020.110576 Kim, S. R. (2012) Evaluation of Vertical and Horizontal Bearing Capacities of Bucket Foundations in Clay. Ocean Engineering, 52, 75–82. https://doi.org/10.1016/j.oceaneng.2012.06.001 Skau, K. S., Jostad, H. P., Eiksund, G., & Sturm, H. (2019) Modelling of Soil-structure-interaction for Flexible Caissons for Offshore Wind Turbines. Ocean Engineering, 171, 273–285. https://doi.org/10.1016/j. oceaneng.2018.10.035 Veritas, D. N., & Lloyd, G. (2016) DNV GL-ST-0126 Support Structures for Wind Turbines. Det Norske Veritas: Høvik, Norway. Wang, J., Sun, G., Chen, G., & Yang, X. (2021) Finite Element Analyses of the Improved Lateral Performance of Monopile When Combined with Bucket Foundation for Offshore Wind Turbines. Applied Ocean Research, 111, 102647. https://doi.org/10.1016/j.apor.2021.102647 Zdravkovic, L., Potts, D. M., & Jardine, R. J. (2001) A Parametric Study of the Pull-out Capacity of Bucket Foundations in Soft Clay. Geotechnique, 51(1), 55–67. https://doi.org/10.1680/geot.2001.51.1.55 Zhang, J., & Wang, H. (2022) Development of Offshore Wind Power and Foundation Technology for Offshore Wind Turbines in China. Ocean Engineering, 266, 113256. https://doi.org/10.1016/j. oceaneng.2022.113256. Zhang, Z.B, Lu, H, Qiu, Y, Zuo, M, et al. (2021) Construction Design and Safety Analysis of Offshore Wind Power Pile Stabilization Platform. J.Thansportation science and technology, 4, 155–160. https://10.3963/j. issn.1671-7570.2021.04.034
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Numerical simulation study on ground subsidence caused by cavity under buried PCCP socket Lin Cheng* Associate Professor, State Key Laboratory of Eco-hydraulics in Northwest Arid Region, Xi’an University of Technology, Xi’an, China
Pengsheng Pan, Yuheng Zhang, Zengguang Xu & Yue Jiang State Key Laboratory of Eco-hydraulics in Northwest Arid Region, Xi’an University of Technology, Xi’an, China
ABSTRACT: In this paper, the numerical simulation method was used to study the influence of the dimensions of the cavity under the buried PCCP socket and the thickness of the covering soil on the upper part of the pipeline on the ground deformation and the stress deformation of the upper pipe body. The results show that the ground subsidence value increases with the increase of the height of the cavity and the span, and when the span reaches a certain critical value, the subsidence value will increase rapidly, and the road surface shows a tendency to be damaged; with the continuous increase of the thickness of the covering soil, the settlement displacement of pavement subsidence centre gradually decreases; under the same span, as the thickness of the covering soil continues to increase, the settlement area of the pavement surface continues to decrease, and with the increase of the span of the cavity, the development and change of the settlement area becomes relatively slow; the surface settlement displacement curve basically conforms to the modified Gaussian distribution law. With the increase of the span of the cavity, although the stress state of the pipeline body changes significantly, the cavity generated at the socket does not cause major structural damage to the PCCP pipeline body. The results of this paper can provide certain guidance for the safety monitoring and structural safety evaluation of buried PCCP.
1 INTRODUCTION In recent years, to satisfy the needs of urban industrial and domestic water, China has built or is building a number of long-distance, large-flow and high-lift pipeline water transfer projects. Steel pipes, concrete pipes, and prestressed concrete cylinder pipes (PCCP) are often used as water transfer pipes in water transfer projects. Among them, since its development in the 1940s, PCCP has been adopted by many foreign water transfer projects due to its strong impermeability, high reliability, good durability, excellent earthquake resistance and low management costs (Wang 2018). In the late 1980s, PCCP was first introduced in China. Since then, the production lines have been expanded continuously, reaching nearly a hundred in recent years. The annual design and production length of PCCP in China has reached more than 5, 000 km, and thousands of kilometres of PCCP are laid every year (Ma 2013). However, due to the generally large water volume and water pressure of the water transfer pipeline, long-term operation of the pipeline may result in pipeline leakage, rupture, and
*Corresponding Author: [email protected]
298
DOI: 10.1201/9781003425823-39
other issues. Many factors can cause PCCP leakage and damage, among which the leakage damage of the socket due to uneven foundation settlement and installation construction is one of the main reasons (Hu 2018; Zhao 2020). The socket is located at the weak point of the pipeline body, and the hidden underground cavity due to the erosion of the soil by the leakage from the socket will not only cause further damage to the socket, but also may lead to the deformation of the pipeline, thereby changing the stress of the soil around the pipeline. If it is discovered in time, major safety accidents, such as pavement surface subsidence, may eventually occur (Li 2020). According to the statistics of Li (2016), the number of pavement surface subsidence or settlement accidents in China can reach as high as 500–1000 times every year. These accidents caused damage to underground pipelines, long-term interruption of road traffic, and major traffic accidents, and brought great threats to people’s lives and properties. Therefore, it is highly important to deeply study the influence of the cavity under the socket on the stress deformation of the ground and the upper part of the pipeline body. Scholars at home and abroad have carried out a series of research on the impact of the cavity under the water transfer pipeline on the ground and the pipeline. Hou et al. (2013) analyzed the ground surface subsidence accidents caused by cavities due to pipeline leakage in recent years. They believed that the underground hidden cavities formed by the leakage of water transfer pipelines were disturbed by external effects, such as upper loads, vibrations, and precipitation, leading to the instability of the critical state, and eventually causing the ground surface subsidence. Davies et al. (Davies 2001) analyzed pipeline construction technology, corrosion damage, and other issues. They found that the pipeline wall and pipe socket joints of the sewage pipeline are relatively more likely to be damaged. Whenever the groundwater level is higher than the pipeline, especially in the rainy season, the groundwater will take away a large amount of soil from the broken part of the seepage pipeline, which in turn leads to the formation of an underground cavity. Ju et al. (Ju 2015) explored the deformation characteristics of underground pipelines and underground surface subsidence through model tests and numerical simulation calculations. Then they put forward engineering suggestions for preventing pipeline accidents caused by ground surface subsidence. Zhou et al. (Zhou 2016) studied the settlement characteristics of the covering soil on the HDPE pipeline and the stress change of the pipeline through large-scale indoor model tests. The above research simplified the water transfer pipeline into a continuous pipeline. It analyzed its dynamic response, without considering the interaction between the pipeline and the soil, and that of the socket between pipelines. In actual engineering, the socket is a vulnerable part of the pipeline, which is often one of the main reasons for pipeline damage (Wang 2018). Therefore, in this paper, ABAQUS finite element software was used to study the ground surface subsidence caused by the cavity under the buried PCCP socket under traffic loads, and discuss the impact of different dimensions of underground cavities formed due to the erosion of the surrounding soil body of the pipeline and different thickness of the covering soil on the ground structure and the PCCP pipeline body, thereby providing certain guidance for the structural safety monitoring of buried PCCP pipelines.
2 FINITE ELEMENT ANALYSIS MODEL 2.1
PCCP model size
In this test, the PCCP pipeline (DN1400) water transfer supporting project of Putian Jinzhong Water Conservancy Pivot was taken as the background, and the buried depth of the pipeline is 3m6m. A finite element model of a PCCP (containing two sets of sockets) and its surrounding soil body was established based on the nonlinear analysis software ABAQUS. PCCP consists of four materials: concrete, steel cylinder, prestressed steel wire and mortar. The specific dimensions of each component are shown in Table 1.
299
Table 1.
PCCP dimensions.
Pipeline inner diameter
Thickness
Steel cylinder thickness
Steel cylinder inner diameter
Mortar thickness
Wire diameter
Winding pitch
1.4 m
0.12 m
0.002 m
1.52 m
0.03 m
0.006 m
0.01241 m
When establishing the finite element model, the foundation, concrete, backfill, and mortar were simulated using the three-dimensional solid element C3D8R. The steel wire was simulated using the three-dimensional rod element T3D2, and the steel cylinder was simulated using the shell element S4R. The modelling range is 15 times the pipeline diameter. The finite element model has a total of 213472 elements and 235656 nodes. The finite element model is shown in Figure 1.
Figure 1.
Finite element model.
In view of the difficulty in measuring the geometric dimensions of the cavity, to simplify the construction of the model, many scholars have found through physical tests that when the gravity flow is present, the test results show that the boundary area of the soil flow is elliptical, which is in agreement with the test results based on the traditional gravity flow theory. During soil loss, the flowing soil will be affected by shear dilation, and the soil body within the flow area will become loose. Therefore, the flow area is also called a “loose ellipse”. That is, the overall area of soil erosion can be approximated as an ellipsoid (Dai 2016; He 2017; Yuan 2014). Based on the above research results, in this paper, the ellipsoid was the underground soil body cavity unit. To accurately express the shape of the underground ellipsoidal cavity, along the Y direction, a is used to represent the long semi-axis, (the long-axis cavity is the span of the cavity); along the X direction, b refers to the width semiaxis; and along the Z direction, c represents the height semi-axis. When selecting variables, the width-to-length ratio, height-to-length ratio, and length are used to control the shape, which are 0.95 and 0.6, respectively. In this paper, a total of six sets of cavity dimensions were formulated, and the “model conversion” function of ABAQUS was used to simulate the generation of cavities. See Table 2 for specific values. 300
2.2
Table 2.
Calculation conditions of different cavity dimensions.
Set
Long semi-axis (m)
Width semi-axis (m)
Height semi-axis (m)
1 2 3 4 5 6
1.5 2.0 2.5 3.0 3.5 4.0
1.4 1.9 2.4 2.9 3.3 3.8
0.9 1.2 1.5 1.8 2.1 2.4
Boundary constraints and interlayer relationships
Apply normal constraints around the model and apply full constraints at the bottom of the model. For PCCP, a separate modeling method was applied without considering the relative slippage or cavity phenomenon of each component. The relationship among mortar and steel wire, concrete and steel cylinder is embedding, and that between the concrete and mortar is binding. For the normal behaviour of the interaction between pipes, hard contact is constrained by the classical Lagrange multiplier method. Coulomb friction is used for the tangential behaviour, and no relative slip is considered for the pipe-soil interaction. 2.3
Constitutive relations and model materials
Considering the nonlinear characteristics of materials, such as concrete and mortar cracking and the yield characteristics of steel wire and steel cylinder in the model, the constitutive relationship of the pipeline body material is determined according to the American AWWAC304 standard (ANSI/A WWA C 304-99) during the calculation. The stress-strain relationship of prestressed steel wire is shown in Formula (1): ( es Ens oðes 0:75fsu =Es Þ s¼ (1) 2:25 fsu 1 ½1 0:6133ðes Es =fsu Þ ðes > 0:75fsu =Es Þ where fsu refers to the tensile strength of prestressed steel wire, which is taken as 1570 MPa; s represents the stress of prestressed steel wire; and es stands for the strain of prestressed steel wire. Where fsg is 0.75 times of fsu, and the steel wire yields when the stress of the steel wire exceeds 0.85 times of fsu. The stress-strain relationship curve of the prestressed steel wire is shown in Figure 2(a). The socket steel ring and steel cylinder are made by welding the Q235 steel plate. The standard value of tensile and compressive strength is 235 MPa. The stress-strain relationship is based on the ideal elastoplastic model. And the stress-strain relationship curve is shown in Figure 2(b). The concrete damage plasticity (CDP) model was used for exploring the stress-strain relationship between concrete and mortar. Since the tensile strength of mortar is ftm’, the corresponding peak tensile strain is etm, the tensile strength of concrete is ft’, and the corresponding peak tensile strain is et, when the tensile strains of concrete and mortar reach 11et and 8etm, respectively, cracks can be observed. The bearing capacity is lost in the meantime. The stress-strain relationship of the concrete core is shown in Figure 2(c), and that of the protective layer mortar is shown in Figure 2(d). In the finite element simulation calculation of ground subsidence caused by the cavity under the buried PCCP socket under the action of traffic load, the material density, elastic modulus, strength and Poisson’s ratio of each component of PCCP are shown in Table 3. The basic parameters of soil body material are shown in Table 4.
301
Figure 2.
Table 3.
Constitutive relationship curve of pipeline body material.
Pipe body material parameters.
Material
Density (kg/m3)
Elastic modulus (MPa)
Compressive strength (MPa)
Tensile strength (MPa)
Poisson’s ratio
Steel wire Concrete Steel cylinder Socket steel ring Mortar
7833 2500 7833 7833 2200
205000 35500 206000 206000 24165
1570 35.5 235 235 45
1570 2.74 235 235 3.49
0.3 0.2 0.3 0.3 0.2
Table 4.
Soil body material parameters.
Soil layer Foundation soil Cushion soil Backfill
Density (kg/ m3)
Elastic modulus (MPa)
Poisson’s ratio
/
80
0.25
/
/
1766 1550
40 20
0.29 0.3
30 25
21 21
302
Cohesion (KPa)
Internal friction Angle/( )
In the simulation calculations to be carried out in this paper, the asphalt concrete layer, cement-stabilized gravel layer, and lime-fly ash layer are usually based on linear elastic layered bodies. Therefore, the simulation of structural layers, such as pavement surfaces, etc., are also based on linear elastic models in this simulation. For foundation soil and backfill soil, the elastic-plastic theory is used for calculation, and the strength criterion is based on the Mohr-Coulomb model. In addition, considering the main structural form of easily damaged pavement that is common in practice, the thickness of the asphalt concrete layer, cement-stabilized gravel layer, and lime-fly ash layer is set to 15 cm, 30 cm, and 25 cm, respectively, to establish the finite element numerical model in this paper. The data is representative of a wide range, and it is a common structural thickness type in cities. The finite element simulation parameters of each layer of material are shown in Table 5.
Table 5.
Pavement structure finite element simulation parameters of asphalt concrete road.
Material
Asphalt concrete
Cement stabilized gravel
Lime-fly ash
Thickness (m) Density (KN/m3) Modulus (MPa) Poisson’s ratio
0.15 25 1200 0.25
0.3 24 900 0.28
0.25 23 380 0.34
3 RESULT ANALYSIS 3.1
Influence of cavity dimensions and soil thickness on pavement stability
To determine the influence of the dimensions of the underground cavity on the stability of the pavement, the cavity dimension data of Set1-6 in Table 2 is selected, and the buried depth is 3 m, 4 m, 5 m, 6 m, and 7 m, respectively. Static and dynamic calculations are carried out for each model, respectively, to analyze the influence of ellipsoidal cavities of different dimensions on ground settlement. ABAQUS is used to calculate the settlement displacement of the pavement and the change of the plastic zone around the cavity to analyze the influence of the span of the underground cavity on the subsidence of the pavement. It is worth noting that when the calculation does not converge, it can be approximately judged that the soil body is damaged. At the same time, the pavement loses its stability, and the soil cavity collapses. The settlement displacement diagram of the pavement under the covering soil in the thickness of 3m is shown in Figure 3. The subsidence displacement curves of the underground cavity roof under different spans and covering soil thicknesses are shown in Figure 4.
Figure 3. The settlement displacement curve of the pavement along the Y direction under the thickness of covering soil of 3 m.
Figure 4. Subsidence displacement curves of underground cavity roofs with different spans and covering soil thicknesses.
303
It can be seen from Figures 3 and 4 that at a span of 8 m, the pavement settlement value increases rapidly, and the pavement loses stability and becomes damaged. At the same time, it can be considered that the pavement has collapsed due to overall damage. Based on the above calculation and analysis, it is considered that the limit span of subsidence damage failure is about 8 m. Regarding the influence of the covering soil thickness, it can be seen from Figure 4 that the settlement displacement at the pavement subsidence centre gradually decreases with the increasing thickness of the covering soil. The analysis shows that when the buried depth is shallow, the covering soil of the cavity is difficult to maintain its stability due to the thin roof of the cavity, and the settlement value is large; the thickness of the covering soil of the cavity continues to increase, which can reduce the impact of the cavity on the pavement structure, and the settlement value decreases accordingly. 3.2
Distribution of ground subsidence
Since the underground cavity is an ellipsoid, when the pavement collapses, the settlement area of the pavement is also approximately elliptical, as shown in Figure 5. By calculating and extracting the pavement surface settlement displacement curve data along the pipeline centre direction (y direction) and the direction perpendicular to the pipeline centre (x direction), the long-axis pavement surface settlement displacement value and the short-axis one of the ellipsoids in the settlement area can be obtained. Based on the existing ground subsidence monitoring technology, such as interferometric synthetic aperture radar, InSAR, global positioning system, GPS, Brillouin optical frequency domain analysis, BOFDA, with the monitoring accuracy of 5 mm or more (Cao 2021; Cheng 2022; Xia 2020), in this paper, the settlement value greater than or equal to 5 mm was taken as the starting point, which can determine the distribution range of the measured value. Then, according to the corresponding coordinate value, the lengths of the long semi-axis and the short lengths of settlement elliptical areas with different spans can be obtained. As per the ellipse area formula, as shown in Formula (2), the elliptical area can be calculated. The resulting distribution range of ground subsidence is shown in Figure 6. S ¼Pab
(2)
where a refers to the long semi-axis of the ellipse, and b stands for the short semi-axis of the ellipse. It can be seen from Figure 6 that with the continuous increase of the span of the cavity, the ground subsidence area is also expanding. Under the same span, the settlement area of the pavement decreases with the continuous deepening of the covering soil thickness, and the development and change of the settlement area are relatively slow with the increase of
Figure 5. The settlement distribution diagram of the pavement with a cavity span of 8m under the covering soil thickness of 3 m.
Figure 6. The distribution range of ground subsidence.
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the span of the cavity. Taking the limit span of ground subsidence damage as 8 m, for example, the settlement range of the covering soil in the thickness of 3 m is 101.3 m2, the deepest part of pavement surface subsidence has reached 8.5 cm, while the settlement range of the covering soil in the thickness of 7 m is 32.1 m2, in the meantime, the settlement value of the pavement surface centre has been reduced to 1.8 cm. It can be seen that the span of the cavity and the thickness of the covering soil of the pipe are important factors affecting the subsidence of the pavement. 3.3
Distribution law of pavement surface settlement displacement curve
According to the calculation results of the finite element model in Section 2.1, it is found that the distribution of the ground subsidence displacement curve caused by the cavity under the buried PCCP socket under the action of traffic load is approximately a Gaussian normal distribution. In this paper, using the modified Gaussian curve model proposed by Klar et al. (KLAR 2005), the Levenberg-Marquardt least squares optimization algorithm was employed to fit the surface subsidence measured by numerical simulation. The expression is as follows: SðxÞ ¼
n n 1 þ eað i Þ
x 2
Smax
(3)
where Smax refers to the maximum displacement of surface settlement; n stands for the curvature parameter of the settlement curve; i denotes the width parameter of the settlement trough n ¼ ea 2a1 2aþ1 þ 1; and a is the correction coefficient. Table 6 shows the fitting parameters of the modified Gaussian curve formula and the maximum value of the ground surface settlement. Figure 7 shows the fitting results of the settlement values under different spans. Table 6. Fitting parameters of the modified Gaussian curve formula and the maximum value of the ground surface settlement. Experimental stage
i/mm
a
Smax/Fitted value
Smax/Measured value
I II III IV V VI
1.06 1.12 1.84 2.39 2.92 3.37
0.00 0.00 0.23 0.41 0.59 0.71
0.72 1.30 2.76 4.48 6.16 8.31
0.82 1.39 2.88 4.61 6.25 8.50
From the above test results, it can be concluded that the ground surface settlement curve fitted using the Klar method is very close to the numerical simulation calculation results regarding curve distribution. The area with relatively large errors is close to the centre of the settlement. That is, the measured value under each working condition is relatively larger compared with the fitted value. The influencing factor is considered to be the result of the traffic load applied in the numerical simulation calculation. To sum up, the modified Gaussian formula can well describe the characteristics of the settlement displacement of the pavement in this paper. Therefore, on the whole, it can be considered that the displacement curve of the ground surface settlement caused by the cavity under the buried PCCP socket under the action of traffic load basically conforms to the modified Gaussian distribution law.
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Figure 7.
3.4
Test results of pavement surface settlement displacement curve.
Effect of the cavity on the PCCP pipeline body
Because the generation of the cavity in the foundation will cause uneven distribution of foundation bearing capacity, the pipe-soil interaction with the existence of the cavity will further cause the pipe stress to change. Therefore, in this paper, the working conditions of cavities under different spans were simulated, obtaining the variation of the stress peak value at the vulnerable characteristic position of each PCCP component under different working conditions. Based on the material stress-strain relationship in Section 1.3, it can be obtained that the concrete core and mortar are compressive materials instead of tensile ones. The circumferential compressive stress of the concrete core bottom, pipe waist outside, pipe top inside and mortar pipe waist is small. When the force is applied, it may cause circumferential tension and cracking. Therefore, in this paper, it is necessary to focus on the analysis of the development of the first principal stress at the characteristic position when the traffic load is just above the underground cavity; the tensile and compressive strengths of prestressed steel wire and steel cylinder are the same, and it is necessary to focus on the analysis of the development of Mises stress at the characteristic position when the traffic load is just above the underground cavity. The result is shown in Figure 8. As shown in Figure 8, (1) With the increase of the span of the cavity, the peak value of the first principal stress at the top of the concrete core increases slowly. The maximum stress value at the top of the pipe is 0.77 MPa at the span of 8 m, but it is still smaller than the tensile strength of the concrete of 2.74 MPa; the stress at the top of the pipe decreases slowly from 0.02 MPa at the span of 3 m to 0.57 MPa. With the continuous expansion of the cavity, the PCCP is not supported at the bottom along the length of the cavity. There is a tendency to sink at the middle pipe joint, and the pipe top is obviously under pressure, and the pressure at the pipe waist increases steadily, but the growth rate is slow, and it is always in a state of compression. (2) With the increase of the span of the cavity, the stress at the pipe waist of the mortar layer increases slowly, and it is always lower than the concrete tensile strength of the mortar layer; the pressure at the bottom of the pipe slowly decreases from 0.136 MPa to 0.022 MPa, and the value change is not obvious; the value of the pipe top increases slowly from 0.007 MPa to 0.09 MPa, and the change value is less than 0.1 MPa, which is far lower than the tensile strength of the mortar layer concrete.
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Figure 8. Variation law of stress peak values at the characteristic positions of PCCP components under different cavity spans.
(3) With the increase of the span of the cavity, the increase rate of the Miss stress at the waist of the steel cylinder tends to 0, and the change is not obvious. The stress at the maximum span of 8 m is 84.308 MPa, which is far less than the tensile strength of the steel cylinder of 235 MPa; when the Miss stress at the bottom of the pipe changes from a 3 m span to a 6 m span, the peak acceleration becomes larger, and when it changes from 6 m span to 8 m span, the peak acceleration slows down, and the maximum value of 61.213 MPa is still much smaller than the tensile strength of the steel cylinder; the stress at the top of the pipe is in a continuous decreasing trend, but the overall change is not obvious. (4) With the increase of the span of the cavity, the Miss stress of the steel bar tends to increase continuously, and when the span changes from 3 m to 7 m, the increase at the pipe waist tends to be gentle. The increased speed at the bottom and the top of the pipe gets faster. When the span changes from 7 m to 8 m, the stress at the bottom and the waist of the pipe increases rapidly, but it is still far less than the tensile strength of the steel bar of 1570 MPa. (5) Because the main material of the PCCP pipe is concrete, according to the American AWWAC304 Standard Concrete stress-strain relationship, the concrete tensile damage dt is used to indicate the cracking degree of the concrete core. From the calculation results, it can be found that under the limit span of the ground subsidence damage, the damage value of the concrete core is 0, and there is no structural damage in the PCCP pipe.
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4 CONCLUSIONS In this paper, by establishing a nonlinear three-dimensional numerical model of PCCP pipesoil interaction under the action of traffic load, the influence of the cavity due to water seepage at the socket on the pavement surface and the in-service PCCP pipeline body under traffic load was explored. The research results show that the cavity has a certain impact on the surrounding soil and PCCP pipeline body, and the main characteristics are as follows: (1) The pavement stability decreases with the increase of the span of the cavity. The larger the span of the cavity, the faster the change rate of the pavement settlement value, and the more obvious the damaging trend. (2) The maximum settlement value of the pavement increases with the decrease in the burial depth of the cavity. The smaller the burial depth of the cavity is, the greater the probability of pavement damage is. (3) When the span of the cavity is small, the range of ground subsidence caused by the cavity is also relatively small. However, with the increase of the span of the elliptical cavity, the settlement change rate increases, and the settlement range increases significantly; under the same span, the settlement area of the pavement surface decreases continuously with the deepening of the thickness of the covering soil, and with the increase of the span of the cavity, the development and change of the settlement area is relatively slow; it has been verified that the displacement curve of the ground surface settlement caused by the cavity under the buried PCCP socket under the action of traffic load basically conforms to the modified Gaussian distribution law. (4) Due to the high strength of PCCP itself and the high design safety factor, the cavity generated at the socket will not cause major damage to the PCCP pipeline body, but with the continuous increase of the span of the cavity, it will further cause the subsidence of the soil body around the cavity. And the resulting secondary damage is the main potential threat to the safe operation of pipelines.
ACKNOWLEDGMENTS Thank you for the support of the following project funds: the Key Scientific Research Project of Shaanxi Provincial Department of Education (Coordination Centre Project) (Grant Nos. 22JY044); Program 2022TD-01 for Shaanxi Provincial Innovative Research Team and the Innovative Research Team of Institute of Water Resources and Hydro-electric Engineering, Xi’an University of Technology (Grant No. 2016ZZKT-14).
REFERENCES ANSI/A WWA C 304-99, Design of Prestressed Concrete Cylinder Pipe. USA: AWWA, 1999. Cao BQ, Liu ZQ, Jian CH, et al. (2021) High Precision Data Processing and Analysis of Land Subsidence Monitoring in Beijing. Journal of Navigation and Positioning, 9: 125–129 + 138. DOI: 10.16547/j.cnki.101096.20210619. Cheng G, Wang ZX, Zhu HH et al. (2022) Research Review of Rock and Soil Deformation Monitoring Based on Distributed Fiber Optical Sensing. Laser & Optoelectronics Progress, 59:51–70. Davies JP, Clarke BA, Whiter JT, et al. (2001) Factors Influencing the Structural Deterioration and Collapse of Rigid Sewer Pipes. Urban Water, 3: 73–89. Dai X. (2016) Model Test and Theory Study on the Urban Underground Engineering Hazards Induced by Loss of Groundwater and Sand. Tianjin University. https://kns.cnki.net/KCMS/detail/detail.aspx?dbname= CDFDLAST2017&filename=1017131106.nh. He YX. (2017) Study on Soil Erosion Around Underground Pipe Leakage. Zhejiang University, https://kns. cnki.net/KCMS/detail/detail.aspx?dbname=CDFDLAST2017&filename=1017270663.nh.
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Hou CQ, Dong MS, Feng HP. (2013) Research on Genesis and Mechanics of Land Collapse in Incompact Soil. Journal of Hefei University of Technology (Natural Science), 36: 63–67. Hu SW, Xue X, Sun YY, et al. (2018) Experimental Study on Mechanical Properties of BCCP Joint Under Foundation Settlement. Yangtze River, 49: 91–96. Ju YW, Wu JY, He WB, et al. (2015) Experimental Study and Numerical Analysis on Influence of Urban Underground Pipelines under the Ground Collapse. Journal of Taiyuan University of Technology, 46: 64– 68. DOI: 10.16355/j.cnki.issn1007-9432tyut.2015.01.013. Klar A, Vorster Te B, Soga K, et al. (2005) Soil-pipe Interaction Due to Tunneling: Comparison between Winkler and Elastic Continuum Solutions. Géotechnique, 55: 461–466. Li HZ, Feng X. (2020) Study on Mechanical Responses Induced by the Soil Voids on PCCP. Hydropower and Pumped Storage, 6: 77–83. Li M. (2016) “Caving in endlessly”, How Does the Safety Under the Sole Guarantee? Investigation of Urban Pavement Collapse. China Construction, 14–16. Ma ZY, Yan CF, Zheng T. (2013) Brief Analysis of the Development and Application of PCCP. Henan Water Resources and South-to-North Water Diversion, 92–93. Wang JH, Chen C, Zhang HP et al. (2018) Application of Prestressed Concrete Tube in Hydraulic Structure. Water Resources Development Research, 18:46–49 + 69. DOI: 10.13928/j.cnki.wrdr.2018.10.012. Wang FM, Fang HY, Li B, et al. (2018) Dynamic Response Analysis of Drainage Pipes with Gasketed Bell and Spigot Joints Subjected to Traffic Loads. Chinese Journal of Geotechnical Engineering, 40(12): 2274– 2280. Xia XY, Wang ZQ. (2020) Accuracy Test and Analysis of PS-InSAR and DS-InSAR Monitoring Urban Land Subsidence. Hydrographic Surveying and Charting, 40: 65–67 + 71. Yuan P. (2014) Study on Collapse Mechanism of SoilSubgrade of City Roads by Water Erosion. China University of Mining and Technology, https://kns.cnki.net/KCMS/detail/detail.aspx?dbname= CMFD201501&filename=1014073738.nh Zhao KJ, Fang HY, Zhang CB, et al. (2020) Mechanical Properties and Failure Mode Simulation Analysis of Joints in PCCP. China Water & Wastewater, 36:10–18. DOI: 10.19853/j.zgjsps.1000-4602.2020.20.002. Zhou M, Du YJ, Wang F, et al. (2016) Physical Modeling of Mechanical Responses of HDPE Pipes and Subsurface Settlement Caused by Land Subsidence. Chinese Journal of Geotechnical Engineering, 38: 253–262.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
The application of BIM technology in urban utility tunnel Chuanhua Xu* & Tiantian Ma* Hefei General Machinery Research Institute Co., Ltd., Hefei, Anhui, China
ABSTRACT: At this stage, the city’s rapid development has brought less and less available space, resulting in greater traffic pressure and fewer living resources. As an important infrastructure and “lifeline” to ensure urban operation, the development of an urban utility tunnel is conducive to solving the above problems. At present, the application of BIM technology in pipe gallery projects is relatively small. Unlike traditional pipe gallery projects, the construction of a pipe gallery based on BIM technology has greater advantages, especially in the design, construction, operation and maintenance stages of the project, which brings high application value to the construction of an urban underground integrated pipe gallery.
1 INTRODUCTION The urban utility tunnel will centralize various pipelines, such as water supply and drainage, electrical system, etc., in the underground tunnel space for unified design, planning, construction, and management to achieve the maximum development of underground space resources (Bai 2015; Lan 2022; Li 2019). Unlike the traditional direct burial method, the urban utility tunnel has avoided the problems of frequent excavation of pavement and repair difficulties, and the economic and social benefits have been significantly improved (Xu 2022). However, due to the complex layout of electrical equipment and pipelines in urban underground space, how to determine the exact location of the conflict has become an urgent problem. The development of Building Information Modeling (BIM) has become the key to solving the above problems. The integration technology of the two has become a strong support for the electromechanical system of urban utility tunnel in the planning, design, construction, operation and maintenance stages (Chen 2022; Huang 2021). BIM technology provides relevant equipment, pipelines, and other information for the electrical system of urban utility tunnels through a detailed three-dimensional space model (Yang 2021). When working cooperatively on the BIM platform, personnel of all types of work involved in the design work can obtain various data in a short time and at a low cost (Zhong 2021). In addition, during the construction process, BIM technology is fully used to conduct BIM layout for the pipeline engineering in the urban utility tunnel, which can solve the collision between the pipeline and the pipeline, the pipeline and the equipment, and the pipeline and the main body of the civil engineering in advance, and ensure the comprehensive utilization rate of the Utility tunnel structure. 2 SIGNIFICANCE OF THE DEVELOPMENT OF URBAN UTILITY TUNNEL The urban utility tunnel refers to the structures and auxiliary facilities built in urban underground space to accommodate two or more types of municipal public pipelines (Li 2022). *Corresponding Authors: [email protected] and [email protected]
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DOI: 10.1201/9781003425823-40
Generally speaking, urban roads are divided into two spaces: aboveground and underground. The aboveground is the main municipal road, and the underground is a municipal common tunnel. The municipal pipelines of power, communication, water supply and drainage, gas, and other disciplines are concentrated in different spaces within a structure, and unified planning, construction and management are implemented to achieve the comprehensive utilization of underground space and resource sharing. The development of an urban utility tunnel is conducive to eliminating the “zippered road” of the city, improving the city appearance, ensuring the need for smooth traffic and maintaining the safe operation order of the city. In addition, in order to promote the harmonious development of humans and nature and promote urbanization, it is an ideal and scientific mode to centralize the management of various pipelines in the way of an integrated pipe gallery, which is conducive to the trend of vertical development of urban space.
3 APPLICATION OF BIM TECHNOLOGY IN THE DESIGN STAGE OF AN URBAN UTILITY TUNNEL 3.1
Line selection
The urban utility tunnel is generally built on urban trunk roads with complex construction environment, major streets with high traffic pressure and irreplaceable, or old urban areas with old and complex facilities. In addition, due to the influence of buildings, river courses and other infrastructure, the linear trend and distribution of the comprehensive pipe gallery need to be reasonably planned and designed (Cai 2022). The introduction of “BIM + GIS” technology into the design of an integrated pipe gallery can effectively avoid unreasonable line planning, compare different construction schemes, and select the best scheme for construction. In the research of promoting the informatization construction of urban utility tunnel, the application of BIM technology and GIS technology in the construction, operation, maintenance and management process of a comprehensive pipe gallery can achieve twice the result with half the effort, as shown in Figure 1.
Figure 1.
3.2
Pipeline collision detection. Before collision optimization; After collision optimization.
Collaborative design
The internal space of the pipe gallery is narrow, including water supply and drainage, power, natural gas, heat, communication cables and other pipelines and lines. At the same time, in order to ensure the safe and secure operation of the pipe gallery, it is necessary to set up some auxiliary facilities systems such as fire protection, ventilation, power supply, lighting, monitoring and alarm, etc., inside the pipe gallery. In addition, different functional entrances and exits, such as personnel entrances and exits, escape exits, air inlets, air outlets, pipeline entrances and exits, and access openings, need to be set in each comprehensive pipe gallery cabin. However, due to the narrow internal space of the pipe gallery and the untimely
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communication between designers of different disciplines, it is easy to produce various design defects and cause pipeline collision. Therefore, when the BIM technology is used to carry out the collaborative design of the integrated pipe gallery project, the pipelines, facility systems and functional entrances and exits designed by various disciplines can be incorporated into the three-dimensional model of the pipe gallery at the same time. The design work of different disciplines can be carried out at the same time, and the design information of different disciplines can achieve real-time correlation, ensuring the interaction of information, improving the efficiency and quality of engineering design, avoiding pipeline collision, and saving costs. 3.3
3D visualization technology
3D visualization technology is the top priority of BIM applications. In the pipeline corridor operation and maintenance platform, 3D visualization is applied in the pipeline corridor 3D model display, analysis data overlay, and 3D-based early warning plan, as shown in Figure 2.
Figure 2.
3D visualization of BIM technology.
Through the 3D model display of the pipe gallery, a large number of complex data of the pipe gallery are expressed in the form of images, which is more intuitive and convenient for the staff to view and use these effective data, clear and easy to read, easy to operate and analyze, and can also provide an effective reference for the management personnel. The 3D visualization technology is used to superimpose various data of the pipe gallery, including asset data, operating status parameters of facilities and equipment, and statistical analysis results in the 3D model of the pipe gallery for the staff to view at any time and anywhere. Three-dimension-based early warning plans can ensure the seamless integration of the system’s three-dimensional geographic information scene and the real facility and equipment scene and realize the non-switching, non-jumping and smooth advance display from the large scene to the fine scene of the facility and equipment. The operator can locate, query and focus on the relevant information of the device in the 3D scene arbitrarily and quickly.
4 APPLICATION OF BIM TECHNOLOGY IN THE CONSTRUCTION STAGE OF AN URBAN UTILITY TUNNEL 4.1
Assembled pipe gallery
If BIM technology is adopted, it can be the same as prefabricated components. Each section of the pipe gallery can be numbered by an accurate BIM model to reduce the investment of workers and equipment, reduce the management difficulty of the construction site, reduce the construction period and realize the investment of reducing costs, and respond to the double-carbon policy promoted by the country. BIM technology can also plan the
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transportation route of prefabricated pipe racks, formulate detailed transportation plans, and maximize transportation efficiency. 4.2
Construction quality control
During the installation of the pipe rack and the electromechanical equipment in the pipe rack, due to the wide variety of electromechanical equipment, the complicated pipeline layout, and the difficulty of on-site management. It is difficult to grasp the actual situation, and it is impossible to make a response mechanism immediately. Applying BIM technology to the construction of a pipe gallery can effectively control the construction quality. Like other projects, the construction of an urban utility tunnel has a wide variety of pipelines and electromechanical facilities. The use of BIM technology can complete the collision detection of electromechanical equipment and pipelines before construction, avoid the waste of the construction period caused by reverse engineering, and improve construction efficiency. At the same time, in the BIM 3D model, holes and expansion joints can be preset in advance to prevent opening during formal construction and damage to the pipe gallery structure. As shown in Figure 3, prevent opening during formal construction and damage to the pipe gallery structure.
Figure 3.
4.3
Pipeline layout and hole reservation.
Construction progress and cost control
Based on the BIM three-dimensional pipe gallery model, the construction progress and cost are accurately controlled, and the plan, information and progress of the whole project are assigned to the three-dimensional model. Before the formal construction, we carry out drills in advance for various construction processes, which is conducive to finding out the possible risks and hidden dangers in the construction process in advance, and formulate corresponding countermeasures in time, so as to ensure that the construction progress does not deviate from the construction plan as much as possible, and complete the construction tasks with high quality and efficiency. In the cost control of the pipe gallery construction stage, various expenditures during the construction process can be centrally managed, the various construction procedures of the project can be accurately controlled, and the amount of manpower, materials and machinery involved in the project can be finely managed, so as to ensure the orderly progress of the pipe gallery project construction, reduce various unnecessary expenditures, visualize the flow of funds, and strictly monitor the project cost.
5 APPLICATION OF BIM TECHNOLOGY IN THE OPERATION AND MAINTENANCE STAGE OF THE URBAN UTILITY TUNNEL 5.1
Daily patrol management
In the daily work management process of the pipe gallery, the patrol inspection management is an important link, which is convenient for the operation and maintenance unit to actively 313
and timely find problems and eliminate potential hidden dangers in the operation process of the pipe gallery. Daily patrol management mainly includes patrol personnel management, patrol plan management, patrol job query, patrol personnel assessment, patrol calendar view, and other functions. According to the maintenance characteristics of the pipe gallery, the management personnel shall formulate the inspection plan and configure several routine inspection routes, so as to arrange personnel for each inspection plan. The staff in charge of patrol inspection shall carry mobile equipment to the site according to the planned route and carry out patrol inspection on the pipelines, cables and equipment in the comprehensive pipe gallery according to the patrol plan, and fill in the patrol inspection record through the mobile client and report the abnormal situation, as shown in Figure 4.
Figure 4. Flow chart of integrated pipe gallery operation and maintenance management system safety management.
5.2
Environmental monitoring and alarm
The environmental monitoring equipment in the integrated pipe rack transmits the real-time detected environmental data in the pipe rack, such as temperature, humidity, H2S, and sump water level, to the monitoring center through the communication link, and the central system stores this part of data. The central system analyzes the temperature, humidity and various gas indicators. When the indicators exceed the set threshold and pose a threat to the health of the staff, the alarm will be given in the center, and at the same time, the alarm will be given in the comprehensive pipe gallery. People are not allowed to enter the comprehensive pipe gallery. The system will start the fan equipment in the area for ventilation. After receiving the on-site event or alarm, the GIS + BIM platform will automatically display the alarm location. The location information can be obtained by mobile terminal positioning and manual input.
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6 CONCLUSIONS At present, the construction of an urban utility tunnel is imminent, and the problem of lack of engineering information can be effectively solved by using BIM technology. Through the BIM technology, after the establishment of an accurate BIM model, the information of various facilities in the pipe gallery will be stored, including the data information of the analysis project, to provide support for the management and construction of the urban utility tunnel project. As an important part of the urban construction project, the urban utility tunnel costs a lot of money. The application of BIM technology can monitor the direction of the fund of the pipe gallery project and reduce the economic disputes of the construction units of the pipe gallery. At the same time, it can also be applied to the disease monitoring and later operation and maintenance stage in the pipe gallery after the completion of the project construction. To sum up, the application of BIM technology to urban utility tunnel projects has high application value.
REFERENCES Bai H. L. Research on the Development Trend of Urban Utility Tunnel [J], Chinese Municipal Engineering, 2015,(6):78–81. Cai L. Y., Zhong C. S., Yi L. Y., et al. Research on the Application of BIM Technology in Green Construction of Urban Comprehensive Pipe Gallery [J]. Juye, 2022 (10): 25–27. Chen Q. S., Li C., Ao J., et al. Application of BIM technology in the Rongdong Integrated Pipe Gallery Project in Xiong’an New Area [J]. Construction Technology (Chinese and English), 2022,51 (07): 74–79. Huang Y. T. Research on Cost Control of Underground Integrated Pipe Gallery Project Based on BIM Technology [J]. Project Cost Management, 2021 (06): 69–73. Lan N. B. Application of BIM Technology in the Construction Phase of Integrated Pipe Gallery [J]. Engineering Construction and Design, 2022 (15): 197–200. Li M. H., Zhang W. Y. Research Status and Construction Method of Urban Utility Tunnel [J]. Infrastructure Management Optimization, 2019,31 (3): 43–45. Li J. F., Li Z. F., Ma X. S. Integrated Pipe Gallery BIM Collaborative Design Mode and Technology Application [J]. Municipal Technology, 2022,40 (09): 220–228. Xu Y.P. Research on Collaborative Management of Intelligent Integrated Pipe Gallery Based on BIM [J]. Building Economics, 2022,43 (S1): 534–539. Yang Z. Y. Research on the Integrated Supervision Platform of Smart Pipe Gallery Based on BIM + GIS + loT [J]. Smart City, 2021,7 (18): 106–108. Zhong W., Li Z. Y., Wan Z. D. Application of Interactive Design and Collaborative Management of Integrated Pipe Gallery based on BIM [J]. China Water Supply and Drainage, 2021,37 (12): 104–108.
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Experimental study on indoor model of water vapor humidification remodeling unsaturated loess under negative pressure conditions Yan-feng Zhang* & Qiong Xia* School of Civil Engineering, Lanzhou Jiaotong University, China
ABSTRACT: In response to the problems of long humidification time, large workload, and poor uniformity in the traditional pre-soaking method of humidifying unsaturated loess, an indoor model study is proposed to cover the surface layer of loess with a sealing film, apply negative pressure and humidify the unsaturated loess with water vapor through an open column at the bottom. The results show that the water vapor humidification process is influenced by three main factors: “pressure difference”, “temperature gradient”, and “moisture content gradient”. The farther the distance of water vapor diffusion is, the greater the vapor pressure dissipation is, and the smaller the soil’s temperature gradient and moisture content gradient are at the same distance. At the end of the test, the water content of the soil near the humidification column in the horizontal direction under negative pressure is close to the optimal water content, and the temperature is close to the vapor temperature, and the farther away from the humidification column, the lower the temperature and water content of the soil. In the vertical direction, upward water vapor transport intensifies under the effect of the negative pressure difference between layers, and the water content of the top layer at the end of the test is significantly higher than that under normal pressure conditions. The water content of the bottom layer was slightly reduced. The moisture distribution in the soil was more uniform after the negative pressure humidification than under the normal pressure condition. The temperature of the top layer at the end of the test was significantly lower than that under normal pressure, and the temperature difference of each measurement point of the bottom layer was significantly increased.
1 INTRODUCTION Natural loess is a porous medium of solid, liquid, and gas phases with developed pores and significant vertical joints. Due to the strong water sensitivity of loess, it is easy to cause uneven settlement of loess foundations, house cracking, and other engineering diseases when wetted by water (Gu et al. 2016; Sun et al. 2022; Wang et al. 2018). To eliminate the wetness of loess, the strong ramming method is often used to treat loess foundations in practical engineering. However, the treatment effect of the strong ramming method is closely related to the water content of the soil (Xu et al. 2022), and the water content of natural loess is low. The strong ramming method cannot achieve a good reinforcement effect. And there are problems such as long humidification time, poor humidification uniformity, and difficulty in controlling the range using the traditional pre-soaked water method to humidify unsaturated loess (An et al. 2017; Yang et al. 2022). Some scholars proposed a new method to humidify unsaturated loess using water vapor (Wang & Lu 2004; Wang et al. 2005). The process of water vapor humidification of unsaturated loess is influenced by various factors such as temperature, pressure, and gravity, etc. Kondo et al. (1992) conducted the first *Corresponding Authors: [email protected] and [email protected]
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DOI: 10.1201/9781003425823-41
study on hydrothermal transport, and Chu and Mariño (2015) proposed a layered soil infiltration model that integrates liquid and gas phase migration. Wang and Su (2010) found that temperature gradients would promote water vapor transport to distant areas through an indoor experimental study of unsaturated loess under different initial conditions. Thomas and Rees (2009) found that temperature transfer and water transport interacted during water transport in soils. Through indoor large-size model tests, Zhang et al. (2015) found that soil density, water content, temperature, and migration time significantly affected gaseous water migration. Lin (2020) conducted a study on water transport in unsaturated loess under the effect of high temperature. They found that the water migration under high-temperature conditions was significantly higher than in a normal-temperature environment. Li et al. (2020), Li et al. (2021), and Li et al. (2022) conducted an indoor experimental study of water vapor humidification of unsaturated loess and found that heat transfer due to vapor and temperature gradient, coupled with water vapor and liquid water migration driven by a pressure gradient, moisture content gradient, and temperature gradient, coexist when water vapor humidifies unsaturated loess. In this paper, based on the existing research on water and gas transport and migration within unsaturated loess, we propose an indoor model experimental study of water vapor humidification through an open-hole humidification column at the bottom under negative pressure conditions to provide a new method and a new idea for water vapor humidification of unsaturated loess. The experimental results of water vapor humidification of unsaturated loess under normal and negative pressure conditions are compared and analyzed to investigate the effect of negative pressure on the temperature and water vapor transport in the soil. The study’s results can provide theoretical support for the new technology of water vapor humidification of unsaturated loess.
2 MODEL TEST OVERVIEW 2.1
Test equipment
The test equipment mainly has a water vapor generator (output maximum water vapor pressure 250 KPa), regulator, bottom opening humidification column, field-shaped negative pressure pipe, pressure stabilization tank, electric contact vacuum meter, electromagnetic relay, water ring vacuum pump, MIKP-300 type air pressure sensor (range 100 KPa +100 KPa, 0.25 level precision), temperature and humidity sensor, and temperature sensor composition. The open bottom column is 80 cm long and 5 cm in diameter, with 6 j1 cm water vapor diffusion holes in each layer, hexagonally symmetrical distribution, each layer interval 5 cm, total 2 layers, as shown in Figure 1. When filling the humidification column at the bottom of the opening with fine wire mesh wrapped, without affecting the water vapor diffusion pressure at the same time, it prevents soil particles from coming into the water vapor diffusion hole. The field-shaped negative pressure pipe is made of 12 PVC pipes 30 cm in length and 3.2 cm in height, with holes opened every 5 cm in the left, right and bottom directions, and the surface is wrapped with
Figure 1.
Bottom open-hole humidification column.
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coarse cloth, which relates to the pressure stabilization tank and vacuum pump through a steel wire hose to form a negative pressure system. The model box is j100 cm and 100 cm high, and the overall installation diagram is shown in Figure 2.
Figure 2.
2.2
Schematic diagrams of the test setup.
Test soil sample
The test soil is Lanzhou loess with developed pores and yellowish-brown color. Its basic physical property index is shown in Table 1. Table 1.
Basic physical property indexes of soil samples.
Initial moisture content/%
Optimum moisture Maximum dry dencontent/% sity/(gcm 3)
Liquid limit/%
plastic limit/%
Relative density of particles
7.01
13.89
27.1
17.3
2.71
2.3
1.78
Experimental protocol
The indoor model soil was filled in 6 layers with a compaction degree of 0.8, of which the thickness of the upper and lower layers was 20 cm, and the thickness of the middle 4 layers was 15 cm. The weight of the required fill was calculated according to the volume of each layer and compacted to the boundary between the layers. The test is designed for two groups of working conditions: 1 group of soil surface layer with an atmospheric pressure of 0 KPa and water vapor of pressure 50 KPa to humidify the unsaturated loess through the open-hole humidification column at the bottom; 2 groups of soil surface layer are covered with two layers of geotextile and sealing membrane and vacuumed through the buried negative pressure pipe. The surface layer pumping and the bottom layer water vapor humidification process are carried out simultaneously to maintain the negative pressure of the upper layer at 10 KPa and water vapor of pressure 50 KPa. Humidifying and remodeling unsaturated loess are through the bottom open-hole humidifying column. The water-thermal-gas transport law and humidification effect of water vapor humidification and remodeling of unsaturated loess under negative pressure conditions were compared and analyzed. 2.4
Sensor arrangement
In this test, there are 3 layers of sensors, buried at a depth of 20 cm, 50 cm, and 80 cm, respectively, in which temperature sensors are buried in 3 directions in each layer, with an angle interval of 120 between adjacent directions. 4 temperature sensors are buried in each direction,
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at 10 cm, 20 cm, 30 cm, and 40 cm from the center. 3 temperature and humidity sensors are buried in each layer, at 15 cm, 30 cm, and 45 cm from the center. The air pressure sensor is buried 1 in each layer, 25 cm away from the center, mainly monitoring the negative pressure transmission in the remodeled loess. The sensor measurement points are shown in Figure 3.
Figure 3.
Test sensor arrangement diagram.
3 ANALYSIS OF EXPERIMENTAL RESULTS 3.1
Negative pressure transfer analysis
Figure 4 shows the air pressure-time curves in the soil at each measurement point under negative pressure conditions at depths of 20 cm, 50 cm, and 80 cm (0 KPa is normal atmospheric pressure). At the beginning of the test, there were obvious changes in the values of the barometric pressure sensor at different depths. The deeper the barometric pressure sensor, the smaller the negative pressure value. This indicates that the negative pressure can be quickly transferred in the remodeled unsaturated loess. Still, the transfer process is hindered by soil particles, and the deeper the soil is, the more difficult it is for the gas to be quickly pumped out.
Figure 4.
Pressure variation with time at different depths.
The air pressure value of the 80 cm depth measurement point reached a stable state after rising rapidly during 50–80 min and 170–230 min, which was analyzed as follows: the 80 cm depth measurement point was close to the bottom water vapor diffusion hole, and the water vapor diffused to the measurement point at 50 min, and the air pressure value of the measurement point rose rapidly under the action of water vapor diffusion pressure. As the test proceeds, the water content of the soil near the measurement point increases after 170 min. The water fills the pores between the soil particles, which hinders the transfer of negative pressure, and the air pressure value of the measurement point rises rapidly again. After the moisture content near the measurement point reached stability, the air pressure value of the measurement points also reached a stable state. 319
The air pressure value of the measurement point at 50 cm depth continued to decline slowly under the effect of vacuum pumping until 200 min when the water vapor diffused to the measurement point, the air pressure value of the measurement point began to rise with the surrounding soil moisture content and reached a stable state after 320 min. 20 cm depth measurement point air pressure value was initially affected by the double influence of the diffusion of water vapor from the bottom layer and the increase of gas pressure caused by the increase of soil temperature, and the air pressure value rose slightly after the increase of the temperature of the soil body. After the water content of the soil and the temperature are more stable, the air pressure value decreases to 10 KPa and remains stable. 3.2
Analysis of water separation field results
The water vapor humidification process soil moisture content and temperature monitoring values were taken as the average value of the measurement points at the same distance from the open humidification column at the bottom for statistical analysis. Figure 5 shows the water content variation with time for each measurement point at 80 cm depth under normal and negative pressure conditions of the overburdened soil on the model test. From Figure 5 (a), the moisture content of the soil at the measurement point of 15 cm rises rapidly after the start of water vapor humidification, and the moisture content reaches a peak of 13.15% at 30 min, after which the moisture content decreases rapidly and remains stable at 12.66%. The moisture content of the soil at the measurement point of 45 cm was slowly increasing since 180 min and reached 9.18% at the end of humidification.
Figure 5.
Variation of water content with time at each measurement point in 80 cm layer.
Analysis of the above results shows that the water vapor humidification range is small at the beginning of the test. Under the initial water vapor pressure and temperature, the moisture content of the soil rises faster the closer to the water vapor diffusion hole, which leads. The accumulation of water at the measurement point at 15 cm is greater than the amount of outward diffusion, and the peak moisture content appears. As the test proceeds, the water vapor diffusion surface further increases. The accumulated water in the soil near 15 cm rapidly diffuses outward under a large water content gradient and temperature and pressure gradient inside and outside. The water content decreases and reaches dynamic equilibrium (the same is true for the peak water content at 30 cm). Due to the obstruction of soil particles, the farther away from the water vapor diffusion hole, the smaller the diffusion pressure and the lower the temperature. Therefore, the farther away from the humidification column, the smaller the temperature gradient and pressure gradient at a certain distance, which results in the phenomenon that the moisture content of the soil rises more slowly at a farther distance from the humidification column, and the moisture content is lower when it reaches dynamic equilibrium. From Figure 5 (b), after the water vapor humidification test under negative pressure, the water content at 15 cm, 30 cm, and 45 cm measurement points rose similarly to that under normal pressure, and the water content at the end of the test was 11.80%, 11.15%, and 320
8.83%, respectively. Compared with the normal pressure test results, the growth and dynamic stability value slope is slightly reduced, and the water content-time curve does not appear to the peak. This is because a negative pressure difference is generated between the soil layers after the negative pressure is applied to the overlying layer. The negative pressure difference between the layers in the diffusion process affects the water vapor. The diffusion tendency to the upper layer is strengthened, affecting the horizontal diffusion rate and the final water content. From Figure 6 (a), the water content at the end of the test was 9.62%, 9.07%, and 7.84% for each measurement point at the burial depth of 50 cm from the humidification column of 15 cm, 30 cm, and 45 cm, respectively. Compared with the water content – time curve of each measurement points at 80 cm depth, the initial water content change point is later, the humidification time is extended, the final state water content is significantly reduced, and the water content peak does not appear at the 30 cm measurement point. The reasons for this are: On the one hand, the upward transport of water vapor is not only hindered by soil particles but also by gravitational potential, which is exacerbated by the adsorption of water vapor into liquid water by soil particles. On the other hand, is the depth of a 50 cm layer, each measurement point is farther from the bottom diffusion hole, the water vapor pressure dissipation is greater, and the temperature transfer is also slower, which causes a certain distance temperature gradient and pressure gradient the lower, further affecting the layer of soil moisture content growth rate and the final moisture content.
Figure 6.
Variation of water content with time at each measurement point in the 50 cm layer.
From Figure 6 (b), each measurement point’s final state moisture content at the burial depth of 50 cm from the humidification column of 15 cm, 30 cm, and 45 cm under negative pressure conditions are 9.86%, 9.67%, and 7.31%, respectively. Compared with the same location under normal pressure conditions, the water content of each measurement point up value, the growth rate was 9.19%, 29.12%, 63.85%. The analysis of the reason is that the negative pressure difference between layers prompted the bottom layer of water vapor upward transport, the same layer of water vapor diffusion effect is weakened, the farther away from the diffusion hole, the more serious water vapor transport in the soil by the negative pressure between layers. This resulted in a 50 cm depth soil body in the bottom layer of water transport upward trend intensified, 15 cm, 30 cm measurement points at the water content than the normal pressure situation, the water content at 45 cm measurement points significantly decreased. Because of the smooth wall of the humidification column, there are more pores on the contact surface with soil particles, the negative pressure is easier to be transmitted along this contact surface, and the resistance of high temperature and highpressure water vapor along this contact surface is also smaller, so the closer the distance to the humidification column, the easier the upward movement of water vapor, which makes the upward movement of water at the 15 cm measurement point greater than that at 30 cm, and the final value of water content of soil at the two places is closer. Figure 7 (a) shows that the buried depth of 20 cm from the humidification column is 15 cm, 30 cm, and 45 cm at each measurement point at the end of the test moisture content of 8.0%, 321
Figure 7.
Variation of water content with time at each measurement point in the 20 cm layer.
7.25%, and 6.93%, respectively. The initial moisture content change point is significantly delayed, and the moisture content rise rate is significantly reduced. The reasons for this are: On the one hand, the soil at 20 cm depth is farther away from the diffusion hole of the bottom layer, and the diffusion of water vapor to the measurement points of this layer is more affected by the obstruction of soil particles, and more gravitational potential energy needs to be overcome, resulting in less water transported to this layer. On the other hand, the soil in this layer is closer to the surface layer. During the test, the temperature of the soil increases, and part of the water in the soil will be evaporated and diffused out, which is the main reason why the water content at the end of the test at the measurement point of 45 cm is less than the initial water content. From Figure 7 (b), the water content at the end of the test at each measurement point was 9.88%, 8.48%, and 7.35% at the burial depth of 50 cm from the humidification column of 15 cm, 30 cm, and 45 cm under the negative pressure condition, respectively. Compared with the rising value of moisture content of each measurement point at the same location under normal pressure conditions, the growth rates were 23.5%, 16.9%, and 6.06%, respectively. Each measurement point’s initial water content change point advanced significantly, and the growth slope increased significantly. This is because the water vapor in the bottom layer and the liquid water adsorbed by the soil particles are transported to the 20 cm layer in large quantities under the negative pressure difference between layers. The moisture content of the soil body is higher the closer to the humidification column. The moisture content gradient of the 20 cm depth layer increases, and under the effect of a larger moisture content gradient, the trend of water transport in the horizontal direction to the distant area increases, which increases the moisture content at the 45 cm measurement point increase. On the other hand, the sealing film covered by the surface layer under the negative pressure condition also weakened the evaporation of water in the upper soil, contributing to the humidification effect. In general, the water vapor at atmospheric pressure and negative pressure conditions remodeled the unsaturated loess through the open column at the bottom, and the water content of the soil increased in a similar pattern in the horizontal direction. The further away from the column, the lower the final water content. In the vertical direction, the water vapor and pore water in the soil under normal pressure are affected by the obstruction of soil particles and gravitational potential, and the water content of the soil decreases from the bottom to the top. Under the negative pressure condition, the soil’s water vapor and pore water are influenced by the negative pressure difference between the layers. The upward migration increases, which increases the water content of the top layer of the soil increase significantly. The test results show that water vapor can quickly achieve a good humidification effect by humidifying and remodeling unsaturated loess through the open-hole humidification column at the bottom. The final moisture content of different depth measurement points 30 cm from the humidification column under normal and negative pressure conditions were compared, as shown in Table 2. The water content of the top layer under negative pressure conditions is significantly higher than that under normal pressure conditions, and the reshaping of the unsaturated loess foundation has better humidification uniformity and can achieve a better humidification effect.
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Table 2. The final moisture content at 30 cm from the humidification column under different conditions. Depth (cm)
Normal pressure water content (%)
Negative pressure water content (%)
20 50 80
7.25 9.07 11.85
8.48 9.67 11.15
3.3
The difference in moisture content (%) 1.23 0.60 0.70
Lift rate (%) 16.97 6.62 5.91
Analysis of temperature field results
Figure 8 (a) is the temperature and time curve of each measurement point under atmospheric conditions at a burial depth of 80 cm, from which it can be seen that the temperature at the measurement points of 10 cm, 20 cm, and 30 cm from the humidification column can reach the same stable value of 96.15 C. The temperature at the end of the test at the measurement point of 40 cm from the humidification column is lower at 84.16 C. The closer the measurement points are to the water vapor diffusion hole, the earlier the initial temperature change point and the faster the temperature growth rate. The analysis of the reasons for this is: The temperature increase of the soil body is the result of the transfer of water vapor in the soil body. The initial water vapor in the high temperature and high pressure is under the action of rapid diffusion in the surrounding soil body. The surrounding soil temperature and moisture content in a short period rise and rapidly reach a stable state. However, because soil particles and adsorption hinder the diffusion process of water vapor, the farther the diffusion distance, the larger the diffusion surface. The more serious the pressure dissipation and temperature reduction phenomenon, the more distant the diffusion hole at the measurement point and the temperature growth rate is slower. Combined with the analysis in Figure 5 (a), the water content of the 80 cm layer of remodeled unsaturated loess is increased significantly, and the pores of soil particles are filled with liquid water converted from adsorbed water vapor. The heat transfer coefficient of the soil layer is increased, which promotes temperature transfer within the soil. Hence, the temperature stability values at the measurement points of 10 cm, 20 cm, and 30 cm are close to the same. Figure 8 (b) can be seen under negative pressure conditions buried depth of 80 cm, from the humidification column 10 cm, 20 cm, 30 cm, and 40 cm temperature stability values at each measurement point were 99.75 C, 97.3 C, 95 C, 89.7 C, compared with the same location under normal pressure conditions monitoring point data, the temperature growth law has no major changes, but the temperature stability value of each measurement point differences are obvious. The reasons for the analysis are, on the one hand, the upward transport trend of water vapor under the effect of the negative pressure difference between layers is strengthened, the horizontal direction of water vapor transport is reduced, the filling of pore water in the soil is reduced, and the temperature heat transfer coefficient is reduced.
Figure 8.
Temperature variation of each measurement point in 80 cm layer with time.
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On the other hand, the negative pressure transfer within the soil constantly pumps gas to the outside. The high-temperature gas is constantly pumped out, affecting each measurement point’s final temperature stability value. As can be seen from Figure 9 (a), the buried depth of 50 cm under normal pressure conditions, 10 cm, 20 cm, 30 cm, and 40 cm from the humidification column at the end of the test moisture content of each measurement point difference is obvious, respectively, 95.0 C, 90.32 C, 69.83 C, 36.7 C. Compared with the 80 cm depth and the same distance measurement point, the temperature rise values were reduced by 1.15 C, 5.83 C, 26.23 C, 47.46 C, and each measurement point temperature initial change point delayed. The growth rate has also been reduced. This is because the measurement points at 50 cm depth are farther away from the water vapor diffusion hole, and the time required for water vapor diffusion to the layer is longer. The farther away from the humidification column, the less water vapor diffusion, the lower the heat transfer coefficient of the soil, and the slower the temperature growth rate, so at the end of the test, the farther away from the humidification column, the lower the temperature of the soil.
Figure 9.
Temperature variation of each measurement point in 50 cm layer with time.
As seen from Figure 9 (b), the initial change point of temperature at the measurement points of 10 cm, 20 cm, and 30 cm from the humidification column under the negative pressure condition is significantly earlier than that under the normal pressure condition. The growth rate is significantly higher, and the stable values of temperature at the final measurement points of 10 cm and 20 cm are close to each other. The temperature-time curve at the 40 cm measurement point from the humidification column has no significant change compared with Figure 9 (a). The reason for this is that: Under the effect of the negative pressure difference between layers, the closer to the humidification column, the greater the upward transport of water and gas. This causes the 50 cm layer closer to the humidification column, soil moisture content elevation increases, and the soil temperature heat transfer coefficient also increases accordingly. Therefore, the temperature of 10 cm and 20 cm are equal, and the temperature of 30 cm increases. 40 cm measurement point is farther away from the humidification column, and the temperature value is increased at the end of the test but not obvious. From Figure 10 (a), the temperature of each measurement points at the buried depth of 20 cm layer under normal pressure conditions at the end of the test decreases further than the temperature of the measurement point at a depth of 50 cm layer and the rate of rise also slows down significantly. Combined with Figure 10 (a), the analysis of the reasons for this is that the transfer of water vapor and liquid water at each measurement point in the 20 cm layer is small, the pores between soil particles are mostly filled by air, the heat transfer coefficient of the soil is small, and the temperature rises more slowly. Hence, the temperature of each measurement point at the end of the test is significantly lower than that of the bottom layer. And all of them did not reach a steady state. From Figure 10 (b), the temperature-time curve of each measurement point in the buried depth 20 cm layer under the negative pressure condition has the characteristics of slowly
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Figure 10.
Temperature variation of each measurement point in 20 cm layer with time.
decreasing first and then slowly increasing. In comparison with Figure 10 (a), it was found that the temperature at the end of the test decreased slightly except for the temperature at the measurement point 10 cm away from the humidification column, and the temperature at all other measurement points decreased slightly. Combined with Figure 7 (b) analysis of the reasons for pumping negative pressure is a process of airflow, the test process of heat in the soil will be constantly pumped out with the air. The initial test is the depth of a 20 cm layer of soil by the diffusion of heat from the humidification column and the dual impact of negative pressure pumping. 10 cm measurement point is closer to the humidification column. The influence of the humidification column transfer temperature is greater. The temperature rises slowly. The rest of the measurement points are greater by the influence of negative pressure pumping. The temperature slowly decreases as the test proceeds. As the test proceeds, a large amount of moisture in the bottom layer migrates to the top under the negative pressure condition. The pores between soil particles are filled with moisture, increasing the soil’s heat transfer coefficient. The influence of negative pressure is pumping to take away the heat in the soil decreases, and the temperature of each measurement point in the 20 cm layer starts to rise more quickly. By analyzing the “water content time curve” and “temperature-time curve” under atmospheric and negative pressure conditions, the temperature is closely related to water content. The higher the water content of the measuring point, the greater the heat transfer coefficient of the surrounding soil, the faster the temperature growth rate during humidification, and the higher the temperature at the end of the test.
4 CONCLUSIONS (1) The negative pressure transfer effect in the remodeling of unsaturated loess foundation decreases along the depth direction. The higher the soil moisture content during the test, the more pore water between soil particles, and the worse the negative pressure transfer effect. (2) The horizontal direction of water vapor humidifying unsaturated loess is mainly affected by water vapor pressure and temperature difference. Because the diffusion process is accompanied by pressure dissipation and temperature transfer, the farther away from the water vapor diffusion hole, the lower the water vapor pressure, the smaller the temperature gradient in the soil, and the lower the final water content. Under normal pressure in the vertical direction, the water content in soil is affected by gravity potential, and the final water content decreases from bottom to top. Under the negative pressure condition, the bottom layer moisture is affected by the negative pressure difference between layers, and the upward migration trend is strengthened. After the test, the moisture content of the upper layer of soil is significantly higher than that under the normal pressure condition, and the uniformity of the moisture content of the soil is better.
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(3) Water vapor humidification process soil temperature and water content are closely related. The higher the water content of the soil, the more pore water between soil particles, the greater the heat transfer coefficient of the soil, and thus the faster the temperature of the soil rises. At the end of the test, except for the higher water content measurement point temperature close to the water vapor temperature, the farther the soil is from the humidification column in the horizontal direction, the lower the temperature, and the vertical direction temperature decreases from the bottom to the top in turn. (4) Vacuum pump pumping negative pressure process will have heat with the air pumped out, the upper layer of the soil is affected by this more obvious, and the end of the test temperature of each measurement point is significantly lower than under normal pressure conditions. The subsoil is less affected, but the temperature gradient at each measurement point is more obvious.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Construction technology of retaining structures for deep foundation pit with H-shaped composite steel anchor piles-precast slab walls by compaction grouting method Luheng Gao* School of Civil Engineering, Jiangsu College of Engineering and Technology, Jiangsu Province, China
ABSTRACT: A compaction grouting method-based construction technology of retaining structures for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls was revealed in this study. The construction process is as follows: a. Surveying, setting out, and pile positioning; b. Pile hanging in place and calibration; c. H-shaped steel pile driving; d. Excavation by layers and sections, and treatment of the pile body; e. Anchor rod driving from top to bottom in a way of “layer by layer, segment by segment, and rod by rod”; f. Hoisting of the precast slab wall in place; g. Insertion and binding of reinforcing bars firmly with the protruding reinforcing bars of the precast slab wall; h. “Layered and segmented” high-pressure grouting construction; j. Joint micro-expansion concrete pouring; k. Construction Step d to Step h are repeated; m. Reinforcing bar binding with the waist beam, formwork support, and concrete pouring; n. Acceptance inspection. In the technical scheme, an integral retaining structure for the foundation pit was formed using H-shaped steel piles, anchor rods and precast slab walls through a reasonable process. Then, the prefabricated construction of the retaining structure system for the foundation pit was implemented based on the principle of “excavation by layers and sections, anchoring layer by layer, section by section, and rod by rod, and grouting immediately after site hoisting”, which was safe and reliable and could effectively save the construction period, accompanied by the dual functions of soil retaining and water stopping.
1 BACKGROUND In recent years, the requirements of the whole society for energy conservation and environmental protection have been continuously elevated, and the development tide of prefabricated concrete structures has been formed due to the aging and shortage of migrant workers. The central government has put forward the goal of “striving to make the proportion of prefabricated buildings in new buildings reach 30% in about 10 years”, which provides a good opportunity for the large-scale development of assembled concrete buildings, with huge market demands and broad development prospects (Gao 2019). Prefabricated concrete buildings represent a revolution in the construction mode, and strengthening the research on key technologies for efficient construction is of decisive significance to promoting the development of prefabricated buildings. A key technology system for the efficient construction of assembled concrete buildings suitable for China’s national conditions should be established. On this basis, advanced technologies can be achieved through continuous improvement in consideration of the construction cost and actual industrialized production capacity, so as to realize the transformation and upgrading of the *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-42
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construction mode of assembled concrete buildings and comprehensively promote the industrialization of buildings (Gao 2019). Prefabricated buildings, as the key development mode emphasized in the “Thirteenth Five-Year Plan” and “Fourteenth Five-Year Plan” in China, have proposed greater challenges to China’s construction industry (Feng et al. 2013). The design and development of new technologies have become an important task to improve the comprehensive level of prefabricated construction in China (Yang et al. 2018). The research and application of the new technology for the construction of prefabricated underground retaining structures for deep foundation pits have pointed out the research direction for promoting the modernization and efficient development of China’s construction industry. In this study, “a retaining structure for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls based on the compaction grouting method” (Figure 1) was publicly designed, which was effectively combined with underground structural construction, realizing prefabricated construction of the foundation pit retaining structure and laying a certain theoretical research basis for the transformation and upgrading of China’s construction industry.
Figure 1.
Finite element model for prefabricated construction of the deep foundation pit.
2 ENGINEERING PROJECT APPLICATION: A CONSTRUCTION METHOD OF THE RETAINING STRUCTURE FOR DEEP FOUNDATION PITS WITH HSHAPED COMPOSITE STEEL ANCHOR PILES- PRECAST SLAB WALLS The project covers a total construction area of 45, 959.74 m2, including 4662 m2 underground construction area and 41, 297.74 m2 above ground construction area, including an 18-storey hotel tower with a structural height of 71.750 m and 24 floors above the ground commercial tower. The height of the building structure is 88.150 m, the form of the building structure is a frame shear wall structure, and the form of the foundation is a pile raft foundation. The thickness of the raft in the basement of the upper main part is 600 mm unless indicated, the thickness of the raft in the lower foundation of the commercial tower is 1200 mm, and the thickness of the raft in other parts of the structure is 1000 mm. The engineering design adopted the construction technology of the retaining structure for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls based on the compaction grouting method following the principle of “excavation by layers and sections, anchoring layer by layer, section by section, and rod by rod, and grouting immediately after site hoisting”. This technology, which realizes the prefabricated construction of the retaining structural system for foundation pits, is safe and reliable and can effectively save the construction period, along with the dual functions of soil retaining and water stopping. 2.1
Design of the retaining structure for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls
The retaining structure for deep foundation pits with H-shaped composite steel anchor pilesprecast slab walls includes foundation pit soil, H-shaped steel piles, precast slab walls, 328
high-pressure grouting holes, soil-permeable slurry, and anchor rods. A number of H-shaped steel piles were respectively hoisted and evenly placed on the surface of foundation pit soil, precast laminated slab walls were arranged between every two adjacent H-shaped steel piles, and anchor rods were set on the flange plate surface of H-shaped steel piles at one side of foundation pit soil to tie and fix the H-beam piles and foundation pit soil. There were highpressure grouting holes at the bottom of precast slab walls, and the soil-permeable slurry was formed, through high-pressure mechanical grouting, in the high-pressure grouting holes between foundation pit soil and H-shaped steel piles and between foundation pit soil and precast slab walls. Several anchor rods effectively tied H-shaped steel piles with foundation pit soil in a firm and reliable fashion, thus realizing the assembly construction of the foundation pit retaining structural system, which could effectively save the construction period with the dual functions of retaining soil and water. The retaining structure for deep foundation pits with H-shaped composite anchor piles-precast slab walls is shown in Figure 2, in which 1 represents foundation pit soil, 2 represents H-shaped anchor piles, 3 represents precast slab walls, 4 represents high-pressure grouting holes, 5 represents soil-permeable slurry, and 6 represents anchor rods.
Figure 2. Schematic diagram of the retaining structure for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls through compaction grouting method.
1. The flange plate of the H-shaped steel pile 2 was 350 mm in width and 20 mm in thickness. The height of the web plate of the H-shaped steel pile 2 was 60 mm greater than that of the precast slab wall 3, accompanied by a thickness of 30 mm. 2. The thickness of the precast slab wall 3 was 100 mm, and the width of each section of the precast slab wall 3 was smaller than the spacing of web plates between two adjacent H-shaped steel piles 2 (0.5 the width of flange plates of the H-shaped steel pile 2 + 50 mm). 3. The high-pressure grouting hole 4, which was a steel sleeve structure, was 18–22 mm in diameter and fabricated in an integrated way with the precast slab wall 3. In addition, the high-pressure grouting hole 4 was arranged at 300 mm above the bottom of each section of the precast slab wall 3. 4. The spacing between two adjacent high-pressure grouting holes 4 was 1.2 m, and at least two high-pressure grouting holes 4 were arranged at the bottom of each section of the precast slab wall 3. 5. The anchor rod 6, which was 12 mm in diameter, was arranged in the middle of each side of the flange plate of the H-shaped steel pile 2. And several anchor rods 6 were evenly arranged along the length direction of the H-shaped steel pile 2 with a spacing of 1.5 m. 6. Anchor rods 6 were densified within 2000 mm of the pile body at the joint between the upper and lower parts of each section of the precast slab wall 3, and the spacing of such anchor roads was 0.6 m in the densified section. 7. The upper and lower joints of each section of the precast slab wall 3 were connected by concrete post-pouring, and the width of the concrete post-pouring section was 2.5 times the thickness of the precast slab wall 3.
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2.2
Construction process
The following steps were included in the compaction grouting method-based construction process of the retaining structure for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls: Surveying, setting out, and pile positioning ! pile hanging in place and calibration Hshaped steel pile driving ! excavation by layers and sections, and treatment of the pile body ! anchor rod driving from top to bottom in a way of “layer by layer, segment by segment, and rod by rod”; hoisting of the precast slab wall in place ! insertion and binding of reinforcing bars firmly with the protruding reinforcing bars of the precast slab wall ! “layered and segmented” high-pressure grouting construction ! joint micro-expansion concrete pouring construction ! Step d to Step h are repeated. ! reinforcing bar binding with the waist beam, formwork support, and concrete pouring ! acceptance inspection. 1. Surveying, setting out, and pile positioning: Before the foundation pit retaining system was constructed, the plane position of H-shaped steel piles was accurately measured, with the plane error controlled within 5 mm. After setting out each pile position, protective measures were taken for the pile position to prevent human factors from affecting the position deviation after the subsequent pile driving (Gao et al. 2020; Liu et al. 2018; Zhang et al. 2015). 2. Pile hanging in place and calibration: Before H-shaped steel piles were driven, firstly, the pile body was lifted into place according to the pile sections designed and manufactured by the factory, and the plane position of the pile site was rechecked, followed by pile driving. Moreover, the deviations of pile sites exceeding design and code requirements were timely corrected. 3. H-shaped steel pile driving: The H-shaped steel piles used were collectively produced in the factory. When the H-shaped steel piles were driven on-site, the static pile pressing method was used for construction. Every time the piles were driven for 2 m, the pile driving plane and perpendicularity deviation were checked, in which the perpendicularity deviation was controlled within 1%, and the piles with perpendicularity deviation exceeding 1% were corrected. If the correction was difficult, the piles were pulled out for hole grouting and re-driving. 4. Excavation by layers and sections, and treatment of the pile body: After the H-shaped steel piles were driven, the soil layer of the foundation pit was excavated by layers and sections, and the excavation depth of each layer was 2–3 m. For sandy soil and silty soil with high water content, the excavation thickness of each layer was controlled within 2 m. During the layered excavation of the soil layer, the pile reinforcing anchor rods were immediately driven in order to strengthen the integrity between the pile body and the soil layer. On the other hand, the soil attached to the pile body should be cleaned up in time before pile driving in the process of soil excavation, and the soil in the reserved anchor hole on the inner and outer flange plates of the H-shaped pile should be cleaned up. Large stones, if any, should be cleaned up, in cooperation with a drilling rig when necessary. 5. Anchor rod driving from top to bottom in a way of “layer by layer, segment by segment, and rod by rod”: Anchor rods were driven based on the principle of “layer by layer, segment by segment, and rod by rod”. During construction, soil excavation and anchor rod driving on the pile body should be well coordinated, the excavation depth of the soil layer should be strictly controlled, and over-excavation was not allowed. 6. Hoisting of the precast slab wall in place: Precast slab walls were prefabricated by centralized production in the factory, which effectively realized the application of prefabricated construction technology in foundation pit retaining. The first section of precast slab walls at the bottom of the foundation pit could be hoisted in place vertically from the top, and other lower sections needed to be artificially nested into two adjacent
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7.
8.
9.
10. 11.
12.
H-shaped steel piles. If the foundation excavation depth was less than or equal to 6 m, the prefabricated slab walls of all sections were hoisted from the top on the premise of ensuring the soil retaining and bearing capacity of H-shaped steel piles, and then subsided in place, during soil excavation, section by section relying upon their dead weight and the load set for upper slab wall hoisting. A concrete post-pouring section of about 250–350 mm was reserved between the left and right sections of precast slab walls, in which two adjacent left and right precast slab walls were fixed and connected using reinforcing steel bars. Insertion and binding of reinforcing bars firmly with the protruding reinforcing bars of the precast slab wall: In order to enhance the integrity of the retaining structure, reinforcing bars were inserted at the connection joint between hoisted precast slab walls and the H-shaped steel piles and bound firmly with the protruding reinforcing bars of the precast slab walls. “Layered and segmented” high-pressure grouting construction: During construction, high-pressure grouting was carried out layer by layer from the high-pressure grouting hole at the bottom of the precast slab wall, and the grouting work of each section of the precast slab wall was performed symmetrically from the middle high-pressure grouting hole at the bottom of the precast slab wall to the left and right directions, that is, from the middle to the two sides, with the grouting pressure not less than 0.6 MPa. The pressure grouting was done by means of multiple supplementary pressure and step-bystep pressing. Finally, the grouting degree should ensure that the slurry between precast slab walls and the soil layer was compact and permeated into the soil layer for not less than 100 mm. Joint micro-expansion concrete pouring: Micro-expansion concrete (concrete strength was not less than C40 and higher by one order of magnitude at least than the concrete strength of precast slab walls) was poured between H-shaped steel piles and precast slab walls (Chen & Li 2019; Liu et al. 2018; Zhu & Pan 2000). Construction Steps (4)-(9) were repeated. Reinforcing bar binding with the waist beam, formwork support, and concrete pouring: After the construction of the upper and lower precast slab walls was completed, a castin-place reinforced concrete waist beam was set between the upper and lower sections of precast slab walls and the waist beam steel bars were connected with the protruding steel bars of the precast slab walls. At least 4 reinforcing steel bars (over Grade 3) with a diameter of 20 mm were adopted as the longitudinal steel bars of the waist beam, and reinforcing steel bars over Grade 2 with a diameter of 8 mm were used as stirrups spaced by not greater than 200 mm. In the process of steel bar binding on the waist beam, the anchoring length of the protruding steel bars of two upper and lower precast slab walls into the waist beam was not less than 1.5 Lae, and the reinforcing steel bars of the waist beam was welded to the web plates of H-shaped steel piles. Acceptance inspection, Table 1 shows the requirements for acceptance indicators.
Table 1.
Main technical indicators.
Technical index
Unit
Limit value
Maximum displacement deformation of pile top Horizontal hole distance of bolt Vertical hole distance of bolt Hole bottom deflection dimension Maximum lateral displacement velocity around the parapet
mm mm mm % mm/d
+40 30 50 3 3
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2.3
Technical essentials
The compaction grouting method-based construction technology of retaining structures for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls has the following characteristics: 1) As mentioned in step (3), on the premise that the bearing capacity of the foundation pit is ensured, the width and thickness of flange plates of Hshaped steel piles should not be less than 350 mm and 20 mm, respectively. In addition, the height of web plates is 60 mm greater than the thickness of precast slab walls to facilitate segment-by-segment hoisting construction of precast slab walls. Meanwhile, the thickness of web plates is not less than 30 mm; 2) pile sections are connected by welding according to Step (3); 3) in Step 5, traditional anchor rods are replaced by steel strands to facilitate anchor rod construction, where the diameter of anchor rods is 10 mm smaller than the reserved holes on the inner flange plate of H-shaped steel piles, or holes are reserved at the relative position of both inner and outer flange plates of H-shaped steel piles, which is convenient for subsequent anchor rod driving. The holes reserved on the inner and outer flange plates should be 10 mm greater than the diameter of anchor rods, which is not less than 12 mm. In addition, the length of anchor rods is determined according to the bearing capacity calculation. Such anchor rods are placed in the middle of each side of the flange plate along the length direction of H-shaped steel piles at a spacing of not more than 1.5 m. Moreover, the anchor rods are densified within 2000 mm of the pile body at the joint between the upper and lower parts of each section of the precast slab wall, and the spacing is not more than 0.6 m in the densified section; 4) as indicated in Step (7), the reinforcing steel bars are not less than 20 mm in diameter and not less than 2 in number; 5) in Step (8), at least 2 high-pressure grouting holes are arranged at 300 mm above the bottom of each section of precast slab wall with a spacing of 1.2 m. Such high-pressure grouting holes are fabricated using steel sleeves in an integrated way with precast slab walls during centralized production in the factory, and their diameter should be 20 mm appropriately; 6) in Step (11), waist beam stirrups are densified at the post-pouring concrete of two adjacent left and right precast slab walls, and the densified section is not smaller than 600 mm, in which the spacing of stirrups is not greater than 100 mm; 7) as mentioned in Step (11), waist beam steel bars are arranged through holes reserved on the web plates of H-shaped steel piles. Perforated reinforcing steel bars with a diameter of not less than 16 mm are connected and fixed with the protruding steel bars of precast slab walls, and their number is determined and calculated according to the specific conditions of the actual project.
3 CONCLUSION A retaining structure design for deep foundation pits with H-shaped composite steel anchor piles-precast slab walls based on the compaction grouting method and its construction process were revealed. Specifically, an integral foundation pit retaining structure was formed using H-shaped steel piles, anchor rods, and precast slab walls through reasonable processes. The construction process follows the principle of “excavation by layers and sections, anchoring layer by layer, section by section, and rod by rod, and grouting immediately after site hoisting”, which realizes the prefabricated construction of the foundation pit retaining structural system. Moreover, this construction process, which is safe and reliable, can effectively save the construction period with the dual functions of soil retaining and water stopping. (1) The R&D and application of the new prefabricated construction technology for deep foundation pit retaining structures should be highly concerned by the industry, and the application of the new prefabricated construction technology for underground structures becomes a higher challenge to the development of China’s construction industry.
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(2) The new prefabricated construction technology of deep foundation pit retaining structures should combine “retaining structures” with “main structures”, aiming to effectively control the construction cost. (3) In the construction of deep foundation pit retaining structures through the new technology, the emphasis should be put on solving the underground waterproof problem during prefabricated construction, and attention should also be paid to key problems such as the connection of components to the underground retaining structure and the “water-stopping” function of the structure. (4) The application of the new prefabricated construction technology for deep foundation pit retaining structures is not only a problem of construction safety but also a problem of process innovation and optimization design. It is expected that this study can extend the application of the new technology in the industry. (5) The application of the new prefabricated construction technology for deep foundation pit retaining structures has opened up a vast space in the field of prefabricated construction in China, and the research and development of the new prefabricated construction technology for underground structures is an opportunity to gradually deepen new ideas, new structures, and new methods of underground structural construction in other fields in China.
ACKNOWLEDGMENT General Natural Science Research Project of Jiangsu Provincial Department of Education in 2021 (21KJD560005), Nantong Science and Technology Planning Project in 2021 (JCZ21075), Nantong Sixth Municipal Training Project of Jianghai Talents in 2022 (T [2022] No.12).
REFERENCES Chen P Y, Li B D. (2019). Study on Waterproof Construction Technology of Open Cut Underground Utility Tunnel [J]. Highway. 64 (01): 175–178. Feng B, Su X, Wang D K. (2013). Construction Technology for Deep Foundation Excavation and Support on the Xinjiaopu River in Shanghai Hongqiao River System [J]. Construction Technology. 42 (12): 24–27. Gao L H. (2019). Development and Outlook on Prefabricated Construction Technologies for Deep Pit Enclosure Structures [J]. Housing and Real Estate. (24): 157. Gao L H, Wang X D, Zhou F. (2020) Research on Key Technologies of Post-tensioned Prestressed Assembly Road Construction with Bilateral Superposed Beams and Plates [J]. Highway. 65(05): 47–50. Gao L H, Yu Z Z, Liu H Y, Wang H X. (2019). Variable-stress Triangle Truss Support System for H-shaped Steel Piles in Foundation Pit Support [J]. Sichuan Building Materials. 45 (12): 79 + 81. Liu X H, Wang H, Xu M L. (2018). Analysis of Supporting Mechanism of Composite Soil Nailing Wall with Inclined and Advanced Micro-piles [J]. Journal of Shaoyang University (Natural Science Edition). (4): 62–68. Liu XX H, Wang H, Xu M L. (2018). Supporting Mechanism Analysis of Composite Soil Nailing Wall with Inclined and Advanced Micro-piles [J]. Journal of Shaoyang University (Natural Science Edition). (4): 62–68. Yang G Y, Li X W, Han B H, Hu J H. (2018). Influence of Construction in CFG Pile Foundation Pit of a Project in Zhengzhou and Treatment Measures [J]. Henan Building Materials. (5): 103–105. Zhang W Y, Yang Z R, Wang W. (2015). Application of Deep Foundation Excavation Supporting Structure and New Construction Technology in High-rise Building Engineering [J]. Construction Technology. 6 (44): 10–13. Zhu G H, Pan H C. (2000). Deep Foundation Pit Construction in Hangzhou Secondary Long-distance Telecommunication Pivotal Building [J]. Construction Technology. 29 (1): 20–21.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Construction method and application of Dynamic UltimateStrength Control Technology (DUSCT) for ultra-shallow buried excavation Zhao Zhang*, Fei Liu & Huanqiu Li Defense Engineering Institute, Academy of Military Science, Luoyang, China
ABSTRACT: The construction of urban underground engineering using the shallow tunneling method has obvious advantages with little impact on surface transportation and underground pipelines. However, this method is greatly affected by geological conditions, surrounding environment, thickness-span ratio, rise-span ratio, and ground vehicle vibration in the ultra-shallow buried environment. The design and construction technology of dynamic ultimate-strength control technology (DUSCT) was used to solve the technical problems of ultra-shallow buried engineering, including dynamic design, ultimate-strength control, and deformation monitoring. The DUSCT was applied to the connecting tunnel construction between a subway line station and a nearby underground commercial street. Engineering tests and application results show that the DUSCT method is safe and feasible and can guide the construction of ultra-shallow buried structures. The deformation monitoring technology reveals the pressure change law of the supporting structure during the excavation.
1 INTRODUCTION Shallow-buried excavation method (Yang 2010) shows apparent advantages, such as little interference with the surrounding environment (Zhao 2009). It has been widely used in urban underground space development recently. The ultra-shallow buried excavation is its application in ultra-shallow buried engineering. It is generally considered that the depth of underground engineering with the ratio of covering soil thickness to structural span not exceeding 0.6 belongs to ultra-shallow buried engineering. Engineering practice shows that the shallow-buried excavation method is greatly influenced by geological conditions, surrounding environment, cover-span ratio, rise-span ratio, and vehicle vibration on the ground. However, the construction environment of ultrashallow buried excavation is even worse than that of shallow-buried excavation, which challenges the safety and reliability of ultra-shallow buried engineering (Wang 2011). Li and Zhao (Li 2018) introduced the critical technologies applied in the underground tunnel of Kangwang Road and cooperated with technical measures to ensure the safety of ultrashallow buried tunnels. Bai (Bai 2020) adjusted the scheme based on settlement control and CRD optimization technology to control the settlement of vault soil in the ultra-shallow buried project effectively. In constructing an entrance tunnel of a subway station in Zhengzhou, Zhao et al. (Zhao 2017) put forward an alternative advanced support and excavation construction scheme, which provided engineering experience for an ultra-shallow buried excavation method in a clay silt layer. The current research focuses on specific
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-43
construction technology, while the systematic and comprehensive construction control methods aiming at the characteristics of ultra-shallow buried excavation need to be further explored and summarized. This paper proposes the DUSCT method to solve the technical construction problems of ultra-shallow underground engineering. The dynamic design and regulation, ultimatestrength control technology, construction principle, and deformation monitoring technology are explained. A series of research results are successfully applied to engineering practice, which can provide similar feasible methods for underground construction in complex environments of cities.
2 DYNAMIC DESIGN AND REGULATION Experience has shown that shallow-buried construction should be monitored using dynamic control information due to many unknown factors, even if the cavern support scheme is theoretically safe and reliable. While strengthening the surrounding rock (soil) and transforming it into a supporting structure, the shallow-buried excavation method should pay more attention to the dynamic design and regulation according to the monitoring information. Through information feedback, the excavation sequence can be reasonably determined, and the excavation scheme can be optimized and refined step by step, which can reduce the stress and energy concentration caused by excavation. Meanwhile, the deformation of surrounding rock (soil) can also be reduced to ensure stability. The critical points of dynamic design and regulation of ultra-shallow buried excavation can be summarized as follows: (1) Dynamic improvement of engineering experience The engineering analogy method is adopted for the preliminary design of underground engineering, relying on various modern analytical methods, such as statistical mathematics. With the help of existing theories, complex engineering problems can be simplified, and mechanical models consistent with engineering practice can be established. The model quantitatively explains the supporting structure and stratum reinforcement mechanism and meets the requirements of supporting design and stratum stability. (2) Dynamic mastery of stratigraphic information Although a detailed geological survey is usually performed, the geological information needs to be more precise due to the large distance between exploration points. Therefore, the changes in stratigraphic information of the survey blind area will be obtained through the construction process. The main methods of geological advance prediction in the construction process include a geological sketch of the working face, geological radar, infrared detection, and advanced horizontal drilling. Then, compare the design conditions according to the advanced prediction information. If the discrepancy is large, the designer should be informed in time to conduct feedback design, check the design parameters, and modify the construction scheme. (3) Environmental dynamic monitoring There are many uncertain influencing factors in the design and construction of underground engineering, so it is difficult for the design scheme to conform to the engineering practice fully. Therefore, monitoring should be strengthened during the construction process, and the changes in the surrounding environment can be grasped at any time through monitoring. By monitoring the feedback information, designers can modify and optimize the support scheme in time, improve the construction technology, and prevent accidents. The monitoring data can also be utilized as the basis for testing and evaluating the stability of supporting structures. (4) Dynamic regulation of excavation and support scheme A scientific, safe, and feasible specific scheme should be formulated according to the structural design before construction. Experts should review the excavation support
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design. The construction process should be strictly organized according to the scheme, and the changes in geology and the surrounding environment should be grasped in time. For example, the excavation scheme of Line 18 of the Comprehensive Transportation Hub Project in Chengdu Expo City was optimized to control settlement in the complex underground space environment (Bai 2020). The problems found in construction should be dealt with in time to ensure the stability of the underground cavern top and slope soil.
3 ULTIMATE-STRENGTH CONTROL TECHNOLOGY OF ULTRA-SHALLOW BURIED EXCAVATION The limit analysis method simplifies the constitutive relation of a material into an ideal rigidplastic stress-strain relation. It solves the limit load by applying the general theory (upper and lower limit theory) of the ideal mechanical body. The limit analysis theory assumes that the soil is the elastic-ideal plastic body or rigid-plastic body. The strength envelope is a straight line and obeys the standard Coulomb material of the orthogonal flow rule. When the load implied on the soil reaches a specific value and remains unchanged, the soil will have an “infinite” plastic flow. The soil is considered to be in the limit state, and the corresponding load is the limit or maximum load. Generally, the covering soil can be regarded as a sliding body without support. After excavation in the lower part, the sliding body will move downwards along the sliding surfaces on both sides. The work done by an external force (including gravity weight and additional load on the ground) is equal to the work consumed by internal force (including resistance on sliding and lateral structural surfaces). The maximum width and footage of the cavern can be determined under the condition of no support and no roof collapse. When adopting preliminary supporting structures, such as pipe sheds, advanced ductwork, and steel arch (Wei 2018), the sum of the upper soil load and additional ground load should be compared with the limit load, and all of them act on the supporting structure. When the sliding body moves, the limit load, the strength of the supporting structure, and the safety and stability coefficient are calculated from the condition that two works are equal in the limit state (Formula 1). The relationship between the supporting and the excavation chamber is: X X W¼ E (1) where W is an external force, such as gravity weight, additional ground load, and supporting force; E is an internal force, such as lateral slip surface resistance. The soil pressure calculation model is shown in Figure 1. Due to the small covering soil thickness, the arch effect is not considered in the pressure load. The pressure is the sum of the overburdened weight and the additional load on the ground. The vertical pressure p is: X p ¼ q0 þ ðrn hn 2Tn Þ (2) where rn is the soil bulk density, hn is the soil thickness, Tn is the shear force of slip surface, q0 is the additional load on the ground. The shear force of the slip surface is generally ignored when the overburden is thin. The lateral earth pressure ea is calculated according to Rankine soil pressure theory: X pffiffiffiffiffiffi ea ¼ sz Ka 2c Ka ; Ka ¼ tan2 ð45o þ F=2Þ; sz ¼ rn hn þ q 0 (3) where sz is vertical stress, including soil gravity weight and ground additional load, Ka is the main dynamic soil pressure coefficient, and c and F are the shear strength index of soil.
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Figure 1.
Calculation model of the soil pressure.
After setting the long pipe shed before excavation, a reinforced concrete (or section steel) supporting arch shall be constructed at the end of the pipe shed. The supporting arch and soil layer in front of the excavation provides two supporting points of the pipe shed. The stress of the pipe shed can be considered as a beam structure. One end can be simplified as a supported condition, and the fixed support is considered an elastic foundation beam (Sun 2018). The calculation diagram is shown in Figure 2.
Figure 2.
Calculation diagram of pipe roof bearing capacity.
4 ULTRA-SHALLOW BURIED EXCAVATION TECHNOLOGY Currently, the supporting control schemes of shallow-buried excavation mostly follow the classic guidelines (Kang 2012; Li 2013). However, the characteristics of ultra-shallow buried projects are distinctive. The dynamic ultimate-strength control of ultra-shallow buried excavation combined with the guideline of “detailed investigation, appropriate arching, and partial excavation” can provide a more systematic method and feasible technical support for construction through theoretical and numerical analysis. It is crucial to grasp the stratum conditions (such as soil properties and groundwater) in time by adopting the shallow-buried excavation method and applying the information feedback method, which is referred to as “detailed investigation”. The change in the risespan ratio affects the stress concentration coefficient in different tunnel parts. When the risespan ratio is minimal, there will be a sizeable tensile stress at the vault and inverted arch of the primary support structure, which is quite adverse to the surrounding rock stability. Therefore, the cavern in soft soil should be arched as much as possible. The larger the risespan ratio is, the smaller the maximum stress at the vault and inverted arch bottom of the 337
same span. For large-span underground structures, the partial or small-section excavation method can improve the coverage-span ratio and reduce the vertical deformation at the cavern top. The implementation plan of partial excavation should be formulated according to geological conditions, section size, construction period, technological level, ground environment, etc.
5 DYNAMIC MONITORING TECHNOLOGY In ultra-shallow buried excavation, the deformation and stress of the ground and supporting cavern structure should be measured dynamically. The influence of excavation, supporting effects, and environmental changes should be grasped in time (Ma 2011). The monitoring contents mainly include deflection and stress of grid arch, vault displacement, horizontal displacement and settlement of supporting system, internal force of secondary lining, surface settlement and deformation, deep soil displacement, settlement, inclination and crack of adjacent buildings, settlement or deformation of underground pipelines and groundwater. Besides instrument monitoring, inspecting the natural changes and stable state of the surrounding rock (soil) near the excavation face is very important. Monitoring the cracking or peeling of sprayed concrete is also necessary. Especially for underground pipelines near the excavation face, detailed exploration should be conducted before construction, and the direction, depth, and inspection-well position should be mastered. Thus, emergency treatment measures can be taken in time in case of engineering danger.
6 ENGINEERING TEST AND APPLICATION 6.1
Project overview
Dehua underground commercial street is located near Erqi Square in Zhengzhou, with two floors underground. In order to realize the connection between the commercial street and the Erqi Square metro station, it is necessary to build an underground connecting passage. Because the heavy traffic above the street cannot be interrupted, the shallow-buried excavation method is chosen. The tunnel in the underground excavation section is 70 m long, 10.23 m wide, and 6.458 m high. The covering soil thickness is 3–4 m. The project crosses West Street, about 60 m wide, surrounded by squares, commercial buildings, and other buildings. By adopting the proposed method, the excavation and primary support are designed, and the construction scheme is worked out. 6.2
Engineering geology and gydrology
From top to bottom, the foundation soils related to the project are miscellaneous fill (2–3 m thick), silt (2–3 m thick), silt (2–3 m thick), silt (dense), silty clay (locally slightly cemented with a thickness of 20–50 cm) and silt (locally slightly cemented with a thickness of 10–50 cm). The underground water level is about 11.50–13.50 m below the ground. The annual variation range of the water level is about 1.5 m. Although the water depth has little influence on the construction, the stagnant water on the upper surface is harmful to the construction and needs special attention. 6.3
Excavation and construction process
The gross cavern span is 10 m, and the covering soil thickness is only 3–4 m. Therefore, the bilateral pilot tunnel method (BPTM) is adopted. The cross-section is divided into three parts; each is about 3–4 m wide. Both side caverns are constructed separately, then the
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middle tunnel is constructed, and the distance behind the side tunnel on the construction face of the middle tunnel is not less than 6 m. The vertical excavation is divided into 2–3 steps because the height of the rough hole is 6.4 m. The construction steps are as follows: Firstly, a large pipe shed and small conduit are constructed, and the vault is reinforced by grouting. After the left cavern excavation, the steel arch frame, side wall anchor pipe, locking anchor pipe, middle partition steel frame, and mesh-reinforced shotcrete are constructed in the left primary support. Secondly, excavate the right cavern, and repeat the same construction work. Thirdly, excavate the middle cavern and make initial support. Fourthly, construct a waterproof layer and reinforcement concrete (RC) lining of the floor and side wall. Lastly, the partition wall support should be dismantled in sections with a length of less than 6 m, according to monitoring. Meanwhile, the waterproof layer and the reinforced concrete lining should be constructed. 6.4
Monitoring and measurement of construction process
The targeted monitoring includes working face observation, vault and surface settlement monitoring, grid stress and temporary support stress monitoring, and earth pressure monitoring at the soil layer interface. The layouts of settlement measuring points on the ground and cave vault are shown in Figure 3. Each section interval is 5 m at the vault, and three measuring points are set at each section. A section is taken every 10 m at the surface, and three monitoring points are set at each section. In monitoring reinforcement stress and soil pressure, two sections with a distance of 6 m are set, and each section is provided with six measuring points, as shown in Figure 4. The initial value is measured first, and the influence of concrete hardening is observed after the concrete spraying.
Figure 3.
The layout of measuring points for subsidence on ground and in tunnel.
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Figure 4. The layout of soil pressure and grille steel bar stress measuring points (Measuring points of reinforcement (E1E6) and soil pressure (P1P6)).
6.5
Measuring results
6.5.1 Monitoring and analysis of land subsidence Figure 5 shows the observed values of 10, 10-1, 30, and 30-1 measuring points during excavation. The accumulated settlement at 10 and 10-1 points are 10.5 mm and 11.2 mm, respectively, and the accumulated settlement at 30 and 30-1 points are 4.4 mm and 6.2 mm, respectively. The side settlement value is greater than the middle, related to the earlier construction of the left tunnel. The excavation method of the BPTM takes a good control effect on the ground deformation. Because the excavation method is reasonable, the total settlement value is within the allowable range, no apparent cracks are found on the ground, and the traffic is unaffected.
Figure 5.
Cumulative ground settlement.
6.5.2 Analysis of soil pressure measurement results According to the soil pressure measurement results (Figure 6), each point pressure gradually changes with the excavation conditions. The compressive stress of P2 increases obviously with time, with a maximum value of 27.4 kPa. The maximum value is still less than the overburdened soil weight (54 kPa), which indicates that even if the overburdened soil is thin, the soil pressure of the arch-supporting structure is less than the pressure value calculated by the soil body. The arch-rising method of shallow-buried excavation is more conducive to the pressure bearing of the vault.
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Figure 6.
Pressure variation with excavation.
Figure 7.
Steel axis force variation.
With the excavation, the soil pressure gradually concentrated on the vault of the side cavern. When the upper step moved forward for more than 1 m (4.25), the soil pressure of P2 first decreased and then increased rapidly, and the soil pressure of P3 changed little, indicating that the pressure concentrated on the right arch cavern. When the lower step arrives (4.30), P2 increases rapidly, while P3 decreases slightly, and the pressure is further concentrated towards the vault of the right cavern. With digging forward continuously, the pressure at P2 is further concentrated, and the pressure at other positions gradually decreases. After a certain distance is reached, the influence of excavation on the pressure field at the measuring point is no longer evident. After the middle cavern is pushed 6 m across the section (5.23), the soil pressure remains unchanged. At this time, the distance between the working face and the measuring point is about twice the span of the side hole. The influence of the working face on the stress field can be ignored. It is worth noting that the soil at the buried part of the pressure gauges is silty soil and clay, which is hard without the influence of groundwater, and the measured lateral pressures (P3, P4, and P5) are relatively small. 6.5.3 Analysis of stress measurement results of reinforcement Figure 7 shows the stress measurement results of steel bars in the first section. When the upper step of the right cavern goes ahead for 1 m (4.25), the vault grille constructed first acts like a simple-supported beam. The lower part is under tension (E3 stress decreases), and the upper part is under compression (E4 stress increases). The axial reinforcement force near the cavern side wall at the arch top is larger than that near the surrounding rock (soil). E1 and E3 are 11 kN and 24 kN on 6.10, and E2 and E4 on the other side are 24 kN and 29 kN, which indicate that the reinforcement at the arch top is mainly subjected to compression and bending deformation. From 6.9 to 6.24, E5 and E6 changed from 19 kN and 15 kN to 17 kN and 17.5 kN, respectively, which indicates that the stress of the partition wall gradually changed from a bending stress state to a positive pressure stress state. The results show that the maximum axial force of reinforcement is 43 kN, which is far less than the compressive yield limit. Therefore, the cavern is safe under the reinforced grid arch support.
7 CONCLUSIONS (1) This paper introduces the concept of DUSCT and summarizes the critical contents of the dynamic design and control process. The DUSCT gives the overall design calculation method of ultra-shallow buried engineering, which guides the calculation of upper bearing capacity and pipe shed bearing capacity.
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(2) According to the construction environment and stress characteristics of urban ultrashallow buried engineering, the DUSCT is put forward. Engineering practice shows that this technology is feasible, which ensures the safety of ultra-shallow buried engineering and reduces the adverse effects of surrounding traffic and other environmental factors on construction. (3) The construction monitoring and measurement results reveal the law of the excavation influence on the pressure of supporting structures in ultra-shallow excavation. Proper arching of a shallow-buried cave is more conducive to bearing pressure; when adopting the BPTM method, the soil pressure gradually concentrates on the vault with the excavation. When the excavation reaches twice the span of the side tunnel, the influence of the working face on the stress field is no longer evident.
REFERENCES Bai P C. 2020 Settlement Control Techniques of the Super-shallow Buried Large-span Mined Tunnel Underneath an Existing Tunnel. Modern Tunnelling Technology, 57(3): 175–181. Kang F Z, Liu M Y, Guo L, etc. 2012 Urban Underground Shallow/ultra-shallow Mining Changed Crosssection of Technology and the Environmental Impacting Monitor. Building Science, 28(S1): 228–231. Li F G, Zhao Y H. 2018 Key Mining Construction Technologies for Super Shallow-covered Extra-large Cross-section Upright Wall Tunnel Crossing Underneath Urban Artery. Tunnel Construction, 38(11): 1868–1877. Li M Z, Wang J, Wu Z Q. 2013 Super-shallow Underground Excavation Technique of Urban Underground Passage Under Complex Environmental Conditions. Architecture Technology, 44(10): 949–953. Ma Y C. 2011 Construction Technique of Multiple-arch Super-shallow Buried Tunnel of Expressway Passing Through Under 104 National Highway. Architecture Technology, 42(8): 754–757. Sun S, Heng C Y, Zhou Z. 2018 Testing Study on Internal Force Analysis of Steel Arch Structure with Ultrashallow Buried Excavation Underground Passage in Rectangular Section. Building Science, 34(5): 137–142. Wang B H. 2011 Surface Settlement Analysis and Construction Technology for Super-shallow Buried Ventilation Tunnel with Large Cross Section Under Complex Geological Condition. Railway Engineering, 1: 46–49. Wei J J. 2018 The Law of Surrounding Rock and Ground Surface Deformation in Loess Tunnel Construction. Science Technology and Engineering, 18(11): 287–292. Yang H. 2010 Analysis of Impact of Construction of Shallow Large-span Tunnels on Environment with Surface Excavation Method. Journal of Railway Engineering Society, 27(5): 43–47. Zhao Y G, Gu Z W, Han C L. 2009 Design and Construction Technology of Shallow and Ultra-shallow Tunnels. Railway, 10: 323–327. Zhao Z Y, Wang Y, Cao W W. 2017 Metro Entrance Tunnel Construction with Supper Shallow-buried Excavation in Clayey Silt Layer. Urban Mass Transit, 20(11): 78–82.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Optimization and safety evaluation of dismantlement scheme for point-supported glass curtain wall in airport terminal buildings Jian Li*, Jian Hong*, Shiyao Liu*, Yuzai Zhou* & Guangbo Wang* The First Construction Engineering Limited Company of China Construction Third Engineering Bureau, Wuhan, China
Chengxiang Xu* School of Urban Construction, Wuhan University of Science and Technology, Wuhan, China
ABSTRACT: Taking the renovation project of the curtain wall of Wuhan Tianhe Airport T2 Terminal as the research object, the SAP2000 software is used to establish the finite element model of the cable structure of the point-supported glass curtain wall, and the rationality of the model is verified by combining construction monitoring data. Based on the construction simulation at the demolition stage, the three different partial demolition schemes of glass panels are compared with the total removal schemes of glass panels. Through the analysis of the displacement of the main truss, the displacement and internal force of the supporting rod, the safety evaluation, and optimization design of the established construction schemes are completed. The results show that the safety performance of the cable structure point-supported glass curtain wall has been improved, and the glass panel, main truss, and support rod meet the requirements of the standard bearing capacity and flexural deformation checking. Compared with the full removal scheme of the glass panel, the peak displacement of the main truss monitoring point along the-X direction of scheme 1 only increases by 1.7 mm, and the peak displacement of the support rod monitoring point along the-Z direction increases by 3 mm. The bending moments of No. 189 and 204 support rods increase by only 40% and 10.86%, which are far lower than the bending bearing capacity of members, and the internal forces of other rods are basically unchanged. With the continuous removal of the glass panel in schemes 2 and 3, the monitoring displacement is further reduced, and the internal force of the support rod is basically unchanged. Therefore, the partial demolition scheme of the glass panel is feasible, and scheme 1 can minimize the demolition cost.
1 INTRODUCTION Building curtain wall refers to the peripheral protective structure of the building or decorative structure composed of a supporting structure system and a panel, which has a certain displacement capacity relative to the main structure and does not share the force of the main structure (JGJ 102-2003). In recent years, research on construction technology and the experimental field of building curtain walls under construction and new building curtain wall has been mainly carried out in China. For example, Chen et al. (2011) used Midas Gen finite element software to verify
*Corresponding Authors: [email protected], [email protected], [email protected], [email protected], [email protected] and [email protected] DOI: 10.1201/9781003425823-44
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the construction safety of the southwest curtain wall column of the Dalian International Conference Center and determined the removal nodes of temporary support. Zheng et al. (2013) analyzed the reasons for the large-area burst of glass ribs in the construction of an allglass curtain wall at the main entrance of the Tianjin Art Museum and reformulated the design and construction scheme to effectively ensure the quality of the project. Xin et al. (2022) discussed the technical problems of the solid wood keel with a super high and large cross-section in the construction of curtain walls. Zhou et al. (2022) adopted reverse construction technology to solve the engineering problem of steel structure assembly of suspended curtain walls supporting cylindrical reticulated shells. Tian et al. (2022) discussed the application prospect of multi-stage synchronous construction technology of unit curtain walls in super high-rise buildings. With the development of new materials, new processes, and new technologies, the building curtain wall industry has entered a large-scale period of renovation, maintenance, demolition, and reconstruction or transformation (Hua 2022). Therefore, it is of great significance to study the partial demolition and reconstruction technology of curtain wall structures, especially the simulation research. Taking the renovation project of the curtain wall of Wuhan Tianhe Airport T2 Terminal as the research object, this paper establishes the finite element model of cable structure pointsupported glass curtain wall based on SAP2000 software. Based on the construction simulation of the demolition stage, three different glass panel partial demolition schemes and all glass panel demolition schemes are compared in terms of main truss displacement, support rod displacement, and internal force. The optimized design of the established construction scheme and the safety assessment before and after the renovation of the point-bearing glass curtain wall are completed.
2 PROJECT OVERVIEW The total construction area of the Wuhan Tianhe Airport T2 terminal reconstruction project is 158, 400 square meters, consisting of the terminal hall, A corridor, B corridor, and corridor channel. Now a new corridor in the T3 terminal is built to connect with the T2 terminal, and the curtain wall at the T2-T3 corridor needs to be demolished and rebuilt (as shown in Figure 1).
Figure 1.
T2-T3 corridor channel.
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2.1
Component information
The combination of civil steel column, top steel truss, column, and self-balancing cable truss constitutes the vertical stress skeleton of the cable-point glass curtain wall structure system. The glass panel adopts 12 (low-E) + 12A + 12 double toughened insulating glass, which constitutes the “trapezoidal” facade of the curtain wall structure as a whole. The selfbalancing cable truss is composed of two horizontal stainless steel cables, the main truss, support rod, and barge jaw. The connection between the curtain wall system and the main structure adopts hinged supports to adapt to the displacement of the main structure. In the design of the curtain wall structure, the variable position deformation of the roof steel frame is fully considered. The vertical load of the curtain wall does not act on the steel frame, and the load of the steel frame does not act on the vertical curtain wall. The material information is shown in Figure 1. The front of the curtain wall is shown in Figure 2, the A-A section layout is shown in Figure 3, and the broad transverse drawing of the self-balancing cable truss is shown in Figure 4. Table 1.
Material information.
Name of the material
Weight density/ kN/m3
Elastic modulus /105MPa
Coefficient of linear expansion /105 C1
Poisson ratio
Aluminium Rolled steel Stainless steel Reinforced glass
28 78.5 78.5 25.6
0.7 2.0 1.3 0.72
2.35 1.2 1.66 0.9
0.3 0.3 0.3 0.2
Figure 2.
2.2
The front of the curtain wall.
Figure 3.
A-A section layout.
Demolition scheme
This project adopts the construction scheme of glass panel removal as a whole. The construction steps are divided into: the preliminary preparation stage, demolition stage, installation stage, and maintenance stage. The construction process is shown in Figure 5, and the construction site is shown in Figure 6.
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Figure 4.
Broad transverse drawing of self-balancing cable truss.
Figure 5.
Construction process.
Figure 6.
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Construction site.
2.3
Monitoring scheme
The layout of monitoring points is shown in Figure 7. It mainly monitors the anchorage position of the main truss and civil steel column and the horizontal displacement and vertical displacement of the connection position of the main truss and vertical tie rod. The monitoring data is shown in Figure 8.
Figure 7.
The layout of monitoring points.
Figure 8.
Monitoring data.
3 MODEL BUILDING AND VERIFICATION 3.1
Element simulation
Based on SAP2000 finite element software, by modifying the bending stiffness of the section of the frame element to 0.05 and setting the tension and pressure failure mode, the stress characteristics of the cable element which only bear tension but not bear pressure and cannot bear bending moment and shear force were approximately simulated. There are three ways to apply the prestress of the cable: cooling method, strain method, and target force method. The target force method is adopted to apply the cable force iteratively within the program to ensure the efficiency and accuracy of calculation (Yang 2022). The prestressed vertical cable is 35 kN, and the prestressed horizontal cable is 40 kN. Combined with the actual monitoring results of the member, the P-Delta and large displacement effect were considered, but the nonlinear behavior of the material was not considered, that is, the plastic hinge was not set. Elastic beam elements were used to simulate non-cable members. 347
3.2
Element subdivision and division
The main trusses, vertical tie rods, support rods, rectangular columns, and box beams were simulated by ordinary beam elements, while the vertical and horizontal stay cables were simulated by cable elements. The unit division method was adopted for detailed processing. Beam elements were divided into 3 sections and cable elements into 4 sections. The glass panel is simulated by shell element with a thickness of 24 mm, and refined by unit segmentation. The structured part is segmented according to the number of 6 6 segments, and the “L” shaped part is segmented according to the maximum size of 800 mm. Under the principle of node sharing, the shell element “point + surface” and “point + line” are used to define the segmentation command (Sap 2000). Divide separately according to the number of divisions 6 6. The division of shell elements is shown in Figure 9.
Figure 9.
3.3
Division of shell elements.
Determination of boundary conditions
Considering the contribution of the glass panel to the overall stiffness of the structure, a thin shell element is used to simulate the glass panel. The barge jaw does not participate in the modeling. It is assumed that the consolidation effect between the support rod and the glass panel is good, and the limited degree of freedom of rotation provided by the ball hinge in the barge jaw is ignored. The structure constraint construction is shown in Figure 10, and the contact and bearing conditions of the model are shown in Table 2.
3.4
Model validation
This project adopts the construction scheme of glass panel removal as a whole. With the implementation of the construction stage under the action of gravity load and prestress, the comparison between the simulated value of peak displacement at the monitoring point and the measured value (absolute value) is shown in Figure 12. As can be seen in Figure 12, the Z-direction (vertical) simulated value is in good agreement with the monitored value, while the Y-direction (horizontal) measured displacement is always higher than the simulated value. The analysis shows that the modeling assumes that the anchorage end is ideally rigid,
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and the tensile deformation of the anchorage end is not considered. The deformation increment of the curtain wall structure is small in the subsequent design service life, which has no control effect on the overall deformation of the structure. Therefore, the influence of anchorage deformation is ignored during modeling. The numerical model is shown in Figure 11.
Figure 10.
Structure constraint construction.
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Figure 10.
(Continued)
Table 2. The contact and bearing conditions of the model. (The direction is the same as the global coordinate system) Contact con- Glass panel + Glass panel Four sides: releasing the bending moment around the edge; dition Glass panel + Support Fixed support rod 1. Top edge: no barge jaw is set, only the out-of-plane Glass panel + Main constraint of the glass panel is considered, without constructure sidering the support along the vertical, and the translation constraint in the direction of UI and U2 is set for the nodes of the divided panel; 2. Residual edge: It is constrained by the barge jaw and cementing material at the same time, but the mechanical bite force provided by the barge jaw is the dominant force, so the intersection position of the glass panel and the barge jaw is set as consolidation; Glass panel + Door In order to ensure the accuracy of the transmission force bucket steel when the glass panel is affected by the external bar, the split node of the glass panel and the connection node of the support rod are considered to be the same node. Main truss + Vertical Tie 1. Top: releasing bending moment M3; 2. Bottom: Releasing M2; rod (first floor) Support rod end: releasing bending moment M2 and M3; Horizontal cable + Support rod Supporting Vertical cable + Border 1. Top: hinge support; condition body 2. Bottom: releasing bending moment M2.
4 OPTIMAL DESIGN OF EXISTING CURTAIN WALL REMOVAL SCHEME 4.1
Partial removal scheme of the glass panel
The actual project adopts the construction scheme of overall removal of glass panels, which undoubtedly consumes more labor and time costs. This paper proposes three partial removal schemes of glass panels, and redivides the working curtain wall area and the dismantling curtain wall area, as shown in Figure 13. Based on the stage construction simulation function 350
Figure 11.
Numerical model.
Figure 12.
Comparison between the data in simulation and monitoring.
Figure 13.
Removal scheme.
of SAP2000 finite element software, the actual construction plan is divided into 13 construction stages, and plan 1–3 is divided into 14, 15, and 16 construction stages respectively, among which No. 1 is the initial stage, No. 2-8 is the glass panel removal stage and No. 8-12 is the vertical cable removal stage. No. 12-16 refers to the removal stage of vertical tie rods. Only considering the combined conditions of gravity load and prestress, the displacement 351
and internal force responses of main truss monitoring points, supporting rod monitoring points, and rod units at different construction stages are compared, so as to realize the optimal design of the existing curtain wall removal scheme.
4.2
DISPLACEMENT MONITORING ANALYSIS OF THE MAIN TRUSS
The layout of the main truss displacement simulation monitoring points is shown in Figure 14(d), and the main truss displacement is shown in Figure 14(a)–(c). As can be seen from Figure 14(a)–(c), the curtain wall structure in the demolition stage mainly generates translational displacement along the X and Z directions. The peak displacement along the Z direction is basically the same as 3.5 mm, while the peak displacement along the -x direction varies greatly among different schemes. However, the maximum peak displacement of schemes 2 and 3 is basically the same as that of all demolition schemes, and the displacement gap is less than 0.5 mm.
Figure 14.
4.3
Main truss displacement monitoring information.
Support rod displacement monitoring analysis
Figure 15 (d) shows the layout of simulated monitoring points for the displacement of the strut, and Figure 15 (a)–(c) shows the displacement of the strut. As can be seen from Figure 15 (a)–(c), since the displacement monitoring point of the supporting rod and the main truss displacement monitoring point are connected through the supporting rod, the curve variation rules of the two along the X direction are basically the same. Among them, the monitoring displacement of different schemes along the Y direction has little difference and is less affected by the construction stage. The monitoring displacement along the Y direction uniformly changes in a straight line. The monitoring displacement in the zdirection ranges from 4 to 3 mm, and the peak displacement in scheme 1 along the -Z direction reaches the maximum of 4 mm, which increases by about 3 mm compared with the total demolition scheme. The peak displacement of scheme 2 and scheme 3 is 3 mm, which increases by about 2 mm compared with the total demolition scheme. The maximum peak displacement in the Z direction is basically the same, both of which are 2.5 mm.
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Figure 15.
4.4
The information about the monitoring displacement of the support rod.
Internal force monitoring and analysis of support rod
The layout of the internal force simulation monitoring unit of the supporting rod is shown in Figure 16 (d). The local coordinate system of the rod is consistent with the global coordinate system, and the X, Y, and Z axes correspond to the local coordinate axes 1, 2, and 3. The internal forces of the supporting rod are shown in Figure 16 (a)–(c). As can be seen from Figure 9, the internal forces (bending moment, shear force, and axial force) of the rods are basically the same in schemes 2, 3, and 0, while in Scheme 1, the bending moment of the rods is slightly higher than that of the other schemes due to a large number of unremoved panels. Among them, the bending moment of No. 189 and No. 204 increases significantly, and their peak bending moments are 250 kNmm and 510 kNmm respectively. They are 40% and 10.86% higher than the total removal scheme, but far lower than the bending bearing capacity of the rod.
5 SECURITY VERIFICATION BEFORE AND AFTER RENOVATION OF THE CURTAIN WALL STRUCTURE 5.1
Engineering design parameters
The seismic fortification intensity of the terminal is 6 degrees, the designed basic seismic acceleration is 0.05g, and the standard value of horizontal seismic action is 0.147 kPa according to Formula (1). For Class B rough ground, the shape coefficient of positive and negative wind pressure is conservatively set at 1.6. According to the provisions of the Load code for the design of building structures (GB50009-2012), the standard value of wind load perpendicular to the curtain wall surface is set at 1.175kPa according to Formula (2). qEAk ¼ bE amax GA
(1)
where qEAk is the standard value of horizontal seismic action (kPa); bE is the dynamic amplification factor, which is determined by 5.0; amax is the maximum value of horizontal
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Figure 16.
Monitoring information of internal force of supporting rods.
seismic influence coefficient, which is 0.04; GA is the standard value of gravity load (kPa) of curtain wall components (including glass panels and aluminum frames) in unit plane area of the glass curtain wall. Wk ¼ bgz ms1 mz W0
(2)
where Wk is the standard value of wind load (kPa); bgz is the gust coefficient at height Z, which is 1.61; W0 is the basic wind pressure, which is 0.35 kPa; ms1 is the shape coefficient of the wind pressure; mz is the height variation coefficient of wind pressure, which is 1.3. 5.2
Safety assessment of curtain wall structural members
5.2.1 The strength and deflection checking of the glass panel Under the action of standard wind load, the x-displacement nephogram of the glass panel of the curtain wall structure is shown in Figure 17. The maximum deflection of the glass panel before the transformation is 28.6 mm, which decreases to 15.4 mm after the transformation, both lower than the standard limit 50 mm (1/60 of the long side span between the support points of the glass panel), with a reduction of 46.15%. According to “Technical Specifications for Glass Curtain Wall Engineering” JGJ 1022003, the most unfavorable combination of dead load, wind load, and seismic load was taken to obtain the stress nemogram (symbol specified in the same global coordinate system) in the glass panel, as shown in Figures 18 and 19. As can be seen from Figures 18 and 19,
Figure 17.
Displacement cloud diagram in the X direction.
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Figure 18.
S11 stress cloud diagram.
Figure 19.
S22 stress cloud diagram.
stress concentration occurs at the connection position of the curtain wall glass panel and support rod, which is the vulnerable part of the structure, and reinforcement measures should be taken. The maximum internal stress of the glass panel is 56 MPa and 52.5 MPa before the glass panel modification, and the maximum internal stress of the glass panel modification is increased to some extent, among which S11 is 59.5 MPa and S22 is 63 MPa, but it is always lower than the allowable large surface strength limit of 84 MPa of the standard tempered glass. Therefore, the glass panel meets the bearing capacity checking requirements. 5.2.2 The deflection checking of the main truss Under the action of the standard value of wind load, the main truss of the curtain wall structure mainly flexes and deforms along the X and Z planes, as shown in Figures 20 and 21. The maximum deflection of the main truss is 24 mm before the transformation, and it
Figure 20.
Displacement cloud diagram in the X direction.
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Figure 21.
Displacement cloud diagram in the Z direction.
decreases to 14 mm after the transformation, with a decrease of 41.66%. In the Z direction, the maximum deflection of the main truss before reconstruction is 0.4 mm, which is increased to 5.5 mm after reconstruction, which is lower than the limit deflection of the main truss member. The minimum deflection (1/250, 30 mm of the span of the member) is 30 mm. 5.2.3 The deflection checking of the support rod Under the action of the standard value of wind load, the deflection of the support rod along the Y and Z direction is shown in Figures 22 and 23. Before and after the transformation, the maximum deflection of the support rod along the Y direction is 2.8 mm, and the maximum deflection along the Z direction is increased from 4 mm to 5 mm. The increase in the deflection is not obvious, and it is always lower than the deflection limit. The minimum deflection (1/200 of the distance from the support point) is 8.2 mm.
Figure 22.
Displacement cloud diagram in the Y direction.
Figure 23.
Displacement cloud diagram in the Z direction.
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6 CONCLUSIONS (1) Compared with the actual plan of complete removal of the curtain wall glass panel, the peak displacement of the main truss monitoring point in plan 1 along the -X direction increased by only 1.7 mm, the peak displacement of the support rod monitoring point along the -Z direction increased by 3 mm, the bending moment of No. 189 and No. 204 supporting rods increased by 40% and 10.86%, which were far below the bearing capacity of the rods, and the internal forces of other rods remained unchanged. In plans 2 and 3, with the continuous removal of glass panels, the displacement of monitoring points is further reduced, and the internal force of supporting rods is basically unchanged. Therefore, it is feasible to partially remove the glass panel, and scheme 1 can reduce the removal cost to the greatest extent. (2) After the reconstruction of the curtain wall structure, the safety performance is improved and meets the standard requirements. After the curtain wall is dismantled and reconstructed, the maximum surface deflection of the glass panel along the X direction is reduced by 46.15%, and the peak stress on the surface is increased, but it is lower than the standard limit of 84 MPa. The deflection of the main truss along the X direction decreased by 41.66%, and the deflection along the Z direction increased to 5 mm, but all of them were lower than the deflection limit of 30 mm. The deflection of the support rod along the Y and Z directions did not change significantly, both of which were lower than the deflection limit of 8.2 mm.
REFERENCES Chen HZ, Wang YL, Liu JG, et al. Analysis of Construction Method of Complex Curtain Wall Column and Roof Linkage in Dalian International Convention Center [J]. Construction Technology, 2011, 42(9):852– 854. Hua YQ, Liu Y, Ma GX. Comprehensive Construction Technology for Rapid Demolition and Renewal of Glass Curtain Wall of Super High-rise Building [J]. Construction Technology (Chinese and English), 2022, 51(15):34–38. JGJ 102-2003, Technical Code for Glass Curtain Wall Engineering [S]. Beijing: China Architecture & Building Press, 2003. (in China) Load Code for the Design of Building Structure: GB 50009-2012 [S]. Beijing: China Architecture & Building Press, 2012. Sap2000 Chinese Edition User Guide [M].2nd Edition. Beijing: China Communication Press. (in China) Tian W, Liu ZL. Multi-segment Synchronous Construction Technology for Unitary Curtain Wall of Super High-rise Building [J]. Construction Technology (Chinese and English), 2022, 51(12):125–130. Xin HJ, Zuo L, Wang CH, et al. Construction Technology of Ultra-high Oversized Section Solid Wood Keel Curtain Wall [J]. Construction Technology (Chinese and English), 2022, 51(24):130–135. Yang Z, He YJ, Q Y. Static Analysis and Simplified Calculation Method of Single-story Suspended Photovoltaic Supports [J]. Science and Technology and Engineering, 2022, 22(21):9252–9259. Zheng YK, Liu GC. Construction Technology of Ultra-high Glass Rib All-glass Curtain Wall of Tianjin Art Museum [J]. Construction Technology, 2013, 42(3):94–97. Zhou F, Chen XM, Yu YY, et al. Reverse Construction Technology of Suspended Curtain Wall Supported Column Face Mesh Steel Structure [J]. Construction Technology (Chinese and English), 2022,51(9):121– 125.
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5D building information modeling status based on bibliometrics Hui Sun* School of Housing, Building and Planning, University Sains Malaysia, Pulau Pinang, Malaysia School of Civil Engineering and Architecture, Beibu Gulf University, Qinzhou, China Guangxi Key Laboratory of Ocean Engineering Equipment and Technology, Qinzhou, China
Terh Jing Khoo, Jiao Wang & Maoying Wang School of Housing, Building and Planning, University Sains Malaysia, Pulau Pinang, Malaysia
ABSTRACT: Building information modeling (BIM) plays a key role in improving productivity and construction quality, cutting costs, and reducing project time. Existing research on 5D BIM lacks a systemic compilation and concentrates on a single application. In this study, 104 papers in this field were collected to understand trends through a bibliometric analysis. The top 19 publications with the highest citation rates were examined. The results showed that 5D BIM has limited application breadth and marketing. Future studies will focus on ways of expanding the application of BIM and integrating it with real engineering.
1 INTRODUCTION The construction industry contributes toward improving the quality of human life while meeting the socioeconomic demands of people, societies, and nations (Goel et al. 2019). However, the industry faces challenges such as ineffective communication, low output, and time and expense overruns (Becerik-Gerber & Rice 2010). To overcome conflicts and keep projects on schedule and within budget, the various disciplinary teams involved in construction projects must form effective collaborations. Inadequate data management and communication cost the construction industry approximately 15.8 billion US dollars yearly or 3–4% of the total revenue (Gallaher et al. 2004). In the construction business, technology is typically used in building projects to improve productivity and quality, reduce project costs, and shorten project duration (Azhar et al. 2008). Building information modeling (BIM) has gained popularity in the construction industry over the past decades. BIM offers the industry several advantages and resource-saving potential throughout the design, planning, and construction phases of structures (Volk et al. 2014). The phrase “building information model” first emerged in a 1992 publication, “Modeling Many Views on Buildings” (Van Nederveen & Tolman 1992). Then, Kymmell (2008) and Azhar (2011) explained its concept and meaning, respectively. The current standard definition is provided by the National Building Information Modeling Standard of the United States (Weldu & Knapp 2012): “BIM is an information integration platform that allows all data generated during a project to be shared and used as a basis for decisionmaking. During the operation of a project, various units and users can use all the data and information integrated in the BIM platform and can work together to complete the project.” Engineering construction is expected to be significantly driven into a new three-dimensional (3D) age by BIM (Cattell et al. 2007). *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-45
BIM is a collection of techniques for gathering, organizing, analyzing, and spreading the data of buildings. From project ideas to building decommissioning stages, BIM serves as the cornerstone for building management. Classical two-dimensional architectural models have arguably transformed into architectural 3D models through BIM technology. The interactivity of BIM has broadened its definition, increasing its utility. Four-dimensional explorations have enabled simulations of building operations through the combination of information and data from 3D object models with scheduling data. Cost data, including quantities, timetables, and pricing, are further integrated into the information to reach the fifth dimension. The as-built model is shown in six dimensions, which may be applied to the facility’s performing phases. Support for facility management is added to the sevendimensional BIM, and safety data, such as emergency plans and security analyses, is added to the eight-dimensional BIM. However, no consensus currently exists beyond five-dimensional (5D) BIM (Charef et al. 2018). 5D BIM technology uses a central database and an independent information model to store and edit engineering information, including the cost aspects. Thus, it improves work efficiency and revolutionizes the construction industry. Therefore, 5D BIM is focused on this study. Figure 1 depicts two revolutions in the building industry through the design and entire life cycle phases (Xu 2017). A key benefit of 5D construction simulation is its capacity to track the project form over time and incorporate dynamic project expenses into the construction model. 5D construction simulation can be used to model the construction design before work begins, reducing variations and costs while optimizing plans. Morrison and Thurnell (2014) emphasize the potential for using BIM as a project management tool and highlight 10 benefits of 5D BIM to quantity surveyors.
Figure 1.
Information changes in the construction industry (Xu 2017).
In this study, several studies on 5D BIM were systematically analyzed. Most prior literature reviews focused on specific BIM areas, but this study was a review of articles from reputable journals in the discipline as well as a wide range of publications. The 5D BIM technique was categorized throughout the paper to ease the comprehension of the literature and to provide a better analysis of trends and research gaps.
2 METHODOLOGY A bibliometric analysis was used to analyze and classify the research on 5D BIM from 2011 to 2023. The analysis comprised three key phases: Step 1, searching for keywords in Scopus; Step 2, selection of the most cited articles; Step 3, classification of articles based on contents. 2.1
Database choice
The Scopus database of journal papers (https://www.scopus.com/sources) was used to select articles relevant to the subject of 5D BIM. Co-authorship patterns and term co-occurrence 359
were identified through bibliographic analysis. Textural strings were used for the search; the strings included terms such as “5D BIM,” “BIM 5D,” and “five-dimensional building information modelling”. Thus, all 104 articles found were relevant to this study. Papers written in English were selected while others were excluded, resulting in a list of 97 articles for analysis. 2.2
Filtering and controlled criteria
To identify the most influential publications on this subject, a second search (in Scopus) was performed for the most cited articles in the 5D BIM literature. The top articles were selected after sorting all articles according to citation rate; only articles with more than 10 citations were selected as a way of ensuring that quality articles were used for analysis. A total of 19 articles were selected from this filtering process. 2.3
Article classification based on content
Each author independently performed the classification procedure using a grounded method. The categories were based on the substance of the articles rather than pre-existing research themes or areas (present in prior studies). All categories were then combined to create a unified classification.
3 BIBLIOMETRIC ANALYSIS According to the bibliometric study, the number of published articles on 5D BIM has increased over the previous decade, from two in 2011 to ten in 2022. The number peaked at 20 in 2020 before dropping again (Figure 2). In terms of the literature source, most 5D BIM papers were published in IOP Conference Series: Materials Science and Engineering (5), followed by Communications in Computer and Information Science, IOP Conference Series: Earth and Environmental Science, and Wit Transactions on the Built Environment (4). The published material spans the years 2012 through 2021 (Figure 3). Nations and territories that show high interest in 5D BIM include China, Australia, the UK, Germany, Hong Kong, the USA, Iran, Canada, Egypt, and Malaysia. In terms of the number of papers on 5D BIM, these nations are among the top 10 in the world (Figure 4).
Figure 2.
Published articles on 5D BIM over the previous years.
360
Figure 3.
The number of 5D BIM papers published yearly by five main publication sources.
Figure 4.
The number of 5D BIM papers published by five main countries or territories.
4 CONTENT ANALYSIS Table 1 lists the 5D BIM journal articles with the most citations. The citation count of each publication in the 5D BIM literature is recorded and compared to gain insight into the trend and important bibliometrics. As shown in Table 1, 2020 had the greatest percentage of articles published (four journal articles), followed by 2014 (three journal articles). However, the works published in 2014 had the most impact and received the most citations. The average number of annual citations of 2020 publications is larger than publications in 2014. Figure 2 and Table 1 indicate that both the quantity and quality of publications published in 2020 were the highest. The top 10 authors (Figure 4) with the most citations are from China, Australia, the UK, Germany, Hong Kong, the USA, Iran, Canada, Egypt, and Malaysia. This result conflicts with the information in Table 1. Australian researchers have the most publications and citations, whereas researchers from the United States and China have more published articles but no more citations than Australian researchers. Among the top selected publications (Table 1) selected from fields and nations, six research axes are identified. Research axes are ranked in sequence according to their citation frequency (Table 2). Among the six research axes identified, contractors have the greatest weightage. Lu et al. (2016) and Forgues et al. (2012) are concerned with planning and investment choice phases.
361
Table 1.
Summary of the top cited papers.
Reference
Published time
(Aibinu & Venkatesh 2014)
2014.07
(Lu et al. 2016)
2016.01
(Elghaish, Abrishami, & Hosseini 2020) (Morrison & Thurnell 2014) (Kehily & Underwood 2017) (Forgues et al. 2012)
2020.06
Journal
Application technology/Related Frequency personnel
Journal of Profes105 sional Issues in Engineering Education and Practice International Jour99 nal of Project Management Automation in Con- 93 struction
2014.02
Construction Eco72 nomics and Building
2017.09
Journal of Information Technology in Construction Construction Research Congress 2012 International Journal of Construction Management Procedia Engineering
2012.05
(Khanzadi et al. 2020)
2020.05
(Xu 2017)
2017.04
(Harrison & Thurnell 2015)
2015.07
(Vigneault et al. 2019)
2019.04
(Mayouf et al. 2019)
2019.06
(Puˇcko et al. 2020.05 2020)
International Journal of Construction Supply Chain Management Archives of Computational Methods in Engineering
5D BIM/quantity surveyors
Application categories Cost modeling process
5D BIM/contractors Cash flow analysis Creating a framework using blockchain & BIM/all project participants 5D BIM/quantity surveyors
Financial automation system
Cost modeling process
43
5D BIM/quantity surveyors
Life cycle costing
42
5D BIM/general contractors
Cost estimation
35
5D BIM/projects managers
Key performance indicators
34
5D BIM/contractors Construction stage taking the central grand project 5D BIM/quantity A large global surveyors practice
24
20
Journal of Engineer- 20 ing, Design, and Technology Energies 19
(Cheng et al. 2017.12 2017)
Visualization in Engineering
17
(Elghaish et al. 2021)
2021.06
(Elghaish, Abrishami, Hosseini, et al. 2020)
2020.04
International Jour16 nal of Construction Management Engineering, Con14 struction and Architectural Management
5D BIM/general Cost managecontractors and BIM ment project managers and consultants 5D BIM/quantity Cost surveyors 5D BIM/energy effi- Decision-making cient designs around optimal energy and economical 4D/5D BIM/offSimulation and shore oil and gas evaluation opplatforms tions of a fixed jacket platform 5D BIM/project sta- Integrated prokeholder ject delivery 5D BIM/stakeholder Integrated project delivery
(continued )
362
Table 1.
Continued Published time
Reference (Soto & Adey 2016)
2016.06
(Marzouk & 2019.09 Enaba 2020) (Wülfing et al. 2014)
2014.08
(Lee et al. 2016)
2016.09
Table 2.
Journal Procedia Engineering
Application technology/Related Frequency personnel 14
Application categories
5D BIM versus arti- Conceptual ficial intelligence/ phase of projects project managers 5D BIM model/con- Construction tractor project performance 5D BIM/small and The lifecycle of a medium crafts enter- construction prises/contractor work
Built Environment 12 Project and Asset Management eWork and eBusi11 ness in Architecture, Engineering, and Construction MATEC Web of 10 Conferences
5D BIM/stakeholders
Conceptual bungalow design
Weight ranking of research axes.
Research axes
Articles
Rank
Contractors Quantity surveyors Stakeholders Project managers Energy-efficient designs Offshore oil and gas platforms
6 5 4 2 1 1
1 2 3 4 5 5
Elghaish et al. (2020) and Wülfing et al. (2014) argue that 5D BIM improves estimate accuracy and balances the cost benefits and drawbacks of various solutions. Xu (2017) proposes a degree of refinement in the building phase of 5D BIM for construction quality assurance. Another key research perspective is the quantity surveyor axis. All authors in this category discuss the benefits of 5D BIM and the challenges associated with its promotion. These challenges are mainly caused by two factors that pose obstacles to the application of BIM. The first factor is the disconnection between people who are knowledgeable about BIM and those who have actual engineering experience. The second factor is the imperfect sharing of building information among stakeholders. Morrison and Thurnell (2014) and Kehily and Underwood (2017) focus on 5D BIM over the entire life cycle of a building, whereas Lu (2016) and Harrison and Thurnell (2015) concentrate only on the early phases of a project. Because of its 3D visibility and information interactivity, BIM offers a collaborative approach to the work of all disciplines and phases of the project. All project stakeholders can be effectively coordinated through the BIM platform to make informed choices, which enhances the implementation of construction project management. A 5D BIM model can incorporate all the qualities of the construction project information. From the perspective of the project manager, Khanzadi et al. (2020) discuss the importance of 5D BIM in business performance and the major influence which BIM has on the latter stages of construction. Soto and Adey (2016) raise concerns about cost projections at the conceptual design stage. Cheng et al. (2017) simulate and evaluate choices of a fixed jacket platform by 5D BIM.
363
Although the contribution of 5D BIM to controlling building costs and business performance is well accepted, there are still barriers to its adoption. Xu (2017) emphasizes that 5D BIM has only been used for major projects in some countries during the past few years. The use of 5D BIM still has limitations, and a research gap exists in how to expand its use and seamlessly incorporate it into engineering projects.
5 CONCLUSIONS BIM technology has evolved from 3D to 5D with the addition of capital cost information and progress plan data. 5D BIM has become an integrated information-sharing platform with schedule, cost, quality, and formation data. After peaking in 2020, the number of publications and citations is now declining, according to the number of papers published recently. Australian authors are most prolific in terms of the number of articles published and citations, whereas Chinese and American authors rank higher for only the number of articles published. The present study on 5D BIM is based on six views and three perspectives: contractors, quantity surveyors, and stakeholders. Although 5D BIM has been accepted for its role in cost containment, more study is required on its engineering applications and application expansion.
ACKNOWLEDGEMENTS The research of the first author is supported by Guangxi Higher Education Undergraduate Teaching Reform Project, Project no.: 2021JGB261.
REFERENCES Aibinu, A., & Venkatesh, S. (2014). Status of BIM Adoption and the BIM Experience of Cost Consultants in Australia. Journal of Professional Issues in Engineering Education and Practice, 140(3), 21–31. Azhar, S. (2011). Building Information Modeling (BIM): Trends, Benefits, Risks, and Challenges for the AEC Industry. Leadership and Management in Engineering, 11(3), 241–252. Azhar, S., Nadeem, A., Mok, J. Y. N., & Leung, B. H. Y. (2008). Building Information Modeling (BIM): A New Paradigm for Visual Interactive Modeling and Simulation Construction Projects. First International Conference on Construction in Developing Countries, 435–446. Becerik-Gerber, B., & Rice, S. (2010). The Perceived Value of Building Information Modeling in the U.S. Building Industry. Journal of Information Technology in Construction, 15(15), 185–201. Cattell, D. W., Bowen, P. A., & Kaka, A. P. (2007). Review of Unbalanced Bidding Models in Construction. Journal of Construction Engineering and Management, 133(8), 562–573. Charef, R., Alaka, H., & Emmitt, S. (2018). Beyond the third dimension of BIM: A Systematic Review of Literature and Assessment of Professional Views. Journal of Building Engineering, 19, 242–257. Cheng, J. C. P., Tan, Y., Song, Y., Liu, X., & Wang, X. (2017). A Semi-automated Approach to Generate 4D/ 5D BIM Models for Evaluating Different Offshore Oil and Gas Platform Decommissioning Options. Visualization in Engineering, 5(1), 12–25. Elghaish, F., Abrishami, S., Abu Samra, S., Gaterell, M., Hosseini, M. R., & Wise, R. (2021). Cash Flow System Development Framework within Integrated Project Delivery (IPD) using BIM Tools. International Journal of Construction Management, 21(6), 555–570. Elghaish, F., Abrishami, S., & Hosseini, M. R. (2020). Integrated Project Delivery with blockchain: An Automated Financial System. Automation in Construction, 114, 103182. Elghaish, F., Abrishami, S., Hosseini, M. R., & Abu-Samra, S. (2020). Revolutionising Cost Structure for Integrated Project Delivery: A BIM-based Solution. Engineering, Construction and Architectural Management, 28(4), 1214–1240. Forgues, D., Iordanova, I., Valdivesio, F., & Staub-French, S. (2012). Rethinking the Cost Estimating Process Through 5D BIM: A Case Study. Construction Research Congress 2012, 778–786.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Hydraulic fracturing test on in-situ stress and its distribution regularities of a flood drainage construction engineering site Ping Lu & Jinliang Zhang* Faculty of Metallurgy and Mining Engineering, Kunming Metallurgy College, Yunnan, Kunming, China
Guoxiang Guo Prospecting Design Institute of China Nonferrous Metals Industry Co., Ltd, Kunming, China
Qi Nie*, Yong Cheng & Yiming Wen Faculty of Metallurgy and Mining Engineering, Kunming Metallurgy College, Yunnan, Kunming, China
ABSTRACT: The influence of in-situ stress on safe construction and the long-term stability of underground engineering is particularly prominent. The stress field distribution of the flood drainage projects is tested by the hydraulic fracturing method. The test results show that: (1) the maximum horizontal principal stress of different buried depths is always the first principal stress. The in-situ stress is dominated by horizontal tectonic stress; (2) the in-situ stress of the test area is dominated by south-north and northeast compression, and the maximum horizontal principal stress direction is N46.8W-N43.5W; (3) the stress distribution regularities in the area meet sH >sV >sh ; (4) the direction of the stress field is close to the surrounding area, and the stress basically increases with the depth. The distribution regularities of the regional stress field can be obtained by linear regression using the least square method.
1 INTRODUCTION The in-situ stress is a natural stress in the strata without engineering disturbance. It is the fundamental force causing deformation and damage to excavation engineering in the underground. It is also the necessary prerequisite for the stability analysis of surrounding rock and the decision of rock excavation engineering (Cai 2013). The analysis of the in-situ stress has become an important basis for underground engineering construction, underground mine construction, gateroad stability analysis, rockburst (coal and gas outbursts) prediction, ground pressure control, etc (Sun 2020). In-situ stress measurement is a prerequisite for geological disaster control (Cai 1993; Pei 2020; Wang 2011). Common methods include the core inclusion relief method, acoustic emission method, inelastic recovery (ASR) method, borehole strain relief method, hole wall deformation stress relief method, hole bottom stress relief method (Meng 2015), three-hole three-dimensional water pressure cracking method, and single-hole three-dimensional water pressure cracking method. Each method has its applicable scope. Many scholars have done a lot of research on the above methods. Cai Meifeng modified the CSIRO method in core elastic modulus, temperature effect, data recording, etc. The measurement results are more accurate (Cai 1991). Sun et al. theoretically proved the feasibility of the ASR method in-situ stress measurement in mines
*Corresponding Authors: [email protected] and [email protected]
366
DOI: 10.1201/9781003425823-46
(Sun 2014, 2020). Ma et al. completed the research and development of surface redirection devices for acoustic emission geological core (Ma 2021; 2020). A large number of studies have improved the accuracy of measurement in theory and practice in different ways, but the distribution of in-situ stress is complicated. It’s impossible to evaluate the measurement accuracy in practice (Hou 2022). In this paper, hydraulic fracturing is selected as the test method. It can determine the magnitude and direction of the maximum and minimum horizontal principal stress in the region. It also gives mechanical parameters such as fracture pressure, retention pressure in-situ tensile strength of rock, and so on. Based on these parameters, the distribution regularities of ground stress, stress level, and engineering characteristics of the rock mass of the project are analyzed. These data provide reliable information for the possibility of disasters and preventive measures in the later construction.
2 ENGINEERING GEOLOGY The flood discharge project is located near the town. Due to the impact of human activities, the flood discharge capacity of the karst pipeline is weakened. In order to solve the local flood disaster, tunnels and flood drains are built. The geological conditions are complicated in the area. The surrounding rock of the tunnels is complex. It is inclined across the Niujiao Mountain watershed, with an elevation of 1855 m, a topographic slope of 35–45 , and locally steep cliffs. Carbonate rocks and clastic rocks are distributed alternately, with karst peak clusters, karst depressions, and erosion landforms. The entrance of the tunnel and the body of the tunnel contain the formation of the Quaternary (Q), Maokou Formation, Qixia Formation, and Liangshan Formation. The lithology in this area is mainly limestone, mudstone, marl, and so on. The region frequently experiences hard rock bursts, collapses, surface subsidence, and soft rock compression deformation and instability.
3 PRINCIPLE OF THE HYDRAULIC FRACTURING METHOD The three assumptions are put forward (Cai 1993). The mechanical model of hydraulic fracturing meeting the above conditions can be simplified as a plane strain problem. When a hole located in an infinite body is subjected to the action of a two-dimensional stress field (s1 , s2 ). The part at a certain distance from the end of the hole is in a plane strain state. At these locations, the stress around the hole is sq ¼ s1 þ s2 2ðs1 s2 Þcos 2q
(1)
sr ¼ 0
(2)
where sq is the tangential stress, sr is the radial stress, and q is the angle between a peripheral point and an axis of s1 . From (1), q = 0 º, sq is the minimum. sq ¼ 3s2 s1
(3)
A section of the hole is sealed off. Then, the high-pressure water is injected into the section of the hole. When the water pressure exceeds the sum of 3s2 s1 and T. At q=0º, the hole wall of s1 will crack. Suppose the water pressure at the time of initial cracking of the hole wall is Pb , then, Pb ¼ 3s2 s1 þ T:
(4)
Due to the effect of geostress, the cracks will close quickly. Generally, the equilibrium pressure is called the instantaneous closing pressure Ps , which is equal to the minimum 367
horizontal principal stress of the vertical fracture surface, then Ps ¼ s2 :
(5)
From Equations (4) and (5), the tensile strength Tis measured. s1 ands2 are obtained by Pb and Ps . There is fissure water in the hole. If the fissure water pressure at the partition is P0 , Equation (4) becomes: Pb ¼ 3s2 s1 þ T P0
(6)
According to Equations (5) and (6), s1 and s2 are obtained. In order to obtain the tensile strength of the rock in the isolation section, a link is added to the hydraulic fracturing test. After the initial crack is produced, the water pressure is removed to make the crack close. Then, pressure is applied to make the crack open again. The pressurePr is denoted as Pr ¼ 3s2 s1 P0
(7)
From Equations (5) and (7), s1 and s2 are obtained without the tensile strength of the rock. When measuring the in-situ stress in the vertical hole, the maximum and minimum horizontal principal stress is written as sH and sh , respectively, and then, the above formula can be expressed as follows: Pb ¼ 3sh sH þ T P0
(8)
Pr ¼ 3sh sH P0
(9)
Ps ¼ s h
(10)
From Equations (9) and (10), the maximum horizontal principal stress sH is obtained. sH ¼ 3Ps Pr Po
(11)
The vertical stress can be calculated according to the weight of the overlying rock. sv ¼ rgd
(12)
where r is the density of the overlying, g is the acceleration of gravity, and d is the depth of the measuring point.
4 FIELD TEST 4.1
Test parameters
The YG-2017 single-loop geostress measurement system is adopted. The system includes a high-pressure pump station (rated pressure 60 MPa), portable data acquisition system, control system, push-pull switch (design strength 70 MPa), electronic orientation instrument (voltage 7.4 VDC, capacity 2200 mAh), packer, impression device (design strength 50 MPa, depth of less than 2000 m, work continuously between temperature (-10 ℃+50 ℃) and high-pressure oil pipe. 4.2
Test procedure
The packer is used to isolate a section of the hole at the selected measuring depth. Then the test section (the fracturing section) is pressurized by pumping fluid. The data acquisition system records the change in pressure with time. The characteristic pressure parameters are obtained by analyzing the measured record curve. According to the theoretical calculation formula, the parameters are obtained. 368
5 DATA ANALYSIS OF TEST RESULTS The rock fracture pressure Pb , instantaneous closure pressure Ps , and fracture reopening pressure Pr can be obtained from the pressure-time record curve, as shown in Figure 1. The maximum sH and minimum sh horizontal principal stress and the in-situ tensile strength of rock Tcan be calculated according to formulas 8 to 10. Fracturing parameters Pbs ,Prs , andPss from the field measurement are surface-recorded values. The relationship between Pb ,Pr ,Ps and Pbs ,Prs ,Pss is as follows: Pb ¼ Pbs þ Pw
(13)
Pr ¼ Prs þ Pw
(14)
Ps ¼ Pss þ Pw
(15)
where Pw is the water column pressure in the drill pipe. 5.1
Principal stress in different test sections
According to the geological conditions of the hole, core RQD, logging data, and the initial data of hole sediment, the 6 sections of stress measurement were carried out from bottom to top. A good pressure curve has been obtained. The pressure recording curve is shown in Figure 1 (a)–(f). The density of rocks is 2.69 g/cm3. The in-situ stress measurement results are shown in Table 1.
Figure 1.
5.2
(a-f) Pressure recording curve of ZK1.
Direction of principal stress
After the fracturing, the electronic directional device with the impression is used in the 407.87 m and 478.00 m test sections. The impression was clear, and the results are shown in Figure 2 (g)–(h). The direction of the maximum horizontal principal stress at 407.87 m is N46.8 W and at 478.00 m is N43.5 W. 369
Table 1.
The in-situ stress measurement results. Fracturing parameters (MPa)
Number Section depth(m) Pb地 1 2 3 4 5 6
389.87 410.27 450.31 468.11 505.09 524.35
Figure 2.
5.3
Pr地
7.64 6.58 8.00 6.82 8.89 7.97 9.40 7.74 10.24 8.52 11.86 10.62
Main stress (MPa)
Fracture direction ( )
Ps地 Pw
P0
T
sH
sh
sV
4.02 5.02 5.62 5.52 5.73 6.75
0.82 0.97 1.37 1.67 1.97 2.13
1.06 1.18 0.92 1.66 1.72 1.24
12.52 15.43 16.48 16.71 16.86 17.98
7.95 9.10 10.10 10.30 10.81 11.99
10.57 10.97 N46.8 W 12.06 12.86 N43.5 W 13.67 14.11
3.93 4.08 4.48 4.78 5.08 5.24
Direction of an impression of ZK1.
The trend of the stress distribution
According to the in-situ stress measurement results, the maximum horizontal principal stress is 12.52–17.98 MPa. The minimum one is 7.95–11.99 MPa in the measured range (392.87–524.37 m). From the relationship between stress and depth, stress basically increases with depth. The stress distribution regularities in the area meet: sH 4sV 4sh . The distribution regularities of the in-situ stress can be obtained by linear regression using the least square method. sH ¼ 0:03109H þ 1:69697
(16)
sh ¼ 0:02519H 1:54372
(17)
sv ¼ 0:0269H
(18)
Based on the above regression formula, the three principal stresses near the tunnel arch (the buried depth of 508.07 m) at the hole location are 16.89 MPa, 10.81 MPa, and 13.67 MPa. 5.4
Lateral pressure coefficient of in-situ stress
As can be seen from the calculation results in Table 2, the in-situ stress near the hole is mainly tectonic stress. The average ratio of maximum horizontal principal stress to vertical stress is 1.29. The minimum horizontal principal stress to vertical stress is 0.81. In this project, attention should be paid to preventing the instability of the side wall caused by large tectonic stress.
370
Table 2.
Lateral pressure coefficient of in-situ stress. Maximum horizontal principal stress
Minimum horizontal principal stress
Vertical stress Number Section depth (m) sv (MPa) sH (MPa)
sH/sV
sh (MPa)
sh/sV
1 2 3 4 5 6
1.18 1.41 1.37 1.30 1.23 1.27
7.95 9.10 10.10 10.30 10.81 11.99
0.75 0.83 0.84 0.80 0.79 0.85
5.5
389.87 410.27 450.31 468.11 505.09 524.35
10.57 10.97 12.06 12.86 13.67 14.11
12.52 15.43 16.48 16.71 16.86 17.98
Characteristics of in-situ stress
According to the regional geological structure data, the compressive stress of the current tectonic movement is mainly north-south and northeast. The average value of the measured maximum horizontal principal stress direction is N45.2 W, which is basically consistent with the stress field direction of the current tectonic movement in the region. 5.6
In-situ tensile strength of rock
Generally, the difference between the two is the in-situ tensile strength of the rock T. By calculation, the in-situ tensile strength of the rock in the project area is 1.30 MPa. The tensile strength of rock measured in a hole has a certain dispersion due to the different test depths and rock structure.
6 CONCLUSIONS AND SUGGESTIONS (1) The survey and analysis show that the tectonic stress in this area still dominates. The maximum principal stress is always the first principal stress in the north-south and northeast directions. The relationship between the three principal stresses is as shown:s H>sV>sk. The maximum horizontal principal stress is 12.52–17.98 MPa.The minimum horizontal principal stress is 7.95–11.99 MPa. The direction of maximum horizontal principal stress is N45.2 W, and the result is basically consistent with the stress field direction of the present tectonic movement in the area. The test results of the hydraulic fracturing method are in good agreement with the field test results. (2) The in-situ stress level in this area is relatively high. In the process of tunnel construction, there may be some engineering problems such as slight rock bursts, side wall rock stripping, falling block, local collapse along the joint, and poor cavity formation. It is suggested to take comprehensive measures such as advanced geological exploration, prereinforcement, and presupport during tunnel excavation to reduce construction risks. (3) According to the relationship between stress and depth, the stress near the surrounding area almost increases with depth. The distribution is regular of geostress and can be obtained by linear regression with the least square method.
ACKNOWLEDGMENTS This research was funded by the “Research Fund of Yunnan Provincial Department of Education, grant number 2022J1305. 371
REFERENCES Cai Meifeng, He Manchao, Liu Dongyan. Rock Mechanics and Engineering[M]. Beijing: Science Press, 2013:129–177. Cai Mei-feng. Review of Principles and Methods for Rock Stress Measurement[J]. Chinese Journal of Rock Mechanics and Engineering,1993, 12(3): 275–283. Cai Meifeng. Studies of Temperature Compensation Techniques in Rock Stress Measurements [J]. Chinese Journal of Rock Mechanics and Engineering, 1991,10 (3): 227–235. Hou Kui-kui, Wu Qin-zheng, Zhang Feng-peng, et al. Application of Different in-situ Stress Test Methods in the Area of 2005 m Shaft Construction of Sanshandao Gold Mine and Distribution Law of in-situ Stress[J]. Rock and Soil Mechanics 2022, 43(04):1093–1102. DOI: 10.16285/j.rsm.2021.1172. Ma Chun-de, Chen Yu-min, Chen Jiang-zhan, et al. T/CGA 026—2021 In-situ Stress Test of Rock Mass in Gold Mine by Borehole Core Acoustic Emission Method[S]. Beijing: China Gold Association, 2021. Ma Chun-de, Liu Ze-lin, Xie Wei-bin, et al. Comparative Study on the in-situ Stress Measurement in Deep Xincheng Mining Area by the Bush Hole Stress Relief Method and Acoustic Emission Method [J]. Gold Science and Technology, 2020, 28(3): 401–410. Meng Nannan. Geostress Measurement Method and Research [D]. Inner Mongolia University of Science and Technology, 2015. Pei Shu-feng, Zhao Jin-shuai, Yu Huai-chang, et al. Inversion Method for Local in situ Stress Considering Stress-induced Damage of Caverns Surrounding and its Application[J]. Rock and Soil Mechanics, 2020, 41 (12):4093–4104. Sun Dong-sheng, Chen Qun-ce, Zhang Yan-qing. Analysis of the Application Prospect of ASR in-situ Stress Measurement Method in Underground Mine [J]. Journal of Geomechanics, 2020, 26(01): 33–38. Sun Dong-sheng. Experimental Study of Anelastic Strain Recovery in-situ Stress Measurement Methods and its Application[D]. Beijing: Chinese Academy of Geological Sciences, 2014. Sun Dong-sheng, Chen Qun-ce, Zhang Yan-qing. Analysis on the Application Prospect of ASR in-situ Stress Measurement Method in Underground Mine[J]. Journal of Geomechanics, 2020, 26(1): 33–38. Wang Bin, Li Xi-bing, Ma Chun-de, et al. Study of Forecast of Rock Burst Based on Three Dimensional insitu Stress Measurement[J]. Rock and Soil Mechanics, 2011,32(3): 849–854.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Cable forces and optimization of construction process on composite girder of cable-stayed bridge Ersen Huang*, Hongjun Ke*, Zhuoyi Chen* & Huanhuan Hu* School of Civil Engineering, Changsha University of Science & Technology, Changsha, China
ABSTRACT: This essay uses the Pingnan Xiangsizhou Bridge, a double-tower, double cable plane steel-concrete composite girder cable-stayed bridge in the Guangxi Zhuang Autonomous Region. Double main girders were employed in the finite element models of the full-bridge spatial rod system to simulate the steel-concrete composite girders. The initial reasonable completion state is determined using the minimum bending energy method. The bridge cable force is then iteratively corrected using the unknown load coefficient method. By pouring three wet joints of the main girder segments at once, the three-section cyclic construction optimizes the building of the Xiangsizhou Bridge. Based on this, the forward iteration method is employed to identify the cable-stayed bridge’s reasonable construction state. The findings indicate that four months can be cut from the bridge’s construction duration when the construction schedule is followed by optimization. Throughout construction and under the most unfavorable load combinations in the state of the completed bridge, the maximum tensile and compressive stresses in the concrete slab and steel main girders did not exceed the specification limitations. The cable force value and the reasonable completion state match excellently, with a maximum error of 1.42%.
1 INTRODUCTION Combining the concrete deck slab and the steel main girders with a shear stud, the composite girder cable-stayed bridge fully utilizes the steel girders’ compressive and tensile properties. Instead of employing an orthotropic plate for the upper portion of the steel girders under pressure, concrete deck slabs are used, reducing costs and improving the bridge deck’s performance (Fragiacomo et al. 2004; Fatemi et al. 2016). The traditional method of the building combined girder cable-stayed bridges involves lifting the main girder section, hanging the stay cables and tensioning them once, pouring the wet joints, and tensioning the cables twice. However, construction is not very efficient, and it takes some time to complete. The combination girder cable-stayed bridge currently uses a double-section cycle construction that is more effective. The stress distribution of the steel main girders and deck slabs of composite girder cable-stayed various construction methods will impact bridges. The multisectional cyclic construction will make the concrete deck slabs more prone to cracking. These factors will also have an impact on the bridge cable force, which will have an impact on the stability and longevity of the structure. Determining the reasonable completion states and reasonable construction state (Zhou et al. 2019) corresponding to the multi-segment cyclic construction scheme while ensuring the safety of the composite girder cable-stayed bridge’s
*Corresponding Authors: [email protected], [email protected], [email protected] and [email protected] DOI: 10.1201/9781003425823-47
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construction and operation periods is, therefore, a critical technical problem for this type of bridge under the tight schedule (Ahmed et al. 2021). The procedure of building the main girders significantly impacts the stress state of the deck slab and steel main girders when the bridge is created. The timing of the deck slab, steel main girders, and the wet joints during casting maintenance time all affect the structure’s ability to withstand stress. The cable-stayed bridge’s construction cable force should be adjusted by the optimization of the main girder erection process during the bridge’s construction (Fabbrocino et al. 2017), as this will have an impact on the reasonable completion state of the cable-stayed bridge (Yan 2001). The impact of construction process optimization on composite girder cable-stayed bridges’ structural forces and cable force has been the subject of extensive investigation in recent years. Wang et al. (1997) investigated the effect of cable preloading on structural deformation and stress in bridges using four different techniques. To compute the initial cable forces of cable-stayed bridges under dead load, Chen (1999) devised the force equilibrium approach, which makes the distribution of bending moments on the bridge deck more logical. To estimate the best post-tensioning force for the structure under bridge-forming conditions, Hassan et al. (2012) suggested a novel approach. This novel method successfully minimizes the deck’s vertical deflection and the pylon’s horizontal deflection. Atmaca et al. (2022) examined how different cable arrangement shapes and cable forces affected the structure’s ability to withstand forces. The influence of concrete shrinkage creep at the post-cast zone of steel-concrete composite beams on the durability of bridge deck slabs was studied by Huang et al. (2019). Hu et al. (2016) investigated how various hysteresis casting techniques affected the structural forces acting on concrete bridge deck slabs. For the combination girder cable-stayed bridge, Qi et al. (2017) investigated using a multi-sectional cycle construction scheme, which may greatly increase construction efficiency and significantly reduce construction time. In conclusion, most current research for the direction of the combined girder cable-stayed bridge concentrates on the effect of the cable force on the structure of the cable-stayed bridge and the research on the procedure for constructing the wet joints of the bridge deck slab. At the same time, there needs to be more research on the three-section cycle construction scheme for the main girder of the cable-stayed bridge. The Guangxi Liyu Expressway Pingnan Xiangsizhou Bridge, which has a main span of 450 meters, is used as the basis for this paper’s analysis of the issues associated with threesection cyclic construction in cable-stayed bridges and the solutions that follow. Studying the effect of the main girder’s three-section cyclic construction processes on the internal force of the bridge structure will help future bridges of the same kind by serving as a benchmark.
2 THE XIANGSIZHOU BRIDGE The main span of the Xiangsizhou Bridge is (40+170+450+170+40) m, and it is a doubletower, double cable plane, semi-floating composite girder bridge. The bridge elevation layout is shown in Figure 1. The bridge deck is 33.5 m wide, the top slope is 2% in both directions, and the main girder is a divided double-box main girder section, as shown in Figure 2. There are 89 girder sections totaling 2042=160 low relaxation high strength main girders, divided into 15 types and numbered AJ. A floating crane lifts the girder sections close to the cable tower and auxiliary pier, while the bridge deck crane lifts the standard girder sections. The fatigue stress amplitude must be at least 200 MPa, and the tensile strength must be 1860 MPa. According to various cable forces, 15.2-37, 15.2-43, 15.2-55, 15.2-61, 15.2-73, and 15.2-85 are used. The main span’s longitudinal anchorage points are spaced apart by 10.8 meters, and the side span’s anchorage points are separated by 10.8 meters plus 57.2 meters. The main tower adopts diamond-shaped towers, and the total height of both towers above the tower base is 147.3 m. The upper tower columns are situated near one another on the 374
Figure 1.
Arrangement of the main bridge of the Xiangsizhou Bridge (Unit: m).
Figure 2.
Cross section of the main girder of the Xiangsizhou Bridge (Unit: cm).
inner side, the upper crossbeam is positioned at the top of the tower, and a middle beam is positioned around 51.4 m from the lower crossbeam. A 500 mm gap is left at each longitudinal end of each section as a post-cast joint between the sections, and C55 expansive concrete is used. The steel used for the main beam is Q345GJ, and the concrete used for the deck slab is C55 high-performance concrete. The steel girders are preassembled to form the composite beam. The precast girder sections needed to be stored for over two months to minimize the effect of concrete creep and shrinkage on the construction.
3 FINITE ELEMENT MODEL In this paper, a finite element model of the spatial rod system of the Xiangsizhou Bridge was established using MIDAS CIVIL finite element software, as shown in Figure 3. The 1884 units in the finite element model are divided into 1724 beam units and 160 tensile-only truss units. Double main girders are used to simulate the steel-concrete composite girder, which
Figure 3.
Spatial finite element model of the Xiangsizhou Bridge.
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means that beam units are used to simulate the steel main girders and the bridge deck, respectively, and stiff jointing in the elastic jointing are used to simulate the shear connection between them. Only the tensile truss unit represents the cable-stayed girder, while the spatial beam unit simulates the main tower, auxiliary pier, and transition pier. Table 1 displays the primary material parameters for the complete bridge model.
Table 1.
Main material parameters.
Structure
Materials
Elastic Modulus (MPa)
Steel beam Tower Bridge deck Cable
Q345GJ C50 C55 Steel strand
2.06 3.45 3.55 1.95
105 104 104 105
Unit Weight (kN/m3) 78.5 26.5 26.5 91
4 DETERMINATION OF REASONABLE COMPLETION STATE For the Xiangsizhou Bridge, a combination of the minimum bending energy method (Liang et al. 2003) and the unknown load coefficient method (Wu 2014) is employed in this research to calculate the reasonable state of the cable-stayed bridge. To get the preliminary results of the bridge state of the cable-stayed bridge, the method first reduces the flexural stiffness of the main beam and main tower by 10, 000 times and calculates integral cast construction. It then iteratively corrects the cable forces using the unknown load coefficient method to get the reasonable completion state of the cable-stayed bridge. Figure 4 depicts the main beam’s bending moment diagrams in a reasonable state.
Figure 4. state.
Bending moment diagram of the main girder of the Xiangsizhou Bridge in a reasonable
The main tower’s maximum bending moment is 6.02104 kN at the tower root, and the main beam’s maximum bending moment is 2.69104 kN at the transition pier, according to Figure 4, which shows that the bending moment distribution of the main tower and main beam is more uniform when using the above comprehensive method. Figure 5 depicts the overall distribution of the stay cable force. The maximum cable force is 6283.3 kN, located at the side-span tail cable B20, to play the anchoring role. The distribution of stay cable essentially follows the law of gradually increasing from the main tower to the far side, and the safety factor corresponding to the maximum cable force of each stay cable under the combined load is close.
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Figure 5.
Cable force distribution at reasonable completion state.
5 FORCE ANALYSIS DURING THE CONSTRUCTION PHASE 5.1
Construction process optimization
The various construction methods for the composite girder cable-stayed bridge not only significantly affect the overall internal structural force of the bridge but also dramatically affect the time required for construction and the number of materials used. The construction period for the entire bridge is constrained due to the location of the Xiangsizhou Bridge, which is constrained by hydrogeological conditions, opening time requirements, and engineering economic aspects. The main girder erection period determines the duration of the bridge. Consequently, a major challenge the bridge must deal with is how to shorten the construction period by optimizing the construction process while still ensuring construction quality, safety, and reasonable structural stresses. There are generally three ways to expedite the construction of a stacked girder cablestayed bridge: First, fewer tensioning and cable transfers once the cable-stayed is tensioned in place; the second one is a shorter maintenance period following the casting of concrete for the wet joints of the bridge deck; and the third one is the use of double-section cycle construction or multi-section cycle construction mode, a one-time casing. As of now, researchers have studied the above three concepts, but there are still flaws in them: For the first method, compared to the stay cable in stages of multiple tensioning, the stay cable in place at one time tensioning method is likely to lead to excessive tension stresses and cracks in the bridge deck, the structure is not conducive to stress and durability; for the second method, the short maintenance time is likely to lead to concrete cracking, and will lead to the bridge deck and the steel main beam cannot be well-formed as a whole; and for the third method, double-section cyclic construction, has been employed in several largespan composite girders cable-stayed bridges in China. The stresses in the steel main girders in the joint section are too high, and the tensile and compressive stresses on the deck plate may be surpassed for the additional three-section cyclic construction. As a result, some appropriate steps should be taken to lessen the stress on the steel main girders and deck plate. According to the Xiangsizhou Bridge’s design specifications, C55 concrete is used in the wet joint of the bridge deck slab. Before continuing the construction process, the wet joint’s concrete strength must be at least 90%. By analyzing the main girder section’s construction cycle, a single-section cycle takes ten days to complete, of which 5-7 days are needed to maintain the wet joints. During this time, no other operations are permitted on the bridge deck to ensure the strength of the concrete. After careful thought and much finite element simulation work, three-section cyclic construction was utilized to optimize the construction approach. Figure 6 depicts the construction steps.
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Figure 6.
Three-section cycle construction steps.
It is necessary to tension the cables on the n+2# girder section for the third time after the first tensioning of the n+3# cable-stayed girders in the steel-concrete composite girder cablestayed bridge to ensure that the tensile and compressive stresses in the deck plates and steel main girders of the completed girder sections are not exceeded. To prevent the lifting of the n +4# girder section from increasing the deck plate tensile stress of the n+2# girder section above the allowable value, tensioning is used to increase the positive bending moment and the corresponding girder section, and to increase the compressive stress reserve of the corresponding girder section. This cable adjustment step depends on identifying a feasible completion state for a steel-concrete composite girder cable-stayed bridge with three section cycles. The three-section cyclic construction (three wet joints between four sections of deck slabs are simultaneously molded, reinforced, concrete poured, and strength pending) can save two sections of main girder construction time, or at least 20 days, compared to the single-section cycle construction for every three consecutive girder sections (one wet joint between two sections of deck slabs is molded, reinforcement tied, concrete poured, and strength pending, which usually takes ten days). The main bridge must be finished on schedule. Thus, it is critical that the entire bridge, which consists of 20 girders sections, saves 120 days of construction time.
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5.2
Reasonable construction state
The main girder sections of large-span cable-stayed bridges are typically tensioned multiple times throughout the construction process and transformed by multiple systems to reach a reasonable bridge state. Large-span cable-stayed bridges are typically constructed symmetrically with double cantilevers. The key to the construction control calculation for this kind of bridge is how to calculate the intermediate cable force. To accomplish this, the forward iteration method (Yan 1999) is used in this paper. The initial cable force value can be calculated using the weight calculation of the main beam section. This group of cable force is then used as the initial tension of the cable to be substituted into the model for the calculation of the normal assembly. The initial tension is then corrected using the least squares method (Shyamal et al. 2018) based on the discrepancy between the result and the target value. Then the obtained cable forces are substituted into the model to calculate the forward iteration, and iterations are repeated until the model converges. Among them, the non-stress cable length (Qin 2003) regulates the final tensioning of each stay cable rather than the tensioning force. 5.2.1 Structural internal force analysis during the construction phase The forward iteration method was used to optimize the steel main girders of the Xiangsizhou Bridge for three-section cyclic construction. The results are shown in Table 2 as the maximum tensile, compressive stresses in each section of the steel main girders and their deck slabs during the whole construction phase. Table 2 shows that the steel main beam’s maximum tensile stress is 86.09 MPa, and its maximum compressive stress is 205.85 MPa, within the specification limits. The maximum tensile stress of the bridge deck slab is 1.65 MPa, and the maximum compressive stress is 13.39 MPa during construction.
Table 2. Maximum tensile and compressive stresses in each section of the steel main girders and deck slabs during the whole construction process. Deck slab stress (MPa) Beam section number
Tensile stress
1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20
0.41 0.39 1.64 1.45 1.36 1.16 1.36 1.57 1.51 1.65 1.25 0.97 0.89 1.24 0.97 0.96 1.21 0.03 1.17 1.46
Compressive stress 0.42 0.26 0.66 0.83 2.75 4.18 5.69 6.74 5.94 6.35 6.72 6.94 7.33 7.53 7.93 12.48 12.23 12.53 12.94 13.39
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Steel beam stress (MPa) Tensile stress 22.29 15.29 74.39 81.86 63.46 85.85 86.09 84.97 64.04 55.62 60.14 64.96 54.49 18.83 56.51 53.61 36.92 55.96 50.86 15.25
Compressive stress 64.98 26.31 124.68 144.10 64.54 138.22 156.98 66.24 135.59 192.24 76.40 149.63 205.85 85.41 153.30 200.00 100.23 158.36 204.11 119.17
5.2.2 Structural internal force analysis in the bridge completion stage Table 3 displays the maximum compressive stresses of the steel main girders and deck slab for each girder section in the bridge completion stage as determined by the forward iteration method.
Table 3. Maximum compressive stresses in each section of the steel main girders and deck slabs at the bridge completion stage. Beam section number 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20
Deck slab compressive stress (MPa) 5.27 5.16 4.92 4.64 4.26 3.99 3.80 3.50 5.29 7.65 8.41 8.27 9.63 5.56 4.21 2.71 3.26 3.79 5.93 4.49
Steel beam compressive stress (MPa) 136.14 149.21 153.06 152.28 177.53 166.60 137.57 162.81 163.59 122.19 181.91 182.68 132.87 161.84 163.40 121.93 156.47 158.47 126.50 189.73
Table 3 shows that the steel main girders and bridge decks slab are only subjected to compressive stresses, with the maximum compressive stress of the steel main girders being 189.73 MPa and the maximum compressive stress of the bridge deck slab being 9.63 MPa. These values meet the specifications. After considering the most unfavorable load combinations during the operation phase, the maximum tensile and compressive stresses of the steel main girders and bridge deck slab also meet the specifications. Table 4 displays the discrepancy between the bridge force during construction and the reasonable completion state force computed using the comprehensive method. Table 4 compares the bridge force values acquired using the forward iteration method and the reasonable bridge force values obtained using the comprehensive method. The maximum difference is 40.11 kN, and the relative difference is just -1.42%. The cable force distribution between the two is reasonable, and the change rule is the same.
6 CONCLUSIONS The composite girder cable-stayed bridge construction process is intricate and fraught with technical challenges, and the reasonable completion state is closely related to the construction process. Based on the double-tower, double cable plane composite girder cable-stayed bridge of Xiangsizhou Bridge in Pingnan, Guangxi, the finite element simulation research of the entire bridge structure was conducted in this work using MIDAS CIVIL, a finite element computation software. 380
Table 4. Difference between reasonable completion state and construction stage cable force comparison. Cable number at side span
Reasonable completion state cable force (kN)
Construction stage cable force (kN)
B1 B2 B3 B4 B5 B6 B7 B8 B9 B10 B11 B12 B13 B14 B15 B16 B17 B18 B19 B20
2826.71 2838.46 2676.58 3463.52 3773.72 3265.97 3414.27 3304.18 3680.29 3954.56 4421.20 5136.19 4905.55 5694.06 5416.82 5665.77 5161.80 5814.35 5737.61 6283.33
2818.18 2832.90 2675.34 3466.69 3780.18 3274.62 3424.41 3318.70 3696.24 3971.91 4441.88 5162.38 4932.81 5726.54 5444.92 5693.99 5190.05 5847.14 5770.19 6315.66
Differences (%) 0.30% 0.20% 0.05% 0.09% 0.17% 0.26% 0.30% 0.44% 0.43% 0.44% 0.47% 0.51% 0.56% 0.57% 0.52% 0.50% 0.55% 0.56% 0.57% 0.51%
Cable number at the main span
Reasonable completion state cable force (kN)
Construction stage cable force (kN)
Z1 Z2 Z3 Z4 Z5 Z6 Z7 Z8 Z9 Z10 Z11 Z12 Z13 Z14 Z15 Z16 Z17 Z18 Z19 Z20
2793.49 2903.02 2820.68 3738.49 4404.27 2666.30 3173.88 3672.86 3513.87 3777.32 4563.95 4472.48 5282.34 5373.30 5535.90 5619.22 6187.87 5452.20 5825.14 5987.37
2820.95 2938.89 2860.79 3781.52 4438.03 2687.98 3183.60 3672.53 3504.05 3761.75 4543.96 4452.65 5264.66 5356.12 5523.17 5610.69 6182.76 5449.41 5823.43 5985.52
Differences (%) 0.98% 1.24% 1.42% 1.15% 0.77% 0.81% 0.31% 0.01% 0.28% 0.41% 0.44% 0.44% 0.33% 0.32% 0.23% 0.15% 0.08% 0.05% 0.03% 0.03%
To establish its reasonable completion state, the finite element model employs a thorough methodology that combines the minimum bending energy method and the unknown load coefficient method. It optimizes the construction of the cable-stayed bridge as it is being constructed and uses the forward iteration method to reach a reasonable construction state for the cable-stayed bridge. The construction process adopted a three-section cyclic construction program, pouring three wet joints at a time, saving two wet joints construction time and speeding up the construction progress. The main girder’s construction shortened the construction timeframe by 120 days, which is crucial for completing the main bridge’s closure on schedule. It is necessary to tension the stay cable on the n+2# girder section for the third time after tensioning the n+3# stay cable for the first time in the three-section cycle construction to ensure that the tensile and compressive stresses in the deck slab and steel main girders of the completed girder section do not exceed the specification limits. To prevent the deck slab tensile stress of the n+2# girder section from exceeding the limit value during the lifting of the n+4# girder section, it is necessary to increase the positive bending moment of the corresponding girder section and the deck slab compressive stress reserve of the corresponding girder section. The determination of a feasible completion state for a steel-concrete composite girder cable-stayed bridge with a three-section cyclic construction depends on this stage of stay cable adjustment. According to the simulation calculation results, the bridge deck slab’s maximum tensile stress during construction was 1.65 MPa, which meets the specification requirements. The bridge has been successfully established and opened to traffic. The measured bridge alignment and internal forces are in excellent accordance with the predicted values, according to the optimized three-section cycle scheme. This paper’s methodology can serve as a reference for projects using the same type of bridge. 381
ACKNOWLEDGEMENT Fund: Program of National Natural Science Foundation of China. Item Number: 51708047.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
The design and optimization of the building electromechanical system based on BIM technology Huize Wu* School of Electrical and Mechanical Engineering, Wuhan University of Technology, China
ABSTRACT: Along with the rapid development of the national economy, China’s urbanization has also accelerated, and under this environment, the construction industry has also entered a rapid development stage. With the development of construction, many branches have gradually emerged, such as mechanical, electrical and mechanical, which is the future development trend of the country; therefore, the construction and development of electromechanical projects have become an urgent need. This paper focuses on the design and optimization of BIM technology in engineering mechanical structure for in-depth discussion and analysis.
1 INTRODUCTION The rise of building electromechanical engineering has accelerated the urbanization process in China, providing great convenience to people’s life and production, and promoting the economic development of China. However, due to the late start of China’s mechanical and electrical engineering construction, there is a gap with the international advanced level, and there are many shortcomings. The emergence of BIM technology has brought new hope for China’s construction mechanical and electrical engineering technology; the reasonable application of this technology can effectively improve China’s construction mechanical and electrical engineering and provide quality products for the general public (Song).
2 RESEARCH BACKGROUND In today’s rapid development of modern technology, Internet figures have been gradually taken into account. The construction industry has experienced thousands of years of development, but so far, the development of the construction industry is far behind this era, rework, delays, waste and other phenomena are common, resulting in the whole project efficiency being very low, and the CAD design level is also very low. It can be used by developers, designers, constructors, and even the final owner to “simulate and analyze” the entire construction process (Chen 2018). The application of this model for the design of buildings, construction and construction of integrated management, is to achieve cost savings and improve the efficiency of the purpose. The information construction of the building is essentially the use of a computer to build a three-dimensional model library, which is a parametric reflection of the material and functional characteristics of the project, which covers the entire life process of the project, including the entire process of the project, including all aspects of the project, including all aspects of the project, including all aspects of the project. The engineering machinery category mainly includes HVAC systems, water supply and drainage systems, and strong electricity and *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-48
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weak electricity system. And in the traditional design system, there is a lack of effective communication between designers in various fields, not to mention collaboration. If in the project, once different pipes are found to be in conflict, it will result in large-scale disassembly and cause the overall project to be reshaped (Wang 2016).
3 RESEARCH OVERVIEW 3.1
Development status
Yangke pointed out that MEP (i.e., HVAC, electrical, water supply and drainage) plays a pivotal role in the design of the building. In the current construction process of construction equipment, the 3 disciplines are independent of each other, and different disciplines make different designs on different pipes, which inevitably encounter some problems, which are usually done by hand. Bim technology is an efficient multidisciplinary collaboration method. By comparing MEP with BIM combined with MEP, the superiority of BIM technology in MEP is illustrated, and the simulation of air conditioning return air using Revit MEP is highlighted. And the importance and development trend of BIM technology applied to construction projects is discussed. Lv Xiaobiao said that the integration of piping in MEP majors (electrical, HVAC, water supply and drainage), due to the type and number of various piping, leads to errors, leaks and touches in the project; due to the existence of serious defects, it is often necessary to solve piping conflicts artificially in the plane, which causes the reduction of project efficiency, thus making the optimization of secondary piping project drawings more tedious and cumbersome. The MEP secondary optimum based on BIM technology is a major technological innovation at present. The purpose of this paper is to explore a new idea of MEP optimization design based on BIM technology, using RevitMEP as an important research platform for this system, and using MEP collision detection combined with multi-professional collaborative design to achieve the optimization of the pipeline as a whole. The quality of work of the product is improved. Based on the MEP collision detection technology, effective conflict points were identified and secondary optimization was carried out through manual screening. Table 1.
Difference between BIM and CAD in design. CAD
BIM
Basic elements
Point, line and plane have no professional Walls, doors, windows, etc. have geometric significance characteristics, architectural physical characteristics and functional characteristics Elements To change the location, size or other Parameterization of building components, and Fainformation of an element, you need to with building properties attached. Under milies draw a picture again or adjust the size the concept of “family”, you can adjust the through the stretch command size, style, material, color, etc. of components only by changing the properties Component There is no correlation between building Components are related to each other relationship components Design If you make a modification on the plane, all If you make a change, the plan, elevation, modification other faces need to be modified manually. section, 3D view, schedule, etc. will be If you do not operate it properly, you will automatically modified, realizing one have a low-level error of inconsistent views change and every change from different angles Plot The construction information provided is It contains all the information of the building limited, and only the paper drawings are and provides visual two-dimensional and electronic. Non-professional people cannot three-dimensional drawings to facilitate the read the drawings communication and collaborative work of all departments of the project
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3.2
BIM concept
An article on Building Information Modeling and Collaborative Work Environments by Antonio Gallo and Ricardo Jadim Goncalves points out that the modernization of the workplace has been a subject of scientific research and innovation. The biggest challenge is to achieve real innovation in the workplace, overcoming a multitude of obstacles and obstacles of all kinds, including human, institutional, social and technological. These so-called next-generation collaborations will transform existing practices by improving communication between people, visualization of creation, knowledge support, and interaction with nature, thus making existing practices more competitive, productive; and creative and collaborative ways of working. Innovation in the workplace is important, and the essence of the practice is to collaborate in a knowledge-filled, functionally diverse work environment (Cai 2022). The important stages in the cycle are as follows: from planning to planning, for example, for development zones in the public sector: design, architects, and engineers working with owners to develop construction projects: construction, the collaboration of many contractors working on different objectives and reporting progress to owners: facility management, the conclusions reached after operators rely on information items. With the development of computer-aided design technology, it is possible not only to process vector information, but also to enrich three-dimensional models of buildings and structures, and to simulate the virtual environment of a project. This is a major trend commonly referred to as BIM. Dr. Eastman refers to the concept of building information modeling in “The BIM Handbook: Guidance for Owners, Building Information Model Managers, Designers, and Contractors,” where he states that the integration of all geometric shapes and information about roles and behaviors into a completely related-life cycle of the construction project. It also contains information about the fabrication process to address the construction process and production flow. On the official website of the National Standard National Industry Management System (NMBMS), it is noted that in the transformation of construction technology from digital electronics using simulated drawings and text, to the application of Building Information Modeling (BIM), and the automotive industry. The early concept, which saw BIM as a simple three-dimensional model of a device, was a potential, interoperable basis for building information modeling processes and tools, as well as modern ways of communication. NBIMSBIM is a definition of a digital representation of the physical and functional characteristics of a device. Therefore, it is a facility associated with a knowledge resource that constitutes a reliable basis for determining its cycle of existence (Wu 2022).
4 BIM TECHNOLOGY IN THE MAIN APPLICATION OF BUILDING MECHANICAL AND ELECTRICAL ENGINEERING (1) In the process of program design, first, we cooperate in designing various disciplines. Construction electromechanical engineering has a great systemic nature and involves many disciplines. Using BIM technology for the scheme design of electromechanical integration, it can make reasonable planning for drainage, HVAC, electrical, etc. at the construction site according to the specific conditions of the site; fire protection, building intelligence, etc. to ensure the scientific nature of the construction and construction of the whole project and make the design of the whole electromechanical integration more reasonable. In practice, we fully collect all kinds of information, figure out the specific application characteristics and conditions of the design, use BIM technology to optimize the design of mechanical equipment, and improve the scientific and reasonable design work of the whole electromechanical integration. The second line of impulse detection and layout of the best configuration. As the integration characteristics of construction machinery and machinery are more prominent, the detection of various pipelines and bridges becomes an important part of the construction. Since various pipeline and bridge sizes cannot be precisely shown on the plane, it is difficult to react to the danger and hazard of collision in time, which increases the difficulty of pipeline layout. And the use of BIM technology 385
means can change the previous work mode, and build a three-dimensional modeling management mode, through the model function to achieve the implementation of the full range of conflict detection; when encountering problems, it can be solved by changing the arrangement of piping, etc. Thus integrating the modeling of each profession into the BIM system; this method can detect problems at an early stage of the project, and can well prevent rework and other situations. Essentially improve the quality of the entire building’s mechanical and electrical engineering construction (Zhang 2019). (2) Engineering construction. The first is to establish the construction deepening model. The construction deepening model is the preliminary work carried out before the official start of construction, construction management personnel according to the construction stage drawings, and model results, combined with construction facilities and equipment, construction organization plan, construction sites, and other factors, the model results for indepth and adjustment, to ensure that the information contained in the model can meet the requirements of the construction process, management processes, etc., and finally by the construction units to verify the model and finally available for construction units’ reference. Second, the schedule of the project. Using BIM technology, the relationship between the construction schedule and construction components can be changed, so that a dynamic connection between the construction schedule and construction components is created. Using the dynamic simulation method, various process methods can be comprehensively compared and feasibility analysis can be conducted so that a scientific plan can be formulated. Using BIM technology, it is possible to track the actual progress of the project in real-time, make full use of the existing construction resources, make reasonable deployment of each process, and after understanding the actual progress of the project; compare it with the planned progress, summarize the reasons for deviations, and take targeted countermeasures to adjust the project progress; ensure that the construction work of all processes is carried out according to the planned progress. Third, the simulation application of construction (Liu 2021). Through BIM technology, construction simulation on site can be realized to optimize each process in advance and establish a scientific construction plan. Simulation is essential in complex pipeline systems to ensure construction quality, improve construction efficiency, and maximize limited engineering resources; to speed up the project, multiple production lines are being built at the same time.
5 DESIGN AND OPTIMIZATION OF BUILDING ELECTROMECHANICAL SYSTEM BASED ON BIM TECHNOLOGY 1. Optimal layout. After accurate BIM design, the cross conflicts between different professions can be greatly reduced, and at the same time, the construction units of each profession adopt the model-based construction plan and construction sequence discussion, which can identify the problems more intuitively and clearly and solve them in advance, and also lay a solid foundation for digital construction and digital building in the later stage. Bim optimizes the space through collision detection in the design stage, and the reasonable layout of pipelines brings great added value to the owner. Without a reasonable space layout, it will result in spatial and visual overcrowding. Using BIM technology, different pipelines are reasonably laid out without affecting the original pipeline function and construction feasibility; the deepening designer can observe the pipelines from any angle and use the space optimization method to increase the original interior net height from 3.1 m to 3.45 m, thus improving the interior space utilization (Ding 2019). 2. Optimal pre-drilled hole arrangement. In the pipeline synthesis, the problem of reserved holes often arises, and the accurate positioning of the holes is usually done by the depth designer according to the imagination of space to draw the approximate hole position, which is prone to omissions and deviations, while the 3D visualization feature of BIM technology allows the BIM model to show the location of the holes to be reserved very 386
well. It can also realize the exact location, which effectively solves many problems encountered by the deepening designers when making holes, and at the same time, improves the quality of the drawings, saves the time of reprocessing, and greatly improves the efficiency of the drawings (Du 2014). 3. Optimize the arrangement of mechanical and electrical pipe supports. In the deepening design of a mechanical structure, the pre-burial and arrangement of the bracket is a very critical links. In areas where the pipeline condition is more complicated, the brackets are difficult to place and difficult to install. In the case that the profile is not cut, the bracket can be installed correctly and reach the ceiling elevation to meet the beautiful and neat construction requirements. From the viewpoint of construction, some brackets need to be preburied steel plates on the ground in construction projects, such as cold storage and other parts where there are more pipes, in order to carry the weight of the pipes, they need to be pre-buried on the ground; however, because the electromechanical pipes are not seriously considered, the positioning of specific parts cannot be precisely controlled, and the ground in a certain area is currently laid out The BIM model can simulate the arrangement of brackets, which can simulate the problems during the construction process in advance and determine the arrangement and placement of brackets precisely (Zhong).
6 CONCLUSIONS In conclusion, China’s construction engineering mechanical engineering construction has huge development potential, therefore, the use of this technology to accelerate the urbanization process in China will play a pivotal role. However, at present, China’s construction engineering technology is still immature and not yet in line with the world. In order to improve China’s mechanical and electrical integration construction process, the relevant departments can apply BIM technology to the whole building’s mechanical and electrical integration project to ensure the smooth implementation of the project and improve the quality of the project; provide better services to the people.
REFERENCES Cai Tian. Study on Optimization of Mechanical and Electrical Deepening Design of Assembly Building Based on BIM Technology [J]. China Construction, 2022(33). Chen Jia, Zhu Hao, Ding You, et al. Development of BIM-based Design Technology and System for Seismic Support Hanger of Buildings[J]. Engineering Construction and Design, 2018(23):3. Ding Qin. Research on Energy-saving Optimization Design of a Building Based on BIM Technology[J]. Science Popular: Science and Technology Innovation, 2019(2):3. Du Juan. Research on the Integration and Interaction Mechanism of Building Information Based on Ontology and Petri Network [D]. Shanghai University of Finance and Economics, 2014. Liu W, Tang Hao. Research on the Optimization Design of Hotel Building Engineering Based on BIM Technology [J]. 2021. Song Yu. Research on the Application of BIM Technology in Mechanical and Electrical Equipment Installation Engineering of Public Buildings [D]. Zhengzhou University. Wang Loyal, Wang Lei, Zhang Qiao. Application of BIM Technology Based on the Integrated Mechanical and Electrical Pipeline Layout of Metro Stations [J]. Information Technology of Civil Construction Engineering, 2016. Wu Cheng. Optimization of Mechanical and Electrical Deepening Design of Assembled Building Based on BIM Technology [J]. Intelligent Building and Urban Information, 2022(001):000. Zhang Sheng, Yue Hao, Chen Wen. Virtual Simulation and Force Research of Building Electromechanical System Based on BIM Technology[J]. Tianjin Science and Technology, 2019, 46(3):4. Zhong Zhi, Zhang Dong, Liu Jian. A BIM-based Prefabricated Production Method for Trench Type Piping System, CN109117510A [P].
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Application of Internet of Things and BIM technology in building intelligent operation and maintenance system Fengyi Han, Dejun Kong*, Kaixiang Wang, Fei Du & Zhihan Zhu BIM Institute of Technology and Industry, Changchun Institute of Technology, Changchun, China
ABSTRACT: With the development of science and technology, many new technologies are being applied to our daily life, including the Internet of Things technology. The new generation of information technology plays an important role in intelligent buildings. Compared with the traditional building operation and maintenance management system, the intelligent building operation and maintenance system has the characteristics of energy saving, active management, rapid response, and so on. This paper mainly discusses the application and development of the Internet of Things and BIM technology in the building operation and maintenance management system through the actual investigation and document analysis.
1 INTRODUCTION With the increase in income, people’s requirements for the living environment are also increasing, and more attention is paid to energy conservation, environmental protection, intelligence, comfort, etc. of buildings. In this context, BIM technology and Internet of Things technology have been applied to the building operation and maintenance management system, and the intelligent operation and maintenance system has developed rapidly. The main difficulty of intelligent building operation and maintenance systems are solving the problems of indoor building visualization and intelligent adjustment of building equipment. In view of these problems, experts and scholars have made certain research achievements. Based on the digital twin theory (Wang 2020), Wang et al. redesigned the building operation and maintenance management system to improve the operation and maintenance management effect. Chu et al. combined BIM technology with GIS technology (Chu 2020) to improve the efficiency of metro operation and maintenance. As a new generation of information technology, the Internet of Things plays an important role in the intellectualization of the building operation and maintenance system. Because it is necessary for us to have a correct understanding of the Internet of Things and BIM, and at the same time strengthen the depth of research on these two aspects to explore deeper value.
2 OVERVIEW OF INTERNET OF THINGS TECHNOLOGY AND BIM TECHNOLOGY 2.1
Internet of Things technology
The arrival of the 5G era has brought great changes to our lives and brought us more doubts. The Internet of Things has entered our lives as a new term. Many people’s first impression of *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-49
the Internet of Things is that Xiao AI Tong Xue. Indeed, smart speakers are part of the Internet of Things, but it is true that the Internet of Things is not a smart speaker that can sing and broadcast the weather. We usually think that the Internet of Things is an extension of the traditional Internet to the physical world. By connecting with real things, the Internet can better serve human beings. The concept of “The Internet of Things” was first put forward in the book “The Road to the Future” in 1995, and the concept of “Internet of Things” was formally put forward by the International Telecommunication Union (ITU) in 2005. Its development process is shown in Figure 1.
Figure 1.
Development history of the Internet of Things.
The Internet of Things is widely used in industry, agriculture, environment, transportation, logistics, and other fields, reflecting the extensive potential of the Internet of Things technology. At the same time, the combination of the Internet of Things and AI+IOT has also become an important development goal of the industry, which has played a great role in the intelligent development of these fields. Through the reasonable allocation of limited resources through the Internet of Things technology, the efficiency and benefits of the industry have been improved, and a series of related application scenarios have been generated, such as smart home, intelligent transportation (Zhang 2021), intelligent health care, and so on, as shown in Figure 2.
Figure 2.
Hotspot applications of the Internet of Things.
389
2.2
BIM technology
The Building Information Model (BIM) is the foundation of digital transformation in the construction, engineering and construction (AEC) industry. The traditional construction industry has also been continuously upgraded and optimized, mainly through two technological revolutions, as shown in Figure 3. BIM improves the accuracy of drawing, reduces the work difficulty of designers, and facilitates the storage and transmission of information. At present, BIM is developing rapidly in the domestic construction field. Through BIM, it is convenient to track the overall project and maximize the use of resources to achieve cost reduction and efficiency increase (Luo 2018).
Figure 3.
Two technological revolutions.
3 MODULE DESIGN OF INTELLIGENT BUILDING OPERATION AND MAINTENANCE SYSTEM The life cycle of building operation and maintenance is long and the amount of information is huge. The management efficiency of traditional operation and maintenance mode is low and cannot work together. Using BIM technology, Internet of Things technology and computer technology to build intelligent building operation and maintenance systems can solve the disadvantages of traditional mode. The mode of “BIM technology+information technology management+operation and maintenance management” is adopted (Bian 2019), as shown in Figure 4. Based on the BIM model, the operation and maintenance framework is designed, and the operation and maintenance management system is used to collect, store and share the data in the building, such as building information, energy consumption information, etc. The new system mainly includes the following four modules: visual display system, equipment monitoring system, safety management system and energy consumption management system.
Figure 4.
Design idea of operation and maintenance platform.
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3.1
Visual display system
The more refined the BIM model is, the larger the amount of model data will be, resulting in slow or even impossible loading of the model. In order to enable BIM to be used smoothly in various web browsers and mobile applications, it is necessary to lighten the model (Bai 2020) and optimize its appearance. In the visualization system, the three-dimensional structure of the model is restored with high-precision wireframe technology, the model is designed in depth with BIM technology, the model is rendered in combination with the calculation engine, and the authenticity of the model is restored. At the same time, the model is rotated and partially enlarged in the browser, which can clearly show the energy, environment, security, equipment, and other conditions in the building. In case of abnormality, the visualization of abnormal events needs to be clear, so the higher the degree of visualization, the lower the difficulty of operation and maintenance, and the higher the efficiency. 3.2
Equipment monitoring system
In the traditional system, the operation information of the equipment cannot be viewed remotely, which is not conducive to fault diagnosis and early warning. Through the combination with the Internet of Things, as shown in Figure 5, the operation status of the equipment can be grasped in real-time, so as to facilitate the recording and statistics of data and improve maintenance efficiency.
Figure 5.
Structure diagram of the equipment monitoring system.
In this module, you can view the status of the equipment in real-time, such as the operation of the elevator and the video picture, as well as the pressure and temperature of the boiler. To minimize losses, create a defect early warning algorithm that will alert the operation and maintenance staff to potential failures. The operation and maintenance personnel can set the parameters of the equipment remotely to save time. Through the Internet of Things technology, the dynamic monitoring of equipment can be realized (Zhang 2021), which provides a guarantee for the safe operation of the system. At the same time, advanced sensors can play their own role in emergencies to ensure the safety of people’s lives and property.
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3.3
Safety management system
Security management mainly includes a security module, access management module and fire control module. In the traditional operation and maintenance system, security, as a branch, operates independently and cannot interact with other systems in real time. When the Internet of Things technology is added, the intelligent operation of the whole system can be realized (Zhang 2019). If the security system detects an illegal intrusion, it will upload the alarm information to the main control unit, and communicate with other sensors to determine the time and location of the intrusion. The fire alarm device is connected to the sensor, which can actively close the gas valve and open the fire extinguishing equipment switch in case of a safety accident. The Internet of Things technology also has unique advantages in wiring (Yang 2022). Traditional security wiring is large and costly. The Internet of Things can use RFID to realize the positioning, identification, and data transmission of equipment, especially in areas with complex engineering environments (Guan 2021). 3.4
Energy consumption management system
With the promotion of the concept of “green economy” and “low carbon economy”, enterprises pay more attention to their own energy consumption level and have taken relevant measures to reduce energy consumption. We apply the Internet of Things technology to energy consumption management, carry out unified monitoring and management of water supply, power supply, gas supply, air conditioning, and other energy sources in the building, build itemized metering modules, write relevant algorithms and build energy consumption models, so as to achieve the goal of building energy refined management, fully grasp the realtime situation of gas consumption, water consumption and electricity consumption in the building, and improve the management quality. The purpose of reducing costs.
4 DEVELOPMENT MEASURES OF INTERNET OF THINGS TECHNOLOGY IN INTELLIGENT OPERATION AND MAINTENANCE The Internet of Things technology has promoted the intelligent process of the building operation and maintenance system and improved the overall management efficiency of the building. However, there are also some problems, such as negligence in personnel management and low stability of the equipment. Such problems and shortcomings cannot be ignored. In order to better promote the iteration of building intelligence, we need to take relevant measures for the application of Internet of Things technology: 1 The whole process of business tracking is as follows. The layout of the system shall be monitored throughout the whole process, and relevant emergency plans shall be formulated to deal with emergencies. 2 We keep the system iterative and upgraded, and adapt to the development of the information age by continuously improving the level of automation. 3 We improve personnel quality, train the personnel, and use the intelligent operation and maintenance system skillfully to reduce the probability of emergencies. 4 Regular maintenance of equipment. In order to ensure the stable operation of the system, the equipment must be inspected to ensure that all hardware equipment can work normally. 5 Information security. Relevant personnel is required to improve their awareness of security protection and account password level. At the same time, the database needs to be backed up regularly to ensure the stability and security of the system.
5 CONCLUSION The advent of the era of the Internet of Things will certainly become a “new inflection point” to promote intelligent architecture. To sum up, the Internet of Things and BIM technology
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have been widely used in the intelligent building operation and maintenance system, helping to improve the efficiency of the daily operation and maintenance management of the building, meet people’s daily needs, and to some extent ensure the safety of people’s lives and property. Therefore, it is necessary for us to focus on the Internet of Things and BIM technology and deeply explore the potential value of the Internet of Things, so as to promote the continuous upgrading of the building’s intelligent operation and maintenance system.
ACKNOWLEDGMENT This work was financially supported by the Science and Technology Development Plan Project of Jilin Province (20220508143RC).
REFERENCES Bai Xue, Ren Chenyu, Zhu Chaoping. Research on Lightweight of BIM Model in Signal Operation and Maintenance System [J]. Railway Communication Signal, 2020,56 (05): 52–54+58. DOI: 10.13879/j. issn1000-7458.2020-05.19538 Bian Jianzhong. Research on the Application of Internet of Things Technology in Building Intelligent Systems [J]. Science and Technology Innovation Guide, 2019, 16 (36): 139–140. DOI: 10.16660/j.cnki.1674098X.2019.36.139 Chu Jingyu, Xiong Ziming, Jiang Fengyu, Li Xianbing, Guo Yujing, Liu Xiaolong. Research on Intelligent Metro Operation and Maintenance System Based on BIM and GIS Data Fusion [J]. Information Technology and Network Security, 2020,39 (05): 75–79+85.DOI: 10.19358/j.issn.2096-5133.2020.05.015 Guan Jing, Guan Qingbao. Application and Development of Integration of Internet of Things and Edge Computing Technology in Smart City Intelligent Buildings [J]. Smart Building and Smart City, 2022 (11): 168–172. DOI: 10.13655/j.cnki.ibci.2022.11.053 Luo Jie. Exploration of the Trend of the Combination of BIM Technology, Internet of Things Technology and Construction Project Management [J]. Green and Environment-friendly Building Materials, 2018 (06): 172– 173. DOI: 10.16767/j.cnki.10-1213/tu.2018.06.126 Yang Yihao, Xiong Wenkang. Application of Internet of Things Technology in Intelligent Building Security [J]. Intelligent Building and Smart City, 2022 (11): 78–80. DOI: 10.13655/j.cnki.ibci.2022.11.023 Wang Yilei, Chen Ye, Wang Wen. Design and Application of Green Building Operation Cost Management System Based on Digital Twins [J]. Building Energy Conservation, 2020,48 (09): 64–70 Zhang Jianjun, Gao Zhigang. Application of 5G edge Computing Gateway in Vehicle Internet of Things [J]. Automation and Information Engineering, 2021,42 (04): 42–45+49 Zhang Yue. Discussion on the Design of Intelligent Park Operation and Maintenance System Based on BIM +IOT [J]. Information and Computer (theoretical version), 2021,33 (24): 146–149 Zhang Jianxun. Automatic Application of Security System in Intelligent Buildings [J]. China New Communications, 2019, 21 (19): 127
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Horizontal top displacement change patterns of the pile foundation in slope under freeze-thaw cycles Gaokai Lu* & Chunxiang Guo School of Civil Engineering, Lanzhou Jiaotong University, Lanzhou, Gansu Province, China
ABSTRACT: This paper analyzes the horizontal frost heaving force on pile foundations in slops in permafrost areas. By using the multi-physics field simulation software comsol, this paper analyzes the horizontal top displacement change patterns of the pile foundation in the slop in permafrost areas under freeze-thaw cycles. The numerical simulation results show that: the pile foundation in the slop in permafrost areas is affected by the cyclic influence of ambient temperature. Its horizontal top displacement change patterns have a shape similar to the periodic sine wave of ambient temperature. And the horizontal displacement of the pile in the downward direction along the slope is more obvious due to the horizontal frostheaving force from the slope foundation to the pile. This paper can provide a reference for the infrastructure construction in the slope area of permafrost areas.
1 GENERAL INSTRUCTIONS Permafrost is a mixture of ice and rocky soil below 0 C. According to incomplete statistics, China is rich in permafrost. There are about 2.15 million km2 of permafrost in China, accounting for more than 20% of the national land area (Zhou et al. 2022). In the QinghaiTibet Plateau region, the percentage of permafrost distribution is higher due to the higher temperature in summer and lower temperature in winter. The foundation in the permafrost zone will be affected by freeze-thaw cycles once a year, which is a severe risk to the safety of the foundation superstructure. Most of the studies on pile foundations in permafrost areas have focused on pile foundations in the flat area, while the research on piles on sloping foundations is less well documented. Under the effect of freeze-thaw cycles, the horizontal top displacement of the combination of pile and soil on the slope will be affected by various factors, including the stability of the pile-soil combination structure, the slope of the ground surface, the pile base and the characteristics of the soil (Campanella R. G. 1994; Campbell J. L et al. 2014; Jia 2021; Saeid EW 2021; Yu 2021). Among them, the stability of the pile-soil combination structure is the most important factor affecting the horizontal top displacement of the pile. If the stability of the pile-soil combination structure is weak, the horizontal top displacement of the combination of pile and soil on the slope will be greatly increased. Besides, the slope of the ground surface will also affect the horizontal top displacement of the combination of pile and soil on the slope. If the slope is steep, the horizontal top displacement of the combination of pile and soil on the slope will also increase. Moreover, the pile base and the characteristics of the soil will also affect the horizontal top displacement of the combination of pile and soil on the slope. If the soil characteristics of the pile base are unsuitable or the soil contains a
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-50
lot of water, the horizontal top displacement of the combination of pile and soil on the slope will also increase. This paper analyzes the horizontal frost heaving force on pile foundations in the slop in permafrost areas. By using the multi-physics field simulation software comsol, this paper analyzes the horizontal top displacement change patterns of the pile foundation in the slop in permafrost areas under the freeze-thaw cycle. This paper expects to provide a reference for the infrastructure construction and maintenance in slopes of permafrost areas.
2 ENGINEERING BACKGROUND OF THE PILE FOUNDATION ON PERMAFROST SLOPE As the main foundation form for buildings in permafrost areas, pile foundation directly determines the safety of upper buildings. With the “large-scale development of the western region” strategy and more permafrost projects on the Qinghai-Tibet Plateau, there will be more infrastructure construction in permafrost areas, which leads to a lot of research on pile foundations in permafrost areas. Most of the studies on pile foundations in permafrost areas have focused on pile foundations in flat areas, while the research on piles on sloping foundations is less well documented. The location of pile foundations in permafrost areas is not all flat, and damage to pile foundations located in slope areas occurs occasionally, which makes the study of pile foundations on slopes located in permafrost areas of great importance. The change of pile top displacement directly determines the safety of the upper building. Therefore, the study of the horizontal top displacement of slope piles is a key issue in the construction of pile foundations on the slope. The background of the permafrost pile foundation in the slop project mainly refers to the application of slope foundation in the special geological environment of permafrost areas, which is characterized by the same direction of surface slope and thawing ablation. As it is highly susceptible to the influence of surface slope and hydrological factors, therefore, a series of technical measures must be taken when constructing a pile foundation in slop. To this end, the pile foundation in slop can resist the effect of freeze-thaw cycles, thus ensuring the stability and slip resistance of the pile foundation in the slop.
3 MECHANICAL CHARACTERISTICS OF PILE FOUNDATION IN THE SLOP IN PERMAFROST AREAS At present, scholars have conducted more studies on the mechanical properties of pile foundation structures on permafrost slopes. The results show that in the study of structures in permafrost areas, the temperature has a great influence on slope foundations, i.e., the soil has a high thermal sensitivity, and temperature directly affects the soil properties and thus changes the pile forces by changing the state of water molecules. The soil temperature equation in the permafrost zone is a mathematical equation used to describe the soil temperature distribution according to the soil temperature distribution law for the soil temperature model divided in the permafrost zone. The equation is: T ¼ T0 þ Kz þ a1 sinð2pl1z þ j1 Þ þ a2 sinð2pl2z þ j2 Þ þ . . . þ an sinð2plnz þ jn Þ
(1)
where T0 = reference point of soil temperature; Kz = longitudinal temperature gradient; a1, a2, . . . , an = amplitude of longitudinal temperature; l1, l2, . . . , ln = longitudinal temperature wavelength; j1, j2, . . . , jn = longitudinal temperature phase difference, respectively. In addition, the temperature changes all the time throughout the year, which further leads to the complexity of the forces on the pile foundation. A pile on a sloping foundation
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exhibits mechanical properties that are different from those on flat foundations. As the soil around the pile is not at the same level in the non-flat area, the soil will have horizontal frostheaving force for the pile, as shown in the figure:
Figure 1.
Frost heaving forces on piles with different foundations.
4 FINITE ELEMENT MODELING AND CALCULATION In this paper, a roadbed section of a road section on the Qinghai-Tibet Plateau is selected as the research object. The soil material is sandy soil containing fine particles on the QinghaiTibet Plateau. According to the finite element theory, the permafrost -2 C soil with finegrained sandy soil is assumed to obey the elastic-plastic model with the Mohr-Coulomb yield criterion. The pile is a linear elastic model, and the pile material is a reinforced concrete pile. The model is shown in Figure 2.
Figure 2.
Geometric model.
This model of piled soil foundation has a length, width, and height of 30 m, 20 m and 40 m respectively, and the slope is 1:1.5. The pile diameter is 0.6 m and the pile length is 18 m, of which 13 m is buried in the ground and the rest is exposed on the surface. All of them are divided by a triangle mesh, and the whole model has 150, 000 cells in total. In the Mohr-Coulomb model, the material assumes that the hardening is determined by the cohesion coefficient, which can be either plastic strain, temperature, or field variables, whose hardening is heterogeneous. The model for the contact between the pile and the soil is the Coulomb friction model. The standard Coulomb friction model assumes that the equivalent friction stress is less than the 396
critical stress without relative motion, and greater than or equal to the critical stress proportional to the contact compressive stress p. teq ¼ mr
(2)
where teq is the equivalent effect force. If the equivalent force is at the critical stress (teq =tcrit ), a slip occurs. m is the friction coefficient, which can vary with other field variables. 5 BOUNDARY CONDITIONS AND INITIAL CONDITIONS Upper boundary conditions According to the boundary layer theory (Zhang et al. 2019; Zhu 1988), the temperature variation data in the permafrost region of the Qinghai-Tibet Plateau is the upper boundary temperature of the roadbed, and the temperature data can be fitted to obtain the following equation. T ðtÞ ¼ 3 þ 12sinð2pt=360 þ p=2Þ
(3)
The unit of t is h. Lower boundary conditions According to the site exploration data, the soil 20 m below the surface is less affected by temperature changes, and the temperature of the soil layer is more constant, generally -1.5 ℃ to -2 ℃, therefore the temperature is taken as -2 ℃ in the calculation. The bottom surface constraint is set to a fixed constraint. Peripheral boundary conditions For the convenience of calculation, the boundary conditions on the four sides of the temperature field are set to be insulated. The four lateral constraints are supported by sticks. Initial conditions Frost heave occurs only when the moisture content of the foundation reaches a certain threshold. The limiting moisture content at this point is called the initial moisture content (Chen 1991; He et al. 2003). The initial moisture content of the soil is set at 18%. The initial temperature of the ground foundation is the corresponding ground temperature of the permafrost layer in the permafrost zone, which is set to -2 C constant. In order to simulate the real foundation soil temperature change in the project more realistically, the soil temperature after 5 complete cycles of freeze-thaw cycles is taken as the initial temperature.
6 ANALYSIS OF NUMERICAL SIMULATION RESULTS OF THE PILE ON THE SLOPE SOIL FOUNDATION In this paper, we will study the change patterns of horizontal top displacement of the pile under the action of three freeze-thaw cycles. The calculation duration is recorded as 3 years, for a total of 36 months. The change patterns of the horizontal top displacement of the pile are shown in the figure below, and the horizontal direction along the slope upward is specified as positive. It can be seen from Figure 3 that the numerical simulation results of the horizontal top displacement of the pile have the same frequency sinusoidal style as the ambient temperature. However, since the slope soil has the effect of horizontal frost heaving force on the pile along the slope downward, the pile displacement is mainly along the slope down, and the maximum displacement along the slope downward is about 5 times the maximum displacement along the slope upward. Therefore, in the actual construction, the soil can be backfilled under the slope behind the pile.
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Figure 3.
The change patterns of horizontal top displacement of the pile.
7 CONCLUSIONS AND PROSPECTS By studying and analyzing the change patterns of horizontal top displacement of the pile under the action of three freeze-thaw cycles, it is concluded that: the displacement of the pile is mainly along the slope down, and the maximum displacement along the slope down is about 5 times of the maximum displacement along the slope up, while the displacement size also has the same frequency “sine wave” relationship with the ambient temperature. So in actual construction, the following measures can be considered: a. It is suggested to install insulation material on the side of the pile so that the soil temperature affects the pile temperature to a lower extent, thus the pile temperature remains a near-visible value, with less internal force and less displacement at the top of the pile b. Grading the slope foundation before pile foundation construction can make the slope foundation flat. Future research on pile foundations on slope foundations in multi-year permafrost areas will focus on the design, construction, stability, and slip resistance of pile foundations on permafrost slopes. The main contents include: a. Geological analysis will be carried out for permafrost pile foundation in slop, and design parameters and construction techniques for pile foundation in slop will be analyzed for different permafrost types; b. The stability and slip resistance mechanism of pile foundation in slop will be explored, the stability and slip resistance of permafrost pile foundation in slop will be studied in depth, and the slip resistance measures of permafrost pile foundation in slop will be discussed; c. The construction technology of permafrost pile foundations in slop will be studied in conjunction with the actual situation, and the advantages and disadvantages of different pile foundations in slop construction technologies will be analyzed to provide effective guidance for the actual construction.
REFERENCES Campanella, R. G. (1994) Field Methods for Dynamic Geotechnical Testing: An Overview of Capabilities and Needs. Dynamic Geotechnical Testing II. Campbell, J. L., Reinmann, A. B., & Templer, P. H. (2014) Soil Freezing Effects on Sources of Nitrogen and Carbon Leached During Snowmelt. Soil Science Society of America Journal, 78(1), 297–308.
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Chen X.B. (1991) Recent Development on Frost Action Study of Soils. Advances in Mechanics, 21(2): 226– 235. He P, Cheng G.D, Yang C.S, et al. (2003) Evaluation Method of Thawing and Sinking Coefficient of Permafrost[J]. Glacial Permafrost, 25(6): 608–613. Jia FJ, Yao Y, Li CC (2021) Preparation and Mechanism Research of Hydration-heat-inhibiting Materials with Microcapsule Sustained-releasing Technology. J Wuhan Univ Technol 36(5):697–705. Saeid EW, Marawan MS, Ahmed MN, et al. (2021) Effect of Soil Improvement Techniques on Increasing the Lateral Resistance of Single Piles in Soft Clay (Numerical Investigation). Geotech Geol Eng 39:4059–4070. Yu JL, Zhou JJ, Gong XN, et al. (2021) Centrifuge Study on Behavior of Rigid Pile Composite Foundation Under Embankment in Soft Soil. Acta Geotech 16:1909–1921. Zhang, M., Wang, J., & Lai, Y. (2019) Hydro-thermal Boundary Conditions at Different Underlying Surfaces in a Permafrost Region of the Qinghai-Tibet Plateau. Science of the Total Environment, 670, 1190–1203. Zhou, Z., Li, G., Shen, M., & Wang, Q. (2022) Dynamic Responses of Frozen Subgrade Soil Exposed to Freeze-thaw Cycles. Soil Dynamics and Earthquake Engineering, 152, 107010. Zhu L.N. (1988) Study of the Adherent Layer on Different Types of Ground in Permafrost Regions on the Qinghai-xizang Plateau. Glacial Permafrost.01: 8–14.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Experimental study on deformation characteristics of cement soft soil under the influence of multiple factors Yujie Zhang, Hongjun Liu* & Zijie Liang Wuyi University, Jiangmen City, Guangdong Province, China
ABSTRACT: The cement mixing pile method is often used to treat the soft foundation of the road. The cement mixing pile is mainly made by adding cement to the soft soil to form the cement soft soil. The deformation characteristics of the cement soft soil are affected by many factors. Therefore, a unidirectional compression test studied the influence of cement dosage, water content, and curing age on the deformation characteristics of cement soft soil. The experimental results show that with the increase of curing age, cement dosage, and water content, the strain of cement soft soil will gradually decrease, and the ability to resist deformation will gradually increase. When the cement dose is 20%, the curing age is 14 days, and the moisture content is small, the cement soft soil will achieve better resistance to deformation. The research results can provide a reference for the design and construction of similar projects.
1 INTRODUCTION With the promulgation and gradual implementation of the Outline of the Development Plan for the Guangdong-Hong Kong-Macao Greater Bay Area, considerable progress has been made in constructing highways and urban roads. Soft soil is widely distributed in the Pearl River Delta region. Soft soil has physical characteristics such as high natural water content, large pore ratio, low shear strength, high compressibility, and poor permeability (Chen et al. 2003, 2022; Hu et al. 2022; Liu et al. 2022). Therefore, a road built on a soft soil foundation often has the problem of a large settlement and easy instability during construction. To ensure the quality of the project, soft soil foundations should be treated generally, and the treatment methods of soft soil foundations are various (He et al. 2021; Wan et al. 2020). Compared with other methods, constructing a cement mixing pile is relatively simple. The soft soil foundation treated by a cement mixing pile has the advantages of small deformation and high overall stability. A cement mixing pile is composed of cement, soft soil, and water, and its ability to resist deformation is affected by many factors (Xiang et al. 2021; Zhang et al. 2021). The influencing factors are the characteristics of the soft soil itself, cement dose, maintenance age, moisture content, etc. Therefore, this paper adopts a unidirectional compression test and considers the influence of various factors to conduct a series of studies on the deformation characteristics of cement soft soil. Finally, the influence law of cement dosage, water content, curing age, and other factors on the deformation of cement soft soil is obtained, which can provide a reference for the design and construction of similar projects.
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-51
2 EXPERIMENTAL STUDY SCHEME 2.1
Test material
The test soil samples were taken from the silt soft soil at K3 + 900 of the west extension of Huasheng Road, Pengjiang District, Jiangmen City, which is widely distributed and has physical characteristics such as high-water content, low permeability, low bearing capacity, etc. The basic physical property indexes are shown in Table 1, and ordinary silicate P42.5 was used for cement. Table 1.
Basic physical property indexes of soil samples.
The natural moisture content/%
Natural void ratio
The specific Soil natural gravity of density/ soil (gcm 3)
Saturability /%
Coefficient of compressibility /MPa 1
Modulus of compression /MPa
Liquidity index
80.99
2.28
2.72
100
2.13
1.54
1.88
2.2
1.50
Factors considered in the experimental study and their composition
An experimental research scheme was designed to study the influence of various factors on the compression deformation characteristics of cement soft soil. The specific factors and their composition are shown in Figure 1.
Figure 1.
Main factors considered in the experimental study and the composition of each factor.
To better conform to the actual working conditions, the test method was a unidirectional compression test (Tang et al. 2022). The samples were prepared manually in a certain proportion. The loading grades of each sample were 25 kPa, 50 kPa, 100 kPa, 200 kPa, 400 kPa, and 800 kPa, and the pressure of each stage was recorded for 24 h according to the set time (Liu et al. 2019).
3 COMPARATIVE ANALYSIS OF COMPRESSION TEST RESULTS 3.1
Effect of cement dosage and curing age on deformation of cement soft soil
Figures 2 to 4 show the effects of cement dose and curing age on the stress-strain relationship of soft cement soil at a soft soil moisture content of 106.1% and a curing temperature of 20 C. From the trend of the curves in the figures, with the increase of cement dose and the 401
Figure 2.
Stress-strain curves for different cement doses at 7 days of curing age.
Figure 3.
Stress-strain diagram for different cement doses at 14 days of curing age.
Figure 4.
Stress-strain curves for different cement doses at 28 days of curing age.
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increase of curing age, the strain decreases gradually with the same applied vertical pressure, which indicates that the ability of cement soft soil to resist deformation increases gradually. Comparing the maximum strains at curing ages of 7, 14, and 28 days, it can be concluded that at the curing age of 14 days, the hydration reaction occurring in the mixture of cement and soil is completed. The strain of cement soft soil does not decrease significantly by increasing the curing time. According to Figures 2–4, the strain decreases significantly when the cement dose increases from 15% to 20% under the same vertical pressure. The strain does not decrease significantly when it increases from 20% to 25%. The optimal cement dose for soft cement soil is 20% considering technical and economic aspects. According to the curve spreading characteristics of Figure 5, overall, the compressive modulus of cement soft soil increases gradually with the growth of the curing age, and the curing age increases obviously within 14 days, and the increase is relatively less obvious and tends to be stable after exceeding 14 days. When the cement dose increased from 15% to 20%, the compressive modulus increased significantly, so the optimal cement dose for soft cement soil was 20%, and the optimal curing age was 14 days.
Figure 5. Variation curve of cement soft soil compressive modulus with the maintenance age at different cement doses.
The compression modulus of soft cement soil with vertical pressure in the range of 100–200 kPa was calculated based on the test data, considering only two influencing factors, namely the age of curing and the cement dose, and the calculation results are detailed in Table 2. From the test results in Table 2, the compression modulus increased by 173% when the age of curing was increased from 10% to 25% at 7 days and increased by 219% at 28 days when the cement dose was increased from 10% to 25%, increasing the compression modulus Table 2. MPa.
Statistics of cement soft soil compression modulus at different curing ages and cement doses/
Curing period
7 days
14 days
28 days
Cement Cement Cement Cement
4.058 4.579 8.040 11.062
4.721 6.562 11.142 14.682
5.208 7.042 13.193 16.611
dose dose dose dose
10% 15% 20% 25%
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by 1.27 times. When the cement dose was 10%, the compressive modulus increased by 28% from 7 to 28 days of curing age, and when the cement dose was 25%, the compressive modulus increased by 50% from 7 to 28 days of curing age, which was 1.78 times greater. The effect of cement dose on the compression modulus is more obvious than the curing age. 3.2
Effect of moisture content and age of maintenance on the deformation of soft cement soils
Figures 6 to 8 show the stress-strain curves of soft cement soils at different curing ages and different moisture contents for a cement dose of 20% and a curing temperature of 20 C. It can be obtained that the strain decreases as the moisture content decreases, and the water content decreases from 106.1% to 88.2%, which is especially obvious.
Figure 6.
Stress-strain curves for different moisture contents at the age of 7 maintenance days.
Figure 7. Stress-strain curves for different moisture contents at 14 days of maintenance age stressstrain curve.
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Figure 8.
Stress-strain curves for different moisture contents at 28 days of maintenance age.
From Figure 9, the compressive modulus of soft cement soils with different moisture contents gradually increases with the curing age. The lower the moisture content and the shorter the curing age, the more obvious this phenomenon is.
Figure 9. content.
Variation curve of compressive modulus with the maintenance age at different moisture
The compression modulus of soft cement soils with vertical pressure in the range of 100–200 kPa was calculated based on the test data considering only two influencing factors: water content and curing age. The calculation results are detailed in Table 3. From the test results in Table 3, the water content decreased from 106.1% to 63.2%. The compression modulus increased by 44% when the maintenance age was 7 days. At 28 days, the water content decreased from 106.1% to 63.2%, and the compression modulus increased by 71%. This shows that the effect of water content on compression modulus is still significant. Table 3. MPa.
Statistics of soft cement soil compression modulus at different curing ages and water content/
Curing period Water Water Water Water
content content content content
106.1% 88.2% 74.4% 63.2%
7 days
14 days
28 days
9.790 11.953 12.939 14.092
13.141 14.233 17.08 22.321
15.040 15.793 18.92 25.557
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4 CONCLUSION 1. With the increase of curing age, cement dose, and water content, the strain will gradually decrease under the same vertical pressure, indicating that the ability of soft cement soil to resist deformation gradually increases. 2. Based on the test results, it is concluded that the soft cement soil achieves a better resistance to deformation when the cement dose is 20%, the maintenance age is 14 days, and the moisture content is small. 3. Considering only the cement dose, curing age, and water content, the degree of influence on the compression modulus of soft cement soils in descending order is cement dose > curing age > water content. The final derived series of influence laws of cement dose, moisture content, curing age, and other factors on the deformation of cement soft soil can provide a reference for the design and construction of similar projects.
ACKNOWLEDGMENT The following projects supported this study: Wuyi University Student Innovation and Entrepreneurship Project, Project No. 2020CX66. School-enterprise cooperation project, project number: HX21147.
REFERENCES Chen F, Tong S-H, Shen S-L. Microscopic Mechanism of Strength Enhancement of Nickel-Iron Slag Powdered Hydroclay in Marine Environment [J]. Journal of Wuhan University (Engineering Edition), 2022, 55(07): 682–690. DOI: 10.14188/j.1671-8844.2022-07-006. Chen X.P., Huang G.Y., Liang C.S. Characterization of Soft Soils in the Pearl River Delta [J]. Journal of Rock Mechanics and Engineering, 2003(01): 137–141. Hu J-L, Gao P-F, Cui H-H, Cui C-Y, Lv X-C. Research on Strength Characteristics and Curing Mechanism of Iron Tailing Sand Hydroclay [J]. Journal of Xi’an University of Architecture and Technology (Natural Science Edition), 2022, 54(03): 338–344. DOI: 10.15986/j.1006-7930.2022.03.004. He Caisheng, Zhu Chengming. Effect of Admixtures on the Compressive Strength of Silt-Clay Hydromulch [J]. Building Structures, 2021, 51(S1): 1373–1376. Jongpradist P, Youwai S, Jaturapitakkul C. Effective Void Ratio for Assessing the Mechanical Properties of Cement-Clay Admixtures at High Water Content [J]. Journal of Geotechnical and Geoenvironmental Engineering, 2011, 137(6): 621–627. Liu Muchen, Cui Zizhi, Song Xin. Effect of Sodium Sulfate-dry-wet-freeze-thaw co-interaction on the Properties of Hydromulch [J]. Science Technology and Engineering, 2022, 22(29): 12977–12984. Tang Changyi, Liu Zhi, Qian Xue, Ma Zuyao. Indoor Experimental Study on the Characteristics of Marine Silt-hydraulic Soil [J]. Highway, 2022, 67(07): 359–367. Wan Z.H, Dai G.L, Gong W.M, Gao L.C. Experimental Study on the Microscopic Erosion Mechanism of Calcareous Sand-hydraulic Soil Plus Solids in Seawater Environment [J]. Xiang Ajia. Cement Soil Barrier Wall to Block and Control Groundwater Pollution[J]. Environmental Engineering, 2021, 39(09): 63–68 + 91. DOI: 10.13205/j.hjgc.202109010. Zhang Xinjian, Tang Changyi, Liu Zhi. Indoor Ratio Test of Silt-Hydroclay and Pile for Mation Effect Analysis [J]. Highway, 2021, 66(06): 81–84.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Application of composite foundation with vibro-crushed stone column in saturated liquefied sand ground treatment of a gas-fired power station in Myanmar Zheng Zheng*, Wei Zhu, Honebo Li, Yafeng Lou & Yang Zhang POWERCHINA Henan Electric Power Engineering Co., Ltd, China
ABSTRACT: Ground liquefaction is one of the main disasters that often happen in earthquakes, so suitable ground treatment is the key to ensuring the safety of engineering. Taking a gas-fired power station project in Myanmar as an example, the feasibility and economy of different ground treatment schemes under saturated liquefied sand ground are discussed in this paper, and a composite foundation with a vibro-crushed stone column is proposed. The results of the test show that the composite foundation can eliminate the liquefaction of the site, and at the same time, it has a good effect on soil compaction, which can effectively improve the bearing capacity of the liquefied sand foundation. This scheme also has a good economy and a short construction period, which can provide references for similar projects.
1 GENERAL INSTRUCTIONS Ground liquefaction is one of the most common disasters in earthquakes. In the liquefaction area, the loss of foundation bearing capacity will cause a large number of buildings to sink and topple. Therefore, selecting a suitable ground treatment scheme is the key to ensuring safety in earthquake areas. WANG et al. (2019) studied the mechanism of packing gravel piles to strengthen liquefied sand by shaking table test. XU et al. (2019) analyzed pile-soilstructure dynamic interaction in liquefaction sites under horizontal ground motion excitation. Hao et al. (2019) compared several schemes of liquefaction identification for silty soil and sandy soil. Yin et al. (2010) proposed an optimal treatment scheme for the elimination of soil liquefaction. Based on a gas-fired power station project in Myanmar, a foundation treatment scheme of composite foundation with a vibro-crushed stone column is proposed in this paper, and the effects of foundation treatment of various detection schemes are tested in the field. At the same time, the economy of different foundation treatment schemes is compared.
2 PROJECT OVERVIEW This project is located in the southwest region of Myanmar. The site belongs to the coastal accumulation landform, the north is low-lying, and the south is slightly higher. According to the seismic safety evaluation report of the site, the basic intensity of the site is 7 degrees, and the peak seismic acceleration is 0.15 g. According to the geological data, the overlying strata
*Corresponding Author: [email protected] DOI: 10.1201/9781003425823-52
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of the site are Quaternary and the underlying is Tertiary mud-stone. The foundation soil is mainly composed of sandy soil, silty clay and mud-stone, in which the silty sand is saturated, the liquefaction grade is medium, the liquefaction depth is about 5 m to 14 m, the compactness is loose to slightly dense, and the bearing capacity is only 100 kPa. At the same time, the site soil compression modulus is low, settlement deformation is large, and it is difficult to meet the needs of gas power plant construction, it is necessary to choose a suitable foundation treatment scheme.
3 COMPARATIVE SELECTION OF FOUNDATION TREATMENT SCHEME At present, the commonly used treatment methods for the liquefied foundation include rigid pile (precast pile, cast-in-place pile), replacement filling, dynamic compaction, vibroflot, etc. In addition to feasibility, the selection of a foundation treatment scheme also needs to consider the economy. The liquefaction depth of this site is deep, and the implementation of large-scale replacement is difficult, so it is not considered. At the same time, the site is close to the hub substation, the vibration requirements of the site are relatively strict, and it is difficult to implement dynamic compaction. Therefore, taking the turbine building as an example, the technical economy of the schemes of precast pipe pile, concrete cast-in-place pile and composite foundation are compared, as shown in Table 1. It can be seen from Table 1 that the economic cost of the vibro-crushed stone column is lower. At the same time, the scheme has high construction efficiency and can effectively shorten the construction period of foundation treatment.
Table 1.
Comparison of different foundation treatment schemes.
Ground treatment scheme
/
Number Total Price (US Dollar)
∅600 mm precast pipe pile
Length of single pile (m) 12 Pile number 130 Length of all piles (m) 1560 Price ($/m) 102 5.6 ∅600 mm concrete cast-in-place pile Volume of single pile (m3) Pile number 94 Volume of all piles (m3) 526.4 310 Price ($/m3) 5.0 ∅800 mm vibro-crushed stone column Volume of single pile (m3) Pile number 550 2750 Volume of all piles (m3) 53.3 Price ($/m3)
159120
163184
146575
4 DESIGN OF COMPOSITE FOUNDATION WITH VIBRO-CRUSHED STONE COLUMN First, the critical base of penetration Ncr is calculated according to the geological section. When the number of penetration hits N63:5 < Ncr , it is determined to be liquefied soil. Ncr can be calculated according to Equation 1 as below: pffiffiffiffiffiffiffiffi (1) Ncr ¼ N0 b½lnð0:6ds þ 1:5Þ 0:1dw 3=r
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According to the following Equation (2), the liquefaction index is calculated according to different point positions: n X Ni IlE ¼ 1 (2) di Wi Ncri i¼1 The calculated liquefaction index was 17, indicating moderate liquefaction. According to the critical base of standard penetration and liquefaction index, the composite foundation with a vibro-crushed stone column is intended to be used in this project test pile, and the pile spacing required for treatment can be calculated as 1.72 m. Considering that there are many factors affecting comparing the actual treatment effect of different pile spacing, ∅800 diameter vibroflot crushed stone pile arranged in an equilateral triangle was used, and three pile spacing of ∅1600 mm, ∅1800 mm and ∅2000 mm were adopted. The replacement rates of foundation soil were 22.7%, 17.9% and 14.5%. The foundation treatment schemes with different pile spacing are shown in Figure 1.
Figure 1.
Composite foundation with pile spacing ∅1600 mm, ∅1800 mm and ∅2000 7mm.
Based on Equations (1) and (2), the test pile penetration was measured. It was found that the site liquefaction could be eliminated when the pile spacing was 1800 mm. Therefore, the scheme of 1800 mm pile spacing was chosen as the foundation treatment scheme of this project.
5 EVALUATION OF FOUNDATION TREATMENT EFFECT To test the treatment effect of composite foundation with the vibro-crushed stone column, the static load test, standard penetration test and gravity penetration test were respectively used to judge the bearing capacity and liquefaction condition of the foundation after treatment. 5.1
The static load test
In the static load test, the slow load maintenance method was adopted, the side length of the bearing plate was 1.7 m, and the relative deformation method was adopted to calculate the bearing capacity. By drawing the P-S curve, the characteristic value of the bearing capacity was determined according to the settlement equal to 0.01 times the side length and diameter of the load, which was not greater than half of the final loading value. Three representative test sites were selected within the range of foundation treatment, and their P-S curves are shown in Figure 2. The static load test results are shown in Table 1. 409
Figure 2.
Table 2.
P-S curve of the static load test.
Result of static load test.
Static load test
TEST NUMBER
Load When Set- 1/2 of the tlement = 0.01 B final load (kPa) (kPa)
Design bear- Bearing caing capacity pacity result (kPa) (kPa) Remarks
Composite foundation load test
CFLT-1 CFLT-2 CFLT-3 AVERAGE
180 182 224 195
180
180 180 180 180
180 180 180 180
Meet the requirement
It can be seen that the bearing capacity of the composite foundation after treatment reaches 180 kPa, which is significantly higher than that of the natural foundation of 100 kPa and can meet the needs of foundation design. 5.2
Standard penetration test
According to the local seismic zoning map of Myanmar and the seismic safety assessment report near the site, it is determined that the seismic intensity of the plant is the VII degree, the peak seismic acceleration is 0.15 g, and the liquefaction degree of the saturated fine sand with poor engineering properties in the upper part of the site is slight to medium. The standard penetration test is used to evaluate the foundation treatment effect of this project. Three representative test holes are selected within the range of the foundation. Because the center of the equilateral triangle has the worst compaction effect, the detection holes are set there, and the rationality of pile spacing can be judged from this. The standard penetration bases for detecting the depth of the liquefied soil layer are shown in Table 3. According to the calculation of Equations (1) and (2), the measured values of the standard penetration number at each monitoring point are all greater than, and it can be determined that the liquefaction of the sand has been completely eliminated.
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Table 3.
N63:5 and Ncr of the detection point.
Test Thickness of liquefied Number Number soil layer (m) of SPT
Critical value of Ncr
Single point liquefaction index IlE
Cumulative liquefaction index IlE
SPT-1
8.79 12.25 14.84 16.92 18.65 8.79 12.25 14.84 16.92 18.65 8.79 12.25 14.84 16.92 18.65
0 0 0 0 0 0 0 0 0 0 0 0 0 0 0
0
SPT-2
SPT-3
Figure 3.
Table 4.
2.1 1.5 1.5 1.5 2.4 2.2 1.4 1.3 1.6 2.4 2.1 1.5 1.5 1.5 2.4
16 19 26 23 23 12 18 17 19 23 13 18 18 21 24
Result of static load test. Compactness
Depth range (m)
DPT-1
Dense Medium dense Dense Dense Dense Medium dense
0.5 7.5 0.5 7.8 0.5 7.3
DPT-3
5.3
0
The curve of the dynamic penetration test.
Test Number
DPT-2
0
– – – – – –
7.5 8.0 7.8 8.0 7.3 8.0
Corrected number of hits >7 >7 >7
Dynamic penetration test
To ensure construction quality, a dynamic penetration test can be used. The test points are all set near the center of the column. The test can be terminated when the number of hammer
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hits is greater than 50 for three consecutive times. 3 test points are selected with representative rows within the range of foundation treatment, and the dynamic exploration curve is shown as follows. It can be seen that all columns are dense, and the dynamic penetration number is greater than 7, so the compactness meets the requirements.
6 CONCLUSIONS Taking the treatment of saturated liquefied sand encountered in a gas-fired power station project in Myanmar as an example, this paper discusses the feasibility and economy of different foundation treatment schemes, analyzes the design of composite foundation with vibro-crushed stone column and tests the treatment effect of the schemes. There are the following conclusions: 1. Vibro-crushed stone column can improve the shear strength of the original soil and form a good drainage channel to eliminate the liquefaction effect of the soil. 2. While eliminating liquefaction of the site, the composite foundation can compress the site soil, thus effectively improving the bearing capacity of the foundation. 3. The composite foundation has low cost and fast construction speed. The scheme applied in this project saves the investment of 4 million dollars and shortens the construction period by 6 months, which has good application significance.
REFERENCES Hao B., Ren Z S., Li C Y. (2019) Comparison of Several Discrimination Criteria for Seismic Soil Liquefaction. Geotechnical Engineering Technique, 33 (5). Doi:10.3969/j.issn.1007-2993.2019.05.007. Li R J., Zhang J., Jiang H. (2018) Shielding Effect and Evaluation of Liquefaction Resistance of Long-short Gravel Piles in Saturated Large-thick Sand Foundation. Journal of Natural Disasters, 27 (6). Doi: 10.13577/j.jnd.2018.0620. Li Y R., Zhang J., Rong X. (2018) Recent Advances and New Problems in Seismic Behavior for Vertical and Batter Pile Foundation in Liquefied Soil. Earthquake Engineering and Engineering Dynamics, 38 (6). Doi:10.13197/j.eeev.2018.06.171.liyr.020. Liu S C., Wang A H., Zhang D W. (2022) Seismic Response of Strength Composite Piles in a Liquefiable Soil. Journal of Civil and Environmental Engineering, 44 (6). Doi: 10.11835/j.issn.2096-6717.2021.014. Wang J B., Song X T., Tian M L. (2019) Experimental Study on the Liquefaction Mechanism of Sandy Soil Enforced by Geo-encased Stone Columns. China Earthquake Engineering Journal, 41 (1). Doi:10.3969/ jissn.1000-0844.2019.01076. Xiong H., Yang F., (2020) Horizontal Vibration Response Analysis of Pile Foundation in Liquefied Soil Under Winkler Foundation Model. Rock and Soil Mechanics, 41 (1). Doi:10.16285/j.rsm.2018.2332. Xu C X., Dou P F., Du X L. (2019) Experimental Study on Seismic Response of Superstructure-piles-soil System – Design of Test Plan for Large-scale Shaking Table Model Experiments. Journal of Disaster Prevention and Mitigation Engineering, 39 (3). Doi:10.13409/j.cnki.jdpme.2019.03.001. Yin F S., Zhou L Q., Yang J. (2010) Optimization of Ground Soils Liquefaction Treatment Scheme for a Main Power-house of the Power Plant in Highly Seismic Intensity Region. Engineering Journal of Wuhan University, 43. Yu J., He Y., Zhang L. (2017) Dynamical Characteristics of Piles in Liquefied Soil Under Horizontal Vibration. Chinese Journal of Geotechnical Engineering, 39 (3). Doi:10.11779/CJGE201703023. Zhao L Y., Shan Z G., Wang M Y. (2022) Analysis of Liquefaction Characteristics of Horizontal Site of Offshore Wind Farm Under Earthquake in the South Yellow Sea. Rock and Soil Mechanics, 43 (1). Doi: 10.16285/j.rsm.2021.0336. Zhao T L., Ren Y P., Xu G H. (2019) Experimental Study on the Trajectory Characteristics of Liquefied Soil Under Wave Action. Periodical of Ocean University of China, 49. Doi:10.16441/j.cnki.hdxb.20180102.
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Application of oblique photography technology in airport site selection Zizhu Zhang Hezhou Major Project Construction Service Center, Hezhou, China
Lan Cheng*, Xinpu Feng, Xiao Wang, Guoliang Zhai, Ke Tang & Shiman Sun Beijing Super-Creative Technology Co., Beijing, China
ABSTRACT: The airport is an important public infrastructure and a key node of air transport, which has a very important impact on the entire civil aviation transport system and regional economic and social development. The first step of airport construction is site selection. The site selection quality will affect the airport construction investment, operation safety, and the driving effect on the local economy. Therefore, the site selection of the airport must take into account a variety of factors. The traditional airport site selection mainly relies on two-dimensional maps for measurement and design, which cannot meet the design requirements of refinement and visualization. Given this problem, this paper uses UAV tilt photography technology to model the site to be selected at Hezhou Airport. It integrates the model into the GIS platform for measurement and analysis, assisting designers in site selection. It is significant for promoting the digital construction of airport site selection.
1 INTRODUCTION In recent years, more and more mountain airports have been selected. The site selection of airports in mountainous areas faces such problems as poor clearance conditions, variable climate, complex geological conditions, and large topographic relief. The traditional airport site selection mainly depends on the site survey and two-dimensional mapping analysis of the designers. However, most of the mountains in the airport site selection area have not been developed, and the terrain is complex. The overall situation of the mountain area cannot be intuitively and vividly understood only by manual work, which has a certain impact on the designers’ decisions. How to quickly and efficiently obtain the high-precision 3D model of the mountain area and assist the designers in selecting the optimal site has become a key issue in civil aviation airport locations. (Gao 2022; Gu 2019) There are many technical means to quickly obtain three-dimensional information in mountain areas, such as remote sensing, aerial survey of unmanned vehicles, threedimensional laser scanning, etc. Remote sensing can quickly obtain a wide range of images and terrain data, but it is not suitable for the needs of refined analysis of airport site selection due to the impact of data accuracy, update frequency, and price; 3D laser scanning has certain advantages in obtaining 3D information of buildings. However, due to the large scope of airport site selection in mountainous areas and the low efficiency of 3D laser scanning, it is not easy to quickly meet the needs of airport site selection to obtain 3D real site data. The advantage of UAV photogrammetry technology is that it can conduct three*Corresponding Author: [email protected] DOI: 10.1201/9781003425823-53
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dimensional modeling with high efficiency and precision, quickly obtain information about the environment and buildings within the scope of aerial survey and restore the plane and three-dimensional information. With the continuous development of aerial survey technology, UAV aerial survey has significantly improved product performance, operation efficiency, work stability, etc. It can realize manual planning and design of routes, and UAV automatically collects data after receiving routes. In this paper, the UAV photogrammetry technology is used to collect the data of the site to be selected for the mountain airport, and after processing and correcting the collected data, the 3D modeling of the site to be selected is carried out, and the model is integrated into the GIS platform to realize the functions of excavation and filling volume analysis, distance measurement, etc., and to assist the designers in airport site selection. Based on the UAV photogrammetry technology, the 3D model of the site to be selected is reconstructed. The designer can use the model to design the scheme, providing low-cost and high-precision data support for the optimization design of the airport site selection scheme. (Liu 2022a, 2022b)
2 OBLIQUE PHOTOGRAPHY FIELDWORK 2.1
Site overview
This paper collects the field data of three sites to be selected at Hezhou Airport. This paper only takes the site of Litushan as an example for analysis. The site of Litou Mountain belongs to the mountain landform formed by folded anticline, and the mountain is in the northeast-southwest direction. The central part of the airfield of the proposed airport is relatively low, and the slope is relatively gentle. There are individual villages and two small ditches, with an elevation of about 460 to 490 meters. The north and south of the flight area are high mountains with steep slopes. The elevation of the mountains at the north end is between 620 meters and 650 meters, and the elevation at the south end is between 580 meters and 610 meters. The terminal area of the proposed site is higher in the north-south direction and lower in the east-west direction. The north-south direction is a mountain with an elevation of 570–596 m, and the east-west side is a hillside with an elevation of about 430–500 m. It isn’t easy to understand the regional situation intuitively and accurately by relying on a manual survey. 2.2
Equipment selection
Huace pure electric vertical fixed-wing UAV is selected as the field data acquisition equipment. The UAV integrates a high-precision airborne GNSS system, multi-redundant IMU inertial navigation fusion algorithm, L1 nonlinear navigation algorithm, TECS total energy control algorithm, high-modulus carbon fiber composite lightweight body, and multiplatform ground control system to achieve stable and reliable use in various complex scenarios and extreme environments. It can quickly complete the data acquisition task, as shown in Figure 1. The UAV fuselage has a battery system, mission equipment, control system, and tail cone power system. The wing is a collection of systems that generate lift for the aircraft. The trailing edge of the wing has an operable control surface. The V-tail controls the pitch angle of the body. 2.3
Route planning
Go to the selected takeoff and landing point, place the UAV at the takeoff and landing point, and start assembly. After assembly, check the battery level, camera parameters, and
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Figure 1.
System composition.
wing connections are secure. After inspection, open the ground station software and the aircraft base station, and connect the UAV and the ground station, and the UAV and the aircraft base station, respectively. After the connection is successful, the flight interface will automatically acquire the current position of the UAV and the satellite image of the current position. According to the requirements of modeling accuracy, the UAV aerial survey range of the site to be selected is 30 square kilometers, the relative height is not more than 400 meters, the course overlap is 85%, and the side overlap is 70%. The route will be generated after checking, shown in Figure 2.
Figure 2.
Generate route renderings.
After the route is generated, check the takeoff point, landing point, and waypoint elevation. To ensure flight safety, the distance between any point and the ground shall not exceed 50 meters. After the elevation check, the route is uploaded to the UAV, and the UAV performs self-inspection. After the self-inspection is successful, the route generated after the plan will be uploaded to the flight control system. The field survey personnel will operate the UAV to take off, and the UAV will start collecting data according to the planned route.
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3 3D MODELING OF THE SITE TO BE SELECTED 3.1
Original photo pre-processing
Due to the large area of the site to be selected, the UAV cannot complete the whole area measurement at one time. During the aerial survey by sorties, the ambient light will vary due to different times. Therefore, the UAV camera will be affected by the ambient brightness in the process of collecting data, making the color of the collected photos often have some differences. If the color difference between the photos and adjacent photos is too large, it will greatly impact the subsequent texture map. It may even lead to the failure of connection point matching during modeling. Therefore, the original photos collected by the UAV are first processed to eliminate distortion and uniform light. 3.2
POS data solution and export
Download the differential data of the aircraft base station, use the CGO-UAV software developed by China Survey to solve the data, set the central meridian and result coordinate system of the survey area during the calculation, import the mobile station data, IMU original data, base station height, and base station coordinates, and the software will automatically solve the problem. After the result is solved, the POS report will be generated, and the POS report data will be matched with the field photos. (Wang 2015, 2021) 3.3
Oblique photographic modeling
The Context Capture 3D modeling software developed by Bentley Company realizes the oblique photography modeling, and the pre-processed photos and POS data successfully solved are imported into the modeling software. Firstly, aerial triangulation is performed. The aerial triangulation calculation can determine the pixels corresponding to the projection of the same physical point in the scene in two or more images. After the aerial triangulation is completed, check whether the results are normal. After the inspection, submit the 3D reconstruction. You can choose different 3D model formats according to different use scenarios. To integrate into the GIS platform and integrate with the BIM model for analysis, the format of the tilt photography results in this paper is OSGB. (Qin 2022; Ma 2022)
4 ACCURACY ANALYSIS AND VISUAL DISPLAY OF 3D MODEL 4.1
Precision analysis
The relative altitude of the drone flight is determined based on the image resolution, and the calculation method is as follows. See, for example, Equation 1 below: H¼
f GSD a
(1)
In the formula, H is the relative altitude. The unit is m, f is the lens focal length, the unit is mm, a is the pixel size, the unit is mm, GSD is the ground resolution, and the unit is m. Six-point coordinates were randomly measured in the survey area to verify the plane and elevation accuracy of the three-dimensional model. The corresponding six-point coordinates were selected from the three-dimensional model. The difference, plane, and elevation errors were calculated, and the model’s accuracy was analyzed accordingly. The error calculation
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formula is as follows, for example, in Equation (2) below: 8 sffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi P > > ðDXÞ2 > > > EX ¼ > > n ffi > > sffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi > < P ðDYÞ2 E ¼ Y > > n ffi > sffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi > > P > > > ðDZÞ2 > > : EZ ¼ n
(2)
where E is the error, and X, Y, and Z are the checkpoint’s longitude, latitude, and elevation. D is the difference between the measured and measured values of the inspection points, n is the number of inspection points, and the accuracy of the inspection points is shown in Table 1. Table 1.
T1 T2 T3 T4 T5 T6 Mid-error
Statistical table of checkpoint accuracy calculation (Unit: m). DX
DY
DZ
0.048 0.059 0.003 0.091 0.012 0.024 0.050
0.1 0.049 0.025 0.032 0.057 0.041 0.056
0.076 0.031 0.067 0.087 0.042 0.065 0.064
Note: D X, D Y, and D Z represent the difference in longitude, latitude, and elevation between the inspection point and the measured value, respectively, while T1-T6 represents the ordinal number of the inspection point.
Through experiments, it can be analyzed that the mid-error of the longitude of the threedimensional model of Hezhou Airport is 0.050 meters, the maximum error is 0.091 meters, and the minimum error is 0.003 meters; The mid-error is 0.056 meters, the maximum error is 0.1 meters, and the minimum error is 0.025 meters; The mid-error of elevation is 0.064 m, the maximum error is 0.087 m, and the minimum error is 0.031 m. Through analysis, it can be seen that the mid-error, maximum error, and minimum error of the six survey points do not exceed 0.15 meters, which conforms to the first-level plane and elevation accuracy standards in the “Production Specification for 3D Geographic Information Model”. The accuracy of this study has reached the centimeter level, which can meet the airport’s requirements for model accuracy. It can be seen that the three-dimensional model reconstructed based on UAV tilt photogrammetry can meet the needs of airports for large-scale models (Wang 2022; Xie 2022) 4.2
3D model display and measurement
The results are shown in the following Figures 3–5.
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Figure 3.
3D model display of the site to be selected.
Figure 4.
Distance measurement of the site to be selected.
Figure 5.
Earthwork calculation of the site to be selected.
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5 CONCLUSIONS UAV tilt photogrammetry technology has changed the traditional airport location method somewhat. It is of great significance to use this technology to quickly obtain the threedimensional model of the site to be selected with a large range, high accuracy, and geographical location attributes and to assist designers in scheme comparison and analysis to improve the digitalization of the airport location and optimize the location scheme. Applying the tilt photogrammetry technology to the airport location business has greatly reduced the field survey work, saved a lot of workforce and material resources, and realized the new mode of digital location.
REFERENCES Gao, Y. Chen, J. (2022). Research on Three-dimensional Modeling of Lighting Change Scenes Based on Virtual Reality Technology. Laser Journal. 43(11):204–209. DOI:10.14016 Gu, L. (2019). A Case Study of General Airport Site Selection in Mountain Areas. Civil Aviation Management. (04):47–50. Liu, J. (2022a). Analysis of Factors Considered in the Site Selection Stage of the Civil Airport. China Science and Technology Information. (09):37–38. Liu, Y. Chen, Y. (2022b). Research and Analysis on Urban 3D Modeling and Singularization Based on Oblique Photography. Geomatics & Spatial Information Technology. 45(09):248–251. Ma, F. Du, S. (2022). Research and Application of 3D Modeling of Tilt Photogrammetry and Ground Laser Scanning Technology. J. Heilongjiang Science. 13(24):94–97. Qin, X. Wang, W. (2022). Research on 3D Digital City Modeling Based on Oblique Photogrammetry. Geospatial Information. 20(09):41–44. Wang, Q. (2021). Application of UAV Tilts Photography Technology in Highway Route Selection in Mountainous Areas. Tianjin Construction Science and Technology. 31(05):10–12. Wang, S. Li, X. (2015). Automatic Acquisition and Management of Electromagnetic Data for UAV Data Link Test. Computer Measurement & Control 22. (04):1276–1278. Wang, W. Pu, J. (2022). Research on Information Extraction of Ground 3D Laser Scanning Technology in Mountain Engineering Mapping. Engineering and Technological Research. 7(04):212–214. Xie, W. Zhang, J. (2022). Hole Repair Method for 3D Modeling of Oblique Photography. Bulletin of Surveying and Mapping. (12):24–28 + 34. cnki.11-2246.2022.0352..
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Satellite positioning and inertial navigation based firefighters positioning system Xiqing Liu, Ansong Feng*, Chonglin Gu & Guozhan Wang Shenyang University of Chemical Technology, Shenyang, China
ABSTRACT: With the development of science and technology, especially the development of intelligent devices, the concept of intelligence is also slowly gaining popularity. Intelligence in the field of firefighting is also valued by various countries including intelligent firefighting in national science and technology development planning and vigorously developing intelligent firefighting systems. In outdoor fire rescue, the GPS (Global Positioning System) outdoor positioning system equipped with firefighters has a slightly lower accuracy and the system module is larger, which is not easy to carry and has a certain impact on the firefighters’ fire rescue, and secondly, in the complex urban environment, it may lead to the interruption of the positioning system signal and the firefighters’ action The trajectory of firefighters is easily lost. To improve the positioning accuracy of the system and to ensure the safety of the firefighters, the positioning system designed in this paper is mainly based on GPS positioning by adding BDS (BeiDou Navigation Satellite System) satellites to constitute dual-mode positioning and adding IMU (Inertial Measurement Unit) modules to the positioning system to further improve the positioning accuracy. Through several experiments, it is shown that after the system data is pre-processed by low-pass filtering for acceleration, the Kalman filter can further eliminate process noise and prevent data scattering while determining the motion trajectory and improving the positioning accuracy to the decimeter level, and the attitude of firefighters can be observed through the attitude angle to meet the requirements of firefighters’ positioning service and guarantee the safety of firefighters’ lives.
1 INTRODUCTION Currently, the most common outdoor positioning method is satellite positioning. GPS (Global Positioning System, usually abbreviated as GPS) single point positioning is also called absolute positioning, GPS is easy to be blocked signal in harsh urban environments (Amini A., 2014; Lim C. 2019), its disadvantage is the low positioning accuracy and realtime data positioning is difficult (Ren C., 2015). Inertial Measurement Unit is widely used in various motion state detection (Chu 2012; Xie G. 2018), relying mainly on kinematics and Newtonian classical mechanics for data operations (Lasmadi L. 2017), as a representative inertial measurement unit (Jin M. 2014), along with accelerometers, gyroscopes and magnetometers (Rajamani R. 2012; Schopp P. 2009). The measurement of velocity and displacement using only the IMU module is subject to drift and bias (Bryne H. 2017). In complex urban environments, GPS and BeiDou dual-mode positioning signals may disappear briefly or be interfered with. In this paper, the IMU sensor is added to the positioning system to ensure the normal operation of the positioning system and the acceleration is *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-54
low-pass filtered and the data from the inertial measurement unit and the satellite positioning unit are passed through Kalman Filtering to improve the positioning accuracy from the meter level to the decimeter level.
2 MATERIALS AND METHODS 2.1
Materials
The STM32F407ZGT6 was selected as the microprocessor, the ATK-IMU901 angle sensor module was selected for the acceleration measurement module, and the ATK-S1216F8-BD GPS/BeiDou dual-mode positioning module was selected for the GPS/BeiDou module. 2.2
Methodology
The STM32F407ZGT6 microcontroller carries out serial communication via Zigbee, and the data read from the ATK-IMU901 inertial measurement unit and the ATK-S1216F8-BD GPS/BeiDou module are processed by Kalman Filtering data, displayed and saved on the PC via serial communication. The system uses inertial navigation to monitor the carrier’s operation status in real time, uses the carrier position and latitude and longitude information obtained by GPS/BDS and combines with IMU901 to solve the carrier’s current speed and position information for fusion solution error correction to achieve decimeter-level positioning accuracy. System design schematics are shown in Figure 1.
Figure 1.
System design schematic.
The IMU module is stationary and the data is recorded continuously for a period, and the difference between the average of the recorded multiple data sets and the ideal value is taken as the zero bias aerror of the IMU. aerror ¼ aavg ai
(1)
where aerror denotes the zero bias value of the IMU, aavg denotes the average of the IMU data measurements, and ai denotes the acceleration ideal.
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Low-pass filtering is a good way to eliminate signal noise (both mechanical and electronic) in accelerometers. The algorithm for low-pass filtering is given by Y ðnÞ ¼ aX ðnÞ þ ð1 aÞY ðn 1Þ
(2)
where a is the filter coefficient, X ðnÞ is the current sample value, Y ðn 1Þ is the last filter output value, and Y ðnÞ is the current filter output value. The Forward-Right-Down (FRD) coordinate system is set by IMU 901, and the translation of the carrier coordinate system to the navigation coordinate system is represented by the matrix Cbn . 2 3 cos qzcos qy cos qxsin qz þ sin qxsin qycos qz cos qxcos qzsin qy þ sin qzsin qx n Cb ¼ 4 cos qysin qz cos qzcos qx þ sin qysin qxsin qz sin qycos qxsin qz cos qzsin qx 5 sin qy sin qxcos qy cos qxcos qy (3) The acceleration components of each axis are: 8 < ax ¼ Ax Cbn ay ¼ Ay Cbn : az ¼ Az Cbn
(4)
where ax is the x-axis acceleration, ay is the y-axis acceleration, and az is the z-axis acceleration. 2.3
GPS/BeiDou module and IMU data fusion
The original GPS output is based on the latitude and longitude information of WGS-84 coordinates, and the GPS coordinate system needs to be projected as a flat map before the heading can be projected. The design uses Mercator projection to project the latitude and longitude. XN ¼ lon Earth:Radius p=180 ð1 þ sin ðaÞÞ YE ¼ Earth:Radius=2 log ð1 sin ðaÞÞ
(5)
a ¼ ðlat pÞ=180
(7)
(6)
where XN and YE are the coordinates of the plane axis of the coordinate system, lon is the longitude and lat is the latitude Earth:Radius is the radius of the earth 6378137 meters. In the dynamic error model, the position and velocity data from GPS/BDS are usually input to a filter, and the filter builds an error model to estimate the IMU errors by comparing the difference between the two. These errors are used to correct the IMU navigation output results to obtain the combined velocity, position, and attitude navigation results. The IMU/GPS combined positioning system error state equation and measurement equation are: xk ¼ F ðxk1 ; uk1 Þ þ wk1
(8)
zk ¼ Hk dxk þ vk
(9)
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where xk is the error state of the system at moment k, uk1 is the input control vector at moment k-1, wk1 is the process noise conforming to the Gaussian distribution, and the covariance is denoted as Qk . zk is the measurement model matrix, vk is the measurement noise conforming to the Gaussian distribution, and the covariance is denoted as Rk . The iterative estimation uses the discrete Kalman filter. 1. The state transfer matrix is brought into the state equation for prediction, and the predicted value A is xk . xk ¼ ðI þ F DtÞxk1
(10)
where F is the error matrix, Dt is the acquisition time interval of 0.02 seconds, I is the unit matrix of order 15, and xk1 is the predicted value of the previous moment. 2. Prediction mean square error Pk . Pk ¼ FPk1 F T þ Qk
(11)
where Qk is the process noise covariance, initialized as a unit matrix of order 15 3. Calculation of the fusion weights of the measurements, resulting in the Kalman gain Kk . 1 Kk ¼ Pk HkT Hk Pk HkT þ Rk
(12)
where Rk is the measurement noise covariance and Hk is the measurement matrix at moment k. 4. Data fusion of the measured values to obtain an updated calibration estimate xk . xk ¼ xk þ Kk ½zk Hk xk1
(13)
where zk is the data fused in the system, zk is the difference between the velocity and displacement data calculated by IMU and the displacement and velocity data calculated by latitude and longitude. The initial velocity in the system is set to 0, i.e., gvx; gvy; gvz is set to 0. 5. Update the estimated mean square error. Pk ¼ ½I Kk Hk Pk
(14)
The xk , which is derived from Equations (2–13), is the desired correction estimate in the system, and each time the IMU is integrated, the correction estimate is subtracted to obtain more accurate data.
3 RESULTS AND ANALYSIS The data collected in this thesis is the data transmitted through the lower computer module, and the data is received and sent to the upper computer at the same time by means of a timer, and the data from the upper computer is received and saved for simulation. To verify that adding KF to the system can better improve the positioning accuracy, this experiment was chosen to run the designed system in a 50*50 m square field for testing, and the received data as well as the actual measured data were saved separately for comparison, and the displacement waveforms obtained after data pre-processing are shown in Figure 2. The real trajectory displacement and the trajectory displacement derived after KF are shown in Figure 3. As can be seen from the waveform display in Figure 2, the blue line indicates the carrier displacement derived from the acceleration integration, and the red line indicates the carrier
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Figure 2.
Tri-axis displacement integration waveform and KF output waveform.
displacement derived from the addition of Kalman Filtering to the system. The displacement derived from the integration of the acceleration data of the inertial measurement unit becomes more inaccurate as time grows and data drift occurs, while the addition of Kalman Filtering to the positioning system clearly shows that the displacement waveform converges to the true value and no data drift occurs. It is obvious from the waveform plot in the Zdirection that this experiment is in a relatively gentle place and no large changes in height occur, and thus the displacement in the Z-direction tends to be 0. The real error values are shown in Table 1. Table 1.
Error values of actual trajectory and KF trajectory.
Actual xDATA direction 1 2 3 4 5 6
4.2 7.4 7.7 8.0 40.2 39.7
Actual ydirection
KF x- direc- KF y- direc- X-direction tion tion error
y-direction error
0.6 1.2 1.2 1.3 6.4 21.6
3.453 6.820 7.141 7.335 39.694 39.049
0.014 0.160 0.117 0.182 0.295 0.776
0.586 1.040 1.083 1.118 6.105 22.376
0.747 0.580 0.559 0.665 0.506 0.651
The displacement trajectories during the experiment are partially extracted and then compared to determine the displacement accuracy that can be achieved after the KF is added to the system. From the data in the table, it can be seen that the KF always has no data domain divergence during the carrier motion, the data error is small, and the position accuracy after KF can reach the decimeter level. 424
4 CONCLUSIONS In this paper, a high-precision miniaturized GPS/IMU combined navigation firefighters positioning system is designed from the perspective of positioning accuracy. The system adopts the GPS/BDS satellite receiving system and inertial measurement unit IMU and uses Kalman Filtering and low-pass filtering technologies to solve the problem of low positioning accuracy caused by the previous use of a single positioning module and achieve high accuracy position and attitude measurement. The combined navigation and positioning system can be used in the future elderly industry, rescue industry, etc. to substantially make up for the defect that GPS cannot be accurately positioned in complex environments.
REFERENCES Amini A., Vaghefi R.M., Garza J.M., Buehrer R., Improving GPS-based Vehicle Positioning for Intelligent Transportation Systems, in IEEE Intelligent Vehicles Symposium, Dearborn, Michigan, USA, January. 2014, pp. 1023–1029. Bryne T.H., Hansen J.M., Rogne R.H., Sokolova N., Fossen T.I., Johansen T.A., Nonlinear Observers for Integrated INS/GNSS Navigation: Implementation Aspects, IEEE Control Syst. Mag. 37 (3) (2017) 59–86. Chu T, Guo N, Backén S, Akos D. Monocular Camera/IMU/GNSS Integration for Ground Vehicle Navigation in Challenging GNSS Environments. Sensors (Basel). 2012;12(3):3162–85. Jin M., Zhao J., Jin J., Yu G., Li W., The Adaptive Kalman Filter Based on Fuzzy Logic for Inertial Motion Capture System, Measurement 49 (March) (2014) 196–204. Lasmadi L., Cahyadi A., Herdjunanto S., Hidayat R., Inertial Navigation for a Quadrotor Using Kalman Filter with Drift Compensation, Int. J. Electrical Comput. Eng. 7 (5) (2017) 2596–2604. Lim C. et al., “Dynamic Performance Evaluation of Various GNSS Receivers and Positioning Modes with Only One Flight Test,” Electronics, Vol. 8, Art. No. 1518, 2019. Rajamani R., Phanomchoeng G., Piyabongkarn D., Lew J.Y., Algorithms for Real-time Estimation of Individual Wheel Tire-road Friction Coefficients, IEEE/ASME Trans. Mechatronics. 17 (6) (2012) 1183– 1195. Ren C., Liu Q., and Fu T., “A Novel Self-calibration Method for MIMU,” IEEE Sensors J., vol. 15, no. 10, pp. 5416–5422, Oct. 2015. Schopp P., Klingbeil L., Peters C., Buhmann A., Manoli Y., Sensor Fusion Algorithm and Calibration for a Gyroscope-free IMU, Procedia Chemistry, Volume 1, Issue 1, 2009, Pages 1323–1326, ISSN 1876 – 6196. Xie G., Gao H., Qian L., Huang B., Li K., Wang J., Vehicle Trajectory Prediction by Integrating Physics- and Maneuver-based Approaches Using Interactive Multiple Models, IEEE Trans. Ind. Electron. 65 (7) (2018) 5999–6008.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Moisture accumulation in wall thermal insulation layer under the action of coupled heat and moisture during heating period in cold regions Weihua Zheng*, Yuan Su*, Yun Gao*, Chao Song*, Caihong Qi*, Zehua Feng* & Wenfei Zhao* College of Civil Engineering and Architecture, Hebei University, Baoding, China
ABSTRACT: In this paper, a typical wall with external thermal insulation in buildings in a cold region city, Baoding, is studied by numerical simulation to investigate the law of moisture content change in the thermal insulation layer during the heating period. In addition, the influence of wall moisture accumulation is analyzed from three aspects: installation of the air layer, setting of the vapor barrier and selection of finish coat to effectively mitigate the problem of moisture accumulation in the wall thermal insulation layer caused by coupled heat and moisture transfer in the wall during the heating period in the cold regions in winter. It is found in the study that long-term moisture accumulation occurs at the material intersection. The setting of the air layer and vapor barrier reduces the moisture content of the thermal insulation layer by 10.6% and 6.81% respectively, and the tile finish and paint finish increase the moisture content of the thermal insulation layer by 5.3% and 3.1% respectively. This provides a reference for better solving the problem of moisture accumulation in the wall thermal insulation layer in the heating period in winter.
1 INTRODUCTION The building envelope is an important part of a building and also a major source of heat and moisture loads to the building (Wei et al. 2020). The materials of most building envelopes are porous media, and the heat transfer relies not only on thermal conductivity driven by temperature gradients but also more on sensible and latent heat changes caused by moisture transfer (Hens 2007). Wang et al. (2014) and Li et al. (2017) demonstrated the non-negligible effect of moisture transfer on the thermal performance of the building in studies on a variety of building materials. Guo established a one-dimensional transient coupled heat and moisture transfer model for multi-layered walls based on the coupled heat and moisture transfer equation of Luikov (1964) and Philip & Devries (1957) with moisture content and temperature as the driving potentials and studied the effect of heat and moisture transfer in walls on the indoor environment (Guo et al., 2014). To summarize the condensation characteristics inside the wall and the moisture level of composite walls under different climatic conditions, Su et al. (2012) and Zhang (2007) investigated the heat and moisture behavior of multi-layered walls under different climates. The climate of cold regions in China is characterized by cold and dry winters (Yu & Xu 2012), where an external thermal insulation structure is used for most of the insulated walls. This structure reduces the building energy consumption by installing a thermal insulation *Corresponding Authors: [email protected], [email protected], [email protected], [email protected], [email protected], [email protected] and [email protected]
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DOI: 10.1201/9781003425823-55
layer to reduce the heat exchange between indoors and outdoors. However, the thermal insulation layer also hinders the transfer of vapor and liquid water. During the heating period in winter, the transfer and accumulation of moisture inside the thermal insulation layer can easily cause the layer to be dampened due to the high indoor temperature and humidity as a result of closed doors and windows and domestic water use. Therefore, it is of great significance to study the moisture transfer and accumulation of the typical external thermal insulation layer in cold regions during the heating period to reduce moisture in the thermal insulation layer, reduce building energy consumption and extend the service life of the thermal insulation layer. Because experimental studies of building envelopes are costly and time-consuming, in this paper, numerical simulation is used to investigate the heat and moisture transfer characteristics of walls in cold regions during the heating period.
2 COUPLED HEAT AND MOISTURE TRANSFER MODEL 2.1
Moisture control equation
According to the law of conservation of mass, the change in moisture content in the control volume can be expressed as the following equation @w @j @w ¼ r dp rPv þ Dl rj @j @t @j where Pv ðj; T Þ ¼ j Psat ðT Þ 4042:39 Psat ðT Þ ¼ exp 23:5771 T 37:58 where w is the moisture content of the material in kg/m3, j is the relative humidity, t is the time in s, Pv is the partial pressure of vapor in Pa, T is the temperature in K, Psat is the partial pressure of saturated vapor related to temperature only in Pa, Dl is the diffusion coefficient of liquid water in m2/s, and dp is the vapor permeability coefficient of the building material in s. In some material libraries, the vapor permeation resistance coefficient m is often given, with the following expression m¼
d0 dp
where d0 is the vapor permeability coefficient in air, in s. 2.2
Heat control equation
According to the law of conservation of energy and the law of conservation of mass, the heat transfer equation for any control volume of the wall is expressed as follows rCp
eff
@T ¼ r krT þ Heva dp rPv @t
where rCp
eff
¼ ðrmat CPmat Þ þ wðjÞCPw
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where r is the density of the control volume in kg/m3, Cp is the heat capacity of the control volume in J/(kgK), rmat is the dry density of the porous material in kg/m3, CPmat and CPw are the heat capacity of the building material and water in J/(kgK) respectively, and Heva is the latent heat of evaporation in J/kg. 2.3
Initial conditions and boundary conditions
The simulation city selected for this study is a cold region city, Baoding, where the indoor temperature is 21 C during the heating period. The relative indoor humidity is high at 60% due to the closed doors and windows and domestic water use. The ASHRAE data is selected as the outdoor air temperature and humidity conditions as shown in Figure 1. The initial temperature of the insulated wall is set to 21 C and the initial relative humidity is set to 40%. The simulation time is from December 5, 2017, to March 5, 2018, for a total of 2160 h.
Figure 1.
Outdoor boundary conditions during the heating period in Baoding city.
The conditions for determining a solution are as follows: qs ¼ hðTamb Ts Þ gs ¼ b PV ;amb PV ;s
gs is the wall moisture transfer rate in kg/(m2-s), qs is the wall heat flux in W/m2; Ts and Tamb are the wall temperature and ambient temperature outside the wall in K; PV ;s and PV ;amb are the partial pressure of vapor at the wall and the partial pressure of ambient vapor in Pa; h is the surface heat transfer coefficient in W/(m2K), which is 8.72 W/(m2K) and 23.26 W/ (m2K) for the inner and outer walls respectively; b is the wall moisture transfer coefficient in s/m, which is 1.85 108 s/m and 14 108 s/m for the inner and outer walls respectively.
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3 MODEL CALCULATION AND RESULT ANALYSIS 3.1
Typical insulated wall structure and thermal parameters
Taking a cold region city, Baoding, in winter, as an example, the external thermal insulation structure of a typical local building is selected, which is composed of a finish coat, a mortar protection layer, an expanded polystyrene board (EPS board) thermal insulation layer and the solid brick substrate from outdoors to indoors respectively, as shown in Figure 2. The parameters of the materials used in the simulation are shown in Tables 1 and 2. In this paper, the thermal and moisture properties of the wall thermal insulation layer are analyzed, as well as the influence of three factors, namely the setting of an air layer, the setting of a vapor barrier, and different finish coat materials, on the moisture accumulation of the thermal insulation layer.
Figure 2.
Table 1.
Schematic diagram of external thermal insulation structure.
Thermophysical property parameters of the materials of different wall layers.
Material Substrate Finishing mortar EPS thermal insulation layer Vapor barrier Air layer Tile finish Paint finish
Thickness (mm)
Density (kg/m3)
Thermal Conductivity Heat Capacity at Constant (W/(mK)) Pressure (J/(kgK))
240 10 100
1923 1805 25
0.5 1.23 0.035
920 840 1380
130 1.205 1905 1315
2.3 0.25 1 0.88
2300 1005.5 848 850
0.1 20 10 5
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Table 2.
Hygric property parameters of the materials of different wall layers.
Material
Vapor Diffusion Vapor Permeation Permeability (s) Coefficient (m2/s) Moisture Content (kg/m3) Resistance
Substrate
2.6 1011
1.16 108
j ð0:5277j2 þ0:9467jþ0:0708Þ
Finishing mortar
5.46 1011
1.3 108
j ð0:022j2 þ0:025jþ0:0001Þ
EPS thermal insulation layer Vapor barrier Air layer
1.1 1011
0
j ð0:0885j2 þ0:096jþ0:0163Þ
4.56 1014 —
0 0
Tile finish Paint finish
3.2
— —
—
0:034j ð5:4j2 þ5:882jþ1Þ
1
1.69 10
9
j ð1:77j2 þ1:92jþ0:326Þ
28
1.72 10
9
j ð0:0785j2 þ0:084jþ0:0189Þ
8
Moisture accumulation in a typical wall with external thermal insulation
Figure 4 shows the change in moisture content in the thermal insulation layer at 90 h, 800 h and 1400 h. At the beginning of the simulation, as there is only one mortar protection layer outside the thermal insulation layer, the outside of the thermal insulation layer is influenced by the change of ambient temperature and humidity, and there is a large temperature gradient and partial pressure gradient of vapor. Therefore, the outside of the thermal insulation layer gets moisture accumulation first. As the simulation proceeds, influenced by the high temperature and high humidity indoors, the moisture content gradually passes through the substrate to the inside of the thermal insulation layer, and the moisture content of the thermal insulation layer rises and forms moisture accumulation in the center of the thermal insulation layer. At the end of the simulation, as the moisture content transfers to the outdoor environment, the moisture accumulation points gradually move closer to the substrate side, and eventually, moisture accumulation occurs at the intersection of the two materials on the side of the thermal insulation layer near the wall. 3.3
Influence of installation of air layer
A sealed air layer is often installed on the exterior walls of buildings in engineering. Because air has good thermal insulation, installation of an air layer on the low-temperature side of the wall thermal insulation layer in cold regions can improve the thermal resistance of the wall and enhance the thermal insulation performance of the wall. However, there is a lack of studies on the influence of moisture accumulation in the thermal insulation layer. From Figures 3 and 5, it can be concluded that after installing the sealed air layer, the maximum relative humidity of the thermal insulation layer during the heating period decreased from 63% to 5%, and the maximum water content decreased from 1.32 kg/m3 to 1.18 kg/m3, which is 10.6% lower. This is due to the good thermal insulation performance of air, which reduces the influence on the thermal insulation layer by changes in outdoor temperature and humidity and plays a good protective role for the thermal insulation layer. Compared with the wall thermal insulation layer without the air layer installed, the temperature and humidity gradients in the insulation layer are reduced, which weakens the driving force to transfer moisture content. As a result, less moisture is transferred to the thermal insulation layer, and the relative humidity and water content of the thermal insulation layer are reduced.
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Figure 3.
Change in moisture content of thermal insulation layer in the heating period.
Figure 4.
Change in moisture content of thermal insulation layer in 90 h, 800 h and 1400 h.
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Figure 5.
3.4
Changes in relative humidity inside the wall after installation of air layer.
Influence of installation of vapor barrier
To prevent the moisture content indoors and in the wall from transferring to the thermal insulation layer and forming moisture accumulation, a vapor barrier is often installed between the substrate and the thermal insulation layer in engineering. From Figures 3 and 6, it can be concluded that after the installation of a vapor barrier, the relative humidity of the substrate and the thermal insulation layer decreases significantly, and the maximum relative humidity of the thermal insulation layer decreases from 63% to 57% during the heating period. At the intersection of the substrate and the thermal insulation material, where condensation is likely to occur, the relative humidity decreases to less than 50%, and the maximum water content decreases from 1.32 kg/m3 to 1.23 kg/m3, which is 6.81% lower. This is due to the large vapor permeation resistance coefficient of the vapor barrier, which hinders the transfer of moisture content from the substrate to the thermal insulation layer, thus reducing the risk of moisture accumulation in the thermal insulation layer. 3.5
Influence of setting of finish coat
When choosing building finish coats in the past, there was a lack of selection methods for building wall facade finish coat materials, and the apparent effect was often taken as the selection criterion only. Since the influence of moisture transfer under the action of coupled heat and moisture in the wall was not considered, the finish coat materials could not be selected properly and the moisture in the wall and the thermal insulation layer was excessive. This seriously damages the thermal performance of the thermal insulation material of the wall and also aggravates the mold and peeling of the thermal insulation layer, etc. In this paper, two commonly used finish coat materials, i.e., tile finish and paint finish, are selected to compare their influence on moisture accumulation in the thermal insulation layer.
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Figure 6.
Change in relative humidity inside the wall after installation of vapor barrier.
From Figures 7 and 8, it can be concluded that the moisture content of the thermal insulation layer is higher in both the case of installing tile finish and paint finish than that in the case of no finish coat. The maximum water content reaches 1.37 kg/m3 and 1.34 kg/m3 respectively, which is 5.3% and 3.1% higher compared to the no-finish coat case. This is due to the use of finish coat materials, which hinders the moisture transfer to the outdoor environment and makes it difficult for the moisture content in the thermal insulation layer to
Figure 7.
Changes in relative humidity inside wall thermal insulation layer with different finish coats.
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Figure 8. Change in moisture content of wall thermal insulation layer with different finish coats in the heating period.
dissipate, resulting in moisture accumulation. Tile finish is more likely to lead to accumulation in the thermal insulation layer than paint finish, increasing the risk of a wall being dampened.
4 CONCLUSIONS In a typical wall with external thermal insulation in a cold region, moisture accumulation tends to occur within the wall thermal insulation layer during the heating period. The accumulation locates first on the low-temperature side near the outdoor environment, and then slowly transfers to the side of the thermal insulation layer near the substrate. Finally, the intersection of the two building materials, i.e., the thermal insulation material and the substrate, is prone to long-term moisture accumulation. The installation of an air layer in a typical wall with external thermal insulation in cold regions raises the temperature of the thermal insulation layer close to the outdoor side, resulting in lower temperature and humidity gradients in the wall. The maximum water content of the thermal insulation layer is reduced by 10.6% and the maximum relative humidity drops to 56%. The installation of the vapor barrier in a typical wall with external thermal insulation in cold regions can effectively prevent moisture in the substrate from entering the thermal insulation layer, thus reducing the water content of the thermal insulation layer. During the heating period, the water content of the wall thermal insulation layer decreases by 6.81% after the installation of the vapor barrier, and the relative humidity at the intersection of the thermal insulation material and the substrate decreases to less than 50%. Compared with a wall without a finish coat, the use of a finish coat makes the overall permeability of the wall worse and tends to aggravate the moisture accumulation in the
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thermal insulation layer. Commonly used tile finish and paint finish increase the moisture content of the thermal insulation layer by 5.3% and 3.1% respectively. Preference should be given to a waterproof and breathable finish coat for engineering use.
REFERENCES Guo X. G., Chen Y. M., Chen G. J., et al. (2014) Effect of Moisture Transfer on the Hydrothermal Performance of Multilayer Wall Subjected to Hot Humid Climate [J] Architecture Technology, 45 (8): 747– 705. 10.3969/j.issn.1000-4726.2014.08.021 Hens H. (2007) Building Physics: Heat, Air and Moisture, Fundamentals and Engineering Methods with Examples and Exercises (Second Edition) [M] Berlin: Ernst & Sohn, 2007: 164.10.1002/9783433608548.ch1 Liu X. W., Chen G. J., Chen Y. M., (2016) Modeling of the Transient Heat Air and Moisture Transfer in Building Walls [J]. Journal of Hunan University (Natural Sciences Edition) 43 (1): 152–156. 10.3969/j. issn.1674-2974.2016.01.020 Luikov A V. (1964) Heat and Mass Transfer in Capillary-porous Bodies [M]. Oxford: Pergamon, 1966: 75–99. 10.1016/S0065-2717(08)70098-4 Li W., Liu F., Chen B. M., (2017) Effects of Moisture Transfer on Heat Transmission of Multilayer Wall [J]. Building Science, 33(8): 117–122. 10.13614/j.cnki.11-1962/tu.2017.08.18 Philip J R, Devries D A. (1957) Moisture Movement in Porous Materials Under Temperature Gradients [J]. Transactions American Geophysical Union, 38 (2): 222–232. 10.1029/TR038i002p00222 Yu S., Xu Z., (2012) The Analysis of Coupled Heat and Moisture Transfer in Building Envelop Based on Numerical Simulation [J]. Advanced Materials Research, 450–451: 1471–1476. https://schlr.cnki.net/Detail/ doi/GARJ2012/SJTTAFDA20F42A7DF48AAF4A1C221B780865 Wei M., Wang B., Liu S. B., (2020) Simulation of Thermo-moisture Coupling Transfer of Deep Buried Engineering Envelope Based on COMSOL Software [J]. Science Technology and Engineering, 20 (14): 5729–5736. CNKI:SUN:KXJS.0.2020-14-041 Wang Y. Y., Liu Y. F., Wang D. J., et al. (2014) Study on the Influence of Moisture Transfer on the Wall Heat Transfer [J] Acta Energiae Solaris Sinica, 35 (7): 1151–1157.10.3969/j.issn.0254-0096.2014.07.008 Zhang L.Y., (2007) Numerical Simulation of Heat and Moisture Transport in the Construction of Building External Walls [D]. Beijing: Beijing Jiaotong University, 2007. 10.7666/d.y1081342
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Bridge strain based on real bridge and indoor experiment Xiaofan Feng, Yu Tang & Lu Peng* Research Institute of Highway Ministry of Transport, Beijing, P.R. China
Bin Li Guizhou Province Quality and Safety Traffic Engineering Monitoring and Inspection Center Co., Ltd., Guizhou, P.R. China
Lin Bai China-Road Transportation Verification & Inspection Hi-Tech Co., Ltd., Beijing, P.R. China
ABSTRACT: Structural Health Monitoring (SHM) system plays an important role in ensuring the safe and stable operation of bridge structures, but there is a lack of evaluation criteria for its various monitoring data. Bridge strain can reflect the local stress state of the structure under load, which is an important indicator for measuring the performance and damage identification of bridges. In this paper, we conduct relevant research on bridge strain and explore the evaluation criteria for the structural state of the bridge through vehicle load design experiments. Finally, the results are verified through indoor experiments. The results show that the stress on the box girder of the bridge increases with the load and is usually less than 100 me. When there is a significant sudden change in the sensor reading, it often indicates that a crack has appeared at the monitoring point. When there is a crack in the structure, if the sensor performance is good, the monitoring work can still be completed. The results of this paper not only provide evaluation criteria for bridge strain, but also provide constructive suggestions for sensor performance evaluation in practical situations.
1 INTRODUCTION Highway bridges are an essential component of the transportation network and play a crucial role in promoting economic development and facilitating travel. However, during their long-term service life, bridges are inevitably subjected to environmental erosion and load fatigue, which poses a great challenge to their long-term stable operation (Chen 2018). After undergoing the process of fatigue and corrosion, some structures still meet the requirements of normal service, while others cannot support transportation work (Rodrigues et al. 2012). Therefore, it is urgently necessary to intervene in reinforcement, repair, and reconstruction work. It is essential to evaluate the current condition of the structure to take effective bridge management measures. Structural health monitoring systems can timely and effectively collect field information, which greatly assists in evaluating the structural condition (Delgrego et al. 2008; Ding & Wu 2022; Sartor et al. 1999). With the development of science and technology, structural health monitoring systems are continuously improving and expanding, making their monitoring data more diverse. Therefore, this provides an opportunity to propose targeted bridge management and
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-56
maintenance suggestions (Dai 2021; Hu 2023). However, currently, there is a lack of unified evaluation standards for various monitoring data (Wu 2015). Bridge strain can reflect the local stress state of the structure under load, which is an important indicator for measuring bridge performance and damage identification. Vehicle load is one of the most important loads for bridge structures, which often leads to bridge collapse or fatigue damage. Vehicle load is an essential basis for bridge design, condition assessment, safety evaluation, maintenance decision-making, and reinforcement (Chen 2018). Therefore, this study focuses on bridge strain, using vehicle load-designed experiments to explore the evaluation criteria for bridge structural status. Finally, indoor experiments were conducted for validation. The research results of this study can help analyze the level of strain load on bridges.
2 FIELD EXPERIMENT 2.1
Experimental site
This paper uses an old bridge spanning a river as the research object. The bridge has a total length of 1819.2 meters and is designed as a separated left and right section. The width of each half-bridge is 11.898 meters, and the main bridge has a span of 50 meters + 100 meters + 160 meters + 160 meters + 100 meters + 50 meters. The upper structure uses a prestressed variable cross-section continuous box girder, which was constructed by using the precast suspension splicing method. The watercourse spanned by the bridge is approximately 1200 meters wide, with water depths ranging from 6 to 28 meters. The navigation standard is Class I for inland waterways, with a clearance height of 22 meters. The overall appearance of the bridge is shown in Figure 1(a), and the schematic diagram of the box girder section is shown in Figure 1 (b). The testing section for the static loading experiment is the maximum positive bending moment section at the edge span.
Figure 1.
Bridge condition.
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Figure 2.
2.2
Schematic diagram of sensor layout.
Sensor layout
The sensors were placed on the cross-section of the box girder as shown in Figure 2. 2.3
Loading method
Structural stress is the intuitive reflection of the bridge’s state of force and an important indicator for measuring the safety performance of component materials. To better grasp the mechanical performance and fatigue characteristics of the bridge, it is necessary to monitor the stress condition of the key stress sections. According to the internal force influence line of the control section, a loading heavy truck was used to distribute the load so that the ratio of the control section’s moment to the design moment under the standard live load met the requirements of the test load efficiency. Eight 35-ton loading trucks were used for the static loading test of this bridge. The loading pattern for the bearing capacity test is shown in Figure 3, the axle load table of the loading truck is shown in Table 1, and the test load grading table (based on the test bending moment grading) is shown in Table 2. 2.4
Experimental loading principles
(1) The efficiency of the test load should meet the requirement of 0.95 h 1.05, where h = Sstat/Sd. Sstat is the calculated value of displacement or internal force at the test location under the test load, S is the calculated value of displacement or internal force under the design live load, and d is the dynamic factor used in the design. (2) The test loading is carried out in several stages, usually 3–5 stages, with one stage for unloading.
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Figure 3.
Table 1.
Schematic diagram of section loading.
Axle load table of loading truck (unit: tons).
Vehicle number
Weight of front axle
Weight of rear axle
Overall weight
1 2 2 4 5 6
5.63 8.61 7.97 8.65 11.11 8.88
29.8 25.73 27.67 27.53 24.05 26.04
35.43 34.34 35.55 36.18 35.22 34.92
Table 2.
Schematic diagram of wheelbase
Grading table of section loads. Designed moment
Load efficiency
Load level
X (m)
Applied moment
Car-S20
Class I highway
Car-S20
Class I highway
Lv Lv Lv Lv Lv Lv
12 10 8 6 4 3
7870 9690 11500 13400 15300 16300
12300
16300
0.640 0.788 0.935 1.089 1.244 1.325
0.483 0.594 0.706 0.822 0.939 1.000
1 2 3 4 5 6
Note: Moment unit kNm; Car-S20 means that the allowable load of the highway vehicle (main vehicle) is 20 tons
2.5
Data preprocessing
The sensors installed on the bridge face two main problems: (1) Service sensors are affected by environmental factors, and their readings may contain large amounts of noise. Figure 4 shows the reading of a sensor when a vehicle passes over the bridge. The sensor can reflect the strain of the bridge, but the noise in the data can greatly affect the experimental results.
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Figure 4.
Sensor data collection.
(2) Multiple sensors simultaneously monitor the strain of the bridge, but the data are not synchronized. Based on the above two problems, the data were processed as follows: (1) The window mean filtering algorithm was used to denoise the data, and the abnormal values and noise were removed. (2) During the static load test, the strain of the bridge will change significantly. Therefore, the collected data were standardized and then symbolized by using Symbolic Aggregate approXimation (SAX) (Jing et al. 2021; Rakthanmanon & Keogh 2013), according to the set threshold to symbolize the data segments. According to the change rule of the symbol sequence, the data of each sensor were matched. 2.6
Experimental results
After the experiment, the measured strain values of the section are shown in Table 3. The results show that the residual strains of each measurement point after unloading are relatively small. Except for the point with a small absolute value, the relative residual strains of the other measurement points meet the requirements of the testing specification (< 20%). This indicates that the section deformation is in the elastic state. As the load level increases, the strain of each measurement point increases linearly. At load level 6, the maximum measured strain value for each point is 70 me. At point A16, the sensor measured a strain of 982 me at load level 2, which is much higher than the values measured at other points. As the load level increases, the change in the strain at this point is very small. Upon investigation, a crack was found at this monitoring point, and the sensor malfunctioned.
3 INDOOR EXPERIMENT To further explore the structural strain, indoor experiments were carried out to verify the results. The concrete beam was placed on a steel pad, and hydraulic jacks were used to apply stress loads at the center of the beam. The experimental setup is shown in Figure 5. The concrete beam was subjected to loads of 0 KN, 10 KN, 20 KN, 30 KN, 40 KN, 50 KN, 60 KN, and 70 KN in sequence, and the values collected by two sensors at a certain measuring point are shown in Figure 6. The results show that when the load is less than
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Table 3.
Measured strain values of the section (me).
Measure point A1 A2 A3 A4 A5 A6 A7 A8 A9 A10 A11 A12 A13 A14 A15 A16 A17 A18
Figure 5.
Lv 1
Lv 2
Lv 3
Lv 4
Lv 5
Lv 6
24 25 26 28 26 19 11 2 12 29 29 31 31 31 23 14 1 8
31 33 31 37 32 25 14 1 14 37 36 37 37 39 28 982 0 10
39 41 39 43 41 32 19 3 16 44 44 45 45 46 36 22 0 11
44 47 46 51 48 38 22 1 17 52 51 53 53 54 41 25 1 14
55 57 55 61 57 44 27 1 21 60 58 60 61 61 46 29 1 14
63 65 64 69 66 52 32 2 23 65 64 65 67 68 51 32 0 15
Unload 1 1 1 1 0 0 0 2 3 2 2 3 1 2 2 0 0 1
Relative residual 1.50% 1.50% 1.50% 1.40% 0.00% 0.00% 0.00% 100.00% 12.50% 3.00% 3.00% 4.50% 1.40% 2.90% 3.80% 0.00% / 6.30%
Indoor experiment.
60 KN, the strain values increase slowly with the load, and the values are stable and less than 100. When the load is 60 KN, cracks appear in the beam. At this point, it can be seen that the strain value suddenly increases to 650 me, but the strain collection value remains stable. When the load is 70 KN, the strain value is 800 me. The above results indicate that when the stress exceeds the bearing capacity of the beam, the strain value will increase significantly. When the beam is in good condition, the strain value generally does not exceed 100 me. When the beam cracks, if the sensor is in good condition, the structural strain can still be monitored stably.
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Figure 6. Sensor collection data. (a) Sensor collection values under various static load levels (b) Mean graph of two sensors under various static load levels.
4 CONCLUSIONS With the acceleration of digital infrastructure construction, the bridge structure monitoring system has become more widespread. However, each monitoring system usually requires the installation of a large number of sensors, which puts great demands on the reliability of sensors. Bridge strain can reflect the local stress state of the structure under load, and is an important indicator for measuring bridge performance and damage identification. However, the current understanding of bridge strain is still relatively unclear. In this study, a static loading test was conducted on an old bridge as an example. It was found that under normal conditions, the stress on the bridge box beam increased with loading, usually not exceeding 100 me. When there was a significant mutation in the sensor reading, it often indicated that a crack had appeared at the monitoring point. In addition,
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this study verified this law through indoor loading tests and found that even if cracks appeared in the beam, if the sensor performance was normal, the structural strain could still be monitored. Therefore, when the load increases but the sensor does not show any significant changes, it indicates that the sensor has malfunctioned. The results of this study not only provide an evaluation standard for bridge structure strain, but also provide constructive suggestions for evaluating sensor performance in practical situations.
REFERENCES Chen, Z.C. (2018) Processing and Modeling Methods For Spatial Monitoring Data of Loads and Responses of Long-span Bridges. Harbin Institute of Technology, Harbin. https://kns.cnki.net/KCMS/detail/detail.aspx? dbname=CDFDLAST2019&filename=1018895491.nh Dai, Y.Y. (2021) Present Situation and Future Development Trend of Highway Bridge Structural Health Monitoring System. Transport Manager World, (29): 79–81. https://kns.cnki.net/kcms2/article/abstract?v= 3uoqIhG8C44YLTlOAiTRKibYlV5Vjs7iJTKGjg9uTdeTsOI_ra5_XWx8MqYL3dO059ZuNIe3pSn3dy gyNQ7hquGv0P_vIwl9&uniplatform=NZKPT Delgrego, M.R., Culmo, M.P., Dewolf, J.T. (2008). Performance Evaluation through Field Testing of Century-Old Railroad Truss Bridge. Journal of Bridge Engineering, 13 (2): 132–138. DOI:10.1061/(asce) 1084–0702(2008)13:2(132). Ding, X., Wu, Z. (2022). Strain Prediction of Continuous Rigid Frame Bridge Health Monitoring Based on Grey Neural Network. China Water Transport, 22 (10): 127–128. https://kns.cnki.net/kcms2/article/ abstract?v=3uoqIhG8C44YLTlOAiTRKibYlV5Vjs7iJTKGjg9uTdeTsOI_ra5_XY7XFz4MTI2gClshia QUTwOg9oFH1YmwfrVYj_An5Gez&uniplatform=NZKPT Hu, Y.L. (2023) Structural Health Monitoring Anomaly Detection Based on Model and Data-Driven. Highway, (03): 278–281. https://kns.cnki.net/kcms2/article/abstract?v=3uoqIhG8C45S0n9fL2suRadTyEV l2pW9UrhTDCdPD64QcJumhXDP-JDzr5a_MThtqvV3pvIYv53NXLfTDCQrhC3zdesBOLYZ&uniplatform= NZKPT Jing, G.Q., Duan, F.J., Peng, L. & Cui, J.J. (2021). On-line Calibration Method for Structural Strain Monitoring System Based on Passive Excitation. Yi Qi Yi Biao Xue Bao/Chinese J. Sci. Instrum. 2021, 41 (08): 137–145. DOI:10.19650/j.cnki.cjsi.J2107610. Rakthanmanon, T., Keogh, E. (2013). Fast Shapelets: A Scalable Algorithm for Discovering Time Series Shapelets. Proceedings of the 13th SIAM International Conference on Data Mining (ICDM2013), Austin, Texas, USA, 668–676. https://www.researchgate.net/publication/289766940_Fast_Shapelets_A_Scalable_ Algorithm_for_Discovering_Time_Series_Shapelets Rodrigues, C., Cavadas, F., Félix, C., et al. (2012). FBG Based Strain Monitoring in the Rehabilitation of a Centenary Metallic Bridge. Engineering Structures, 44: 281–290. http://dx.doi.org/10.1016/j. engstruct.2012.05.040. DOI:10.1016/j.engstruct.2012.05.040. Sartor, R.R., Culmo, M.P., & Dewolf, J.T. (1999). Short-term Strain Monitoring of Bridge Structures. Journal of Bridge Engineering, 4 (3). DOI: 10.1061/(ASCE)1084–0702(1999)4:3(157) Wu, S.M. (2015) Status and Development of Structural Health Monitoring in Civil Engineering. Scientific and Technological Innovation, (23): 218. https://kns.cnki.net/kcms2/article/abstract?v=3uoqIhG8C44YLTlO AiTRKibYlV5Vjs7ir5D84hng_y4D11vwp0rrtdJsKfjlLGF2fickZEVzZzMgqQp6C2i5d2K5NUUBaHZC& uniplatform=NZKPT
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Design of test platform for early fire detection of power transmission and transformation equipment Guo-qiang Liu*, Guo-chun Li & Ya-nan Hao State Grid Shandong Electric Power Research Institute, Jinan, China
Zhi-peng Zhao Institute of Automation, Qilu University of Technology (Shandong Academy of Sciences), Jinan, China
ABSTRACT: Electric energy is one of the basic energy to ensure social production and life, and the stable operation of the power grid is particularly important for the development of smart cities. In recent years, frequent fire accidents with power equipment have seriously affected the safe operation of power transmission and transformation systems. In this paper, the aspirated smoke fire detector based on the principle of cloud and mist chamber is used to design a fire monitoring test platform for power transmission and transformation equipment. The early fire risk of switch cabinets, cable trenches, power equipment rooms and other places can be identified by scientifically arranging the sampling pipe network. The monitoring and early warning platform can send out fire warnings in real time, collect monitoring data, and display the changing trend of fire information in the field environment in the form of curves. The test shows that the platform can accurately warn of fire and locate the fire source, has high sensitivity, reliability and convenience, and can play an important role in the safe operation of power transmission and transformation equipment.
1 INTRODUCTION Electric power is the lifeblood of the development of the national economy. At the same time, the development level of the electric power industry is an important indicator of a country’s economic development. A safe and efficient electric power system is a strong support for the healthy development of the national economy. China’s “Fourteenth FiveYear Plan” and the 2035 long-term goal outline mentioned that we should strengthen the construction of important transmission projects such as power transmission channels. The whole power transmission and transformation system has a huge structure, complex equipment operation environment, closely interconnected power grid links, transformers, reactors, capacitors and other equipment, and many combustibles, making its fire safety face serious challenges. Through the research and test of fire monitoring and early warning of power transmission and transformation equipment in typical scenarios, it can further deepen the understanding of fire characteristics and fire prevention and control of power systems, form targeted fire prevention, control technical schemes and fire accident prevention measures and help to enhance the security guarantee of power transmission and transformation system. In terms of early fire monitoring of power equipment, the main technical means at home and abroad at present include point-type smoke detection, point-type temperature detection, *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-57
aspirating smoke detection, infrared ultraviolet flame detection, thermal imaging, optical fiber temperature measurement, etc., and various detection technologies have different emphasis and different application scenarios. Experts and scholars in relevant fields at home and abroad have also carried out a lot of research on fire occurrence mechanisms, detection mode, recognition effect and other aspects. Xie and others analyzed the defects of several electrical fire monitoring technologies from the aspects of national standard requirements and practical application effects and proposed an edge computing technology system based on multi-dimensional process parameter calculation and sensor-level microelectronic structure. Pan studied the research and development of high-sensitivity detection systems based on particle counting and light scattering principles at home and abroad in terms of the importance of ultra-early fire detection and the development trend of detectors and analyzed the necessity of multi-parameter combination and interference factor elimination in fire detection to accurately judge the fire. Based on the technology of fire particle detection and time-sharing cyclic sampling and positioning, this paper designs a test platform for the fire monitoring system of power transmission and transformation equipment, tests the effectiveness of the aspirated smoke fire detector used for the fire detection of the relatively enclosed space of the power transmission and transformation equipment, collects and analyzes the test data, and verifies the applicability of the aspirated smoke fire detector to the fire of the power transmission and transformation system.
2 FIRE CHARACTERISTICS OF POWER TRANSMISSION AND TRANSFORMATION EQUIPMENT As an important safety threat to power equipment, fire is closely related to the electric heating effect in the case of a short circuit and an overload of equipment. When the temperature of the equipment rises abnormally, the number of fire particles will increase rapidly. At the contact points of some electrical equipment, such as switch, contact terminal, etc., if there is a fault or when power is on or off, there will be an electric arc and electric spark due to the instantaneous change of load, which will generate high temperature locally to ionize the surrounding air to form plasma and generate a large number of ionized particles. With the gradual increase of the concentration of fire particles, the particles continue to spread around. Without the influence of airflow, the particles will spread in an “inverted cone” upward and diagonally upward direction. In the process of diffusion, if there is flowing airflow, it will affect the diffusion direction and speed of fire particles, making the fire particles accelerate along the airflow direction, and the concentration will be diluted to a certain extent. When the fire continues to intensify thermal decomposition particles and visible smoke particles, even gray and black smoke particles will be generated, further increasing the diffusion range. When an open fire finally occurs, the smoke generated will fill the protective space. In the early stage of fire, smoke particles are generated. Because only smoke particles do not generate open fire, temperature sensing detection, infrared ultraviolet detection, etc. cannot be identified. Through the aspirated smoke detection technology aiming at the characteristics of early fire, the concentration change of small smoke particles in the air can be identified and the fire risk can be perceived in advance.
3 ASPIRATED SMOKE DETECTION PRINCIPLE 3.1
The detection principle of the cloud chamber
Under normal conditions, the number of invisible particles floating in the air is only about 20000/cc and only 25000/cc–30000/cc in high dust areas. In the early stage of fire, after the
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cable insulation layer or plastic shell of power equipment is overheated, its surface will release tiny thermal decomposition particles, which can reach 500000/cc–1 million/cc in a short time. The change of particle concentration can quickly identify the fire risk.
Figure 1.
Pyrolysis particles and smoke particles.
The aspirated smoke fire detector uses the light scattering principle to detect the particles generated in the early stage of the fire. The light emitted by the light source in the photoelectric detection darkroom is reflected or scattered by smoke particles, and the photosensitive device converts the received optical signal into an electrical signal for output. When the particle concentration in the early stage of fire is low, the optical signal received by the photosensitive device is very weak, and the detection circuit needs to filter and amplify the signal before accurate identification. In signal processing, weak signals are easy to be impacted. The interference will also be amplified, affecting the accuracy of detection. To strengthen the effective optical signal and avoid the interference of dust particles, small particles are condensed into small water droplets through the cloud chamber, and the volume is increased to increase the scattered light intensity. The cloud chamber is a device filled with saturated steam of air and alcohol. When charged particles enter, the volume of the container expands rapidly, the temperature decreases rapidly, and the steam is in a supersaturated state. The above particles are contained in the center of each small water droplet by using the condensation characteristics of water droplets so that the small smoke particles to be measured can be converted into small water droplets. The light-reflecting surface formed by foggy water droplets is larger, and the reflected light intensity also increases in response. The quantity of pyrolytic particles in the air is calculated by the reflected light intensity received by the photosensitive device. The number of dust particles is relatively small compared with pyrolysis particles (about 1:25 or more), so it can accurately distinguish between dust interference and fire smoke particles to avoid false alarms. The aspirating smoke detection process using a cloud chamber is shown in Figure 2. 3.2
Time-sharing circulating sampling fire location device
The aspirated smoke fire detector extracts the air in the protected space through the sampling pipe and sampling hole for monitoring. One detector can correspond to multiple sampling pipes to protect a large space. Once the detector detects a fire alarm, it needs to locate the specific location of the protected space. The time-sharing cyclic sampling fire
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Figure 2.
Cloud chamber detection model.
location device is a method to realize early fire location. According to the distribution of power equipment in the protected space, different detection areas are divided. The sampling pipe branch of each detection area is equipped with a pipeline control valve that can be remotely controlled by the network. The air sampling airflow in each detection area is controlled by the time-sharing and alternate start and stop of the pipeline control valve. When the detector detects that any branch has an early fire risk, the pipeline control valve will not switch, but will continue to monitor the detection area with fire risk until the smoke concentration reaches the alarm threshold to send a fire alarm, or the fire risk is eliminated, and continue to monitor the multi-zone cyclic sampling. The aspirated smoke fire detector performs air sampling, identification analysis and fire alarm location for all detection areas in a time-sharing cycle sampling mode. The layout of the fire location device is shown in Figure 3.
Figure 3.
Schematic diagram of time-sharing cyclic sampling location.
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4 FIRE MONITORING SYSTEM PLATFORM FOR POWER TRANSMISSION AND TRANSFORMATION EQUIPMENT Based on the aspirating smoke detection system using the principle of cloud and fog chamber and time-sharing cyclic sampling and positioning technology, the early fire detection test of power transmission and transformation equipment was carried out, and the test platform was built for different scenarios of power transmission and transformation equipment such as Switch cabinet, cable trench and power equipment room. Through the analysis of fire characteristics in different scenarios of the power transmission and transformation system, the fire monitoring and early warning system are tested to verify the performance of early fire detection. By analyzing and studying the test results and data, the fire detection scheme and system are optimized. 4.1
Sampling pipe layout
The aspirated smoke fire detector draws air through the sampling pipe network arranged in the protection space for monitoring. Different protection space and equipment layouts adopt different sampling pipe layout methods. The layout of sampling pipes for scenarios such as switch cabinet, cable trench and power equipment room is designed below. 4.1.1 Layout of sampling pipe of the switch cabinet Circuit breakers, relays, capacitors, etc. are installed in the high-voltage switch cabinet. The electrical equipment is connected through cables. The early fire risk caused by a short circuit, overload, excessive contact resistance, etc. of the internal equipment is not easily detected. The thermal decomposition particles and smoke generated by the fire are not easy to spread out because of the enclosure of the cabinet, so the sampling pipe should be arranged inside the cabinet. The sampling pipe parallel is extended to the top of the cabinet along the top of the cabinet, the pipeline control valve is installed on the sampling pipe, and a capillary sampling pipe is added behind the pipeline control valve. The capillary sampling pipe has an outer diameter of 7 mm and an inner diameter of 5 mm. The end is inserted into the switch cabinet and fixed on the top surface of the cabinet to directly extract the air inside the cabinet. The sampling pipe layout of the switch cabinet is shown in Figure 4.
Figure 4.
Layout of switch cabinet.
4.1.2 Layout of sampling pipe of cable trench The internal structure of the cable trench is narrow and long. The cable tray is arranged in layers up and down along both sides. Cables are laid on the cable tray. The fire risk inside the 448
Figure 5.
Layout of cable trench.
cable trench is mainly caused by the thermal decomposition of the insulation layer of the cable due to overload, short circuit, fault heating, etc., which releases a large number of thermal decomposition particles, and the heat accumulation leads to fire. The sampling pipe inside the cable trench is laid parallel to the cable laying direction along the outer edge of the upper cable tray and is fixed on the cable tray with stainless steel ties. In the setting of sampling hole spacing, GB 50116-2013 Code for Design of Automatic Fire Alarm System requires that the protection area and radius of each sampling hole should meet the protection area and radius of the point-type smoke fire detector. The specification requires that the protection area of a single smoke detector is 80 m2 and the protection radius is 6.7 m in a space with a floor height of 12 m and an area of no more than 80 m2. The space of the cable trench is narrow and long, the interior is relatively closed, and the number of air changes is low. After a comprehensive analysis, one sampling hole is set every 12 m on the sampling pipe. 4.1.3 Layout of sampling pipe of the switch cabinet Circuit breakers, relays, capacitors, etc. are installed in the high-voltage switch cabinet. The electrical equipment is connected through cables. Large power equipment is installed in the power equipment room (including box-type transformer, transformer room, etc.), which will generate a large amount of heat energy due to voltage conversion and energy consumption in daily operation. When the power equipment in the equipment room is seriously overloaded (including an external short circuit) and fails, it is easy to bring fire risk, and a large number of thermal decomposition particles will be released under abnormal heating, electric arc, electric spark, etc. The size of the equipment room selected for the test is 4.5 m 4.5 m 5 m, the internal space volume is 101.25 m3, and the floor area is 20.5 m2. The protective area of the sampling hole of the aspirated smoke fire detector is affected by the number of air changes in the space. According to the standard requirements, the larger the number of air changes, the smaller the protective area of the sampling hole is. The space with air changes is 30 freq/h, the protection area of a sampling hole is 23 m2, and the protection radius of the sampling hole is 4.8 m; The space with air changes is 80 freq/h, the protection area of a sampling hole is 9 m2, and the protection radius of the sampling hole is 3 m. The minimum protection area of each sampling hole is 9 m2, setting two sampling holes in each equipment room. The sampling pipe is fixed by a metal suspender, and the distance between the two suspenders is 1.5 m to 2 m. The sampling pipe extends horizontally above the electrical equipment in the equipment room and is more than 1.5 m from the top of the electrical equipment. A large arc elbow at the corner of the sampling pipe is used to reduce the airflow resistance. The opening direction of the sampling hole is vertically downward, and two sampling holes are arranged on both sides of the power equipment.
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Figure 6.
4.2
Layout of the equipment room.
Transmission network layout
The transmission network is mainly equipped with the gateway, switch, serial communication module and other data transmission equipment. The aspirating smoke fire detector is connected to the monitoring host and the pipeline control valve through the transmission network. The RJ45 network cable is connected from the detector to the switch through the pipe, and connected to the monitoring host by the switch. By configuring an independent code and network IP address for the detector, the upload of monitoring and early warning data is achieved, realizing remote control of the monitoring host. The RS485 data line from the pipeline control valve is connected to the serial port gateway, and the RJ45 network line and the detector network line are connected to the switch and monitoring host in parallel to realize a communication connection. 4.3
Fire monitoring and early warning system
By deploying the “relatively enclosed space pre-fire monitoring and warning system” on the monitoring host, including databases, software programs, etc., on the interface of the monitoring and early warning system, the detector information (attribute information, parameter information, operation status, monitoring zone), real-time monitoring data, positioning and
Figure 7.
Signal transmission network.
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Figure 8.
Monitoring and early warning system interface.
early warning information, as well as equipment account information and data analysis, etc. can be clearly and intuitively queried. The changing trend of fire information in the field environment can be displayed in the form of curves. The parameters of the detector and sampling control device can be set and adjusted.
5 FIRE MONITORING SYSTEM PLATFORM TEST The system test can be conducted after the fire monitoring system platform is built. In the switch cabinet test, the lower part of the switch cabinet is installed with trunking and cables. The cable overload tester is connected with the 1 m long test cable in the trunking. In the cable trench test, the cable overload tester connects the test cable with a length of 5 m on the cable trench bridge. In the test of the power equipment room, the cable overload tester is placed on the test table below the sampling hole, and the two ends of the test cable with a length of 2 m are respectively fixed on the two terminals of the overload tester. For the above tests, stainless steel ties are used to fix the K-type thermocouple and the test cable together to monitor the temperature change during the test. Before the test, the time of the detector, temperature recorder and monitoring host shall be checked and consistent. The camera shall be used to record the changes of time, current and temperature. The monitoring host terminal can display the change curve of smoke particle concentration and early warning conditions (time, alarm level, monitoring value, positioning information, etc.) in real time. The automatic power-off time of the cable overload tester is set to 600 s so that it can automatically have power-off protection when the time is up. Turning on the power supply of the cable overload tester for the test and loading a large current (70–150A) on the cable, the temperature of the cable will gradually rise with the increase of the current, and reach the heat resistance limit. Cable generates thermal decomposition particles. The particle concentration gradually increases and is inhaled by the sampling hole and extracted to the aspirating smoke fire detector for data analysis. The monitoring data is sent out from low to high until reaching the set threshold. The data and change curve of fire particle concentration can be viewed at any time in the early warning system at the monitoring host. During the whole process of cable heating, overheating and smoking, the smoke concentration value, fire warning time and warning level of the fire monitoring system are tested. In addition, through the cable overload test in different detection areas, the air sampling, monitoring, and early warning efficiency in different environments, and the control effect of time-sharing cyclic sampling and positioning device are tested to verify the positioning and early warning function of the fire monitoring system. 451
Figure 9.
Smoke generated by cable overload.
6 CONCLUSIONS In this paper, based on the principle of cloud and mist chamber, the aspirated smoke fire detection and time-sharing cyclic sampling and positioning method, the early fire monitoring and prealarm system for power transmission and transformation equipment are designed. According to several typical equipment locations and fire characteristics, different sampling pipes and sampling hole layout methods for different protection spaces are designed, such as switchgear, cable trench and power equipment room. A test platform was built for the cable overload smoke detection test. The test verifies that the aspirating smoke fire detector can effectively identify a large number of invisible thermal decomposition particles generated by the early fire of electrical equipment, and can locate the fire area, quickly and accurately identify the early fire risk.
ACKNOWLEDGMENTS This paper was supported by the Independent Research and Development Project of State Grid Shandong Electric Power Company (2021A-077)
REFERENCES Bakhoum E G. 2012. High-sensitivity Miniature Smoke Detector [J]. IEEE Sensors Journal, 2012, 12 (10): 3031–3035. Guoqiang Liu, Guihai Li, Zhipeng Zhao, et al. 2022. Application of Perfluorohexadone Fire Extinguishing Agent in Fire Prevention and Control of Power Transmission and Transformation Systems [J]. Shandong Electric Power Technology, 2022, 49 (01): 36–40. Jia J, Wang W, Liu H, et al. 2021. Fire Ignition Mechanism and Extinguishing Method of Cable in High Voltage Switchgear [C] 2021 IEEE 2nd China International Youth Conference on Electrical Engineering (CIYCEE). IEEE, 2021: 1–5. Jia-sheng L I. 2009. Residual Current Detector Design for Fire Supervisory [J]. Electrotechnics Electric, 2009. Krüll W, Tobera R, Willms I, et al. 2012. Early Forest Fire Detection and Verification Using Optical Smoke, Gas and Microwave Sensors [J]. Procedia Engineering, 2012, 45: 584–594. Liu X, Hou D, Ji J, et al. 2021. Experiment and Numerical Simulation of Cable Trench Fire Detection [J]. Case Studies in Thermal Engineering, 2021, 28: 101338. Martin D. 1993. Safety against fire risks [C] Proceedings of Intelec 93: 15th International Telecommunications Energy Conference. IEEE, 1993, 2: 332–335. Qian Tang, Qizhi Pang, Chao Wang. 2008. Analysis of Cable Trench Fire Accidents and Preventive Measures [J]. Industrial Safety and Environmental Protection, 2008 (01): 53–55. Wentao Dai. 2017. Research on Fire Detection and Alarm Technology for Cable Tunnels and Integrated Pipe Corridors [J]. Fire Science and Technology, 2017, 36 (01): 89–92. Yereance R A, Kerkhoff T. 2010. Electrical Fire Analysis [M]. Charles C Thomas Publisher, 2010.
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Landscape conservation planning based on forest fire prevention: A case study of Lushan Forest Park in Sichuan Province Meixin Qiu & Zigang Yao* Department of Landscape Planning and Design, East China University of Science and Technology, Shanghai, China
ABSTRACT: To explore the ecological sensitivity of mountain scenic spots and how to combine forest fire prevention with ecological conservation planning, the GIS spatial analysis method was used to study Lushan Forest Park in Liangshan Prefecture, Sichuan Province. Different from the indexes and entry points selected by previous studies on sensitivity, this paper considers the effects of site topography, stand structure, and climate factors on forest fire occurrence, and studies forest fire factor as one of the evaluation factors for ecological sensitivity. The study shows that the ecological sensitivity of Lushan Forest Park is high on the whole. According to the ecological sensitivity classification, the ecological planning area is divided, and relevant suggestions are put forward based on forest fire prevention.
1 INTRODUCTION In recent years, forest fires occur frequently all over the world, causing serious ecological and economic losses. Examples include extreme fire in California in 2017, Liangshan Prefecture in China’s Sichuan Province in March 2019, the Amazon rainforest in Brazil in August 2019, and Australia in September 2019. Facing the great challenge of forest fire, scholars have made a lot of research on forest fire protection. Fire has been an important medium for landscape and resource management since prehistoric times (Schwartzman et al. 2013). Fire also has an ecosystem regulatory function, helping to mitigate the negative potential threat of extreme fire to biological communities (Sil et al. 2019). However, overly strict fire prevention policies, economic interest-oriented felling, invasive alien species, and other factors (Fischer et al. 2016) have suppressed natural fires, and forest fuel accumulation (Ager et al. 2016) has greatly increased the probability of major fires. In recognition of the importance of fire in ecosystems, many countries and regions have begun to adopt measures such as planned burning and rational forest restoration plans to reduce fire risk and improve the resilience of ecosystems. For example, fire protection landscape planning in California, the United States, combines vegetation, climate, topography, and human activities in the region and successfully reduces the incidence and loss of fire by adopting a variety of fire prevention measures (Gobster et al. 2004). However, there are few studies on the conservation planning of scenic spots considering the factors of the forest fire. A sub-theme of the 2020 IFLA (International Alliance of Landscape Architects) International Student Landscape Design competition is natural disasters, which put forward the question of how landscape architecture practice can help reduce the occurrence of natural disasters. Therefore, this paper proposed making an empirical analysis from the perspective of landscape planning and design to explore landscape conservation planning based on forest fire defense. *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-58
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2 METHODOLOGY 2.1
Research area
Lushan-Qionghai Scenic Spot is located in the suburb of Xichang City, capital of Liangshan Yi Autonomous Prefecture, Sichuan Province, China, which is about 8 km away from the city center. Geographical coordinates are 102 16’–102 20’E and 27 47’–27 52’N. Scope of the Lushan Tour area is defined as follows: (1) The Front Mountain of Lushan Mountain (Mountains facing Qionghai Lake east of Ridge Line) is from 4.5 km along the ridge line of Lushan Mountain to the top of the main peak, the top of the White Cloud Temple, along the hillside to Guayao village, retreat outward 80–500 M, the east along the No. 108 National road to the west about 200 meters, partially to the No. 108 national road as the boundary. (2) Back Mountain of Lushan (mountain west of the ridge line): 200–400 meters on both sides of Wuxian Temple and deep gully, 200–400 meters on both sides of Xiangshuigou. This paper takes this area as the planning area of Lushan Forest Park, with a planned area of 11.67 km2.
Figure 1.
2.2
Location map of the study area.
Data source
The data in this study were mainly from Google Qionghai-Lushan Scenic Spot in Xichang with a 1:1000 digital topographic map, Xichang Qionghai-Lushan Scenic Spot Planning in Xichang City (2011–2030), description of Qionghai Scenic Spot Planning in Xichang City (2005–2020), and other relevant data released by Liangshan Bureau of Statistics. 2.3
Experimental procedure
There is abundant research on the evaluation of ecological sensitivity. To weaken the relationship between various ecological factors and consider the representativeness, comprehensiveness, and operability of data, based on previous studies (Guoyu et al. 2019) and from the perspective of respecting the site and natural disaster, we selected the five most sensitive factors for ecological spatial planning, namely elevation, slope, aspect, water confluence, and forest fire. Elevation, slope and aspect catchment can be analyzed based on DEM data obtained by ArcGIS. Forest fire factors can be determined according to the comprehensive judgment of terrain, climate, wind direction and other factors to determine the prone location of hill fire and analyze its spreading range by ArcGIS. 454
Figure 2.
2.4
Experimental flow diagram.
Spatial analysis model
Kernel Density Estimation is a method to identify and analyze the clustering characteristics, which can determine the spatial agglomeration of sample points in the region. For data x, x, x . . . , the kernel density estimation formula is as follows: fn ð x Þ ¼
x x 1 Xn i K i¼1 r nr
(1)
where xi is the position coordinate of sample point I (I = 1, 2, 3, . . . n), n is the number of samples, r is the search radius calculated by kernel density, and K is the kernel function (Jingyao et al. 2016). Hydrological analysis tools in ArcGIS were used to generate catchment basins and flow networks in the study area. The principles of flow direction, confluence accumulation and river network calculation are shown in Figure 3 (Yong et al. 2018).
Figure 3.
Flow direction, confluence accumulation and river network calculation.
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The weighted superposition method was adopted to construct the ecological sensitivity evaluation model of Lushan Forest Park. The calculation formula is as follows: Xn E¼ B Wk (2) k¼1 k In Formula (1), E is the comprehensive evaluation value of ecological sensitivity; n is the total number of factors affecting ecological sensitivity; k is the number of factors affecting ecological sensitivity; Bk is the sensitivity evaluation value of factor numbered k; Wk is the weight of the factor numbered k, and W1 + W2 + . . . Plus Wk is equal to 1 (Huali et al. 2005).
3 RESULTS AND DISCUSSION 3.1
Weight analysis and single factor evaluation
Through expert consultation and demonstration, and referring to relevant studies (Dongguo et al. 2015; Yuanyuan et al. 2017), each factor is assigned 5, 4, 3, 2, and 1, respectively, indicating its ecological sensitivity. In ArcGIS 10.4 platform, the data on ecological factors were standardized, and the reclassification tool was used to reassign every single factor. The AHP method is used to analyze the weight of these 5 ecological factors. After constructing a random judgment matrix and consistency test, the consistency ratio CI/RI value is 0.008928571 ( 0.1), which has a satisfactory consistency (Zhi et al. 2002).
Table 1.
Criteria for evaluation and classification of ecological sensitivity factors.
Serial number
Factors
Classification
Value
Weight
1
Elevation
Slope
3
Aspect
4
Water
5
Fire
1 2 3 4 5 1 2 3 4 5 1 2 3 4 5 1 2 3 4 5 1 2 3 4 5
0.1544
2
< 1700 m 1700 – 1800 m 1800 – 1900 m 1900 – 2000 m > 2000 m First-grade (< 10 ) Second-grade (10 –19 ) Third-grade (20 –29 ) Fourth-grade (30 –40 ) Fifth-grade (40 –90 ) First-grade (south) Second-grade (southwest, southeast) Third-grade (east, west) Fourth-grade (northwest, northeast) Fifth-grade (north) Level 1 catchment area Level 2 catchment area Level 3 catchment area Level 4 catchment area Level 5 catchment area First-grade (< 250 m Spread range) Second-grade (250–500 m Spread range) Third-grade (500–750 m Spread range) Fourth-grade (750–1000 m Spread range) Fifth-grade (1000–1250 m Spread range)
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0.3087
0.0738
0.1544
0.3087
3.2
Comprehensive evaluation of ecological sensitivity
According to Model (2), the weighted superposition tool in ArcGIS spatial analysis module was used for each single factor layer to obtain the single factor evaluation diagram and comprehensive evaluation diagram of the ecological sensitivity of Lushan Forest Park (Figure 4). In the comprehensive evaluation, because the fourth and fifth grades are very close, the fourth and fifth grades are combined into the same grade, which can be divided into four sensitive zones.
Figure 4.
3.3
The area and proportion of the four zones.
Ecological conservation planning of Lushan Forest Park
Based on the comprehensive analysis of the social, economic, and cultural status and ecological sensitivity of Lushan Forest Park, the planning of Lushan Forest Park should be based on the protection of the ecological system of the scenic spot with ecological sensitivity analysis as the core basis, ecological protection, and recreation as the main functions of the scenic spot, and it should be divided into four zones; Ecological conservation zone (ECZ), an area of 198.95 hectares, accounting for 17.04%; Restricted development zone (RDZ), an area of 541.18 hectares, accounting for 46.36%; Moderate development zone (MDZ), an area of 376.12 hectares, accounting for 32.22%; Tourism development zone (TDZ), an area of 51.16 hectares, accounting for 4.38%.
Figure 5.
The area and proportion of the four zones.
3.3.1 Ecological Conservation Zone (ECZ) In the highly sensitive areas, most of the elevations were above 1900 m. The steep slope is all > 30 ; There are certain catchment areas, most of them are in the starting area of the primary catchment area, and the radiation range of easy ignition point is the highest, so they should be set up as an ecological conservation zone. It is necessary to pay attention to the protection of ecology, and implement a forest fire prevention policy, regularly stipulating 457
workers enter the forest operation in the season when the fire is not easy to remove the accumulation of combustible materials, such as dead leaves and weeds; It is strictly prohibited to encroach on the habitats of animals and plants. Adequate living and active sites must be provided for the animals and plants in the area to prevent the fragmentation of woodland patches and create multi-level and species-rich plant communities. No construction facilities (except a forest fire observation tower) shall be constructed, and no disturbance to the ecological environment shall be caused. 3.3.2 Restricted Development Zone (RDZ) The sensitive area is with a high elevation and steep slope. There are certain catchment areas, mainly distributed in the west, north and south of the study area. The restricted development zone is close to the ecological conservation zone, and there is the famous Lu Shan temple group on the site, which contains many immovable cultural relics and movable cultural relics, for example, Yi Slave Society Museum, Sanjiao Nunnery, Guangfu Temple, etc. Due to historical reasons, these ancient temples were built here, but the site is highly sensitive to ecology. Therefore, ecological protection should be considered while protecting these cultural heritages, and encroachment of land within the area should be prohibited. The building materials of these ancient temples require a higher level of fire protection, so attention should also be paid to fire prevention. To make use of the geographical advantages of the site to carry out cultural heritage popularization education activities, delimit the scope of construction and development, define its use, restrict the recreational behavior of tourists in sensitive areas, adopt flow control management, prohibit nearby residents from occupying and constructing the land in the area, and prevent the destruction of the habitat of organisms. 3.3.3 Moderate Development Zone (MDZ) The moderately sensitive zone, with the flat terrain and gentle slope, is mainly distributed in the east and southeast of the study area along the lake and the areas near the west of the south with a gentle slope and strong catchment. The ecological sensitivity of this region is low, and human activities are more complex and rich. It is close to Qionghai Scenic Spot and is close to National Road. Convenient transportation advantages and cultural elements can be used to build tourist attractions with local cultural characteristics. In terms of forest fire prevention, the site is far from the fire-prone area and near the water source. It is suggested to plan a demonstration area for controlled fire and a multistage fire-resistant forest belt which plans to burn in the season when forest fires are not easy to happen. On the one hand, it is conducive to the renewal of the forest ecosystem; On the other hand, it can provide popular science education for tourists. 3.3.4 Tourism Development Zone (TDZ) In the less sensitive area, the terrain has no obvious fluctuation, the slope is gentle, the altitude is low, the catchment is more, and the area is close to the tourist area and villages, which is mainly distributed in the southeast corner of the study area and the central part along the lake. Close to the national road and with convenient transportation, most of the buildings are B&Bs, hotels, and surrounding villages in Qionghai Scenic Spot. They have low ecological sensitivity and weak ecological functions. Some supporting service facilities are available for construction and development to a high degree to provide a site for the service center of the scenic spot to carry out some recreational activities. It is suggested to set up the Liangshan fire memorial landscape to commemorate the martyrs who died in the fire, carry out science popularization education of forest fire defense and set up life-saving experiences and other activities.
4 CONCLUSIONS By studying the ecological sensitivity and ecological conservation planning of Lushan Forest Park in Liangshan Prefecture, Sichuan Province, this paper draws the following conclusions: 458
(1) The ecological sensitivity of Lushan Forest Park is relatively high on the whole, and the ecological sensitivity level can be divided into four levels, which are less sensitive area, accounting for 4.38%; moderately sensitive area, accounting for 32.22%; sensitive area, accounting for 46.36%; high sensitive area, accounting for 17.04%; (2) According to the classification of ecological sensitivity, ecological planning suggestions are put forward for Lushan Forest Park, which can be divided into four zones: ecological conservation zone, restricted development zone, moderate development zone and tourism development zone; (3) Considering the natural ecology, cultural relics protection, scenic spots and human life safety in the adjacent urban areas of Lushan Forest Park, it is suggested to set up a reasonable fire prevention system, but more scientific fire prevention methods should be used, such as setting up a controlled fire demonstration area in the area with low ecological sensitivity, far away from the fire-prone area and near the water source, and carrying out a proper plan to burn combustibles under the forest. At present, some achievements have been made in the research and practice of ecological sensitivity, but there is still a lack of ecological planning for fire-prone scenic spots. How to better plan the forest area requires the cooperation of landscape planning experts, forestry planning experts, and fire prevention experts to make more thinking and more effective practice for the harmonious coexistence of humans and nature.
REFERENCES Ager, A.A., Day, M.A., Vogler, K. (2016). Production Possibility Frontiers and Socioecological Tradeoffs for Restoration of Fire-adapted Forests. Journal of Environmental Management, 176,157–168. doi:10.1016/j. jenvman.2016.01.033. Dongguo Z., Binggeng X., Yonglin C. (2015). Tourism Land Strategy of Mountain Tourism city Based on Ecological Sensitivity Evaluation: A Case Study of Zhangjiajie City. Economic Geography, 35 (6), 184–189. doi:10.15957/j.cnki.jjdl.2015.06.026. Fischer, A.P., Vance-Borland, K., Jasny, et al. (2016). A Network Approach to Assessing Social Capacity for Landscape Planning: The Case of Fire-prone Forests in Oregon, USA. Landscape and Urban Planning, 147, 18–27. doi:10.1016/j.landurbplan.2015.10.006. Gobster P. H., Westphal L. M. (2004). The Human Dimensions of Urban Greenways: Planning for Recreation and Related Experiences. Landscape and Urban Planning, 68 (2–3), 147–165. doi:10.1016/ s0169-2046(03)00162-2. Guoyu W., Weilan B. (2019). Research and Practice Progress of Ecological Sensitivity Assessment of Scenic Spots. Chinese Landscape Architecture, 35 (2), 87–91. https://kns.cnki.net/kns8/defaultresult/index. Huali C., Yonghua W., Guoping D., et al. (2005). Analysis of Ecological Sensitivity in Landscape Ecological Planning of Tourism Destination: A Case Study of Fenghuang Nanhuashan National Forest Park in Hunan Province. Landscape Architecture, 2, 39–40. https://kns.cnki.net/kns8/defaultresult/index. Jingyao K., Jinhe Z., Huan H., et al. (2016). Spatial Distribution of Traditional Villages in China [J]. Progress in Geography, 35 (7): 839–850. https://kns.cnki.net/kns8/defaultresult/index. Schwartzman, S., Boas, A.V., Ono, K. Y., et al. (2013). The Natural and Social History of the Indigenous Lands and Protected Areas Corridor of the Xingu River Basin. Philosophical Transactions of the Royal Society B: Biological Sciences, 368 (1619), 20120164–20120164. doi:10.1098/rstb.2012.0164 Sil, Â., Fernandes, P. M., Rodrigues, A. P., et al. (2019). Farmland Abandonment Decreases the fire Regulation Capacity and the Fire Protection Ecosystem Service in Mountain Landscapes. Ecosystem Services, 36, 100908.doi:10.1016/j.ecoser.2019.100908 Yongzhong T., Wenjian W., Yaobin S., et al. (2018) GIS Spatial Analysis Basics Tutorial. Science Press, Beijing. pp. 252–257. https://www.sciencep.com Yuanyuan S., Lin W., Jin W. (2017). Ecological Sensitivity Assessment of Yellow River Delta Nature Reserve. Journal of Ocean University of China (Natural Science Edition), 47 (11), 96–102. doi:10.16441/j.cnki. hdxb.20160329 Zhiguo H., Yan L., Zhihua F., et al. (2002). Calculation of High-order Mean Random Consistency Index (RI) in Analytic Hierarchy Process. Computer Engineering and Applications, 12, 45–47 + 150. https://kns.cnki. net/kns8/defaultresult/index.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Analysis of the widening reconstruction of existing bridge based on grillage analysis theory Bing Li* & Jinhui Chen School of Civil Engineering, Fujian Chuanzheng Communications College, Fuzhou, China
ABSTRACT: The traffic demand in our country is increasing constantly, and many bridges have been designed with lower load limits and longer ages. The bridge deck width is insufficient and the traffic capacity tends to be saturated. It is an important and urgent task to increase the traffic capacity of existing roads. The cost of bridge reconstruction and reinforcement can save about 60% 90% compared with the new bridge. Bridge widening is an effective way to solve the traffic pressure. In this paper, the mechanical performance of the bridge is studied from three aspects: the whole force, the lateral force, and the antioverturning ability. The original bridge model, the single beam model, and the grillage model are established respectively by using the finite element software. The force difference between the rigid-jointed model and the hinged model is analyzed and compared. The main research results are as follows: there is a big difference in torque between the rigid-jointed model and the hinge-jointed model. Widening the bridge under dead load and live load can improve the bearing capacity of the old bridge. Compared with the bridge, the antioverturning capability of the span bridge is improved, and the anti-overturning coefficient of the rigid-jointed model is greater than that of the articulated model. For the original bridge, the width bridge increased the torsional stiffness of the bridge, anti-overturning ability to meet the requirements. The results of this paper can provide technical guidance and theoretical support for the reconstruction of existing bridges.
1 INTRODUCTION Because of the limitation of the design concept at that time, the early-built highway could not make an accurate forecast of the future traffic volume. With the development of the national economy and the increase in traffic volume, many bridges are facing the problem of insufficient traffic capacity. The widening of bridges that do not meet the need for access can improve the transport capacity of national trunk lines and enable the entire national transport network to play its role in promoting further economic development. In order to adapt to the development of traffic volume under the economic premise, the widening of existing bridges is the first choice, in line with the requirements of national sustainable development (Zhang 2020; Zhou 2020a, 2020b). There are many ways to widen the bridge. When widening existing bridges, it is necessary to select appropriate widening methods according to existing bridge structure type, structure status, construction conditions, and technology (Chen 2016; Zhang 2021). From the point of view of structural forces, different widening methods and techniques of splicing will lead to different stress systems after splicing. Therefore, the joint widening method is the primary factor to determine the mechanical performance of the bridge structure (Maarten 2018; Song *Corresponding Author: [email protected]
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2021; Wang 2020). At present, the existing bridge splicing and widening methods mainly include the following. Liao (1990) proposed adding a side beam widening method in which he pointed out that the lower support structure also needs to be widened; the specific widening methods include increasing the cover beam and adding an inclined bracing system. Li (2015) proposed adding a cantilever beam widening method where he pointed out that a tbeam bridge by adding a cantilever beam is a relatively simple widening add sidewalk. Kenneth (2006) proposed the strutted box widening method (SBWM) to widen the existing box girder by adding inclined bar support. Wang (2013) proposed the SCWCBM method for adding a steel cantilevered widening concrete box girder and verified its effectiveness through full-scale model tests. This method not only build a new pier column and increase the clearance under the bridge but also possess the advantages of good structural integrity and beautiful shape after widening. Xu (2011) proposed a new bridge widening method to build a new bridge on the outside of the existing bridge with the same or similar structure form as the existing bridge, and then connecting to the existing bridge. This paper mainly studies the mechanical performance of the bridge, mainly from the overall force, lateral force, and anti-overturning capacity of three aspects of analysis. The original bridge model, the single beam model, and the grillage model are established respectively by using the finite element software. This paper mainly studies the influence of the structure connection form and lateral moving load on the structure and analyzes and compares the force difference between the rigid-joint model and the hinge-joint model. The different schemes of bridge widening are evaluated and analyzed.
2 GRILLAGE ANALYSIS THEORY 2.1
Grillage theory
The grillage method is the equivalent grillage analysis which divides the structure section into discrete parts. The advantage of the grillage method is that it can take into account bending, torsion, distortion, and other factors, and can more accurately reflect the lateral stress state of the structure. Therefore, the grillage method can be used to obtain more accurate results when calculating the transverse forces of the bridge. The center line of every main beam should be kept at the center line of webs as far as possible when the beam-grid method is used for section division. The length of grillage division should be moderate as far as possible, and the neutral axis of each longitudinal main beam section should be consistent with the neutral axis of the original structure section. The breakdown is shown in Figure 1.
Figure 1.
Schematic diagram of beam grillage division.
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Under the consideration of the shear lag effect, the torsional stiffness of the longitudinal section can be calculated as Formula 1: Id ¼
2h2 t1 t2 t1 þ t2
(1)
The calculation of the transverse bending stiffness of the box section is shown as Formula 2: Iy ¼
h 2 t 1 t2 t1 þ t 2
(2)
where Id Torsional stiffness Iy Bending stiffness t1 Thickness of top plate t2 Base plate thickness hCenter distance between the top and bottom plates In calculating the torsional stiffness of the converted section, only the deformation caused by rigid torsion is considered, and the influence of transverse distortion is neglected. When the box girder is torsional, the shear force is transmitted through the top and bottom plates and the side webs, and part of the shear force is transmitted through the middle webs. The total torque on the cross-section is composed of the opposite shear forces between the longitudinal beams. The distribution of shear force in the torsion section is shown in Figure 2.
Figure 2.
Schematic diagram of the shear distribution of grillage.
Through the balance equation, the torsional stiffness of the longitudinal member can be calculated as follows: GJx ¼ G
(3)
GJx ¼ 2Gðb1 h02 t1 þ b2 h002 t2 Þ
(4)
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Figure 3.
Schematic diagram of internal force balance of grillage joint.
From the above formula, the torsional stiffness per unit width of the section is calculated in Formula 4. GJx ¼ 2Gðh02 t1 þ h002 t2 Þ ¼
2Gh2 t1 t2 ðt1 þ t2 Þ
(5)
where b1 Width of top plate b2 Base plate width 0
h Distance from=mml : mi > the top plate to section centroid 00
h Distance=mml : mi > from the bottom plate to section centroid
2.2
Virtual beam theory
The simulation of a virtual crossbeam is brought about by the division of beams and lattices above. The height of the virtual crossbeam h is the average thickness of the cantilever or the thickness of the roof, and the width of the virtual crossbeam is the sum of half the distance between the left and right crossbeams. The virtual beam schematic is shown in Figure 4:
Figure 4.
Virtual beam.
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The stiffness of the virtual beam is calculated as follows: Vertical stiffness of virtual edge members: half of the width of the cantilever section. Bendingstiffness : I ¼
B d 03 B d 03 B ; Torsional stiffness : C ¼ ; Shear area : I ¼ d 0 2 12 2 6 2
(6)
Lateral stiffness (unit width) of virtual edge members: only the roof is considered. Rigidity : I ¼
d3 B d 00 3 ; Torsional stiffness : C ¼ ; Shear area : I ¼ d 12 2 6
(7)
where 0
d Average thickness of cantilever plate dAverage thickness of the roof Lateral stiffness (unit width) of virtual edge members: only the roof is considered.
3 STRUCTURAL STRESS ANALYSIS 3.1
The finite element model
(1) Finite element beam model In this paper, the Midas Civil model is used, and the new and old bridges are simulated by beam elements, respectively. In the middle, the virtual beam is used to simulate the transverse stiffness, and the models are established respectively, as detailed in Figure 5. One is a rigid-jointed model, the other is a hinged model. The hinge model uses Midas Civil to release the beam-end constraint and eliminate the constraint on bending moment to simulate the hinge. The bearing arrangement is simulated by double supports, and the torsion effect is considered.
Figure 5.
A sketch of the finite element model of the bridge.
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(2) Finite element grillage model In this paper, the transverse force of the bridge should be analyzed. When the singlebeam model is used, the simulation of the lateral force is not precise enough, and it can not reflect the distribution of the internal force of the structure. Therefore, it is necessary to establish a refined model, using the grillage method to establish the model and the use of vehicle load for comparative analysis. In this paper, Midas civil is used to build a grillage model, as shown in Figure 6. The boundary of the model and the value of the load are the same as above. After the grillage model is established, the cross-section characteristics of each longitudinal beam and cross-beam are calculated and modified. Two models are established, one is the rigid connection model and the other is the hinge connection model.
Figure 6.
Finite element model of whole bridge lattice.
(3) Anti-overturning finite element model Due to the rapid development of China’s transportation in recent years, there have been several bridge capsizing accidents, and thus bridge capsizing is widely concerned. In the process of bridge reconstruction, it is very necessary to analyze the anti-overturning ability of the bridge in the widening reconstruction. In this section, the original bridge model is established, and the rigid-jointed model and hinge-jointed model of the widened bridge are reconstructed. The anti-overturning analysis of these three models is carried out respectively. This paper studies the influence of the split width reconstruction on the anti-overturning ability of the original bridge and the influence of the different connection forms on the anti-overturning ability. This paper mainly studies the influence of the anti-overturning ability of the original bridge. The finite element model is modeled by Midas Civil, which is shown in Figure 6. 3.2
Results of structural stress analysis
(1) Results of stress analysis on superstructure In order to compare the results of the analysis model, the typical cross-section internal forces are selected for comparative analysis. The internal force results of the middle
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Figure 7.
Torque diagram under dead load.
section of each span and the top section of the pier of the continuous beam are extracted and analyzed. A sketch of the pier top and mid-span section is shown in Figure 7. Figure 7 shows that the calculation results of internal forces of the Broadside Bridge under dead load are better than those of the old bridge. It shows that the force of the old bridge can be improved and the integral stiffness of the old bridge can be increased after the bridge is widened. Figure 8 shows that the bending moment of the rigid-jointed model is slightly lower than that of the articulated model under automobile load. Because the virtual beam shares part of the transverse moment, the shear forces are basically the same. Under automobile load, the torque of the rigid-jointed model is obviously lower than that of the articulated model, and the torque of the rigid-jointed model is also slightly lower than that of the articulated model under overall heating. This is because the rigid-jointed model lateral stiffness and torsional strength are greater. The influence of the two different connection modes on the torsional stiffness of the structure should be taken into account in the design and construction of the bridge. (2) Transverse force analysis The calculated positions in this section are the mid-span sections of transversely moving loads, and the results are shown in Figure 9. From Figure 9, the moment diagrams of the rigid-jointed model and the hinge-jointed model show that the data of the longitudinal beam at both ends are close, and the moment of the hinge-jointed model of the middle longitudinal beam is larger than that of the rigid-jointed model. The bending moment of the hinged model is 10.47% and 11.47% higher than that of the rigid model respectively. Because the moment can not be transferred when hinged, the moment of the longitudinal beam at the middle connection of the bridge increases, and the increased load value is about 10% compared with the rigid
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Figure 8.
Sketch of internal force of cross-section under automobile load.
Figure 9.
Internal force diagram of transverse moving load.
connection model. Under the action of transverse moving load, the axial force diagram of the rigid-jointed model and the hinged model has a large difference between the two ends on the transverse section. The axial force of the articulated model is larger than that of the rigid-jointed model in general and increases by about 11% in the old side beam, about 50% in the new side beam, and about 2% 7% in the middle longitudinal beam. (3) Results of anti-overturning analysis The overturning axis of the original bridge is chosen to connect the left and right supports of the Central line of the road. The detailed overturning calculation results are shown in Table 1.
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Table 1.
Results of anti-overturning analysis. Project
Widen the bridge Original Bridge model
Models Results
Articulated model
Characteristic pedestal
Li (m)
4
The vertical force
Dead load standard value
RGi
(kN)
Standard value of the most unfavorable live load
RQ11
Feature State I check
1.0RGi + 1.4RQ11
Feature State II check
Just connect the model
4
4
2006.02
2006.22
1270.66 141.24
272.42
2019.29
240.36
Checking conclusions P Stability effect RGiLi (KN.M) P Instability effect (KN.M) RQ11Li P P Stability coefficient RGiLi/ ( RQiLi)
1808.28
1624.84
Meet the requirements 8077.14
8024.1
8024.87
5082.64
564.98
1089.66
7
128.5
33.2
From the above calculation results, the original bridge and the bridge meet the specifications of anti-overturning requirements, because the bridge is a double-bearing, so the bridge’s anti-overturning capacity is high. The safety of the bridge structure meets the safety requirements, and the anti-overturning ability of the rigid-jointed model is greater than that of the articulated model.
4 CONCLUSION In this paper, based on the analysis and research of the structural stress of the span bridge, the old bridge model, the single beam model, and the grillage model are established respectively by selecting the typical bridge; moreover, the overall force, lateral force, and anti-overturning of the bridge are analyzed in detail. The main conclusions are as follows: (1) The results of the whole force analysis show that there is little difference between the rigid-jointed model and the hinged model in the moment and shear force. The torsional stiffness of the rigid-jointed model is better than that of the rigid-jointed model. The stress of the bridge can be improved under dead load and live load, but it is more disadvantageous under temperature load. (2) The influence of the connection forms of the new and old bridges on the structure is mainly reflected in the following aspects: the bending moment mainly affects the position of the middle connection of the transverse section, and the moment effect of articulation is about 10% larger than that of rigid connection; the axial force mainly affects the two ends of the transverse section; The influence of the old bridge side is smaller than that of the new one, and the shear effect of hinge is 11% and 50% larger than that of rigid connection respectively. (3) Compared with the Starý most bridge, the anti-overturning capability of the span bridge is improved, and the anti-overturning coefficient of the rigid-jointed model is greater than
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that of the hinged model. In view of the original bridge, the span bridge is equivalent to the original bridge with additional bearings, increasing the torsional stiffness of the bridge. (4) The results of this paper can provide technical guidance and theoretical support for the reconstruction of existing bridges. The corresponding reasonable theory method is put forward to provide a reliable basis for drawing up relevant design codes for bridge widening and reconstruction, and the research results can also provide a reference for existing bridge widening design.
ACKNOWLEDGMENTS Research on deviation modeling and coordinated design of steel truss girder for highway and iron (Fujian Provincial Transportation Science and Technology Project 202029).
REFERENCES Chen Kang-ming, WU Qing-xiong, Chen Bao-chun, ZHANG Gang. China Journal of Highway and Transport, 2016, 29(11):99–107. Kenneth W. Shushkewich. Transverse Analysis of Strutted Box Girder Bridges[J]. Journal of Bridge Engineering, 2006, 11(1). Liao Shao-xian. Some Characteristics of Highway Bridge Widening [J]. South-central Highway Engineering, 1990(02):51–52. Li Ze-tao, Wang Qian. Beam Bridge Widening Method and its Application [J]. Northern Communications, 2015(02):9–14 + 17. Maarten Rikken, Daan Tjepkema, David Gration. Using Fracture Mechanics Principles in Steel Bridge Renovation Projects[J]. IOP Conference Series: Materials Science and Engineering, 2018, 419(1). Song She-xian. Construction Technology Analysis of Expansion Joints in Road and Bridge Engineering [J]. Heilongjiang Science, 2021, 12(04):112–113. Wang Xiao-min, Wang Chao, Li Jiang. Key Technology of Steel Anti-collision Wall Construction of Urban Overpass [J]. Engineering Construction,20,52(09):70–73. Wang Qian. Theoretical Analysis and Experimental Study on Widening Method of Orthotropic Steel Cantilever Plate of Concrete Box Girder [D]. Dalian University of Technology, 2010. Xu Qiang, Splicing Technology of Bridge and Culvert Structures in Expressway Reconstruction and Expansion Project, Beijing, People’s Communications Press, 2011, 1–5 Zhang Lei. Research on Key Technologies and Problems of Highway Bridge Foundation Construction [J]. Transportation World, 2020(22):101–102. Zhou Xin-ping, Tang Chen-hao, Ding Wen-bo, Zong Xue-mei. Mechanical Properties Analysis of Cross-hole Widened Continuous Box Girder [J]. Road Machinery & Construction Mechanization, 2020a, 37(07):11–16. Zhou Jianhua. Construction Technology Analysis of Bridge Reinforcement and Bridge Widening [J]. Housing and Real Estate, 2020b(09):220 + 252. Zhang Guo-hui, Yang Zhen-dong, Yan Yi-zhi, Wang Ming-ming, Wu Liang, Lei Hong-jun, Gu Yan-shuang. Experimental and Theoretical Prediction Model Research on Concrete Elastic Modulus Influenced by Aggregate Gradation and Porosity[J]. Sustainability, 2021, 13(4).
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Acoustic design comparison for one auditorium indoor environment: A case study of bell-shaped University hall in China Ziqing Tang* Architecture College, the Taiyuan University of Technology, Taiyuan, People’s Republic of China
Zhengguang Li Department of Architecture, Zhejiang University of Science and Technology, Hangzhou, People’s Republic of China
Meijun Jin Architecture College, the Taiyuan University of Technology, Taiyuan, People’s Republic of China
ABSTRACT: In the procedure of acoustic design for this chosen auditorium with a bellshaped plane in China, there appear several apparent differences indicating that a listener, particularly in the audience area in the designed acoustical configuration, would have a subjective perception of greater sound source strength, slightly higher reverberation, higher sound clarity but poorer lateral sound energy disturbance on the whole. Furthermore, concerning the audience area characterization, it is observed that persons on the platform would appreciate a stronger sensation of reverberation than in the typical positions of the audience area configuration. Besides, the subjective aspects that remain common for both configuration situations are the perceived lateral sound energy distribution in the middle audience area, as well as the apparent sound width, but some stronger differentiation in the back corners of upper and lower seats, elaborating a quieter sound environment there when one participating in an activity or enjoying a performance on the stage.
1 INSTRUCTIONS In one auditorium, the multifunctional acoustic performance is usually characterized by several auditory elements (lecturing, reporting, meeting, symphony orchestra, singing, word, and music listening), and the geometry and materials complexity of the auditorium contribute to the formation of sub-spaces, which can be evaluated concerning the activities at that location (Alberdi 2021; Autio 2021; Berard 2016) [1–3]. It is essential to discover the optimization differentiation before and after this hall’s careful acoustic design and configuration to provide suggestions in these spaces with similar acoustic questions.
2 A CASE STUDY 2.1
Description of the Auditorium
This auditorium at Fuzhou Foreign University aims to contribute to the student’s daily meeting reports and performance usage. The auditorium has 820 full seats, including 653 *Corresponding Author: [email protected]
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pool seats, 2 disabled seats, and 165 balcony seats. The auditorium has a volume of about 5800 m3, with the volume of each seat about 7.07 m3/seat. The plane pattern of the auditorium is approximately bell-shaped, with a maximum size of 23.8 m in width, and the back wall of the pool is far from the platform’s front edge line, with a distance of 28.7 meters and a total area of about 580 m2. The stage has a maximum size of 8.2 m in height and 12.6 m in width. An emphasis on the diversity of hall functions characterizes the auditorium. It is about creating a stage and an intimate space between the audience, increasing communication, and creating a good performance environment. Despite the latter, a sound amplification system will be installed. 2.2
Design requirement
For an auditorium, its successful sound quality is quite dependent on the subjective perception of the audience. One hall recognized as having excellent sound quality must have the best physical acoustic parameters. Optimal acoustics parameters need to be realized by building measures to be suitable for multifunctional purposes. This acoustic optimization measure can be carried out using Odeon software to predict the acoustic behavior of the auditorium and the design of the indoor finishes. The auditorium hall should achieve low background noise in the auditorium. Noise isolation ability for this hall has to be strong, especially for the internal air conditioning outlet noise control, a key element for this question. Enough loudness, speech clarity, and music clarity are also selected for the plan due to the same utilization consideration. The strong direct sound from the sound source on the stage can be obtained in the audience. The reflective surface of the auditorium has good diffusion characteristics, and the sound field is evenly distributed to avoid improper body shape and defective sound. Meeting the purpose of variety shows and other functions, the auditorium also considers the functions of movies and conferences. The sound absorption and frequency characteristics should be controlled at a reasonable value to obtain the ideal reverberation time and frequency properties. The former is a main control acoustic parameter regarding sound design construction. Generally, if the reverberation time is too short for the room’s high clarity, the sound will be dry and silent; otherwise, the speech clarity and intelligibility will get loud and confused. The reverberation time is chosen as one index for refurbishment in the multifunction hall. Each audience obtaining abundant early reflection sound, especially early side reflection sound, can improve listening intimacy and presence. Ultimately, the stage persons must have good self-hearing and mutual hearing conditions.
3 MATERIALS AND METHODS 3.1
Optimization parameters
Considering the optimization requirements above, it is stipulated in GB/T 50356-2005 “Code for Architectural Acoustic Design of Multi-purpose Halls of Theaters and Cinemas” in China. Considering the auditorium modulus volume can be determined at an indoor intermediate frequency (500 Hz–1000 Hz), the best reverberation range is 0.9 s–1.3 s. Since the auditorium also needs to support artistic performance, film and conference, speech and music sound listening, the middle frequencies (500 Hz–1000 Hz) determine the auditorium; Otherwise, the reverberation time in low frequency (center frequency 125 Hz band) should be increased by 10%–30%, and the high frequency (center frequency 2000 Hz– 4000 Hz, band) should be 10%–20% reduction. Ensure that the persons operating in the hall can know all the information the sound broadcasting system sends.
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Figure 1. Relationship between volume and RT in a room for the meeting, reporting, and multifunctional usages.
Background noise control is also to China’s current standard GB/T 50356-2005” Code for Architectural Acoustic Design of Theater, Cinema and Multi-purpose Hall.” The auditorium room is determined to apply with the NR-30 noise evaluation values (Table 1). Table 1.
NR-30 noise evaluation values. Octave frequency band central frequencies (Hz)
NR values
31.5
63
125
250
500
1000
2000
4000
8000
NR-30
76
59
48
39
34
30
26
25
23
The index describing the uniformity of sound field distribution in each position in the hall and expressing the difference of sound pressure level in different positions under the condition of the natural sound source, the inequality of the sound field in the hall should be 4 dB. The difference between the maximum and the lowest voice level should be 8 dB. Fast speech transmission index STI represents the signal change degree of the original sound signal at the receiving point under the action of the indoor sound field. In this design, take STI 0.6, excellent speech clarity. 3.2
Objectives of sound quality evaluation
The results of the simulation of the unused sound-absorbing materials in the audience can be predicted that the appropriate intermediate frequency reverberation time is 1.89 s, and the average value of speech transmission index STI is 0.51. The sound absorption materials should be reasonably arranged to reduce the acoustic Reverberation Time of the environment due to too long RT to improve the indoor sound absorption volume, especially at medium-low frequencies (100 Hz–1000 Hz). The panels are absorbent resonant perforated systems, working as Helmholtz resonators. Knowing the optimal RT of the auditorium, the frequency with maximum acoustic absorption was identified, and the geometric parameter values of the perforated resonant panel were calculated. It is suggested several correspondent methods for the decorative effect that the depressed parts of the side walls of the auditorium are arranged with 15 mm round hole sound absorbing silicon ceramic plate (the perforation rate is not less than 12%, the pore diameter is less than 8 mm), the rear cavity is greater than 100 mm installed. The sound insulation cotton is filled inside. The speech transmission index STI is related to speech intelligibility or speech articulation, and there is a good correspondence between them, as shown in Table 2.
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Table 2.
Relationship of STI and speech intelligibility.
STI
> þ þ þ gH þ ¼0 < 1=3 @t @x @y @x Hffiffiffiffiffiffiffiffiffiffiffiffiffiffi ffi p 2 2 2 > > : @N þ @ðuNÞ þ @ðvNÞ þ gH @Z þ g n u u þ v ¼ 0 1=3 @t @x @y @x H
(2)
where H is the water depth, m; Z is the water level, m; M, N are the single-width flow in x, y direction, m2/s; u, v are the components of the average flow velocity in x, y direction, m/s; n is the roughness coefficient; g is the acceleration of gravity, take 9.8 m/s2; t is the moment, s; q is the source-sink term. 2.3
Flood risk zoning
The flood risk zone is defined by the degree of integrated risk, which reflects the spatial distribution of flood risk and the variability of flood risk levels between regions, under the combined influence of multiple magnitudes of flood inundation. The “Combined Risk (R)” value for each calculation unit is determined using the following formula. Hi þ Hiþ1 R¼ ðpi piþ1 Þ 2 i¼1
(3)
H ¼ a1 a2 h
(4)
4 X
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where i ranges from 1 to 5, corresponding to 5a, 10a, 20a, 50a, and 100a, respectively. The frequency of flooding in a certain year is denoted by p, where p = 1/i. The maximum inundation depth is represented by h, while the equivalent depth value is denoted by H, dm. The correction coefficient for maximum travel velocity is a1, and the correction coefficient for maximum inundation duration is a2. When considering the influence of “Maximum inundation depth” as the main factor, it is necessary to take into account the risk elements of “Maximum travel speed” and “Maximum inundation duration” and adopt the “Equivalent water depth”. In areas with insufficient information and where the flood risk is mainly characterized by the inundation depth, a1 and a2 are both set to 1. The risk level of the calculation unit is determined based on the integrated risk degree (R) index, which is divided into four categories: low, medium, high, and extremely high. The range of risk differentiation levels is shown in Table 1. (Wang 2021) Table 1.
Risk zoning level delineation table.
R
R KoYuLun. Hence, the Aksu region was the strongest, and KoYuLun was the weakest by comprehensively considering the hail disaster intensity and disaster rate. Comparing the hail disasters in 20 counties and cities, it was found that the annual occurrence frequency was the largest in Wenquan (2.7 times), followed by that (2.5 times) in Aksu. The annual disaster-affected area was the maximum in Aksu (9383 hm2), followed by that (6400 hm2) in Shawan. The annual economic losses were the maximum in Aksu (13.65 million yuan), followed by those (9.01 million yuan) in Awat. As for the annual disaster index, Aksu (4.46) and Awat (2.75) ranked top two. 3.2
Monthly variability of hail disasters
The monthly distribution of hail disasters on both sides of the Central Tianshan Mountains exhibited obvious variability, and hail disasters were concentrated from May to September, accounting for 96% of the whole year (Figure 3). The peak occurrence frequency of hail disasters in the four regions—Bozhou, Kuima Basin, Aksu region, and KoYuLun region— appeared in July, July, June, and May, respectively, the peak disaster-affected area appeared in July, May, May, and May, respectively, the peak economic losses appeared in July, May, June, and May, and the peak disaster index appeared in July, May, May, and May, respectively. 3.3
Annual variability of hail disasters
The interannual changes in the occurrence of frequency, disaster-affected area, and disaster index of hail disasters all showed a significant linear growth trend, indicating that the
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Figure 3. Monthly distribution of hail disasters on both sides of the Central Tianshan Mountains during 1961–2020.
occurrence frequency and intensity of hail disasters increased year by year (Figure 4). During 1961–2020, the occurrence frequency of hail disasters grew by 6.8 times every ten years, the disaster-affected area by 17400 hm2 every ten years, the economic losses by 24.30 million yuan every ten years, and the disaster index by 8.1 every ten years.
Figure 4. Interannual variation of hail disasters on both sides of the Central Tianshan Mountains during 1961–2020.
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4 DISCUSSION Climate warming and humidifying are conducive to the occurrence and development of hail clouds, which will lead to more and stronger hail disasters. During the period of 1961–2020, the interannual variation of precipitation and the average temperature in the warm season (May–September) on both sides of the Central Tianshan Mountains showed a significant linear increase trend, with the precipitation increasing by 5.3 mm every ten years. The average temperature is increasing by 0.14 C (Figure 5a, 5b) every ten years. Therefore, evident warming and humidifying take place in the warm season with frequent hail disasters, which will certainly aggravate the occurrence frequency and intensity of hail disasters.
Figure 5. Interannual variation of precipitation, temperature, seeding area and hail days on both sides of the Central Tianshan Mountains during 1961–2020.
The continuous expansion of crop planting areas will also cause the increasing frequency and intensity of hail disasters. During 1961–2005, the crop planting area in the Central Tianshan Mountains increased linearly by a small margin, increasing by 4370 hm2 (Figure 5c) every year. However, the linear increase amplitude was evidently enlarged during 2006–2020, increasing by 77750 hm2 every year, and the linear growth amplitude of the latter was 17.9 times that of the former. This revealed that agricultural production developed rapidly, and a vast area of wasteland was reclaimed and improved during 2006–2020. The occurrence frequency and intensity of hail disasters are jointly affected by the changes in climate factors and crop planting area. In this study, the influencing degrees of the precipitation (RR), average temperature (TT) in the warm season, and crop planting area (S) on the occurrence frequency (C) and disaster index (Z) of hail disasters were discussed through the MLR method. Next, the MLR formulas of C and Z were acquired through the least square method as C = 0.404RR + 0.309TT +0.209S (R = 0.64, P Bozhou (2.5) > KoYuLun region (1.4). The disaster rate was sorted in descending order as Aksu region (8.7%) > Bozhou (6.9%) > Kuima basin (4.1%) > KoYuLun region (3.3%). Among the 20 counties and cities, the maximum occurrence frequency, disaster-affected area, economic loss, and disaster index appeared in Wenquan (2.7 times), Aksu (9383 hm2), Aksu (13.65 million yuan), and Aksu (4.46), respectively. The monthly distribution of hail disasters on both sides of the Central Tianshan Mountains displayed apparent variability. They were mainly concentrated from May to September, accounting for 96% of the year. The most intense hail disaster in Bozhou appeared in July, which occurred in May in the Kuima basin, Aksu region, and KoYuLun region. The occurrence frequency and intensity of hail disasters on both sides of the Central Tianshan Mountains were increasing year by year. Specifically, the occurrence frequency of hail disasters, disaster-affected areas, economic loss, and disaster index presented significant interannual linear growth trends, which grew by 6.8 times, 17400 hm2, 24.3 million yuan, and 8.1, respectively, every ten years during 1961–2020. On both sides of the Central Tianshan Mountains, the precipitation, temperature in the warm season (May-September), and crop planting area increased year by year, with their contribution rates to the occurrence frequency (intensity) of hail disasters being 44% (46%), 33% (15%), and 23% (39%), respectively. Due to the climate warming-humidifying progress and reclaiming wasteland, the occurrence frequency and intensity of hail disasters on both sides of the Central Tianshan Mountains grew year by year. 580
ACKNOWLEDGMENTS This work was supported by the Research Project of Macao Polytechnic University (RP/ FCHS-03/2022).
REFERENCES Cao, Z. (2008) Severe Hail Frequency Over Ontario, Canada: Recent Trend and Variability. Geophys Research Letters, 35: 1–3. https://doi.org/10.1029/2008GL034888. Changnon, S.A., Changnon, D. (2000) Long-term Fluctuations in Hail Incidences in the United States. Journal of Climate, 13(3): 658–664. https://journals.ametsoc.org. Chen, B.X., Chen, K., Wang, X., et al. (2022) Spatial and Temporal Distribution Characteristics of Rainstorm and Flood Disasters Around Tarim Basin. Polish Journal of Environmental Studies, 31(3): 2029–2037. DOI:10.15244/pjoes/143579. Eccel, E., Cau, P., Riemann-Campe, K., et al. (2012) Quantitative Hail Monitoring in an Alpine Area: 35-year Climatology and Links with Atmospheric Variables. International Journal of Climatology, 32(4): 503–517. DOI:10.1002/joc.2291. Kunz, M., Sander, J., Kottmeier, C. (2009) Recent Trends of Thunderstorm and Hailstorm Frequency and Their Relation to Atmospheric Characteristics in Southwest Germany. International Journal of Climatology, 29 (15), 2283–2297. DOI:10.1002/joc.1865. Ma, Y., Wang, X., Zhao, B. K., et al. (2002) Spatial and Temporal Statistical Character of Hail in Xinjiang. Bimonthly of Xinjiang Meteorology, 25(1): 4–5. http://smylzqx.cnjournals.com Malkarova, A.M. (2011) Estimation of Physical Efficiency of Hail Protection Accounting for Changes in Hail Climatology. Russia Meteorology Hydrology. 36(6): 392–398. DOI:10.3103/S1068373911060057. Mladjen Ćuri´c, Dejan Janc. (2016) Hail Climatology in Serbia. International Journal of Climatology, 36(9): 3270–3279. DOI:10.1002/joc.4554. Punge, H.J., Kunz M. (2016) Hail Observations and Hailstorm Characteristics in Europe: a Review. Atmospheric Research, 176–177: 159–184. http://dx.doi.org/10.1016/j.atmosres.2016.02. Simeonov, P., Bocheva, L., Marinova, T. (2009) Severe Convective Storms Phenomena Occurrence During the Warm Half of the Year in Bulgaria (1961-2006). Atmospheric Research, 93 (1–3), 498–505. DOI: 10.1016/j.atmosres.2008.09.038. Wang, X., Chu, C.J., Mou, H. (2020) Spatial Pattern and Interannual Variation Characteristics of Snow Disaster in Xinjiang. Arid Zone Research, 36(6): 1488–1495. DOI: 10.13866/j.azr.2020.06.13. Wang, Z., Zhang, X., Wang, X., et al. (2016) Accumulated Impact of Operating Conditions on the Specific Cake Resistance in Dead-end Microfiltration Mode. Desalination and Water Treatment, 57, 1967–1976. DOI:10.1080/19443994.2014.983178. Xie, B.G., Zhang, Q.H., Wang, Y.Q. (2008) Trends in Hail in China During 1960-2005. Geophysical Research Letters, 35, L13801. DOI:10.1029/2008GL034067. Zhang, C.X., Zhang, Q.H., Wang, Y.Q. (2008) Climatology of Hail in China: 1961-2005. Journal of Applied Meteorology and Climatology, 47, 795–804. DOI:10.1175/2007JAMC1603.1.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Study on the scheme of lowering the flood control section Ting Zhang* Zhejiang Institute of Hydraulics and Estuary (Zhejiang Institute of Marine Planning and Design), Hangzhou, Zhejiang Province, China Zhejiang Provincial Key Laboratory of Hydraulic Disaster Prevention and Mitigation, Hangzhou, Zhejiang Province, China
Junbo Yao* Shangyu District Water Conservancy Bureau, Shaoxing, Zhejiang Province, China
ABSTRACT: The flood control section of the reservoir is crucial to the flood control operation of the reservoir and the flood control safety of the downstream river. In recent years, with the development of the social economy and the completion of river regulation, the downstream flood control section has also changed accordingly. Taking a reservoir as an example, this paper evaluates the necessity and feasibility of the downward movement of the flood control section by analyzing the flow capacity of the downstream channel of the reservoir and provides strong support for the actual operation of the reservoir.
1 INTRODUCTION The construction of reservoirs, especially large reservoirs, often bears the important responsibility of improving the downstream flood control capacity. The main purpose of setting up the flood control section is to provide an important basis for flood control dispatching decisions of the basin by monitoring the flow and water level of the flood control section and combining it with the hydrological forecasting system when the reservoir or the flood control protection object encounters different storm floods and to guide each reservoir to carry out flood dispatching with higher quality (Gou 2015). With the rapid development and progress of water conservancy in China, the improvement of the regulation and flood discharge capacity of the downstream river section of the reservoir has laid a good foundation for the implementation of the reservoir control section.
2 EXISTING PROBLEMS Due to historical reasons, there are various problems in the downstream flood control sections of many reservoirs. For example, the location of some flood control sections is far from the reservoir, and the catchment area of the section is large, so it is difficult for the reservoir to carry out compensation regulation and operation; The location of the flood control section selected by some reservoirs does not meet the conditions for flow monitoring, and the flow control operation of the control section cannot be implemented in practice; The location of control section of some reservoirs is improper, the upstream and downstream flow is disordered, and the flow monitoring is difficult (Zhou et al. 2022). Therefore, it is necessary *Corresponding Authors: [email protected] and [email protected]
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DOI: 10.1201/9781003425823-74
to carry out relevant research on the setting of the reservoir flood control section to give full play to the practical role of the flood control section (Hu et al. 2001).
3 CASE ANALYSIS 3.1
Research background
A reservoir in Zhejiang Province is a large (2) water conservancy project that focuses on water supply and flood control and takes into account the comprehensive utilization of power generation and irrigation. The total reservoir capacity is 168.4 million m3, and the flood control capacity is 12.4 million m3. The reservoir is designed for a 100-year return period and checked for a 2000-year return period. The rainwater collection area above the reservoir dam site is 254 km2, the main river length is 28.4 km, and the average slope of the main river is 11.6‰. The hydroproject structure is mainly composed of a barrage, spillway, power generation and diversion, water supply, and vent tunnel, powerhouse and downstream reverse regulation pool retaining dike, and side weir gate. The crest of the concrete face rockfill dam is 398.0 m long and 10.0 m wide, the crest elevation is 177.40 m, the crest elevation of the wave wall is 178.60 m, and the maximum dam height is 124.4 m. The 100-year design flood level is 174.30 m, and the corresponding storage capacity is 160.4 million m3; The check flood level in 2000 is 176.20 m, and the corresponding storage capacity is 168.4 million m3. At present, the control section of the reservoir is located at the bend of the river, and the flow is disordered. The upstream and downstream do not have the conditions for flow monitoring, so the flow control operation of the control section cannot be implemented. Therefore, in order to improve the operability of compensation regulation in reservoir flood control operation, it is proposed to adjust the reservoir flood control section and move it down to the section with flow measurement conditions. By measuring the river flow of the section, we can understand the actual flow of the downstream river, provide more reliable flow data for the reservoir operation, and optimize the reservoir operation. In combination with relevant planning requirements, the regional flood control standard within the scope of reservoir flood control protection is a 20-year return period. After years of urban development and construction and river regulation in the basin where the reservoir is located, the original flood control section and flood control dispatching mode of the reservoir can no longer meet the regional flood control requirements. Therefore, it is necessary to study the possibility and feasibility of adjusting the flood control setting plane of the existing reservoir without changing the flood control capacity and control dispatching, so as to improve the regional flood control capacity to a 20-year return period. 3.2
Research purposes and ideas
1. Research objective According to the actual situation of the downstream river of the reservoir, we analyze the rationality of the current setting of the downstream flood control section of the reservoir, study the necessity and feasibility of adjusting the control section, and recommend the appropriate dispatching scheme according to the model calculation results. 2. Research ideas According to the flood control capacity of the project and the importance of the protected objects, the reservoir adopts the flood control dispatching mode of graded control flood discharge in the flood season and determines the safety standards, safety discharge and corresponding dispatching authority at all levels. At the same time, it is also necessary to clearly define the criteria for determining whether the reservoir will change to dam
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protection and increase discharge after encountering floods that exceed the downstream flood control standards. The flood control section of the reservoir is proposed to be moved down this time, and its feasibility shall be analyzed and judged according to the following principles: 1. The flood control standard of the reservoir itself will not be affected; 2. We do not change the scope of submerged immigrants; 3. We define the controlled discharge flow of reservoir compensation regulation; 4. With comprehensive consideration of reservoir risks and benefits, the research ideas are as follows: 1 We collect and sort out a long series of rainfall data, and calculate and review the design storm or corresponding storm in the reservoir area, the upstream of the control section, and the dam site to the control section; 2 We calculate and recheck the design flood or corresponding flood of each subarea according to the rechecked design storm and the basin characteristic value of each subarea; 3 According to the upstream and downstream flood control requirements and the design flood at the Qiantong section, the allowable controlled discharge of the reservoir at different levels is determined, and the flood regulation calculation is carried out accordingly to determine the final recommended dispatching scheme; 4 We determine the implementation plan for moving down the control section. 3.3
Research analysis
3.3.1 Basin characteristics According to the 1:10000 aerial topographic map, the watershed characteristic values of each subarea are as follows: 1. Characteristic value of the reservoir basin: the catchment area is 254 km2, the mainstream length is 28.4 km, and the average gradient of the river is 11.6‰. 2. The characteristic value of the basin above the control section: the catchment area is 315 km2, the mainstream length is 42.9 km, and the average gradient of the river is 9.8‰. 3. Dam site to control section: rainwater collection area is 61 km2, mainstream length is 14.5 km, and the average gradient of the river channel is 3.1‰. 3.3.2 Design storm Based on the comprehensive analysis of the starting and ending time of the data of each station, the rainstorm series from 1961 to 2021 is selected for a total of 61 years, and the P-III theoretical frequency curve is used to visually estimate the fitting line. Table 1 shows the design storm results at various frequencies (Yu et al. 2010). Table 1.
Design storm results of reservoir basin.
Period Annual maximum
Mean value Time (mm) Cv
H1d H24h H3d Plum flood H1d H24h season H3d Typhoon flood H1d H24h season H3d
153.6 176.6 231.1 66.8 76.9 112.4 148.3 170.6 220.8
0.58 / 0.56 0.48 / 0.44 0.58 / 0.56
Each frequency (P%) Design storm (mm)
Cs/ Cv
0.02 0.05 0.1 0.2 0.5 1
3.0 / 3.0 4.0 / 4.0 3.0 / 3.0
778 705 649 593 519 462 406 330 895 810 746 682 597 532 466 379 1123 1019 940 861 756 675 594 486 303 275 253 231 203 181 159 131 348 316 291 266 233 208 183 150 461 420 389 358 316 285 253 210 751 681 627 573 501 447 392 318 864 783 721 659 576 514 450 366 1073 974 898 823 722 645 568 464
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2
5
10
20
271 312 403 109 125 178 262 301 385
212 244 317 87 100 145 205 235 303
3.3.3 Design flood The simple loss deduction method is adopted for the calculation of runoff production. Assuming that the maximum soil moisture content is 100 mm and the initial soil moisture content is 75 mm, the initial loss is 25 mm. The maximum after-loss of 24-hour rainfall is 1 mm/h, and the after-loss of the rest days is 0.5 mm/h. The instantaneous unit hydrograph method of Zhejiang Province is used to calculate the concentration (Deng & Li 2020). The formula of the instantaneous unit hydrograph method of Zhejiang Province (that is, using the Nash instantaneous unit hydrograph method) is generally as follows: UðtÞ ¼ ðt=KÞn1 et=k =½K GðnÞ UðtÞ—Instantaneous unit hydrograph longitudinal height at time t t; K—Parameters reflecting the catchment time; n—Adjustment times; GðnÞ—N-order incomplete gamma function; t—time. Tables 2 and 3 show design flood results. Table 2.
Design flood results of reservoir. Each frequency (%)
Period
Item
Unit
Peak discharge Flood peak modulus 3days flood volume 24hours flood volume Plum flood sea- Peak discharge Flood peak modulus son 3days flood volume 24hours flood volume Typhoon flood Peak discharge season Flood peak modulus 3days flood volume 24hours flood volume
3
Annual maximum
Table 3.
m /s m3/(s km2) Billion m3 Billion m3 m3/s m3/(s km2) Billion m3 Billion m3 m3/s m3/(s km2) Billion m3 Billion m3
0.05 0.2
1
2
5
10
20
5240 20.6 2.36 1.93 2052 8.1 0.93 0.72 5166 20.3 2.32 1.89
3551 14.0 1.52 1.25 1189 4.7 0.60 0.44 3475 13.7 1.50 1.22
3122 12.3 1.32 1.09 996 3.9 0.52 0.38 3051 12.0 1.31 1.06
2525 9.9 1.06 0.87 746 2.9 0.41 0.30 2449 9.6 1.05 0.85
2018 7.9 0.86 0.71 667 2.6 0.33 0.24 1945 7.7 0.85 0.68
1471 5.8 0.66 0.53 459 1.8 0.25 0.18 1406 5.5 0.65 0.51
4482 17.6 1.97 1.62 1652 6.5 0.78 0.59 4408 17.4 1.94 1.58
Table of corresponding flood results from the dam site to control section. Each frequency (%)
Period Annual maximum
Item
Peak discharge Flood peak modulus 3days flood volume 24hours flood volume Plum flood season Peak discharge Flood peak modulus 3days flood volume 24hours flood volume Typhoon flood sea- Peak discharge son Flood peak modulus 3days flood volume 24hours flood volume
Unit 3
m /s m3/(s km2) Billion m3 Billion m3 m3/s m3/(s km2) Billion m3 Billion m3 m3/s m3/(s km2) Billion m3 Billion m3
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0.05 0.2
1
2
5
10
20
1014 16.6 0.43 0.35 413 6.8 0.18 0.13 996 16.3 0.41 0.34
659 10.8 0.27 0.22 249 4.1 0.11 0.08 646 10.6 0.26 0.21
576 9.4 0.24 0.19 206 3.4 0.09 0.07 565 9.3 0.22 0.18
451 7.4 0.18 0.14 157 2.6 0.07 0.05 445 7.3 0.17 0.14
349 5.7 0.14 0.11 110 1.8 0.06 0.04 347 5.7 0.14 0.11
255 4.2 0.11 0.08 70 1.1 0.05 0.03 248 4.1 0.10 0.08
859 14.1 0.36 0.29 332 5.4 0.15 0.11 846 13.9 0.34 0.28
3.4
Flood discharge capacity of downstream river
The model (Su 2022) is established according to the measured topographic data (Figure 1) (Ji et al. 2015), and the relationship between the river flow capacity and water level at the control section is shown in the following table (Lin et al. 2019). It can be seen from Table 4 that when the water level at this section is 28.52 m, the discharge capacity reaches 3443 m3/s, which is far greater than the discharge capacity of the 20-year return period determined at the river training stage, 1750 m3/s. Table 4.
Relationship between water level and flow capacity at the control section.
Water level (m) 3
Overcurrent capacity (m /s)
Figure 1.
3.5
18.47
26.80
28.52
0
1623
3443
Measured drawing of the control section.
Recommended scheme for downward movement of the control section
In this study, it is proposed to move the current control section down to the section with measured flow conditions, which is about 2 km downstream of the current control section and about 14.7 km away from the reservoir dam site. The positions of the two sections are shown in the figure below. According to the above analysis, the 20-year return period control flow at the proposed control section is 1750 m3/s, and the control water level is 27.00 m. 3.6
Flood level review
The flood routing is carried out according to the scheme of lowering the flood control section, and the calculation results are compared with the original design results and the flood control characteristic water level (Hu & Chen 2013). Through comparison, it can be seen that after the reservoir control section is moved down to the section with measurement conditions, the reservoir can be operated according to the current dispatching principle, and
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the current characteristic flood level can be maintained unchanged (Yang et al. 2005). There is no adverse impact on the dam safety and downstream flood safety of the reservoir, and the reservoir control section is feasible to move down.
4 CONCLUSIONS 1. If real-time flow monitoring cannot be carried out at the current control section of the reservoir, it is necessary to move the control section down to the section with flow measurement conditions. 2. In the selection of the control section, the flow capacity of the downstream river should be fully analyzed to determine the reasonable safe flow and control water level. At the same time, the hydrological data series should be extended, the design flood should be reviewed, and the flood regulation calculation should be carried out to compare whether the determined control section and various indicators are reasonable. 3. The reservoir shall strictly comply with the operation and dispatching principles approved by the competent authorities at higher levels, strictly obey the orders of the flood control headquarters, and ensure the safety of the reservoir during the flood season; At the same time, do a good job in the reservoir control water level transition period, and effectively use water resources.
REFERENCES Deng, Y., Li, J.H. (2020) Calculation and Analysis of Design Flood at Downstream Flood Control Section of Reservoir. Journal of Yellow River Conservancy Technical, 32(3): 11–13. https://doi.org/10.13681/j.cnki. cn41-1282/tv.2020.03.003. Gou, R.P. (2015) Analysis on Determination of Characteristic Water Level of Flood Control Section. China High Tech Enterprises, 346(31): 125–126. https://doi.org/10.13535/j.cnki.11-4406/n.2015.31.062. Hu, L.L., Chen, H.B. (2013) Analysis of the Influence of the Upstream Dike Construction of Puyang River on Flood Return. Zhejiang Hydrotechnics, 41(6): 23–26. https://doi.org/10.13641/j.cnki.33-1162/ tv.2013.06.014. Hu, R.W., Sun, X., Wang, H.Q. (2001) Analysis of Influence of Reservoir Operation on Downstream Flood Control. Zhejiang Hydrotechnics, 4: 55–56. https://doi.org/10.13641/j.cnki.33-1162/tv.2001.04.028. Ji, C.H., Zhang, X.N., Xie, R. (2015) Influence of Water Depth and Beach Roughness on the Flow Capacity of Compound Cross-section Channel. Yellow River, 37(1): 42–45. https://doi.org/10.3969/j.issn.10001379.2015.01.011. Lin, F.F., Xiao, J.Q., Zhou, M.R., Deng, S.S. (2019) Response of Channel Shape and Flow Capacity Adjustment of the Lower Jingjiang River to Upstream and Downstream Boundary conditions. Journal of Hydraulic Engineering, 50(05): 641–649. https://doi.org/10.13243/j.cnki.slxb.20180800. Su, Y.M. (2022) Comparison and Analysis of HEC-RAS One and Two Dimensional Models in Water Surface Profile Calculation. Water Sciences and Engineering Technology, 6: 12–15. https://doi.org/10.19733/j. cnki.1672-9900.2022.06.04. Yang, K.J., Cao, S.Y., Liu, X.N., Zhang, Z.X. (2005) Comparison and Analysis of Discharge Calculation Methods for Compound Channels. Journal of Hydraulic Engineering, 5: 563–568+574. https://doi.org/ 10.13243/j.cnki.slxb.2005.05.009. Yu, S.H., Lin, C.D., Hu, L.L. (2010) Recheck of Flood Control Capacity of Hecun Reservoir dam. Zhejiang Hydrotechnics, 5: 43–44. https://doi.org/10.13641/j.cnki.33-1162/tv.2010.05.012. Zhou, R., Zhang, Y.P., Liu, L. (2022) Discussion on Drawing Flood Control Dispatching Diagram of Large Reservoirs When the Downstream River Flow Capacity Does Not Meet Flood Control Standards. Shaanxi Water Resources, 7: 69–71+76. https://doi.org/10.16747/j.cnki.cn61-1109/tv.2022.07.058.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
The effect of traffic blockage on fire smoke spreading and safety escape in highway tunnel Zubin Ai, Zhensheng Cao, Shengjun Hou & Chenchen Jiang Power China Road Bridge Group Co., Ltd., Beijing, China
Jianbin Zang* Tongji University, Shanghai, China
ABSTRACT: When a traffic accident causes a fire in a highway tunnel, traffic congestion may hinder the spread of smoke and the ability to escape safely. In this paper, FDS and Pathfinder software are used to examine the impact of a vehicle blocking on critical wind speed, tunnel vault temperature, and safe tunnel egress. Using FDS software, the effect of different blocking vehicle sizes on the critical wind speed was analyzed when the surface area of the fire source was held constantly. It was discovered that the lower the critical wind speed is, the greater the tunnel’s blocking area is. For a given fire scale, the higher the tunnel’s blocking ratio is, the greater the critical wind speed reduction ratio is. The higher longitudinal ventilation velocity and the greater shielding effect of a higher vehicle on the fire source can reduce the vault’s temperature, reduce the tunnel’s heat accumulation, and safeguard the tunnel’s structural integrity. Pathfinder was utilized to investigate the effects of stair pitch, resting platform, and stair spacing on evacuation time. The results indicate that the shortest evacuation time occurs when the stair slope is set to 30 . When compared with the original design of 80 m stair spacing, the 60 m spacing reduces evacuation times by more than 10%. It is more conducive to the safety of personnel. This paper enriched the research on tunnel smoke dispersion, which is useful in designing tunnel fire escape systems.
1 INTRODUCTION In recent years, the number and length of highway tunnels in China have increased rapidly because of higher driving speed and population density, which results in a rise in safety issues, such as tunnel fire smoke exhaust, evacuation, and firefighting. The structure of a highway tunnel is unique. It is a narrow underground space with a lengthy evacuation passage, poor ventilation, and low visibility. If the fire cannot be contained effectively, there will be unforeseeable losses of life and property. According to existing research, tunnel fires will result in the following problems (Wu et al. 2013)—large smoke production, high temperature, low visibility, and rapid spread, too many vehicles, channel blocking. All these factors lead to secondary accidents, difficult evacuation of trapped individuals, and difficult firefighting, as well as communication issues and poor leadership. Some experts and scholars have analyzed the mass deaths and mass injuries caused by tunnel fire accidents, and their findings indicate that the fire smoke is the primary cause of casualties, and its harmfulness is primarily reflected by the following three aspects (Lai et al. 2017; Qu 2015; Zhao et al. 2016)—the fire smoke containing carbon monoxide, carbon dioxide, and other toxic gases, *Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-75
severely damaging the respiratory systems of those who are trapped. (2) As a result of the strong light reduction caused by the smoke, the visibility in the tunnel will be drastically reduced, making it more difficult to evacuate the trapped individuals. (3) As fire smoke emits intense and high-temperature radiation, it will damage tunnel structures and endanger the lives of those trapped inside. With the advancement of computer technology in the 21st century, computer simulation is widely employed in the study of tunnel fire smoke flow. Throughout history, simulation software has continuously advanced in terms of professionalism and maturity, starting with the first subway environment simulation (SES) computer program developed in the United States, to the JASMINE mathematical model established by Cox et al. in the United Kingdom (Miles & Gox 1996; Yuana & Gox 1996), to a large number of large-scale commercial simulation software for computational fluid dynamics and fire specific software (FDS). In terms of the mathematical model, Haghighat et al. (2018) proposed a method for selecting boundary conditions under fire simulation conditions by conducting multi-group studies on multi-scale highway tunnels. Thomas (1958, 1968) in the United Kingdom was the first to propose the concept of critical wind speed and derive its semi-empirical formula. Wu & Bakar (2000), Lee (2005), Oka and Atkinson (1995), Chow et al. (2015;,2010, 2012), and Li et al. (2018) enriched the research on the critical wind speed calculation formula and the relationship between the Froude number and the critical wind speed at different angles. Xu et al. (2013) conducted a series of studies on the critical wind speed of horizontal tunnel fires and obtained an undetermined critical wind speed coefficient of 0.92. Wang et al. (2015) used the numerical simulation method to study the stratification characteristics of fire smoke in tunnels under different wind speeds. Their results showed that when the wind speed was lower than the critical wind speed, flue gas backflow in tunnels was evident, and the thickness and length of the flue gas backflow layer increased significantly with decreasing ventilation speeds. The research on tunnel fires has also progressed from a focus on fire characteristics to a comprehensive system comprising accident probability, fire consequences, human response, structural response, ventilation system response, and numerous interrelated topics (Qiu 2005). Being similar to fire simulation, personnel evacuation research focuses on the development and application of simulation software. Approximately 30 types of evacuation simulation software are currently available, including EVACNET, Building EXODUS, STEPS, FIRECAM, SIMULEX, and PATHFINDER (Gwynnes et al. 2011). Ronchi et al. (2012) used the Lantueno tunnel in Spain as an illustration. Three types of evacuation model software (FDS+Evac, STEPS, and Pathfinder) and the Fire Prevention Society’s analysis and calculation engineer (SFPE) manual were studied and compared, and there was no significant difference between the models of evacuation time obtained. Wang (2014), Yang (2012), and Yang et al. (2007) used numerical simulations in conjunction with the effects of temperature and carbon monoxide on personnel to determine the escape location selected by personnel as well as a reasonable escape passage spacing. Zhang et al. (2017) used FDS software to study the influence of different vehicles and fire scales on the critical wind speed in the highway tunnel. The results indicated that the lower the critical wind speed was, the closer the traffic jam was to the fire upstream, while there was no obstruction when it was far from the fire. Liang et al. (2017) used FDS to study the influence of different fire scales and blocking types of different vehicles (cars, vans, trucks, and trucks) on the critical wind speed of fire, and discovered that on the same fire scale, the greater the blocking ratio in the tunnel is, the greater the reduction ratio of critical wind speed is, and proposed the reference value of critical wind speed in the case of blocking. The temperature distribution is one of the primary characteristic parameters of tunnel fire smoke, which not only affects the mechanical properties of the tunnel lining structure but also jeopardizes the safety of individuals trapped inside the tunnel. Because the degree of tunnel lining damage during a fire is largely dependent on its temperature, it is of utmost importance to investigate the maximum temperature near the tunnel vault during a fire. Through scale storage model testing, Yi et al. (2011) and Yuan et al. (2010) confirmed the 589
accuracy of Kurioka’s model and modified it. Using numerical simulation, Guo et al. (2020) examined the influence of the heat release rate of the fire source, the tunnel width, and the height of the tunnel on the temperature distribution of the tunnel fire vault. They developed a model that can better predict the temperature distribution of small-scale tunnel fires or weak plumes of smoke. A high-drop tunnel is defined by the Code for Highway Tunnel Design JTGD 70-2004 as a tunnel with a slope greater than 3%. When a mountain tunnel encounters the obstacle of a steep drop, a spiral structure is typically used to overcome it. Spiral tunnel construction cases at home and abroad are on the rise, but the number of spiral tunnels in the world is still relatively small. Consequently, spiral tunnel structures are difficult to design and build, as well as the presence of large gaps, which increases the likelihood of fire accidents. As a result, this type of tunnel’s firesmoke propagation law and safety escape reality should be studied in depth. This paper investigates the impact of vehicle obstruction on smoke diffusion and evacuation time in high-drop highway tunnels, it includes, (1) A simulation analysis of the effect of different heights, widths, and numbers of blocking vehicles located upstream of the fire on the critical wind speed and vault temperature of the tunnel (2) A simulation study of the effect of different evacuation stair slopes, with or without a resting platform, and evacuation stair spacing on the evacuation time of a single- hole, double-deck urban tunnel under conditions of vehicle congestion after a fire.
2 METHODOLOGY 2.1
Simulation software
In this study, FDS and Pathfinder simulation software were used. The Fire Dynamic Simulator (FDS) is a computational fluid dynamics (CFD) simulation program developed by the National Institute of Standards and Technology (NIST). Typically, a mixed fraction combustion model is used. The combustion chemical reaction is approximately completed when the fire source mixes with air, and the components involved in combustion in the fire source region can be expressed as a fraction of the mixing fraction. In this paper about the tunnel fire, FDS numerical simulation can give more accurate results. Pathfinder, a program frequently used to study the evacuation of tunnels and underground spaces, is an evacuation and movement simulation-based emulator. Pathfinder offers two modes of mobile simulation—steering mode and SFPE mode. In the steering mode, individuals should maintain a safe distance from one another. Contrary to the SFPE model, people will be crowded together, but the door will restrict the flow of people. In this simulation, the steering mode was utilized to represent the speed and evacuation time of the individuals to be evacuated through the complex passage more accurately. 2.2
Model settings for fire smoke spread simulations
2.2.1 Tunnel and fire source model The geometric model utilized in this study is depicted in Figure 1, which is 50 m long, 10 m wide, and 5 m tall. The fire source is located in the center of the model at a height of 1 m, and the surface area is 2 m1 m, for a total of 2 m2.
Figure 1.
Schematic diagram of the tunnel and fire source model.
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Temperature and smoke monitors are installed every one meter at a height of 4.8 m along the centerline of the tunnel. Figure 2 illustrates how temperature and smoke monitors are installed every 0.5 m above the fire source in order to detect vault temperatures and smoke concentrations.
Figure 2.
Location diagram for temperature and smoke monitoring stations.
The tunnel wall’s thermal thickness boundary and “CONCRETE” attribute were set. Both tunnel entrances were designed as natural openings and velocity entrances, respectively. The ambient temperature was set to 293 K, the fire source’s HRR was set to 20 MW, and the duration of the simulation was 900 s. The effect of the surface area of the fire source on smoke diffusion in the event of a vehicle obstruction was examined and the power of the fire source to 20 MW in numerical simulations was set. The critical wind speed refers to the minimum mechanical ventilation wind speed after a fire, which prevents the backflow of high-temperature flue gases. It is an important factor in tunnel fire flue gas control. Now, the occurrence of backflow is contingent on a number of variables, including fire scale, tunnel slope, ventilation, and wind speed. To study the change rule of the critical wind speed, the size of the fire source’s surface will be altered. When the size of the blocked vehicle is 4 m 2 m 2.5 m and the surface size of the fire source is 1 m 1 m, 2 m 1 m, and 2 m 2 m, the heat release rate per unit area varies, but the critical wind speed is 2.7 m/s, indicating that, in the case of the blocked vehicle, the critical wind speed is independent of the surface size of the fire source. 2.2.2 Vehicle-blockage conditions To examine the effect of upstream blocked vehicles on fire smoke diffusion and critical wind speed, 8 groups of cases with varying numbers and sizes of blocked vehicles are compared, and 3 to 4 longitudinal ventilation velocities are determined for each group to determine critical wind speed. There are a total of 27 numerical simulation cases performed. The surface area of the fire source is 2 m1 m, and the rate of heat emission is 20 MW. Table 1 details the particular conditions. Table 1.
Cases Group Group Group Group Group Group Group Group
1 2 3 4 5 6 7 8
Conditions of blocked vehicles’ dimensions and number. Fire Heat Release Rate
Number of Blocked Vehicle Vehicles Dimension
20 20 20 20 20 20 20 20
1 1 1 1 1 2 2 2
MW MW MW MW MW MW MW MW
4 4 4 4 4 4 4 4
m2 m3 m4 m4 m4 m2 m2 m2
m1.5 m1.5 m1.5 m2.0 m2.5 m1.5 m2.0 m2.5
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m m m m m m m m
Blocking Ratio
Ventilation Velocity
6% 9% 12% 16% 20% 6% 8% 10%
3.0 2.9 2.8 2.6 2.6 2.6 2.4 2.4
m/s, m/s, m/s, m/s, m/s, m/s, m/s, m/s,
3.1 3.0 3.0 2.7 2.7 2.8 2.6 2.6
m/s, m/s, m/s, m/s, m/s, m/s, m/s, m/s,
3.15 m/s, 3.2 m/s 3.1 m/s 3.1 m/s 2.8 m/s, 2.9 m/s 2.8 m/s, 3.0 m/s 3.0 m/s 2.8 m/s 2.8 m/s
To study the effect of blocking vehicles on smoke diffusion and critical wind speed under constant firepower conditions, the size, number, and position of blocking vehicles were varied. Figure 3 illustrates a tunnel model.
Figure 3.
2.3
Traffic congestion.
Model settings for personnel escape simulations
2.3.1 Hypothesis Currently, the distance between longitudinal evacuation stairs at home and abroad is set at 80 m-150 m (Ronchi et al. 2012). Because of the traffic volume in 2039, the distance between stairs is set to 80 m to ensure safety, and evacuation stairs are evenly distributed throughout the tunnel. Based on the characteristics of the tunnel, the following hypotheses can be formulated for the fire scene—there is only one fire in the tunnel; there is no ventilation; the fire is located at the exit of the lower tunnel. Due to the blocking of the vehicles, which prevents the passage of vehicles upstream, personnel will need to exit the building through the next exit and evacuate to the upper level via the evacuation stairs. 2.3.2 Settings of traffic characteristics Table 2 illustrates the comprehensive proportion of tunnel models in this example. The length of a car is 3 m, that of a medium bus is 6 m, and that of a bus is 12 m. A distance of 1.5 m separates the vehicles in the tunnel. There are 47 vehicles in the tunnel fire influence area of 80 m, and they are evenly distributed throughout the tunnel, requiring a total of 258 people to be evacuated. Table 2.
The proportion of vehicles in tunnels.
Type
proportion
Size (m)
Number (person)
utilized
Car Medium bus Bus
85% 10% 5%
3.0 1.6 6.0 2.0 12 2.5
4 20 80
50% 60% 80%
2.3.3 Setting of personnel characteristics The external environment, race, evacuation route, gender, and other factors will have a direct impact on the walking speed of individuals, and the walking speed will have a direct impact on the walking time. When a tunnel fire occurs, PIARC (PIARC Committee on Road Tunnels 1999) suggests that the walking speed range should be between 0.5 m and 1.5 m per second. According to a large number of studies and data collections, the average walking speed of Westerners is between 1.2 m and 1.5 m per second. Due to racial differences, some experts and scholars have estimated that the evacuation walking speed ranges
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from 1.0 m to 1.4 m per second. The value should be reduced by 20% if people are affected by smoke from a fire. The number of people to be evacuated in the model is 258. It is assumed that the population is composed of 40 percent adult males, 40 percent adult females, 10 percent elderly, and 10 percent adolescents. Markers for shoulder width, height, walking speed, and color are set accordingly. Table 3 lists parameter settings in detail. Table 3.
Composition of evacuees.
Personnel Type
The Proportion
Adult male 40% Adult women 40% Male elderly 5% Elderly women 5% Male adolescent 5% Female adoles5% cent
The Number Shoulder of/a Width/m
Height/ m
Speed/ (m/s)
Color
105 105 12 12 12 12
1.75 1.65 1.68 1.56 1.62 1.57
1.19 1.1 0.9 0.8 0.85 0.75
Blue Red Black Green Grey Yellow
0.48 0.4 0.44 0.38 0.43 0.41
2.3.4 Vehicle and personnel distribution In the worst-case scenario, the accident occurred at the exit, which was unavailable for evacuation. Figure 4 depicts the schematic diagram of vehicle and personnel distribution. Assuming that the closest vehicle to the fire source is a bus, the fireplace is in front of the bus, and personnel in the bus can detect the fire source directly and quickly. According to the Technical Specifications for Fire Control in Highway Tunnels, the response time of the fire detector in the tunnel should not exceed 60 s, and the response time of personnel at various positions in the tunnel should be selected based on the distance from the fire source (Jiang et al. 2015)—the response time of a person within 0 m to 30 m is 5 s, within 30 m to 60 m, 30 s, and beyond 60 m, 60 s. The time required for fire detection, confirmation, and notification is 60 s.
Figure 4.
Distribution of vehicles and personnel.
2.3.5 Case settings and boundary conditions for evacuation simulation The net height of the tunnel is 5.8 m, the width of the stairs is 0.8 m, the net height of the stairs with a rest platform is 2.9 m, the rest platform is 1.2 m by one meter, and the width of the fire door is 1.1 m. In building architecture, the stair slope should be greater than 20 but less than 50 . Because slopes are greater than 50 , a ladder should be constructed. In this paper, the inclination of evacuation stairs is specified as 60 , 45 , or 30 , with or without a rest platform. The Pathfinder evacuation software is used to model and calculate the required evacuation time for personnel. The evacuation simulation conditions are illustrated in Table 4.
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Table 4.
Case settings and boundary conditions for evacuation simulation.
Working Conditions
Stair Slope
Stair Width (m)
Fire Door Width (m)
Rest Platform (m)
Condition 1 Condition 2 Condition 3
60 45 30
0.8 0.8 0.8
1.1 1.1 1.1
1.2 1 1.2 1 1.2 1
3 RESULTS AND DISCUSSION 3.1
Influence of blocked vehicles on smoke diffusion
3.1.1 Influence of blocking area and blocking ratio on critical wind speed After simulation, each working condition’s critical wind speed is displayed in Table 5. In conditions 1, 2, and 3, as the width of the blocked vehicle increases, the critical wind speed decreases, assuming that the length and height of the blocked vehicle remain constant. In conditions 3, 4, 5, 6, 7, and 8, it is known that the higher the vehicle height is, the lower the critical wind speed is under identical vehicle length and width. In conditions 1 and 6, it can be concluded that when the size of blocking vehicles remains constant, the critical wind speed decreases as the number of vehicles increases. According to conditions 3 and 6, 4 and 7, 5 and 8, when the blocking area remains the same, the critical wind speed decreases as the number of vehicles increases. According to the above analysis, the critical wind speed decreases as the blocking area increases. From the perspective of the blocking ratio, it can be concluded that when the number of blocking vehicles is constant, the critical wind speed decreases as the blocking ratio increases. According to conditions 3 and 6, 4 and 7, 5 and 8, it is possible to conclude that when the blocking ratio is constant, the critical wind speed is lower when the number of blocking vehicles is higher. The results confirmed the variation rule of critical wind speed of fire obtained by Liang et al. (2017) via FDS numerical simulation under the blocking conditions of four vehicle types (cars, vans, trucks, and trucks) with different fire sizes: for a given fire size, the higher the blocking ratio in the tunnel is, the greater the reduction ratio of critical wind speed is. Table 5.
Cases Group Group Group Group Group Group Group Group
1 2 3 4 5 6 7 8
Effects of blocked vehicles dimensions and number on critical wind speed. Fire Heat Working Condi- Release tion rate
Number of Blocked Vehicles
Condition Condition Condition Condition Condition Condition Condition Condition
1 1 1 1 1 2 2 2
1 2 3 4 5 6 7 8
20 20 20 20 20 20 20 20
MW MW MW MW MW MW MW MW
Vehicle Dimension
Blocking Ratio
Critical Wind Speed
4m2m 4m3m 4m4m 4m4m 4m4m 4m2m 4m2m 4m2m
6% 9% 12% 16% 20% 12% 16% 20%
3.15 m/s 3.1 m/s 3.0 m/s 2.8 m/s 2.7 m/s 2.8 m/s 2.6 m/s 2.4 m/s
1.5 m 1.5 m 1.5 m 2.0 m 2.5 m 1.5 m 2.0 m 2.5 m
3.1.2 Influence of height of blocked vehicles on vault temperature Conditions 3, 4, and 5 in Table 5 were chosen for comparison, and the same longitudinal ventilation speed of 2.8 m/s was used to examine the effect of blocked vehicles at varying heights on the temperature of the vault close to the fire source. As depicted in Figure 5, the change curves of vault temperature with different vehicle heights reveal that at the same 594
ventilation speed of 2.8 m/s, the vault temperature in the upwind direction near the fire source is lower when the vehicle height is 2 m and 2.5 m, but it is higher when the vehicle height is 1.5 m. However, the critical wind speed at the height of 2 m is exactly equal to the wind speed of longitudinal ventilation, and the critical wind speed at the height of 2.5 m is less than the wind speed of longitudinal ventilation. Consequently, under these two conditions, the hightemperature flue gas will float downstream under the influence of longitudinal ventilation, causing a rise of the temperature of the vault in downstream. When the height of the vehicle is 1.5 m, the critical wind speed is 3.0 m/s, which is greater than the longitudinal ventilation wind speed. Consequently, some flue gas will backflow. In addition, the low height of the vehicle renders it incapable of effectively shielding the heat radiated from the fire source. Thus, the temperature of the vault will increase near the fire source. When the height of the vehicle is 1.5 m, the tunnel vault temperature is generally high, with a maximum temperature of 520 C and instability. When the vehicle reaches a height of 2 m, the temperature of the tunnel vault directly above the fire source decreases, and the upstream temperature is effectively regulated. When the vehicle height is 2.5 m, the vault temperature is too high at 900 s. At other times, the downstream temperature is lower when vehicles are between 1.5 m and 2 m in height. And the upstream temperature is managed effectively. It demonstrates that the vertical ventilation velocity is greater than or equal to the critical wind speed and the shielding effect of higher vehicles on fire sources can lower the vault’s temperature, reduce heat accumulation in the tunnel, and protect the tunnel’s structural integrity.
Figure 5.
Variation curves of vault temperature for various vehicle heights.
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3.2
Analysis of personnel escape influencing factors in the event of traffic blockage
According to the standard for evaluating the safety of an evacuation, the basic safety criterion for evacuation is (Hu et al.2014; Yang 2012): evacuation time (REST) time of fire hazard (AEST). Figure 6 illustrates the various phases of tunnel fire development and evacuation behavior. Evacuation time consists of fire detection and alarm time (Ta), personnel response time (Tr), and evacuation movement time (Te). It is measured from the moment of fire to the moment of evacuation to the safety area (Tm). The evacuation movement time (Tm) consists of the evacuation time (TA) on the bus, the evacuation time (Tp) in the tunnel, and the time (TM) spent entering the safety zone via the escape door. Fire danger time (AEST): the time between the start of a fire and when the fire poses a threat to personnel safety. Several tunnel fire cases and research data (Xiao 2013) indicate that a fire can endanger human life within 10 minutes. This paper specifies 10 minutes as the fire danger time, which is also the optimal time for fire rescue. If the evacuation time exceeds the time at which the fire reaches a dangerous state, it will be difficult to successfully rescue the injured.
Figure 6.
Diagram of different stages of tunnel fire development and evacuation behavior
[10]
.
3.2.1 Influence of stair slope on personnel escape Figure 7 depicts the evacuation time for simulated stair slopes of 30 , 45 , and 60 when the stair spacing is 80 m.
Figure 7.
Evacuation time at various slopes with 80 m stair spacing.
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It can be seen that as the stair slope increases, so does the required evacuation time. The evacuation time required by a stair slope of 60 is significantly longer than the fire danger time of 600 s. However, the evacuation time for a stair slope of 45 is close to the fire danger time. Figure 8 depicts the evacuation situation of various stair slopes at various times. Within 120 s, there is almost no difference in the evacuation of stairs with different slopes, but as time passes, the steeper the slope is, the fewer people could be evacuated per unit of time and the longer it takes to evacuate. When the slope is 60 , after the internal evacuation time of the bus and the tunnel evacuation time reached 240 s, everyone gathers at the exit door. Due to the steep slope, few people are carried, and the problem of crowding and queuing is severe.
Figure 8.
Evacuation conditions compared across various slopes and periods.
The stairs, with a 60 slope and no rest platform, require the longest evacuation time. Figure 9 depicts the evacuation situation of people in the tunnel at various times: at 60 s, people were exiting from the two buses in a crowded manner, but there was no congestion on the evacuation stairs. At 120 s, nearly all of the personnel from the two buses disembarked, and some personnel in the tunnel did not reach the evacuation staircase. Four minutes later, a large number of individuals were still in line at the exit of the escape staircase. At 600 s, there were still over 40 individuals who had not reached the safety zone, and their lives were in grave danger. The duration of severe crowding is approximately 9 minutes, which is susceptible to psychological panic and stampedes, and the evacuation and rescue time exceeds the fire danger time. According to the results of the above calculations, the stair slope of 60 is too steep and the evacuation time is too long, which is consistent with the building architecture requirements. It is not acceptable to incline the stairwell beyond 50 .
Figure 9.
Personnel escaping the tunnel at different times on a 60 slope without a rest platform.
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3.2.2 Influence of resting platform on personnel escape As shown in Figure 10, when the slope is 30 , the presence of a resting platform has little effect on personnel escape time. The stairs with a resting platform required less time to evacuate at 45 and 60 than those without. Figure 8 demonstrates that beginning at 300 s, when the stair angle is 30 , the number of people evacuated by stairs with a resting platform is less than that of stairs without one. Starting at 512 s, when the stair angle is 45 , more people are evacuated with a resting platform than without. Beginning at 540 s, when the stair angle is 60 , more people are evacuated with a resting platform than without. When the slope is 60 , the presence of a resting platform increases the number of people carrying the stairs, alleviates the crowded waiting situation at the exit, and facilitates the faster evacuation of trapped people. When the slope is 45 , there is no significant difference between with and without a resting platform, but the total evacuation time is slightly shorter with a resting platform. When the slope is 30 , the presence of a resting platform increases the time spent on the stairs, and the evacuation time with a resting platform is longer than without one. The evacuation times for slopes of 30 and 45 are within a safe range, and the plan with a slope of 30 and no resting platform has the shortest evacuation time. Figure 10 depicts the evacuation situation of personnel in the tunnel at various times. The evacuation situation within 180 s is identical to the situation without a resting platform when the slope is 60 ; the only difference is that 69 people escaped safely at 180 s, whereas only 50 people escaped safely at 180 s when the slope was 60 and there was no resting platform. During this time, voice broadcasts were conducted for guidance and psychological guidance in the event of an emergency to prevent psychological panic among personnel. At 480 s, the tunnel was nearly empty, and there were 21 individuals on the evacuation staircase. In 512 s, everyone was evacuated.
Figure 10.
Personnel escaping from the tunnel at a 30 slope without a rest platform at different times.
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3.2.3 Influence of stair spacing on personnel escape When the stair spacing is 80 m and the stair slope is 45 and 30 , the simulated evacuation time falls within the safe evacuation time range but is too close to the fire danger time. To optimize the design for personnel safety, we need to reduce the stair interval to 60 m, set the stair slope to 30 and 45 , and build the rest platform. The personnel settings will remain unchanged, and 234 individuals will be evacuated. Figure 11 illustrates the simulation’s results. Clearly, there is little difference between the stair slopes with and without a resting platform in terms of evacuation time. When the stair slope is set to 30 , the evacuation time is reduced by more than 60 s compared with 45 . Compared with the original design of 80 m stair spacing, the evacuation time has been drastically reduced by more than 10%. Combining the principle of safety ergonomics with the optimal stair slope of 30 –45 , the optimal plan of escape stair setting is the stair spacing of 60 m, the slope of 30 without a resting platform, and the shortest evacuation time.
Figure 11.
Evacuation time of stairs with or without resting platforms at different slopes.
4 CONCLUSION In this paper, FDS and Pathfinder software are used to examine the impact of a vehicle blocking on critical wind speed, tunnel vault temperature, and safe tunnel egress. Research conclusions are as follows: 1) In the event of a vehicle obstruction, altering the size of the fire source’s surface and setting the power of the fire source to 20 MW, the critical wind speed is independent of the surface size of the fire source. 2) The lower the critical wind speed is, the greater the tunnel’s blocking area is. In the given fire scale, the higher the blocking ratio in the tunnel is, the greater the reduction ratio of critical wind speed is. 3) The higher longitudinal ventilation velocity and the greater shielding effect of a higher vehicle on the fire source can reduce the vault’s temperature, reduce the tunnel’s heat accumulation, and safeguard the tunnel’s structural integrity. 4) The evacuation time is much longer than the fire danger time when the stair slope is 60 , and the shortest evacuation time is 512 s when the stair slope is 30 . The evacuation time without a resting platform is shorter than that with a resting platform, but it is closer to the fire danger time. When compared with the original design of 80 m stair spacing, the 60 m spacing reduces evacuation times by more than 10%. This is more conducive to the safety of personnel. 599
REFERENCES Chow W K, Gao Y, Zhao JH, et al. Smoke Movement in Titled Tunnel Fires with Longitudinal Ventilation [J]. Fire Safety Journal, 2015, 75:14–22. Chow W K. Scale Modeling Studies on Smoke Control Using Smoke Screens in a Tilted Tunnel Fire [J]. Journal of Applied Fire Science, 2012/2013, 22(2):165–178. Chow WK, Wongk Y, Chung W Y. Longitudinal Ventilation for Smoke Control in a Tilted Tunnel by Scale Modeling [J]. Tunnelling and Underground Space Technology, 2010, 25(2):122–128. Guo Qinghua, Yan Zhiguo, Zhang Yao, et al. Study on the Temperature Distribution of Highway Tunnel Fire Vaults Under Natural Ventilation, By [J]. 57(z1): 667–675 in Modern Tunnel Technology, 2020.10.1002/ mtt.2020.10.13807/j.cnki.mtt.2020.S1.089. Gwynnes, Galea ER, Lawrence P J. Simulation Occupant Interaction with Smoke Using Building EXODUS [C]. The 2nd International Symosium on Human Behavior in Fire. Massachusetts Institute of Technology, Boston. The USA. 2011. Haghighat A, Luxbacher K, Lattimer B Y. Development of a Methodology for Interface Boundary Selection in the Multiscale Road Tunnel Fire Simulations[J]. Fire Technology, 2018, (1):1–38. Hu Longhua, Peng Wei, and Yang Ruixin. Fundamentals of Tunnel Fire Dynamics and Prevention Technology [M]. Science Press, Beijing, 2014. Jiang Xuepeng and Xi Xuedong. Longitudinal Evacuation Port Spacing for Highway Shield Tunnel [J]. 31(8), pp. 254–258 (2015). Jinxing Lai. Zhou Hui, Cheng Fei, Wang Ke, Feng Zhihua. Statistical Analysis of Highway Tunnel Fire Accidents and Disaster Prevention and Mitigation Measures[J]. Tunnel Construction. 2017(04). Li Y Z, Ingason H. Discussions on Critical Velocity and Critical Froude Number for Smoke Control in Tunnels with Longitudinal Ventilation[J]. Fire Safety Journal, 2018. Miles S D, Cox G. Prediction of Fire Hazards Associated with Chemical Warehouses [J]. Fire Safety Journal, 1996, 27(4): 265–287. Niu Yi and Zhang Xiaocui. Simulation Study on the Effect of Different Vehicles on Critical Wind Speeds in Highway Tunnel Fires [J]. Fire Protection Technology and Product Information, 2017, 04: 25–26. Oka Y, Atkinson G T. Controlof Smoke Flow in Tunnel Fires[J]. Fire Safety Journal, 1995, 25 (4): 305–322. Pang Liqin, Xie Bingxue, and Liang Huagang, A Highway Tunnel Fires at Critical Wind Speed [J]. 2017, 36 (11): 1515–1517. PIARC Committee on Road Tunnels. Fire and Smoke Control in Road Tunnels [R]. France: PIARC, 1999. Qiu Yuliang, Complex Ventilation Network Analysis Technology for Highway Tunnel Research [D], Chang’an University, 2005. Qu Jianrong. Research on the Evacuation and Rescue of Personnel in Single-hole, Two-way Traffic Highway Tunnel Ffires [D]. Master’s thesis, Chang’an University, 2015. Ronchi E,Colonna P, Capote J, et al. The Evaluation of Different Evacuation Models for Assessing Road Tunnel Safety Analysis[J]. Tunneling and Underground Space Technology. 2012, 30: 74–84. Sung Ryong Lee, Hong Sun Ryou. An Experimental Study of the Effect of the Aspect Ratio on the Critical Velocity in Longitudinal Ventilation Tunnel Fires[J]. Journal of Fire Sciences, 2005, 23(2): 119–138. Thomas, R H. 1958.The Movement of Buoyant Fluid Against a Stream and the Venting of Underground Fires [R]. Borehamwood, England, Fire Research Station. Fire Research Note #351. Thomas, RH. 1968. The Movement of Smoke in Horizontal Passages Against an Airflow[R]. Borehamwood, England, Fire Research Station. Fire Research Note#723. Wang Feng, Dong Guohai, and Wang Mingnian. Study of Critical Wind Speed for Smoke Control in Curved Tunnel Fires [J]. Modern Tunnel Technology, 2015, 52(5): 84–89. Wang Winbin. Research on Fire Escape and Safe Evacuation Facilities in Highway Tunnels [D]. Xi’an: Chang’an University, 2014. Wu Dexing, Xu Zhisheng, and Li Weiping. Highway Tunnel Fire Smoke Control [M]. Beijing: People’s Traffic Publishing House, 2013. Wu Y, Bakar M ZA. Control of Smoke Flow in Tunnel Fires Using Longitudinal Ventilation Systems-a Study of the Critical Velocity [J]. Fire Safety Journal, 2000,35(4): 363–390 Xiao Rong. Risk Analysis of Highway Tunnel Fire Accidents [J], Fire Science and Technology, 32(04), 444– 445, 2013. Xu Zhisheng, Zhao Hongli, Li Hong, Jiang Xuepeng, and Li Dong. A Model of Critical Wind Speed for Horizontal Tunnel Fires [J]. Journal of Central South University (Natural Science Edition), 2013,44(3): 1138–1143.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Simulation of sunshine temperature field of steel box girder under marine environment Bin Zhou*, Linghua Zeng*, Dong Zhang* & Yong Zhang* China Communications Construction Second Highway Engineering Co., Ltd, Xian, China
Yuewei Yang* Chongqing Special Paving Engineering Technology Co., Ltd, Chongqing, China
Zhaoting Liu* & Xiangfan Cao* School of Civil Engineering, Chongqing Jiaotong University, Chongqing, China
ABSTRACT: In this paper, the sunshine temperature field of the steel box girder of the Peljesac Bridge is studied under the Marine environment condition. The temperature field is the premise of calculating the temperature effect and the basis of bridge structure analysis and design calculation. Based on the steel box girder of the Peljesac Bridge, the temperature of the steel box girder at different positions under sunshine conditions was measured. The finite element model of the steel box girder is established, and the thermodynamic parameters are selected to simulate the temperature field of the steel box girder under sunshine conditions. The validity of the temperature field model is verified by comparing the calculated results with the measured temperatures. The temperature field model in this paper can predict the temperature variation and peak value of the key points of the bridge, which provides an important basis for the reliability of the temperature calculation of the structure. This paper’s temperature field simulation method provides an important bridge temperature field simulation reference.
1 INTRODUCTION This paper works for the construction of the Peljesac Bridge. The Peljesac Bridge is located in Croatia, which connects the Croatian mainland with the Peljesac Peninsula in its southernmost Dubrovnik-Neretva County, linking the country’s south to the contiguous north. In the bridge construction process, the influence of environmental temperature will be superimposed with the load effect, which will have an important impact on the bridge. For long-span steel Bridges, most of them adopt the form of a steel box girder. Due to the material’s strong temperature conductivity, the steel plate’s relative thickness is small, and the local structure is complex. They are more sensitive to temperature. The long-span steel bridge is usually a statically indeterminate structure many times, and the temperature stress caused by temperature change will reduce the structure’s safety. To accurately predict and evaluate the temperature and its influence, it is necessary to establish a suitable temperature field model using the measurement and calculation simulation method. Through long-term data collection, the temperature variation rule of Bridges under natural conditions using statistics is obtained, and the temperature variation rule model of different bridge structures is provided (Ding et al. 2012; Erik et al. 2018; Zhu et al. 2021). Fan (2021) studied the temperature field of a steel-concrete bridge structure through an
*Corresponding Authors: [email protected], [email protected], [email protected], [email protected], [email protected], [email protected] and [email protected]
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DOI: 10.1201/9781003425823-76
indoor controllable temperature source and established a more accurate numerical model. Choi et al. (2011) predicted the evolution of temperature and relative humidity of the composite bridge based on the theoretical model and compared it with the measured data. The influence of each factor on temperature and relative humidity was evaluated quantitatively. The temperature response caused by different bridge structures under natural conditions, such as daily temperature through field monitoring data, is analyzed (Lin et al. 2020; Niu et al. 2020; Rodriguez et al. 2014 Zhou et al. 2020). Zhao (2019) analyzed the deflection variation of the beam structure of the Nanjing Dashengguan Yangtze River Bridge under the action of temperature and traffic load. Cai (2012) studied the influence of temperature change on the in-plane stability of arch structures under compression, and the results showed that temperature had a significant influence on arch structures. Xu (2009) analyzed the temperature response of the whole structure of the bridge through the monitoring system on Tsing Ma Bridge and established the statistical relationship between temperature and bridge displacement to predict the temperature effect of Tsing Ma Bridge. To study the distribution law of the temperature field of steel box girder under sunshine conditions and other environmental factors, taking the side span steel box girder section of Peljesac Bridge as an example, the temperature field of steel box girder under sunshine conditions was established by combining the method of measurement and model. 2 MEASUREMENT AND RESULTS OF SOLAR IRRADIATION TEMPERATURE ON STEEL BOX GIRDER The following part describes the generation of the sunshine temperature field model. Initially, the temperature of the steel box girder was measured under sunshine conditions. Then the temperature field model is established, and the measured data is compared with the model. 2.1
Temperature measurement
The steel box girder at the approach bridge of Peljesac Bridge was measured, and the temperature variation of the top and bottom surface of the steel box girder was mainly measured. The environmental conditions are from November 13th to 14th, 2020, with the natural sunshine. The weather with a light breeze is fine on November 13th and 14th. Temperature measuring instruments: The main measuring instruments used to measure the temperature of the steel bridge deck are DT-1310 and NR-81532B, whose function is specially used to measure the solid surface temperature. The measuring range is 50 – 500 C, and a mercury thermometer measures the air temperature of the bridge structure. Measuring point arrangement: The measuring point location is located in the side-span section of the Peljesac Bridge. Ten measuring points are arranged in each section, and two sections are measured. The measuring point of section A is the same as section B’s, and the section distance is 5 m. All measuring points are arranged on the surface of the steel box girder. The measured cross-section is shown in Figure 1.
Figure 1.
Schematic diagram of measuring section A and section B.
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As shown in Figure 2, the measuring points of section B are B1-B10. The measuring points are arranged symmetrically according to the center of the section of the steel box girder. B4 is arranged at the middle point of the steel bridge deck, B1 and B7 are arranged at the edge of the pavement on the upper surface, B2 and A6 are arranged 9.3 m away from the center point, B3 and B5 are arranged at 4.8 m away from the symmetrical section point and the middle position between B2 and B4. The lower surface measuring points B8 and B10 are symmetrical and arranged at the edge of the bottom surface, while B9 is at the bottom surface’s center. The section A measuring point is the same as section B.
Figure 2.
Layout of section B.
Measuring method and process: First, we select the relatively flat position of the steel box girder measuring point, arrange to measure point after determining the position, circle the measuring point with a marking pen, and circle each measuring point with a circular radius of 5 cm. We determine the measuring point by a circle. During the measurement, we record the temperature of the measuring point every 1 H or so and record the measuring time of each measuring point, as well as the atmospheric temperature and wind speed from 7:00 to 21:00. 2.2
Measurement result
This part mainly introduces the measurement results of atmospheric temperature and measuring point temperature. 2.2.1 Atmospheric temperature As shown in Figure 3, we can see that the atmospheric temperature rises from 7:00. To 14:00. But there is a temperature fluctuation during the period. The maximum temperature is 22 C
Figure 3.
Scatter plot of atmospheric temperature measurement for November 13–14.
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around 14:00. Then the temperature decreases slowly, with little change after 18:00. Compared with the temperature of the steel deck top surface, the maximum temperature of the air reaches 22 C around 14:00, and the maximum temperature of the steel deck top surface is higher than the maximum temperature of the air. 2.2.2 Measuring point temperature According to the analysis of statistical data results, some locations have the same change rule and temperature value, so points with obvious performance characteristics and boundary points are selected for the introduction. The following figure shows the temperature scatter plot of points A1 and B1. As shown in Figure 4, it can be seen that the temperature at the edge of the pavement plate (measuring points A1, B1) under sunshine conditions is about 12 C at 7:00, which increases quickly from 7:00 to the maximum of 24 C at 11:00–12:00 and slowly decreases at 12:00–21:00. At 21:00, the temperature is about 16 C.
Figure 4.
Temperature scatter plot of measuring points A1 and B1.
The measuring points A1 and A4 on the top surface reach the maximum at about noon. In contrast, the measuring points A9 and A10 on the bottom surface are not directly affected by solar radiation, and their maximum temperature occurring point lags the top surface. Due to the influence of wind speed, there is a certain gap between the maximum temperature on Nov. 13 and Nov. 14. Meanwhile, the atmospheric temperature also shows a certain nonlinear change, which is more sensitive than that of the bridge. The temperature data of sections A and B are almost the same, i.e., the bridge temperature is the same in the longitudinal direction of the bridge. The maximum temperature difference on the same day is not more than 1.5 C. The measured results are in good agreement. Under sunlight, the longitudinal temperature change of the bridge structure is small. The maximum temperature of the top surface is 27 C, the maximum temperature of the bottom surface is 20 C, the temperature of measuring point A4 is higher than A1 by about 2 C, and the maximum temperature is reached at about noon. The maximum temperature of the bottom surface is reached at 15:00. The maximum temperature change of the top surface in a single day is within 16 C, while the maximum temperature change of the bottom surface is within 7 C. The temperature change of the top surface is greater than that of the bottom surface, and the maximum temperature value appears faster than that of the bottom surface.
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3 SUNSHINE TEMPERATURE FIELD SIMULATION OF STEEL BOX GIRDER Due to the limited location of measuring points, the temperature field distribution of the steel box girder needs to be fully reflected. Therefore, a temperature field model of steel box girder under sunlight temperature is established using finite element theory and compared with the measured results. 3.1
Structural model parameters
The temperature field measurement of the steel box girder is mainly carried out in the side span section of the Peljesac Bridge. For this reason, this section is selected for finite element temperature field simulation analysis. The steel box girder is an orthotropic structure with a single box and three chambers. The width of the bridge deck is 22.5 m. The height is 4.5 m, the width of the bottom plate is 8.1 m, and the two-way transverse slope is 2.5%. The structural design parameters are shown in Table 1. Table 1. Size diagram table. The first column gives the names of the bridge components. The second column gives the thickness of the component. Position
U rib
Bottom plate
Top plate
Curb plate
Diaphragm plate
Thickness
8 mm
16 mm
16 mm
14 mm
20 mm
Because the longitudinal temperature of the bridge under the action of sunlight is almost the same, the model structure is simplified, and the steel box girder model of the local beam segment is established for analysis. First, the temperature changes in the longitudinal direction are almost the same under the environmental temperature, such as sunshine conditions. To a certain extent, the temperature field simulation of local beam segments under sunshine conditions can reflect the temperature field distribution law of steel box girder under sunshine conditions. Secondly, the thermal conduction calculation of the temperature field does not need to set mechanical boundary conditions but only thermal boundary conditions. The bridge structure span of 16 m can be selected to simulate the heat transfer process of the bridge structure under the action of environmental factors such as sunshine. 3.2
Physical and thermal parameters of the material
To solve the temperature field of a steel bridge deck under the effect of sunshine temperature, the finite element temperature field calculation does not involve mechanical calculation. It does not need mechanical parameters such as elastic modulus and Poisson’s ratio. It only needs to determine the material’s density, specific heat capacity, and thermal conductivity. The material type and thermal parameters are important factors affecting temperature transmission. For the steel box girder studied in this paper, S355 is used as the steel deck material, and S355 is used as the U rib material. The physical and thermal parameters are shown in Table 2. Table 2. Physical and thermal parameters of materials. This table provides the physical parameters of the component material. Material
Density (kg/m3)
Specific heat (J/kg K)
Heat conductivity (W/m C)
S355
7850
460
58.2
606
3.3
Finite element model of sunshine temperature field
According to the bridge structure, a 16 m steel box girder model is established to calculate the temperature field under sunshine using the finite element method. First, a threedimensional steel box girder model is established using ABAQUS finite element analysis software. DC3D20 solid element simulates the steel box girder, and the grid attribute is heating transfer. When calculating the temperature field, this paper uses the ABAQUS heat transfer analysis step function, selects the transient heat transfer step, and sets the analysis step time to 48000 s (13 hours), and the analysis step length to 360 s (6 minutes). The sunshine condition is equivalent to the external heat source providing heat for the steel box girder, uses the heat exchange condition to input the heat load, and uses the film layer heat dissipation condition to define the convection heat transfer effects as structural heat change caused by airflow. Radiant heat transfer is defined by surface radiation. Concerning the measured results of atmospheric temperature, the initial temperature is set to 14 C. The daily temperature change is defined by amplitude, which reflects the change rule of ambient air temperature, to simulate the temperature change of the bridge structure in the natural environment.
4 ANALYSIS OF TEMPERATURE FIELD ON STEEL BOX GIRDER UNDER SOLAR RADIATION This part mainly analyzes the variation law of the sunshine temperature field and verifies the accuracy of the temperature field model. 4.1
Temperature field distribution law in different periods
The sunshine temperature field model is obtained by solving the sunshine temperature model with ABAQUS software. The temperature distribution at 600 s, 21000 s, and 48000 s is shown in the following figure. It can be seen from Figure 5 that when the steel box girder is exposed to sunshine for 600 s (7:00), the temperature distribution is relatively uniform. The temperature is about 14 C. From the distribution, the maximum temperature is concentrated in the center. Currently, the sunshine temperature effect changes with the sunshine radiation intensity curve. At 600 s (7:00), the steel box girder receives low solar radiation intensity, and the ambient temperature is also about 14 C. Small solar radiation intensity causes small temperature changes.
Figure 5.
Sunshine temperature field distribution diagram of steel box girder at 600 s (7 points).
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It can be seen from Figure 6 that the temperature distribution of the steel bridge deck is relatively uniform when the steel box girder is exposed to sunlight for 21000 s (13:00). The temperature is about 26 C. The temperature of the steel bridge deck is significantly higher than the temperature of the bottom plate. At this time, the effect of sunlight temperature changes with the curve of sunlight radiation intensity, the solar radiation intensity gradually becomes the strongest, and the atmospheric temperature rises, causing the top plate of the steel box girder to rise faster than the bottom plate.
Figure 6.
4.2
Sunshine temperature field distribution diagram of steel box girder 21000 s (13 points).
Top plate temperature change rule
To study the temperature variation law of the steel box girder, according to the obtained temperature field structure, the top plate midpoint (A4) measurement point data is selected for temperature time history data analysis. As shown in Figure 7, the measured results on the 13th and 14th days are compared with the calculated results of the model. The maximum calculated temperature of the model is 27.2 C, which is less than 0.5 C compared with the measured data. The maximum temperature appeared at 22420 s (13:00). During the process of gradually strengthening the solar radiation intensity from 600 s (7:00) to 22420 s (13:00). Under the influence of wind speed and radiation heat transfer, the temperature showed an overall rising trend. The rising rate decreased from large to small and gradually became flat when approaching the maximum value. After 22420 s (13:00), the solar radiation intensity gradually weakened, and the airflow increased. It caused the temperature to drop rapidly, but the temperature change rate slowed when it stayed flat with the atmospheric temperature. After 48000 s (21 o’clock), the
Figure 7.
Comparison of top plate center measuring point (A4) and measured results.
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temperature remains at about 15.5 C, and the temperature change value is small, consistent with the atmospheric temperature change. The temperature change of the top plate of the steel box girder is consistent in the longitudinal direction. During the horizontal temperature rise, the temperature in the middle is the highest. At 7:00, the temperature of the top plate of the steel box girder is about 14 C. At 13:00, the maximum temperature of the top plate of the steel box girder is about 26 C. At 21:00, the temperature of the top plate of the steel box girder is about 16 C. From 7:00 to 13:00, the temperature gradually rises. From 13:00 to 21:00, the temperature gradually decreases. The temperature change of the top plate of the steel box girder is similar to that of the atmosphere. Still, because the steel box girder has good thermal conductivity and heat absorption capacity, the maximum temperature of the top plate of the steel box girder structure is higher than the atmospheric temperature. Still, the time when the maximum temperature occurs is like the atmospheric temperature change. 4.3
Temperature variation rule of the bottom plate
To study the temperature variation law of steel box girder, the midpoint (A9) measuring point data are selected for temperature time history data analysis according to the obtained temperature field structure. The measuring point data of the midpoint (A9) of the bottom plate is shown in Figure 8. The temperature at the center of the bottom plate changes slightly from 7:00 to 8:00 and then gradually increases with the increase of solar radiation intensity and the change of atmospheric temperature, reaching the maximum of 19 C at 14:00, and then slowly decreases to 15 C and tends to remain unchanged.
Figure 8. Comparison of the central measuring point (A9) of the bottom plate of the sunshine temperature model with measured results.
The model calculation curve generally agrees with the measured data, and the temperature change at night is small. Under the effect of solar radiation, the top plate temperature rises by 14 C, and the bottom plate temperature rises by 5 C. The bottom plate is in a shadow state, and its temperature change is relatively consistent with the atmospheric temperature change. In contrast, the top plate’s temperature changes greatly because it has been exposed to sunlight for a long time. At night, the temperature of the top and bottom plates is roughly the same. In the daytime, the temperature of the top plate rises rapidly with the change of the
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solar incidence angle from 7:00 (i.e., 0 s). The temperature reaches the maximum at 14:00, while the temperature of the bottom plate fluctuates due to the solar radiation scattering and the temperature rise of the surrounding atmosphere. The temperature of the bottom plate reaches the maximum of 19 C at about 15:00. 5 CONCLUSIONS The temperature of the characteristic points of the steel box girder of the Peljesac Bridge in the marine environment is measured on the spot, which provides a practical basis for model verification. The temperature calculation model of the steel box girder is established, and the temperature field of the steel box girder is calculated. The correctness of the calculation model is verified by comparing the temperature variation of key points. The temperature calculation supports the alignment control of the installation of Peljesac Bridge segments. It provides a reference for the temperature prediction and temperature effect calculation of bridges in the same type of marine environment. REFERENCES Cai, J., Xu, Y., Feng, J. & Zhang, J. (2012), “Effects of Temperature Variations on the In-Plane Stability of Steel Arch Bridges”, Journal of Bridge Engineering, 17(2), 232–240. 10.1061/(ASCE)BE.19435592.0000228. Choi, S., Cha, S. W., Oh, B. H. & Kim, I. H. (2011), “Thermo-hygro-mechanical Behavior of Early-age Concrete Deck in the Composite Bridge Under Environmental Loadings. Part 1: Temperature and Relative Humidity”, Materials and Structures, 44(7), 1325–1346. 10.1617/s11527-011-9723-8. Ding, Y., Zhou, G., Li, A. & Wang, G. (2012), “Thermal Field Characteristic Analysis of Steel Box Girder Based on Long-term Measurement Data”, International Journal of Steel Structures, 12(2), 219–232. 10.1007/s13296-012-2006-9. Erik, G., Oskar, L. I., Miklós, M. & Mario, P. (2018), “Validation of Temperature Simulations in a Portal Frame Bridge”, Structures, 15, 341–348. 10.1016/j.istruc.2018.05.006. Fan, J., Liu, Y. & Liu, C. (2021), “Experiment Study and Refined Modeling of Temperature Field of Steelconcrete Composite Beam Bridges”, Engineering Structures, 240, 112350. 10.1016/j.engstruct.2021.112350. Lin, J., Briseghella, B., Xue, J., Tabatabai, H., Huang, F. & Chen, B. (2020), “Temperature Monitoring and Response of Deck-Extension Side-by-Side Box Girder Bridges”, Journal of Performance of Constructed Facilities, 34(2). 10.1061/(ASCE)CF.1943-5509.0001431. Niu, Y., Wang, Y. E. & Tang, Y. (2020), “Analysis of Temperature-induced Deformation and Stress Distribution of Long-span Concrete Truss Combination Arch Bridge Based on Bridge Health Monitoring Data and Finite Element Simulation”, International Journal of Distributed Sensor Networks, 16(10). 10.1177/1550147720965036. Rodriguez, L. E., Barr, P. J. & Halling, M. W. (2014), “Temperature Effects on a Box-Girder IntegralAbutment Bridge”, Journal of Performance of Constructed Facilities, 28 (3), 583–591. 10.1061/(ASCE) CF.1943-5509.0000423. Xu, Y. L., Chen, B., Ng, C. L., Wong, K. Y. & Chan, W. Y. (2009), “Monitoring Temperature Effect on a Long Suspension Bridge”, Structural Control and Health Monitoring, 17 (6), 632–652. 10.1002/stc.335. Zhao, H., Ding, Y., Nagarajaiah, S. & Li, A. (2019), “Behavior Analysis and Early Warning of Girder Deflections of a Steel-Truss Arch Railway Bridge under the Effects of Temperature and Trains: Case Study”, Journal of Bridge Engineering, 24 (1). 10.1061/(ASCE)BE.1943-5592.0001352. Zhou, L., Chen, L., Xia, Y. & Koo, K. Y. (2020), “Temperature-induced Structural Static Responses of a Long-span Steel Box Girder Suspension Bridge”, Journal of Zhejiang University. A. Science, 21(7), 580– 592. 10.1631/jzus.A1900696. Zhu, Q., Wang, H., Xu, Z., Spencer, B. F., Mao, J. & Gong, Z. (2021), “Mapping Temperature Contours for a Long-span Steel Truss Arch Bridge Based on Field Monitoring Data”, Journal of Civil Structural Health Monitoring, 11(3), 725–743. 10.1007/s13349-021-00507-6.
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Comparison and analysis of three common ground treatment methods for the treatment effect of airport thick fill foundation Bin Yan*, Jie Ma & Yecheng Li CAAC Central Southern Airport Design & Research Institute (Guangzhou) Co., Ltd., Guangzhou, China
ABSTRACT: To compare and discuss the treatment effects of the “traditional dynamic compaction method”, “column hammer dynamic compaction method,” and “DDC composite foundation” commonly used in the thick fill foundation treatment of airport engineering, this paper takes the thick fill foundation of Shiyan Wudang Mountain Airport as the foundation, and takes the PLAXIS finite element numerical simulation as the means to carry out numerical simulation analysis of three foundation treatment methods. Through analysis, it is found that the critical filling thickness of the DDC composite foundation is the largest, followed by column hammer dynamic compaction, and the traditional dynamic compaction is the smallest. In the range of foundation treatment depth and the range of foundation treatment influence depth, the post-construction settlement increases linearly with the thickness of the filled foundation, and the growth slope is the same under the three types of foundation treatment. When the thickness of pile foundation is small, the effect of the three kinds of foundation treatment tends to be similar.
1 PREFACE With the continuous development of China’s civil aviation transportation industry, civil airport reconstruction and expansion projects are increasing. For some airport expansion projects with complex original landforms, it is necessary to treat the original thick-fill foundation. In this case, selecting an economical and efficient foundation treatment method according to local conditions has become the key to the effect and cost control of the foundation treatment for the reconstruction and expansion of such airports. Therefore, the industry has conducted some research on such issues. Yao et al. (2016) found that DDC piles have a good compaction effect on collapsible loess by in-situ testing the foundation treatment of an airport in the Loess Plateau. Through the analysis of the foundation treatment effect of Chongqing Wulong Xiannvshan Airport, Chen et al. (2019) found that the DDC pile has the advantages of low cost and good treatment effect. Niu et al. (2017) found that the DDC pile and column hammer dynamic compaction method has a good treatment effect in constructing mountain airports. Hu et al. (2022) found that a DDC pile can significantly improve the original foundation soil’s physical and mechanical indexes, improving the foundation’s bearing capacity, and accelerating the foundation soil’s consolidation and settlement. It has good applicability for improving the treatment of high-fill slopes of deep soft soil foundations. Huang et al. (2018) found that under the same area replacement rate, the cost of a DDC pile is much lower than that of immersed gravel pile. This paper takes the thick fill foundation with different fill thicknesses under the maximum fill height (10 m) of Shiyan Wudang Mountain Airport as the research object and, *Corresponding Author: [email protected] DOI: 10.1201/9781003425823-77
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through the PLAXIS numerical simulation, compares and analyzes the foundation treatment effects of “traditional dynamic compaction”, “column hammer dynamic compaction” and “DDC composite foundation”, which has certain reference value for future engineering design.
2 THEORETICAL ANALYSIS The dynamic compaction method uses powerful tamping energy to give an impact force to the foundation and generate a shock wave in the foundation. Under the impact force, the rammer will punch the upper soil, destroy the soil structure, form a tamping pit, and dynamically compress the surrounding soil. It repeatedly lifts the new column hammer and ordinary rammer to a high place to make them fall freely. It gives impact and vibration energy to the foundation, forms pits, fills the hole with topsoil, and compacts it layer by layer to form a small and large undisturbed soil compaction reinforcement. It forms a composite foundation treatment method with the laterally compacted soil between piles. It repeatedly lifts the new column hammer and ordinary rammer to a high place to make them fall freely. It gives impact and vibration energy to the foundation, forms pits, fills the hole with topsoil, and compacts it layer by layer to form a small and large undisturbed soil compaction reinforcement. It forms a composite foundation treatment method with the laterally compacted soil between piles. The mechanism of these two foundation treatment methods is similar, both of which can improve the structure and properties of the original soil by tamping. The most intuitive difference is the influence depth of foundation treatment corresponding to different tamping energy. According to the empirical values proposed by the relevant industry codes, the differences between the two foundation treatment methods in the depth of influence are shown in Table 1 and Table 2. Table 1.
Effective reinforcement depth of dynamic compaction (m).
Single-rammer energy (kN•m)
Gravel soil, sand, and other coarse-grained soil
Silt, cohesive soil, collapsible loess, and other fine-grained soil
1000 2000 3000 4000 5000 6000 8000
5.06.0 6.07.0 7.08.0 8.09.0 9.09.5 9.510.0 10.010.5
4.05.0 5.06.0 6.07.0 7.08.0 8.08.5 8.59.0 9.09.5
Table 2. Effective reinforcement depth of the new type of column hammer ultra-deep compaction (m). Single-rammer energy (kN•m)
Gravel soil, sand, and other coarse-grained soil
Silt, cohesive soil, collapsible loess, and other fine-grained soil
900 1000 1100 1200 1300 1400 1500
8.0–9.0 9.0–10.0 10.0–11.0 11.0–12.0 12.0–13.0 13.0–14.0 14.0–15.0
7.0–8.0 8.0–9.0 9.0–10.0 10.0–11.0 11.0–12.0 12.0–13.0 13.0–14.0
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Post-construction settlement is calculated theoretically. The layered summation method recommended in the Code for Design of Building Foundation is adopted, and the calculation formula is as follows (Shen 2020): Pi Hi Ei X S¼ mi Si Si ¼
(1) (2)
where S is the total settlement; Pi is the average value of the additional stress on the top and bottom of the i-layer soil, mainly composed of the self-weight load of the fill, the self-weight load of the pavement structural layer, and the aircraft load. Since the self-weight load of the fill and the self-weight load of the pavement structural layer have a large area of action, the attenuation of the additional stress is not considered; Hi is the thickness of layer i of soil, generally with additional stress = 0.2 . The depth of the self-weight stress of the soil layer is taken as the calculation depth. Esi is the compression modulus of layer i, and m is the empirical coefficient of settlement calculation specified in the Code for Design of Building Foundation, which is determined by the equivalent value of the compression modulus. As it is speculated that the groundwater level in the project site is high, the single-side drainage model is considered when calculating the consolidation settlement. The degree of consolidation is calculated as follows (Meng 2020): u¼1
8 p2 Tv e 4 p2
Tv ¼
Cv t H2
(3) (4)
where Cv is the vertical consolidation coefficient; H is the thickness of the soil layer. The calculation formula for post-construction settlement of original foundation is as follows: S0 ¼ ðUn Um ÞS
(5)
where S0 is a post-construction settlement; Um is the corresponding consolidation value (%) after construction; Un is the degree of consolidation corresponding to the design service life (%), which is taken as 1 for this project. The soil mass structure can be improved by tamping, increasing the drainage path, and making the soil denser. It can effectively increase the compression modulus and consolidation coefficient, thus improving the consolidation degree of soil mass and reducing the postconstruction settlement. According to the above formula, it can be inferred that both traditional dynamic compaction and column hammer dynamic compaction have good effects on the control of the post-construction settlement of the foundation. However, because the impact depth of column hammer dynamic compaction is significantly higher than that of traditional dynamic compaction, the treatment effect of the column hammer dynamic compaction method is better than that of the traditional dynamic compaction method. DDC composite foundation method is first to impact the hole and then carry out dynamic compaction and impact extrusion on the filler in the pile hole so that the filler is compacted and compressed towards the bottom and around the hole to form a pile and at the same time. It impacts the soil layer to eliminate various adverse characteristics of the foundation soil and improve the bearing capacity of the foundation. The principle of this method is like that of compaction pile composite foundations. At the same time, because the DDC pile material is often coarse gravel, slag, and construction waste, the DDC pile also has the function of the drainage channel in the composite foundation. In this case, the degree of consolidation is
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calculated using the ideal shaft foundation consolidation formula. The specific formula is as follows (Yang 2015): 2 8 8CFhn d 2 þp4HC2v t e U rz ¼ 1 2 e (6) p n2 3n2 1 lnðnÞ Fn ¼ 2 (7) n 1 4n2 (8) n ¼ de =dw where U rz is the average degree of consolidation; Ch is the transverse consolidation coefficient; Cv is the longitudinal consolidation coefficient; N is the caliper ratio; de is the diameter of the influence range of the shaft, and dw is the diameter of the shaft; H is the thickness of soil layer; t is the consolidation time. On the one hand, in the case of a certain pile arrangement scheme, compared with dynamic compaction and traditional dynamic compaction, DDC composite foundation method increases the horizontal drainage path, which is more conducive to increasing consolidation degree. On the other hand, the compression modulus of DDC pile material and the compression modulus of the soil between piles after compaction are far greater than that of the undisturbed thick fill foundation. According to Formula (1), it can be inferred that the DDC composite foundation method can effectively reduce the total settlement. 3 PARAMETER SETTING AND MODEL ESTABLISHMENT The object of this study is a thick layer of fill foundation of Shiyan Wudang Mountain Airport. The maximum fill height above the original ground is 10 m. According to the previous geological exploration data, the thickness and properties of each layer are shown in Table 3. Table 3.
Stratum statistics.
Geotechnical and material Average thickness (m) Constitutive relation Drainage type Natural gravity (kN/m3) Saturation gravity (kN/m3) Compression modulus (MPa) Poisson’s ratio Cohesion (kPa) internal friction angle ( ) Permeability coefficient (m/s)
Filling body
Thicklayer fill
Strongly weath- Moderately ered rock weathered rock
Slightly weathered rock
10
Variable
17
70
Drainage 23 23 30 0.2 6 30 2 10-6
Drainage 18 18.76 5 0.2 5 19 2 10-6
22 Mohr-Coulomb Drainage Undrained 25 26 26.9 26 44 Incompressible 0.15 0.1 72 120 25 22 2 10-6 0
Undrained 26 26 Incompressible 0.1 120 22 0
Based on the indoor test and in-situ test results in the process of foundation treatment related to the preliminary project, the influence depth of the foundation treatment under the three methods and the average value of the foundation property index within the influence depth range are summarized, as shown in Table 4. The step-by-step construction method is adopted in the modeling process to restore the construction process. The first step is to establish the model below the original ground elevation. The second step is setting the filling, which lasts 10 days. The third step is to set a one-year settlement period for the filled model, apply the external load (50 kPa), and finally we set a one-year settlement period to obtain the post-construction settlement. The specific typical model is shown in Figure 1. 614
Table 4. The influence depth of foundation treatment under three methods and the nature of foundation within the influence depth range. Foundation treatment method
Traditional dynamic compaction
Column hammer dynamic compaction
DDC composite foundation method
Maximum influence depth (m) Geotechnical and material
10
15
30
Constitutive relation Drainage type Natural gravity (kN/m3) Saturation gravity (kN/m3) Compression modulus (MPa) Poisson’s ratio Cohesion (kPa) internal friction angle ( ) Permeability coefficient (m/s) Consolidation time (d)
Foundation soil within the Foundation soil within influence range of traditional the influence range of dynamic compaction column hammer dynamic compaction Mohr-Coulomb Drainage Drainage 18 18 18.76 18.76 25 28 0.2 0.2 6 6 35 38 2 10-6 2 10-6 365 365
DDC pile rubble
Soil between piles
Drainage 18 20.1 150 0.2 6 30 1 10-5 365
Drainage 18 18.76 20 0.2 10 19 2 10-6 365
Figure 1. Schematic diagram of a typical model of foundation treatment (taking 30 m filling thickness as an example).
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Figure 1.
(Continued)
4 SIMULATION RESULTS To compare the treatment effects of the three foundation treatment methods in this study on the different thicknesses of the fill foundation, this study summarizes the post-construction settlement of the three foundation treatment methods within the range of 1040 m thickness of the fill foundation and compares them with the specification limit value and draws the “post-construction settlement - fill thickness” curve as shown in Figure 2. As can be seen from Figure 2, with the increase of the stacking thickness, the postconstruction settlement curve of the three foundation treatment methods is approximately a piecewise function curve composed of two linear functions with large slope differences, and the piecewise inflection points are located at a depth of influence of the three foundation treatment methods. These three foundation treatment methods have an obvious improvement effect on the compressibility of foundation soil. To summarize the rules more accurately, this study carried out linear fitting for the settlement curve of each foundation treatment method. The fitting results are shown in Table 5.
Figure 2. methods.
“Post-construction settlement - fill thickness” curve under three foundation treatment
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Table 5. Linear fitting results of “post-construction settlement - fill thickness” curve under three foundation treatment methods. Within the scope of influence Foundation treatment method
Fitting formula
Fitting coefficient
Outside the scope of influence Fitting formula
Fitting coefficient
y = -0.0503x + 0.2926 R2 = 1 Traditional dynamic y = -0.011x - 0.1 R2 = 1 compaction Column hammer dynamic y = -0.0096x - 0.1006 R2 = 0.9995 y = -0.0503x + 0.5096 R2 = 1 compaction DDC composite y = -0.0043x - 0.1075 R2 = 0.9799 y = -0.0499x + 1.2674 R2 = 0.9998 foundation
It can be seen from Table 5 that under the three foundation treatment methods, the linear fitting degree of each subsection of the “post-construction settlement - fill thickness” curve is high, and the linear fitting coefficient is above 0.9. At the same time, it can be seen that for the three different foundation treatment methods, when the thick fill foundation is within the influence depth of the foundation treatment, the DDC composite foundation has the highest degree of improvement in the compressibility of the foundation, followed by the column hammer dynamic compaction, and the traditional dynamic compaction is the smallest. However, outside the influence range of foundation treatment, the slope of the “post-construction settlement - fill thickness” curve under the three foundation treatment methods is almost the same. When the fill thickness exceeds the influence depth of the foundation treatment, the settlement of the fill foundation outside the influence depth will play a leading role in the post-construction settlement. In this case, the growth rate of the post-construction settlement with the fill thickness under different foundation treatment methods is the same. At the same time, according to the comparison between the post-construction settlement and the specification limit value, the critical thickness of the thick fill foundation under the maximum fill height (10 m) in Shiyan Wudang Mountain Airport under various foundation treatment methods can be summarized, as shown in Table 6. Table 6.
Critical fill thickness under different foundation treatment methods.
Foundation treatment method Regional division
The critical thickness of filling the foundation in the runway influences the area
Critical thickness of landfill foundation in taxiway influence area and apron influence area
Traditional dynamic compaction Column hammer dynamic compaction DDC composite foundation
11.78 m 16.10 m 31.41 m
13.77 m 18.08 m 33.41 m
It can be seen from the above table that the critical filling thickness of the thick fill foundation under the maximum filling height (10 m) in Shiyan Wudang Mountain Airport under the three foundation treatment methods is the largest for the DDC composite foundation, followed by the column hammer dynamic compaction. The traditional dynamic compaction is the smallest. To compare the treatment effects of the three foundation treatment methods more intuitively in this study, the numerical model calculation results of the three different foundation treatment methods under the three landfill thicknesses of 5 m, 10 m, and 15 m are selected
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Table 7. Comparison of settlement deformation nephogram results of different foundation treatment methods.
for comparison. The settlement deformation nephogram results simulated by PLAXIS software are summarized in Table 7. It can be seen from the above table that when the thickness of the fill foundation is within the range of the depth affected by the foundation treatment, the treatment effects of the three foundation treatment methods are similar. However, once the thickness of the piled foundation treatment exceeds the range of the depth affected by the foundation treatment, the post-construction settlement of the foundation will increase rapidly. From the perspective of investment cost, the simulated DDC composite foundation has a pile diameter of 2.4 m, a pile spacing of 4 m, and is arranged in a regular triangle. When the average pile length is 10 m, the average cost per square meter is about 90 yuan. When the average pile length is 15 m, the cost per square meter is about 135 yuan. Compared with the cost per square meter of traditional dynamic compaction, which is 40 yuan, the cost per square meter of column hammer dynamic compaction is 100 yuan (Wang 2012; Zhao 2011), and the construction cost of DDC composite foundation is relatively high. This method applies to areas with large stacking thicknesses and high settlement requirements. When the stacking foundation thickness is small and the foundation treatment effect is guaranteed, the traditional dynamic compaction and column hammer dynamic compaction foundation treatment methods can save costs. 5 CONCLUSIONS Through the simulation calculation and comparative analysis of the numerical models of the thick fill foundation under the maximum fill height (10 m) in Shiyan Wudang Mountain Airport under three foundation treatment methods, this paper has formed some conclusions that can have reference significance for future projects as follows: (1) Regarding the critical filling thickness of the foundation, DDC composite foundation is the largest, followed by column hammer dynamic compaction, and traditional dynamic
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compaction is the smallest. According to the specific calculation results, it can be inferred that the DDC composite foundation method is more applicable when the thickness of the piled foundation exceeds 16 m, and the column hammer dynamic compaction method is more applicable when the thickness of the piled foundation exceeds 11 m but does not exceed 16 m. When the filling thickness is less than 11 m, the traditional dynamic compaction method can make the foundation meet the specifications and engineering requirements. (2) For the three different foundation treatment methods, with the increase of the filling thickness, the “post-construction settlement - filling thickness” curve under the three foundation treatment methods is approximate to a piecewise function curve composed of two linear functions with a large slope difference. The inflection points of the curve are the corresponding influence depth of each foundation treatment method. It can be inferred from the slope of the curve within the range of the depth of foundation treatment that the deformation modulus of the foundation within the range of the depth affected by the foundation treatment is the largest in the DDC composite foundation, followed by the column hammer dynamic compaction method and the traditional dynamic compaction method. It can be inferred from the slope of the curve outside the depth range of the foundation treatment that when the thickness of the fill exceeds the influence depth of the foundation treatment, the settlement of the fill foundation outside the influence depth range will play a leading role in the post-construction settlement. (3) Through the comparison of the settlement deformation program of three foundation treatment methods under the same fill foundation thickness, it can be intuitively seen that under the condition of small fill foundation thickness, the difference of foundation treatment effect between DDC composite foundation, column hammer dynamic compaction method, and traditional dynamic compaction method is very small. This difference tends to be obvious with the increase in foundation fill thickness. By comparing the cost of the three foundation treatment methods, it can be concluded that the DDC composite foundation is less economical when the thickness of the fill foundation is smaller (less than 15 m).
REFERENCES Chen Jian, Li Zhicheng, Rao Jianqiang. Research on the Application of the Deep Dynamic Compaction Method (DDC Method) in the Foundation Treatment of the Plateau Airport — Taking the Pilot Section Project of Chongqing Wulong Xiannvshan Airport as an Example [J]. Chongqing Architecture, 2019, 18 (07): 46–50. Hu Gang, Wang Yuxuan, Ma Jianrui. Application of DDC Pile in Foundation Treatment of a High-fill Airport in Southwest China [J]. Journal of Civil Aviation, 2022, 6 (03): 38–41. Huang Jin, Qiu Cunjia, Wang Rui, Yang Biao, Wang Shuang, Zhou Chunfeng Comparison of the Application of DDC Pile and Vibro-immersed Gravel Pile in the Foundation Treatment of the Airport in the Southwest Mountain Area [C]//. Proceedings of the 2018 National Engineering Geology Academic Annual Conference, 2018: 217–222. Meng Jing. Research on Reinforcement Technology of Collapsible Foundation in Construction Engineering [J]. Building Materials and Decoration, 2020 (20): 36+38. Niu Wenli. Application of Column Hammer Dynamic Compaction and DDC Pile in Airport High Fill Foundation Treatment [J]. Shanxi Architecture, 2017, 43 (31): 55–56. DOI: 10.13719/j.cnki.cn14-1279/tu.2017.31.029. Shen Binbin. Research on the Change Law of Embankment Under Dynamic Compaction [J]. Fujian Transportation Science and Technology, 2020 (04): 39–45+56. Wang Xuelang Collapsible Deformation Mechanism, Foundation Treatment and Experimental Study of Large Thickness Collapsible Loess [D]. Lanzhou University of Technology, 2012. Yang Peng Research on Bearing and Settlement Characteristics of Loess Foundation Strengthened with DDC Piles for High-speed Railway [D]. Lanzhou Jiaotong University, 2015. Yao Xuegui, Yao Zhihua, Zhou Lixin, Lei Yuanfeng. Experimental Study on Ground Treatment of an Airport Selfweight Collapsible Loess Site [J]. Building Technology, 2016, 47 (03): 213–217. DOI: 10.13731/j.issn.10004726.2016.03.005. Zhao Shibin Analysis and Research on Settlement and Deformation of Dynamic Compaction Foundation and Pile Foundation of Large Storage Tank [D]. Ocean University of China, 2011.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Online optimization for enhanced Tunneling Boring Machine (TBM) attitude control during the tunneling process Kunyu Wang & Limao Zhang* School of Civil and Hydraulic Engineering, Huazhong University of Science and Technology, Hongshan District, Wuhan, Hubei, China
ABSTRACT: Tunneling boring machine (TBM) is a widely used approach to tunnel construction worldwide and stable attitude control is a key factor for its high-quality construction. To solve the problem of inconsistency between the actual attitude and the target attitude during the TBM tunneling process, this study proposes an online optimization method based on the genetic algorithm (GA), which, combined with the TBM virtual model, can continuously optimize the PID parameters during the tunneling process to achieve the actual TBM attitude stable around the target value. First, the input and output of the proportional-integral-derivative (PID) controller are recorded to construct an updatable transfer function for simulating the thrust cylinder groups. On this basis, a virtual model of the TBM is built by MATLAB Simulink to visualize the tunneling process and interact with the TBM. Afterward, the online optimization method is deployed on the virtual model to obtain the optimal combination of the PID parameters which are finally assigned to the PID module on the thrust cylinder group. The online-GA method is applied to a TBM experimental platform to verify its superiority and the optimization results show that the proposed method reduces the over-tunning by an average of 91.8% and increases the fitness by an average of 27.6%, compared to the state-of-art offline optimization algorithms, achieving reliable performance in TBM construction.
1 INTRODUCTION Tunneling boring machine (TBM) is one of the most important means of underground tunneling construction, with features such as safety and efficiency. High-quality TBM engineering construction requires precise control of the TBM attitude, while the current control of the TBM attitude mainly relies on the operator’s experience, which is unreliable in the complex and changing underground environment. Therefore, a method that can continuously optimize the TBM attitude is urgently needed for engineering applications. In the past, the research on TBM attitude pairs mainly had theoretical analysis, but it was gradually replaced by machine learning methods (He et al. 2020) due to its difficulty in handling massive TBM data. Although better results can be obtained by using these methods, the problem that insufficient nonlinear expression capability of traditional machine learning methods is gradually exposed as the TBM dataset becomes increasingly complex. After that, scholars began to study the application of deep learning methods (Zhou et al. 2019) in TBM attitude analysis and prediction. As the research progresses, the characteristics of the TBM dataset as a kind of time-series data are explored and utilized continuously.
*Corresponding Author: [email protected]
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DOI: 10.1201/9781003425823-78
However, most of the current research on TBM attitude is limited to data-driven theoretical studies, ignoring how the TBM itself can accurately follow the ideal TBM attitude in theory. Whether a TBM can accurately follow the designed trajectory is dependent upon the control accuracy of the thrust system. While conventional TBMs are driven directly by cylinders to drive the cutter plate for tunneling, a PID controller is used to control the TBM attitude in this study. PID controller has been widely used in various systems due to the advantages of simple structure and high adaptability. Typical applications of the PID controller are autonomous driving (Soe Paing et al. 2021), electrical system control (Gecer et al. 2021), and hydraulic system control. To make the PID controller with a better control effect, optimization algorithms are introduced to optimize the PID control parameters. The commonly used optimization algorithms in the field of TBM are mainly particle swarm optimization (PSO) (Mustafa & Hashim 2020), genetic algorithm (GA) (Ren et al. 2022), and their improvement algorithms. However, these algorithms, when applied to TBM attitude control, often require a large amount of tunneling data to construct an agent model, which is not realistic in actual TBM engineering anymore. In TBM engineering applications, it is often necessary to obtain the optimal PID control parameters immediately after the start of TBM construction and to continuously update the optimal PID control parameters as the TBM continuously tunneling. The main technical problems of this research are: (a) how to simulate the TBM thrust system; (b) how to obtain the optimal parameters of the PID controller with limited TBM tunneling data; and (c) how to update the PID controller under complex engineering conditions. Based on the above three points, an online optimization method of PID controller parameters is proposed based on the GA, and an experiment is designed to verify the effectiveness of the method. The innovations of the method are: (a) a virtual model of the TBM thrust system is constructed, (b) an online PID controller parameter optimization method is proposed for TBM construction applications, and (c) a field experiment method is designed to verify the effectiveness of online optimization. The subsequent part of this paper is structured as follows: Section 2 briefly reviews the research on the application of different types of optimization algorithms in engineering; Section 3 describes the flow of the proposed method in this study in detail; Section 4 shows the test results of the proposed method on an experimental platform; Section 5 discusses the advantages of online optimization methods compared with other advanced optimization methods; Section 6 summarizes the conclusions obtained in this research.
2 RELATED STUDIES Optimization algorithms can be divided into two main categories, which are offline optimization methods and online optimization methods. The offline optimization method has already been applied in the field of TBM construction, which is mainly based on the data collected from existing projects to construct an agent model and seek the global optimal combination of parameters according to the model. The application of offline optimization algorithms in the field of TBM is mainly to optimize the hyperparameters of prediction models. For example, the performance of the LSTM model in predicting the penetration rate of TBMs is improved by using the Gray Wolf model (Mahmoodzadeh et al. 2022). A hybrid approach based on fuzzy logic, PSO, and harmony search is proposed to predict the TBM penetration rate (Afradi et al. 2022). Besides, ant colony optimization, bee colony optimization, and PSO are compared in the prediction of the TBM penetration rate (Afradi et al. 2020). The online optimization method, on the other hand, is still less used, as the optimal combination of parameters is continuously updated during the TBM construction process according to the changes in the construction environment. Although the offline optimization 621
method has achieved satisfactory theoretical results, the generalization of the application is insufficient. The underground environment during TBM construction is often complex and variable, and the offline algorithm has difficulty coping with this situation without prior knowledge. Currently, there are few engineering applications of online optimization methods in the field of TBMs, but many in other fields (Garber 2021; Nazari et al. 2021; Pan et al. 2022). Therefore, it is very urgent and significant to study the application of online optimization methods in the field of TBM attitude control. 3 METHODOLOGY The online GA method for enhanced TBM attitude control is composed of virtual model construction and online PID parameters optimization. The virtual model of the TBM thrust system provides visual interaction capability and physical parameters calculation for the TBM experiment platform, while the optimization algorithm improvement is based on GA and PID to propose an online optimization method for PID parameters and TBM attitude. 3.1
Virtual model construction of TBM thrust system
This study constructs the physical simulation model of the TBM thrust system based on MATLAB Simulink, which is used to update the physical state of the TBM thrust system. MATLAB Simulink is a graphical physical system simulation platform, which can well simulate the physical quantity transfer of the thrust cylinder system. The structure of the virtual model is shown in Figure 1. The input signal is coupled by 4 step signals (as shown in Figure 2), which fully simulates the TBM attitude from horizontally straight to “raised or low head” tunneling, restoring
Figure 1.
Simulink model of TBM thrust system.
Figure 2.
Target attitude of TBM in vertical.
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horizontally straight after “raised or low head” tunneling, and keeping stable in three attitudes. the PID controller is responsible for the process of controlling the cylinder pressure difference through the cylinder stroke difference, and the control result is input into the transfer function to get the thrust cylinder deformation difference after each control. 3.2
Online genetic optimization for PID parameters
The GA is a global optimization algorithm that simulates population crossover, variation, evolution, and natural selection (Li & Shi 2022). The core of PID controller optimization is to obtain the optimal combination of three control parameters (KP, KI, and KD), and a PID controller is considered as an individual in a population, where PID parameters are the three genes of the individual. Population crossover is reflected in the interaction of parameters of different PID controllers, while population mutation is reflected by randomly floating the PID parameters in a certain proportion during the updating process. The final natural selection of the population is screened by the fitness function, which is shown in Equation (1). The real meaning of the fitness function adopted in this study is the area enclosed by the control curve and the target curve. On the one hand, the fitness can be a good measure of the amount of overshoot of the PID controller, and on the other hand, it can evaluate the degree of floating of the PID controller after stabilization. ð the fitness ¼ errorðtÞ2 dt (1) The traditional offline optimization algorithm requires a large amount of data to construct the agent model to calculate the global optimal solution, which is difficult to apply in the construction of TBM engineering. In actual TBM construction, it is necessary to seek the optimal combination of PID parameters in the present moment based on limited tunneling data and constantly update the optimal PID parameter combination as the tunneling process proceeds and the environment changes, so this study improves the traditional GA algorithm to be an online optimization approach. The pseudocode of the online GA algorithm is shown as follows. Begin While tunneling start: Calculate the transfer function; If the transfer function changed: Randomly generate the initial PID parameter populations; Calculate the fitness of the initial individual; While fitness value changed in the last 10 generations: Select n individuals according to the fitness function; Individual crossover; Individual mutation; End while; If The current optimal individual is better than the previously optimal individual: Output the optimal individual to the experimental platform; End if; End if; End while; End. The online GA algorithm initializes the PID controller based on previous TBM experience and starts the tunneling immediately thereafter. While the tunneling process starts, the TBM experimental data is recorded at fixed time intervals and the conversion relationship between the TBM thrust cylinder pressure difference and the thrust cylinder deformation difference is 623
started to be fitted. The conversion function is then continuously updated in the Simulink model, while the PID parameters are continuously optimized in the virtual model, and the online optimization results are fed back to the TBM experiment platform until the TBM tunneling is completed. It is important to note that the TBM experiment process is not ideal; as the experiment proceeds, the noise generated during the data acquisition process will affect the online optimization process, and when the online optimization results can be maintained within the allowable accuracy, the real-time feedback can be suspended and the subsequent TBM process can continue with the current results until the error is close to the error limit.
4 CASE STUDY This study tests the optimization method on the TBM experimental platform. The validation process is mainly divided into 2 parts, namely, the verification of the performance of the virtual model and the verification of the online optimization method. 4.1
Case background
The experimental platform is built according to the real TBM with the proportion of 1:1, and two sets of PID controllers were used, one of which controlled the thrust force difference between the upper and lower cylinders according to the attitude difference of the upper and lower cylinders, while the other is responsible for controlling the force difference between the left and right cylinders. Under the initial conditions, the PID control parameters are set according to the previous engineering project. This study takes the TBM attitude control in the vertical direction as an example and the PID control effect is verified by three tunneling methods in the vertical direction: upward tunneling, downward tunneling, and straight ward tunneling. 4.2
Virtual thrust system verification
The virtual model of the TBM thrust system is built based on the MATLAB Simulink platform to transfer and update the physical quantities. The performance of the model is shown in Figure 3. Since noise is bound to exist in the real tunneling data, the fitting of the transfer function in this study is more lenient and only requires that the fitting results can reflect the data changes in the tunneling process in terms of trend. It can be seen from Figure 4 that the trend and results of the conversion function are consistent for different stages of TBM construction.
Figure 3. Performance of the virtual thrust system. (a) Stage 1; (b) Stage 2; (c) Stage 3; (d) Stage 4; (e) Stage 5.
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4.3
Result analysis
After verifying the TBM virtual model, the experiment of PID online optimization is carried out. First, the initial population of PID parameters is randomly generated based on fixed boundaries, and the upper ([30, 30, 30]) and lower ([0, 0, 0]) boundaries are determined according to the actual engineering environment. The crossover rule for the PID parameters is that an individual randomly selects 1 or 2 PID parameters to be exchanged with other individuals. After that, the mutation of the individual follows a Gaussian distribution centered on the individual itself, while the standard deviation of the Gaussian distribution decreases according to Equation (2). g sg ¼ sg1 (2) G where sg and sg1 are the standard deviation of the current generation and the last generation, respectively, and g and G are the current generation and total generation, respectively. It should be noted that after considering the engineering reality, the abort condition of the online GA method will also be that the current optimization loop stops when the optimal fitness of continuous 10 generations no longer changes. The online optimization processes of 5 stages are shown in Figure 4. The first stage is shortly after the start of the tunneling process, and the transfer function of this stage is also relatively simple, so the optimization process converges quickly and ends at the 16th generation. The optimization process of the remaining four stages is all stopped until the 50th generation.
Figure 4.
Online optimization of each stage. (a) Stage 1; (b) Stage 2; (c) Stage 3; (d) Stage 4; (e) Stage 5.
In the optimization results of the first stage, a combination of manually set PID parameters is added as a reference. Although the online optimization results are not satisfactory in the first stage, the optimization results are acceptable compared to the manually set PID parameters that can only remain stable in the initial stage. In other words, in the initial stage of the tunneling process, the optimization results with slightly larger error but stable attitude are more favorable to the TBM construction compared with the uncertainty of manually adjusted parameters. The optimization results of the subsequent four stages are all substantially improved compared to the first stage, and the detailed information is summarized in Figure 5 and Table 1. 625
Figure 5. Results of online GA for different stages. (a) Stage 1; (b) Stage 2; (c) Stage 3; (d) Stage 4; (e) Stage 5. Table 1.
Detailed information on online optimization results for each stage. Best fitness value
Stage
Manual control (105)
Online-GA
Max over-tunning value (mm) Online-GA
1 2 3 4 5
312.79 5.99 6.12 6.53 6.53
95876 730 1260 661 1997
– 23.2 37.7 11.3 1.1
* ‘-’ denotes no data available.
5 DISCUSSIONS To further verify the superiority of the proposed method, this study compares the online GA method with traditional offline methods such as GA and PSO. Figure 6 shows the optimization process of both methods, and it can be seen that the PSO method has a faster convergence speed compared to the GA method. The converged optimization results are presented in Figure 7, and the optimization results are summarized in Table 2 after calculating the optimization degree according to Equation (3). The results show that the max over-tunning value of PSO is smaller than that of GA but the objective function value is larger. The online GA method outperforms the GA and PSO methods in both metrics, where the max over-tunning improvement is up to 94.8% and 88.8%, and the fitness value improvement is 21.1% and 34.0%, respectively. IM ¼
Figure 6.
FMC FOGA 100% FMC
Optimization process of different methods. (a) GA; (b) PSO.
626
(3)
Figure 7.
Optimization results of different methods. (a) GA; (b) PSO.
Table 2.
Comparison results of different optimization methods.
Methods
Max over tunning value (mm)
IM
Best fitness value
IM
Online GA GA PSO
1.1 21.3 9.81
– 94.8% 88.8%
661 838 1001
– 21.1% 34.0%
6 CONCLUSIONS In this paper, an online GA is proposed for improving TBM attitude control. The method mainly simulates the TBM thrust system through a virtual model and then feeds back to the TBM experimental platform after optimizing the PID parameters based on the virtual model. The conclusions obtained are as follows: (1) compared with manual control, the online optimization algorithm can significantly improve the stability and accuracy of TBM attitude control; (2) compared with the offline optimization algorithm, the online GA method still has significant advantages, with an average reduction of 91.8% in max overtunning value and an average increase of 27.6% in fitness. Despite the good performance of the online GA method, improvements are still needed in future research. First, more influencing factors should be considered in the simulation of the TBM thrust system to improve the simulation accuracy. In addition, more optimization objectives should be set to further improve the overall performance of the optimization results.
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Civil Engineering and Disaster Prevention – Zende, Ran & Gao (Eds) © 2024 The Author(s), ISBN 978-1-032-54618-6
Research and application of a mountain flood forecasting and early warning system based on Xinanjiang model Hongri Zheng* ZhejiangIinstitute of Hydraulics & Estuary (Zhejiang Institute of Marine Planning and Design), Hangzhou, Zhejiang Province, China Zhejiang Provincial Key Laboratory of Hydraulic Disaster Prevention and Mitigation, Hangzhou, Zhejiang Province, China
Shun Yu* Jindong District Water Bureau, Jinhua, Zhejiang Province, China
Yong Luan* Zhejiang institute of hydraulics & estuary (Zhejiang institute of marine planning and design), Hangzhou, Zhejiang Province, China
ABSTRACT: In order to realize the fine prediction and early warning of mountain flood disasters, a mountain flood forecasting and early warning system based on the Xinanjiang model was developed. The system consists of a monitoring system, forecasting system, early warning system, monitoring center, and so on. Through the practice of mountain flood disaster prevention such as “20210816”, the system has given full play to its functions of advance prediction and accurate early warning, effectively avoiding casualties and reducing property losses.
1 INTRODUCTION Faced with the more and more serious mountain flood disasters in the world, many countries have or are developing effective mountain flood disaster monitoring and early warning system, and strive to minimize the disaster degree. The United States established the Flash Flood Guidance System (FFG) in 1969. Georgakakos (2006) applied FFG in the early warning of mountain floods, analyzed the characteristics of surface runoff and carried out real-time flood simulation calculation. Seo et al. (2013) and Reed et al. (2007) established an improved grid FFG system and DFFG method based on FFG respectively. After nearly 50 years of research and development, FFG has been well applied in many parts of the world (Looper & Vieux 2012). Canada established a flood control command system based on the GIS data analysis function (Liu & Todini 2010). In Japan, the distributed hydrological model is adopted to calculate the production and confluence, and the mountain flood disaster warning system is developed in small and medium-sized river basins (He et al. 2016). In addition, Javelle et al. (2010) increased the accuracy of mountain flood disaster warnings by considering the influence of soil moisture content on runoff production in the early stage. Martina et al. (2006) proposed a mountain flood disaster warning method based on critical rainfall. Douvinet et al. (2015) proposed a mountain flood disaster warning method based on cellular automata. Montesarchio et al. (2011) used radar monitoring data and the entropy *Corresponding Authors: [email protected], [email protected] and [email protected] DOI: 10.1201/9781003425823-79
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decision method to solve the critical rainfall optimization model, which improved the accuracy of the model. In the past decade, China has intensified its efforts to prevent and control mountain flood disasters and has initially established a mountain flood disaster prevention system suitable for our national conditions. Mountain flood disaster monitoring and early warning systems have also been established throughout the country, achieving a breakthrough from scratch and bringing significant disaster prevention and mitigation benefits into play. However, at present, mountain flood disaster prevention still has some problems, such as the low precision of early warning, the low level of monitoring and forecasting and the low degree of guarantee of operation and maintenance. In this paper, a mountain flood forecasting and early warning system based on the Xinanjiang model was developed. The system was named as the radio monitoring (forecasting) and warning system of the Xiaoshun watershed. This study made a preliminary exploration of the fine forecasting and early warning of mountain flood disasters, in order to improve the accuracy of mountain flood early warning and the level of mountain flood forecasting technology.
2 SYSTEM INTRODUCTION The radio monitoring (forecasting) and warning system of Xiaoshun watershed in Jindong District is mainly composed of a monitoring system, forecasting system, early warning system, monitoring center, and so on. 2.1
Monitoring system
A large number of automatic monitoring stations, including automatic rain measuring stations, automatic water level stations, all-factor weather stations, and video image stations, have been built in the Xiaoshun watershed of Yuandong Township. Among them, there are 13 automatic rain stations, 14 automatic water stations, 4 full-factor weather stations, and 9 video image stations. The data acquisition equipment of this system uses the wireless integrated low-power acquisition equipment, namely LoRa terminal equipment. Meanwhile, based on the wireless high bandwidth AD hoc network and wireless LoRa technology, it realizes the high-speed transmission of data and video signals, which has the characteristics of safety and control, maturity and stability, long distance, and low power consumption. It can realize real-time positioning, remote wireless control, wireless data transmission, and other functions. Eleven key villages of mountain flood disaster prevention within the Xiaoshun watershed of Yuandong Township are associated with water and rain monitoring stations. 2.2
Forecasting system
The mountain flood disaster prediction model was established by collecting rainfall, water level, meteorology, field images, and other data by automatic monitoring equipment and combining with the underlying surface features such as topography, river network characteristics and vegetation coverage in Xiaoshun watershed of Yuandong Township. Based on the model calculation, the variation trend of the water situation of Xiaoshun Creek in Yuandong Township was predicted. According to the flood rise rate, the remaining time (overflow time) of reservoir overflow and river embankment was predicted, and the probability of mountain flood disaster at each control node was dynamically analyzed. This time, the mountain flood disaster prediction model is established on the basis of Xinanjiang model of three water sources. 630
The structure of the Xinanjiang model is designed to be decentralized and is divided into four hierarchies: evapotranspiration calculation, runoff production calculation, water source calculation and confluence calculation. 2.2.1 Evapotranspiration calculation The evapotranspiration calculation of the Xinanjiang model divides the soil layer into three layers and adopts the three-layer evaporation model for calculation. The specific calculation method is as follows. WM ¼ UM þ LM þ DM
(1)
W ¼ WU þ WL þ WD
(2)
E ¼ EU þ EL þ ED
(3)
EP ¼ KC EM
(4)
where W is the total tensional water storage, WU is the upper tensional water storage, WL is the lower tensional water storage, WD is the deep tensional water storage, E is the total evapotranspiration, EU is the upper evapotranspiration, EL is the lower evapotranspiration, ED is the deep evapotranspiration, and EP is the evapotranspiration capacity. 2.2.2 Runoff production calculation After a lot of practice and experience summary, the water storage capacity curve of the basin generally adopts the form of a parabola. The line of the water storage capacity curve is as follows. B 0 f W ¼1 1 F WMM
(5) 0
where f is the area of runoff production, F is the area of the whole basin, W is the water storage at a single point in the basin, WMM is the maximum water storage at a single point in the basin, and B is the index of the distribution curve of water storage capacity-area. The following formula can be obtained according to the relationship between the water storage capacity-area distribution curve and rainfall runoff conversion. PEþA ð
R¼ A
f 0 dW ¼ F
PEþA ð "
1 1
0
W WMM
B # dW
0
(6)
A
When a portion of the basin produces runoff, PE + A